Concrete Repair, Rehabilitation and Retrofitting: 2nd International Conference on Concrete Repair, Rehabilitation and Retrofitting, ICCRRR-2, 24-26 November 2008, Cape Town, South Africa

CONCRETE REPAIR, REHABILITATION AND RETROFITTING II

PROCEEDINGS OF THE 2ND INTERNATIONAL CONFERENCE ON CONCRETE REPAIR, REHABILITATION AND RETROFITTING (ICCRRR), CAPE TOWN, SOUTH AFRICA, NOVEMBER 24–26, 2008

Concrete Repair, Rehabilitation and Retrofitting II

Editors

Mark G. Alexander Department of Civil Engineering, University of Cape Town, South Africa

Hans-Dieter Beushausen Department of Civil Engineering, University of Cape Town, South Africa

Frank Dehn Universität Leipzig/MFPA Leipzig, Germany

Pilate Moyo Department of Civil Engineering, University of Cape Town, South Africa

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2009 Taylor & Francis Group, London, UK Typeset by Vikatan Publishing Solutions (P) Ltd., Chennai, India Printed and bound in Great Britain by Cromwell Press Ltd, Towbridge, Wiltshire All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by: CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail: [email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl ISBN: 978-0-415-46850-3 (Hbk + CD-rom)

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Table of contents

Preface

XV

ICCRRR Committees

XVII

Keynote papers Design for service life: How should it be implemented in future codes J.C. Walraven

3

Importance of microstructural understanding for durable and sustainable concrete K.L. Scrivener

11

Advanced NDT methods for the assessment of concrete structures H. Wiggenhauser

19

Efficiency control of electrochemical repair techniques C. Andrade, I. Martínez, M. Castellote & P.G. de Viedma

31

External strengthening of continuous beams with CFRP L. Taerwe, L. Vasseur & S. Matthys

39

Diagnostic analysis and therapy of severely cracked bridges H.-P. Andrä & M. Maier

49

Invited papers Research in the field of repair – Actual approaches and future needs M. Raupach

61

Reflections on future needs in concrete durability research and development Y. Ballim, M.G. Alexander, H.D. Beushausen & P. Moyo

67

Reinforcement corrosion: Research needs C. Andrade

75

Theme 1: Concrete durability aspects Innovative materials and influences of material composition High performance concrete for extreme applications B. Hillemeier, R. Wens & L. Hänisch

85

Durability and microstructural development during hydration in ultra-high performance concrete B. Möser, C. Pfeifer & J. Stark

87

Concrete ability to resist chloride ions using ternary blended cement I.S. Oslakovic´, R. Roskovic´ & D. Bjegovic´

89

Sulphate resistance of high volume fly ash cement paste composites E. Aydin

91

Durability of light-weight concrete with expanded clay aggregate M. Hubertova, R. Hela & R. Stavinoha

93

V

On the potential of rubber aggregates obtained by grinding end-of-life tyres to improve the strain capacity of concrete A.C. Ho, A. Turatsinze & D.C. Vu

95

Investigations on the PCC-Microstructure after Mechanical Load A. Flohr, A. Dimmig-Osburg & K.A. Bode

97

Composites in structures – Use of special strength theories based on the form of anisotropy V. Lackovic´, J. Krolo & M. Rak

99

Concrete containers for containment of vitrified high-level radioactive waste: The Belgian approach B. Craeye, G. De Schutter, H. Van Humbeeck & A. Van Cotthem

101

Performance of different types of Pakistani cements exposed to aggressive environments A.R. Khan

103

Durability requirements in self-compacting concrete mix design A. Ioani, J. Domsa, C. Mircea & H. Szilagyi

105

Chemical attack of SCC: Immersion tests and X-ray CT V. Boel, K. Audenaert & G. De Schutter

107

Study of chloride penetration in self-compacting concrete by simulation of tidal zone K. Audenaert & G. De Schutter

109

Transport of water and gases in crack-free and cracked textile reinforced concrete R. Barhum, M. Lieboldt & V. Mechtcherine

111

Application of FRC and FRSC in structural elements of St Rok tunnel E. Seferovic & E. Barisic

113

Durability of Strain-Hardening Cement Composites (SHCC) – An overview G.P.A.G. van Zijl

115

A novel durability design approach for new cementitious materials: Modelling chloride ingress in strain-hardening cement-based composites F. Altmann, V. Mechtcherine & U. Reuter

117

A two component bacteria-based self-healing concrete H.M. Jonkers & E. Schlangen

119

Self healing properties with various crack widths under continuous water leakage A. Hosoda, S. Komatsu, T. Ahn, T. Kishi, S. Ikeno & K. Kobayashi

121

Using natural wood fibers to self heal concrete M.R. de Rooij, S. Qian, H. Liu, W.F. Gard & J.W.G. van de Kuilen

123

The effect of geo-materials on the autogenous healing behavior of cracked concrete T.H. Ahn & T. Kishi

125

Surface protections to prevent Alkali-Aggregate Reactions (AAR) in concrete structures L.F.M. Sanchez, S.C. Kuperman & P.R.L. Helene

127

Micro-scale alterations of cementitious surfaces subjected to laser cleaning process and their potential impact on long-term durability A.J. Klemm, P. Sanjeevan & P. Klemm

129

Evaluation of anti-fouling strategies on aerated concrete by means of an accelerated algal growth test W. De Muynck, A. Maury, N. De Belie, J. De Bock & W. Verstraete

131

Corrosion testing of low alloy steel reinforcement M. Serdar, I.S. Oslakovic´, D. Bjegovic´ & L. Valek

133

Influence of the geometry and reinforcement content of specimens on frost de-icing salt resistance of concrete F. Dehn, M. Orgass & K. Pistol

VI

135

Service life modelling and prediction of durability Life cycle management of concrete structures Part I: Birth certificate T.F. Mayer & P. Schießl

139

Life cycle management of concrete structures Part II: Monitoring Ch. Dauberschmidt, G. Kapteina & Ch. Sodeikat

141

A Spreadsheet Model for Service-Life Predictions E.A.B. Koenders, M. Ottelé & B. Obladen

143

Modeling of the influence from environmental actions on the durability of reinforced concrete structures A. Lindvall

145

Studies on prediction models for concrete durability R. Heiyantuduwa & M.G. Alexander

147

Critical chloride content in reinforced concrete – State of the art U. Angst & Ø. Vennesland

149

Use of durability indexes in probabilistic modelling for durability design of RC members R.N. Muigai, P. Moyo & M.G. Alexander

151

A simplified and practical approach regarding design for durability of reinforced concrete structures based on probabilistic modeling of chloride ingress J.J.W. Gulikers

153

Assessment of the behaviour of concrete in the initiation period of chloride induced corrosion of rebars M. Castellote & C. Andrade

155

Corrosion propagation in cracked and uncracked concrete M.B. Otieno, M.G. Alexander & H.D. Beushausen

157

Effect of temperature on transport of chloride ions in concrete Q. Yuan, C. Shi, G. De Schutter & K. Audenaert

159

Repair of architectural concrete and related modelling of carbonation-induced corrosion H.S. Mu¨ller, E. Bohner & M. Vogel Determination of critical moisture content for carbonation of concrete M.O. Mmusi, M.G. Alexander & H.D. Beushausen Effect of environmental parameters on concrete carbonation. DURACON collaboration (Mexican results) E.I. Moreno, C. Vinajera-Reyna, A. Torres-Acosta, J. Pérez-Quiroz, M. Martínez-Madrid, F. Almeraya-Calderón, C. Gaona-Tiburcio, P. Castro-Borges, M. Balancan-Zapata, T. Pérez-López, M. Sosa-Baz, E. López-Vázquez, E. Alonso-Guzmán, W. Martínez-Molina, J.C. Rubio-Avalos, L. Ariza-Aguilar, B. Valdez-Salas, D. Nieves-Mendoza, M. Baltazar & O. Troconis-Rincón

161 163

165

Improvements in durability design of concrete bridges in Croatia P. Sesar, S. Pastorcic & Z. Banic

167

Fire engineering in Croatia D. Bjegovic´, M. Jelcˇic´, M. Carevic´, M. Drakulic´ & B. Peroš

169

Modelling of reinforcement corrosion DFG Research Group 537: Modelling of reinforcement corrosion – An overview of the project P. Schießl & K. Osterminski

VII

173

DFG Research Group 537: Modelling reinforcement corrosion – Investigations on the mechanism of cracking and spalling E. Bohner, N. Soddemann & H.S. Mu¨ller

175

DFG Research Group 537: Modelling of reinforcement corrosion – Macrocells and time dependence M. Raupach, J. Warkus & J. Harnisch

177

DFG Research Group 537: Modelling of reinforcement corrosion – Validation of corrosion model S. von Greve-Dierfeld, C. Gehlen & K. Menzel

179

DFG Research Group 537: Modelling reinforcement corrosion – Corrosion measurements on cracked reinforced concrete beams K. Osterminski, A. Volkwein, W. Tian & P. Schießl

181

DFG Research Group 537: Modelling reinforcement corrosion – Numerical modelling of bond strength of corroded reinforcement C. Fischer, J. Ožbolt & C. Gehlen

183

DFG Research Group 537: Modelling of reinforcement corrosion – Modelling loss of steel cross sectional area and design for durability K. Osterminski & P. Schießl

185

DFG Research Group 537: Modelling reinforcement corrosion – Observation and monitoring of self-corrosion processes in chloride contaminated mortar by X-ray tomography M. Beck, J. Goebbels, D. Meinel & A. Burkert

187

Theme 2: Condition assessment of concrete structures Degradation assessment and service life aspects An Austrian experience with identification and assessment of alkali-aggregate reaction in motorways E.K. Fischboeck & H. Harmuth

191

Chloride ion propagation in onshore zone of Recife-PE R.B. Pontes, E.C.B. Monteiro, R.A. de Oliveira & S.C. de Paiva

193

Investigation of the deteriorations of educational institutes M.M. Alshebani & S. Azhari

195

Durability evaluations on the bridge over the Ruwais lagoon at Jeddah M. Arici & M.F. Granata

197

Assessment of post-fire reinforced concrete structures. Determination of depth of temperature penetration and associated damage C. Alonso

199

Computer integrated knowledge systems for the assessment and diagnosing distress in concrete structures in Persian Gulf F. Moodi

201

Mechanical behaviour of corroded rebars and influence on the structural response of R/C elements S. Imperatore & Z. Rinaldi

203

NDE/NDT and measurement techniques Nondestructive testing methods for hardly accessible parts of structures H. Anton, I. Netinger & E. Evic´

207

Vibration based assessment of shear connectors in concrete composite bridges B. Sibanda, P. Moyo & H.D. Beushausen

209

VIII

New approach to in-situ evaluation of watertightness of reinforced concrete underground structures A. Zakorshmenny

211

Experimental evaluation of indirect sonic wave transmission technique in the diagnosis and monitoring of concrete slabs G. Concu, B. De Nicolo, F. Mistretta & L. Pani

213

Ultrasonic imaging of post-tensioned concrete elements: New techniques for reliable localization of grouting defects M. Krause, B. Gräfe, F. Mielentz, B. Milmann, M. Friese, H. Wiggenhauser & K. Mayer

215

Practical experience of geo-radar technique application in the course of an integrated inspection of historical buildings A. Kirilenko & A. Baukov

217

Wireless monitoring of structures including acoustic emission techniques C.U. Grosse, M. Kru¨ger & S. Bachmaier

219

OSSCAR – Development of an On-Site SCAnneR for automated non-destructive bridge testing A. Taffe, T. Kind, M. Stoppel & H. Wiggenhauser

221

Development of a portable LIBS-device for quality assurance in concrete repair A. Taffe, D. Schaurich, G. Wilsch & F. Weritz

223

High speed chemical analysis of concrete surfaces using the LIBS method within the ILCOM project M. Bruns, M. Raupach, C.D. Gehlen, R. Noll, G. Wilsch & A. Taffe

225

On the possibility of rapid nondestructive determination of compressive strength of cement materials N. Kamenic´, T. Matusinovic´ & J. Šipušic´

227

Locating reinforcing bars in concrete with Electrical Resistance Tomography K. Karhunen, A. Seppänen, A. Lehikoinen, J.P. Kaipio & P.J.M. Monteiro

229

Electrical resistance tomography imaging of concrete A. Seppänen, K. Karhunen, A. Lehikoinen, J.P. Kaipio & P.J.M. Monteiro

231

Materials and structural assessments Quality assessment of a 40-years old tunnel E.A.B. Koenders, M. Ottelé, O. Çopurog˘lu & K. van Breugel

235

Deficiencies in concrete structures “The Sherlock Holmes” factor A.N. van Grieken

237

Contribution to diagnosis of alkali-silica reaction in a bridge structure S.O. Ekolu

239

Condition assessment and repair alternatives for Lucko bridge overpasses of Zagreb K. Mavar, S.S. Palic & A. Balagija

241

Service life extension of existing precast concrete girders D.I. Banic, D. Tkalcic & Z. Banic

243

Approaches for the assessment of the residual strength of concrete exposed to fire E. Annerel & L. Taerwe

245

Residual strength of R.C. buildings after a fire: A case study I. Venanzi, A.L. Materazzi & M. Zappia

247

Assessment of fire-damaged concrete structures and the corresponding measures Y. Anderberg

249

Different methods to investigate and to repair the damages of a post-tensioned deck of a high-way bridge after a fire incident due to a traffic accident A.-W. Gutsch & F. Herschelmann IX

251

Structural behaviour of beams under simultaneous load and steel corrosion G. Malumbela, P. Moyo & M. Alexander

253

BETOSCAN – An instrumented mobile robot system for the diagnosis of reinforced concrete floors M. Raupach, K. Reichling, H. Wiggenhauser, M. Stoppel, G. Dobmann & J. Kurz

255

Accelerated high water corrosion K.P. Mackie

257

Condition assessment of prestressed concrete structures with static and dynamic damage indexes X.Q. Zhu & H. Hao

259

Debond Detection in RC Structures using piezoelectric materials X.Q. Zhu, H. Hao, K.Q. Fan, Y. Wang & J.P. Ou

261

Development of a new test method for mineral based composites – Wedge splitting test K. Orosz, B. Täljsten & K. Orosz

263

Evaluation of fibre distribution in concrete using AC impedance technique B. Srinath & M. Santhanam

265

Simple hydration equation as a method for estimating water-cement ratio in old concrete S.O. Ekolu

267

Selection of an optimal method for humidity elimination in masonry buildings I. Netinger, H. Anton & D. Bjegovic´

269

Repair of guard gallery columns, Robben Island Maximum Security Prison H.G. Smith & M.G. Alexander

271

Theme 3: Concrete repair, rehabilitation and retrofitting Repair methods and techniques Concrete repair: Research and practice – The critical dimension A.M. Vaysburd, P.H. Emmons & B. Bissonnette

275

Performance based rehabilitation of reinforced concrete structures S.L. Matthews & J.R. Morlidge

277

Repair of bridge IB42 over the Limpopo river at Stockpoort border post E.J. Kruger & W.S. Humphries

279

The aging of concrete sewers: Replace or rehabilitate V.A. da Silva & A.M. Goyns

281

Concrete retrofitment solutions utilized at the Van der Kloof Dam spillway bridge P.D. Ronné & R.S. Maliehe

283

Rehabilitation of major bridges over the spillways of three of South Africa’s largest dams P.D. Ronné, A.A. Newmark & E.J. Viljoen

285

Protection of concrete with water repellant agents – What is required to achieve a sufficient penetration depth? A. Johansson, M. Janz, J. Silfwerbrand & J. Trägårdh

287

Accelerated transport of corrosion inhibitors as complementary methodology for electrochemical chlorides extraction method M. Sánchez & C. Alonso

289

Crack repair in concrete using biodeposition N. De Belie & W. De Muynck

291

Cracked concrete repair with epoxy-resin infiltration R. Felicetti & V.H. De Domenico

293

X

Concrete repair with ultra ductile micro-mesh reinforced abrasive resistant and impervious mortar C. Flohrer, M. Tschötschel & S. Hauser

295

Electrochemical treatments of corroded reinforcement in concrete C. Christodoulou

297

Electrochemical impedance monitoring of carbon fiber as an anode material in cathodic protection M. Chini, B. Arntsen, Ø´. Vennesland & J.H. Mork

299

CP of the rear reinforcement in RC-structures – Numerical modelling of the current distribution M. Bruns & M. Raupach

301

The application of electrochemical chloride extraction to reinforced concrete bridge members P.E. Streicher, G.E. Hoppe, V.A. da Silva & E.J. Kruger

303

The corrosion protection of embedded steel reinforcement in reinforced concrete structures using galvanic anodes H. Bänziger, J. Vogelsang & G. Schulze

305

Re-alkalisation technology applied to corrosion damaged concrete G.K. Glass, A.C. Roberts & N. Davison

307

On pathology and rehabilitation teaching of concrete structures: A case study N.G. Maldonado, R.J. Michelini, N.F. Pizarro & M.E. Tornello

309

Comparative study of cleaning techniques to be used on concrete indoors M. Bouichou, E. Marie-Victoire & D. Brissaud

311

Repair materials and systems Material Data Sheet Protocol – From anarchy to order F.R. Goodwin, A.M. Vaysburd & P.H. Emmons

315

Utilization of high performance fiber-reinforced micro-concrete as a repair material M. Skazlic, D. Bjegovic´ & M. Serdar

317

Properties of modern rendering systems based on mineral binders modified by organic admixtures J.C.-M. Capener

319

Low shrinking self-compacting concretes for concrete repair C. Pistolesi, C. Maltese & M. Bovassi

321

Effects of fiber and silica fume reinforcement on abrasion resistance of hydraulic repair concrete Y.-W. Liu, C.-C. Lee & K.S. Pann

323

Use of calcium sulfoaluminate cement to improve strength of mortars at low temperature J. Ambroise & J. Péra

325

Development of Strain-Hardening Cement-based Composites for the strengthening of masonry A.-E. Bruedern, D. Abecasis & V. Mechtcherine

327

Air permeability of hardener-free epoxy-modified mortars as repair materials M.A.R. Bhutta, K. Imamoto & Y. Ohama

329

Performance of corrosion inhibitors in concrete exposed to marine environment I.N. Robertson & C. Newtson

331

Hydrophobic treatments on concrete – Evaluation of the durability and non-destructive testing M. Raupach & T. Bu¨ttner

333

Application of the NMR-technique to concrete-coatings J. Orlowsky, M. Raupach, M. Baias & B. Blu¨mich

335

Durability of adhesion of epoxy coatings on concrete; causes of delamination and blistering M. Raupach & L. Wolff

337

XI

Protection and rehabilitation of chemically high stressed concrete surfaces by thin glass K. Schubert, B. Hillemeier & P. Osselmann

339

Bonded concrete overlays and patch repairs Adhesion – A challenge for concrete repair L. Czarnecki

343

Concrete repair and interfacial bond: Influence of surface preparation B. Bissonnette, A. Nuta, M. Morency, J. Marchand & A.M. Vaysburd

345

Correlation between the roughness of the substrate surface and the debonding risk F. Perez, M. Morency, B. Bissonnette & L. Courard

347

Performance of spall repair materials D.W. Fowler, D.P. Whitney & D. Zollinger

349

Innovative concrete overlays for bridge-deck rehabilitation in Montréal R. Gagné, B. Bissonnette, R. Morin & M. Thibault

351

Mortar mix proportions and free shrinkage effect on bond strength between substrate and repair concrete R. Abbasnia, M. Khanzadi & J. Ahmadi

353

Evaluation of saturation and microcracking of the superficial zone of concrete: New developments L. Courard & J.-F. Lenaers

355

Spalling of sprayed concrete under tunnels fire conditions C. Féron, C. Larive & G. Chatenoud

357

Repair of slab surface with thin SFRC overlay J. Šušteršicˇ, I. Leskovar, A. Zajc & V. Dobnikar

359

SHCC repair overlays for RC: Interfacial bond characterization and modelling G.P.A.G. van Zijl & H. Stander

361

Modelling the performance of ECC repair systems under differential volume changes J. Zhou, M. Li, G. Ye, E. Schlangen, K. van Breugel & V.C. Li

363

Prevention of damages in industrial floors with screed layers R. Breitenbu¨cher & B. Siebert

365

Evaluation of the effect of load eccentricity on pull-off strength L. Courard, A. Garbacz & G. Moczulski

367

Proposal of an experimental programme for determining tensile relaxation in bonded concrete overlays C. Masuku, H.D. Beushausen & P. Moyo

369

Structural repairs and strengthening Analytical modelling of retrofitted reinforced concrete members with flexible bond E. Raue, H.-G. Timmler & H. Schröter

373

Experimental and analytical investigation of concrete confined by external pre-stressed strips H. Moghaddam, S. Mohebbi & M. Samadi

375

Flexure and shear behavior of RC beams strengthened by external reinforcement F. Minelli, G.A. Plizzari & J. Cairns

377

Simplified verification of the bond resistance of externally bonded reinforcement at the area of the support moment C. Muehlbauer, R. Niedermeier & K. Zilch

XII

379

Loading test on a retrofitted pretensioned concrete girder after fire L. Taerwe & E. Annerel

381

Anchorage failure of RC beams strengthened with FRP at the bottom face F. Arifovic & B. Täljsten

383

Strengthening and verification of the prestressed road bridge using external prestressing M. Moravcik & I. Dreveny

385

Numerical simulation of continuous RC slabs strengthened using NSM technique J.A.O. Barros, G. Dalfré & J.P. Dias

387

Non-linear finite element modeling of epoxy bonded joints between steel plates and concrete using joint elements L. Hariche & M. Bouhicha

389

Increase in the torsional resistance of reinforced concrete members using Textile Reinforced Concrete (TRC) F. Schladitz & M. Curbach

391

Rejuvenating our aging structures: Cumberland street parking garage rehabilitation P. Sarvinis

393

Mountain pass slope failure retrofitted with half viaduct bridge structure E.J. Kruger, A.A. Newmark & M. Smuts

395

Rapid repair of RC bridge columns subjected to earthquakes A. Vosooghi, M.S. Saiidi & J. Gutierrez

397

Rehabilitation problems regarding an over pass in Timisoara – Romania A. Bota & Alexandra Bota

399

Seismic retrofit and rehabilitation An approach of service life in repair of structures with concrete due to natural hazards N.G. Maldonado, N.F. Pizarro, R.J. Michelini & A.M. Guzmán

403

Seismic behavior of FRP-upgraded exterior RC beam-column joints Y.A. Al-Salloum, S.H. Alsayed, T.H. Almusallam & N.A. Siddiqui

405

3D analysis of seismic response of RC beam-column exterior joints before and after retrofit R. Eligehausen, G. Genesio, J. Ožbolt & S. Pampanin

407

Seismic assessment of a RC school building retrofitted with innovative braces L. Di Sarno & G. Manfredi

409

Seismic response of FRP-strengthened corner RC beam-column joints T.H. Almusallam, S.H. Alsayed, Y.A. Al-Salloum & N.A. Siddiqui

411

Experimental investigation on the seismic behaviour of SFRC columns under biaxial bending M. Palmieri, G.A. Plizzari, S. Pampanin & J. Mackechnie

413

Retrofitting techniques and FRP systems Feasibility study of a novel prestressing system for FRP-laminates W. Figeys, L. Schueremans, D. Van Gemert, K. Brosens, L. Van Schepdael & J. Dereymaeker

417

FRP tendon anchorage in post-tensioned concrete structures J.W. Schmidt, B. Täljsten, A. Bennitz & H. Pedersen

419

Durability of CFRP strengthened concrete structures under accelerated or environmental ageing conditions K. Benzarti, M. Quiertant, C. Aubagnac, S. Chataigner, I. Nishizaki & Y. Kato

XIII

421

Durability of GFRP composite made of epoxy/organoclay nanocomposite H.G. Zhu, C.K.Y. Leung & J.K. Kim Performance of Reactive Powder Concrete (RPC) with different curing conditions and its retrofitting effects on concrete member T.P. Chang, B.T. Chen, J.J. Wang & C.S. Wu

423

425

Residual strength of SBR-latex modified high strength concrete under repeated impact loading S.R. Shashikumara, S. Ali Khan Zai, K. Muthumani & N. Munirudrappa

427

Modeling the behavior of insulated FRP-strengthened reinforced concrete beams exposed to fire W.Y. Gao, K.X. Hu & Z.D. Lu

429

Fracture optimization of RC beams strengthened with CFRP laminates with finite element and experimental methods A.A. Ramezanianpour & A. Gharachorlou

431

Comparing the behaviour of reinforced HSC beams with AFRP bars and confined HSC beams with AFRP sheets under bending R. Rahgozar, M. Ghalehnovi & E. Adili

433

Corroded RC beam repaired with near-face mounted CFRP rods A. Kreit, F. Al-Mahmoud, A. Castel & R. Francois Effect of damaged concrete cover on the structural performance of CFRP strengthened corroded concrete beams A.H. Al-Saidy, A.S. Al-Harthy & K.S. Al-Jabri

435

437

Role of U-shaped anchorages on performance of RC beams strengthened by CFRP plates A.R. Khan

439

Shear-flexure interaction of RC circular columns strengthened with FRP sheets P. Colajanni & A. Recupero

441

Strengthening of R/C existing columns with high performance fiber reinforced concrete jacket A. Meda, G.A. Plizzari, Z. Rinaldi & G. Martinola

443

Retrofitting of a precast industrial building M. di Prisco, M. Lamperti & S. Lapolla

445

Strength and ductility of unreinforced concrete columns confined with CFRP composites under uniaxial loading P. Sadeghian, A.R. Rahai & M.R. Ehsani Experimental investigation on repair of RC pavements with SFRC G. Boscato & S. Russo Experimental research of performance and strength of damage – Concrete beams strengthening with CFRP using two different epoxies A.R. Rahai, M.R. Saberi & R. Mahmoudzadeh

447 449

451

Strengthening of shear wall with high performance RC jacket A. Marini & A. Meda

453

Author index

455

XIV

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Preface

These Proceedings represent papers presented at the Second International Conference on Concrete Repair, Rehabilitation and Retrofitting (ICCRRR 2008), Cape Town, South Africa, November 2008. The conference Proceedings contain papers presented at the conference, and classified into a total of 12 sub themes which can be grouped under three main themes: − Concrete durability aspects − Condition assessment of concrete structures − Concrete repair, rehabilitation and retrofitting A large number of papers discuss performance and assessment of innovative materials for durable concrete construction. Interesting fields, some quite new, are covered, such as self-healing techniques, high performance concretes, and strain hardening composites. The amount of papers submitted on the topic of service life modelling and prediction of durability confirms the positive international developments towards performance-based methods for durability design and specification. Another fact that is evident from the paper submissions is that large advances have recently been made in the fields of non-destructive testing and condition assessment of concrete structures. The papers in the proceedings cover interesting new techniques for the assessment of reinforcement corrosion and their interpretation. Further, vibration-based evaluation of the structural capacity of reinforced concrete members is discussed, representing a relatively new and promising technique for the assessment of corrosion- or fire-damaged structures. The majority of papers discuss recent developments in concrete repair, rehabilitation and retrofitting techniques. An important research area lies in the field of specifications for repair materials and systems. Here, an integrated approach is needed, linking assessment techniques and service life modelling to appropriate repair methods. A number of papers deal with these important issues, confirming that the industry is on the right track towards efficient and durable repairs. Based on research reports and case studies, latest developments on repair strategies and materials are presented, ranging from surface protection techniques to full-scale repairs. Bonded concrete overlays and patch repairs remain important fields for most repair projects. Techniques and materials for crack-free overlays with sufficient bond strength are discussed. Numerous papers were submitted on the topic of strengthening and retrofitting, highlighting the need to cope with increasing loads and deteriorating structures and showcasing latest developments in FRP strengthening systems. The Second International Conference follows the highly successful First International Conference on Concrete Repair, Rehabilitation and Retrofitting (ICCRRR 2005), also held in Cape Town, in November 2005. At that event, many expressed the opinion that a following event of a similar nature was needed—hence this second conference. Judging by the papers received for this conference, both the interest in this topic and the quality of the presentations remain high. This conference, as was the previous one, is a collaborative venture by researchers from the South African Research Programme in Concrete Materials (based at the Universities of Cape Town and The Witwatersrand) and the Material Science Group at Leipzig University and The Leipzig Institute for Materials Research and Testing (MFPA) in Germany. The organisation and implementation of the conference continues to embody a strong South African – German link, reflected in the excellent support given to the conference by researchers and practitioners from these two countries. However, the range of presenters at the conference continues to indicate its truly international nature, with authors being drawn from 42 countries and numerous research and industrial organisations. This continues to fulfil an aim of these conferences, to strengthen relationships not only between Africa and Europe but also between countries and regions from all over the world. The background, in industry and the state of national infrastructures, continues to be highly challenging and demanding. The facts remain that much of our concrete infrastructure deteriorates at unacceptable rates, that we need appropriate tools and techniques to undertake the vast task of sound repair, maintenance and rehabilitation of such infrastructure, and that all this must be undertaken with due cognisance of the limited budgets available for such work. New ways need to be found to extend the useful life of concrete structures cost-effectively. Confidence in concrete as a viable construction material into the 21st Century needs to be retained and sustained, particularly considering the environmental challenges that the industry and society now face. This second conference also continues to seek to extend a sound base of theory and practice in repair and rehabilitation, through both theoretical and experimental studies, and through good case study literature. It also XV

seems to the organisers that two key aspects need to be addressed currently: that of developing sound and easily applied standard practices for repair, possibly codified, and the need seriously to study the service performance of repaired structures and repair systems so as to inform future actions. In fact, without substantial effort at implementing the latter goal, much of the effort in repair and rehabilitation may prove to be less than economic or satisfactory. All papers submitted for ICCRRR 2008 were subjected to a full process of peer review, and the Proceedings contain only those papers that were accepted following this process. The review of manuscripts was undertaken by members of the International Scientific/Technical Advisory Board and other identified leading experts, acting independently on one or more assigned manuscripts. This invaluable assistance, which has greatly enhanced the quality of the Proceedings, is gratefully acknowledged. Special acknowledgements are due to the following organisations: − − − − − − − − − − − − −

National Research Foundation of South Africa Deutscher Beton- und Bautechnik-Verein Cement and Concrete Institute of South Africa Concrete Society of Southern Africa Deutscher Ausschuss für Stahlbeton The South African National Roads Agency Deutsche Bauchemie RILEM American Concrete Institute Sika South Africa BASF Construction Chemicals Afrisam Fosroc

Finally, the editors wish to thank the authors for their efforts at producing and delivering papers of high standard. We trust that the Proceedings will be a valued reference for many working in this important field and that they will form a suitable base for discussion and provide suggestions for future development and research. M.G. Alexander H.D. Beushausen F. Dehn P. Moyo Editors

XVI

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

ICCRRR Committees

Local organizing committee M.G. Alexander, University of Cape Town (Co-chairman) H.D. Beushausen, University of Cape Town (Co-chairman, secretary) F. Dehn, Universität Leipzig / MFPA Leipzig (Co-chairman) P. Moyo, University of Cape Town (Co-chairman) L. Alexander, University of Cape Town E. Yelverton, University of Cape Town R. Heiyantuduwa, University of Cape Town S. Ekolu, University of the Witwatersrand J. Sheath, Cement and Concrete Institute, South Africa E. Kruger, South African National Roads Agency P. Ronne, BKS Pty Limited P. Adams, Sika South Africa International scientific and technical advisory board Professor M.G. Alexander, University of Cape Town, South Africa Professor C. Andrade, Instituto de Ciencias de la Construccion Eduardo Torroja, Madrid, Spain Dr. H. Andrae, LAP Stuttgart, Germany Professor Y. Ballim, University of the Witwatersrand, South Africa Professor A. Bentur, Israel Institute of Technology, Haifa, Israel Dr. H.D. Beushausen, University of Cape Town, South Africa Dr. B. Bissonnette, CRIB Laval, Canada Professor D. Bjegovic´, University of Zagreb, Croatia Professor J.M.W. Brownjohn, University of Sheffield, United Kingdom Professor E. Bruehwiler, EPFL Lausanne, Switzerland Dr. P. Castro Borges, CINVESTAV-Mérida, Mexico Professor L. Courard, Universitè de Liège, Belgium Professor L. Czarnecki, Warsaw University of Technology, Poland Dr. F. Dehn, Universität Leipzig/MFPA Leipzig, Germany Dr. E. Denariè, EPFL Lausanne, Switzerland Professor M. Di Prisco, Polytechnico di Milano, Italy Dr. S. Ekolu, University of the Witwatersrand Professor D. Fowler, University of Texas at Austin, USA Prof. C. Gehlen, University of Stuttgart, Germany Mr. Michael Grantham, Concrete Solutions, UK Dr. G. Grieve, Cement and Concrete Institute, South Africa Dr. C. Grosse, University of Stuttgart, Germany Professor H. Hao, University of Western Australia, Australia Professor P. Helene, Escola Politécnica of University of São Paulo, Brazil Professor B. Hillemeier, TU Berlin, Germany Mr. E. Kleen, MC Bauchemie, Germany Professor K. Kovler, Israel Institute of Technology, Israel Mr. E. Kruger, South African National Roads Agency Professor K. Leung, UST, Hong Kong Professor M. Limbachiya, Kingston University of Technology, United Kingdom Dr. H.-U. Litzner, Deutscher Beton- und Bautechnik Verein, Germany Professor M. Lopez, Ponticica Universidad Católica de Chile, Chile Dr. J. Mackechnie, University of Canterbury, New Zealand XVII

Dr. U. Maeder, Sika Schweiz AG, Switzerland Mr. Richard Morin, Canada Professor P. Moyo, University of Cape Town, South Africa Professor H.S. Müller, Universität Karlsruhe, Germany Professor A. Nanni, University of Miami, USA Mr. A. Newmark, BKS Pty Limited, South Africa Professor Y. Ohama, Nihon University, Japan Professor G. Plizzari, University of Bergamo, Italy Professor A. Ramezanianpour, Amir Kabir University of Technology, Iran Professor M. Raupach, RWTH Aachen, Germany Professor P. Schiessl, TU München, Germany Professor G. De Schutter, Ghent University, Belgium Mr. N. Schröter, Deutsche Bauchemie, Germany Professor J. Silfwerbrand, CBI Stockholm, Sweden Dr. K. Stanish, Walker Restoration Consultants, USA Dr. R. Torrent, Holcim Group Support, Switzerland Dr. A. Turatsinze, LMDC Toulouse, France Professor D. Van Gemert, KU Leuven, Belgium Professor G. Van Zijl, University of Stellenbosch, South Africa Dr. A. Vaysburd, Vaycon, USA Professor Øystein Vennesland, Norway Professor J.C. Walraven, TU Delft, The Netherlands Professor F. Wittmann, Aedificat Freiburg, Germany Professor A. Zingoni, University of Cape Town, South Africa

XVIII

Keynote papers

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Design for service life: How should it be implemented in future codes J.C. Walraven Delft University of Technology, Delft, The Netherlands

ABSTRACT: Design for durability has gained the same level of importance as design for safety and design for serviceability. Many individual contributions to design for service life have been noted in recent years. Such initiatives should, after some time, result in a consistent approach. When trials are made to develop such an approach, inevitably “blind spots” are discovered. This paper gives an overview of the most important developments and moreover traces areas were knowledge is still inadequate and further research is needed.

1

INTRODUCTION

strategy. Design for service life requires, that all relevant deterioration mechanisms are identified and that those, which are relevant for the structure to be designed are selected. Structures can be prone to deterioration due to:

In the light of the developments in the field of concrete structures the increasing attention for “design for service life” is remarkable. The background of this development is clear. Damage due to insufficient attention for durability, or due to the use of inadequate criteria, have led to expensive repair and even demolition and replacement. De Sitter (1984) made a statement which perfectly touches the core of the matter. This statement is known as the Rule of Fives: “If no maintenance is carried out the later repair costs will be five times the saved maintenance costs. If no repair is carried out, the cost of renovation will be five times the money saved by not repairing”. Meanwhile we have experienced the truth behind this saying: expensive maintenance and repair measures are necessary there, where in the past the need for the aspect durability was ignored. The following question is how maintenance can be minimized by appropriate design. It is clear that this may result in appreciable cost savings in time. Now it also becomes clear that in the design stage not only the costs of the new building, but most of all the integral costs, including maintenance, adaptation and demolition have to be considered. In order to develop a suitable approach to design for service life it is necessary to know which aspects play a dominant role and where still open questions stand out. This paper gives a survey of developments and needs for further research.

− Chloride penetration − Carbonation − Frost-thaw effects, whether or not in combination with de-icing salts − Chemical influences (attack by acids or sulphates) − Ettringite formation − Alkali silica reaction, Fig. 1 In this list damage due to the penetration of chlorides and carbonation are by far the most important

2 THE DEVELOPMENT OF A STRATEGY FOR SERVICE LIFE DESIGN 2.1

Deterioration mechanisms and their influence on service life

A treatment will be given of the various methods to extend the service life by means of a rational design

Figure 1.

3

Damage due to ASR.

causes for damage to concrete structures. It is therefore understandable, that a substantial number of research projects have been focused on a better description of those mechanisms. With regard to the prediction of durability with mathematical models the Duracrete models are well known. These models describe the penetration of chloride and carbon dioxide on the basis of diffusion models. In this way the deterioration of concrete structures in time can be calculated. Fig. 2 shows the reduction of R (Resistance) in time due to deterioration. In the same diagram the increasing loads (S) on the structure are represented. If as well the scatter of R(t) and S(t) are regarded, it can be calculated until when the structure can carry the load with specified reliability. It is then as well clear when repair and/or strengthening of the structure should be carried out. 2.2

Figure 3. Methods for extending the service life of a concrete structure.

process is delayed. In this respect two possibilities are distinguished:

Practical methods for the extension of service life

The deem to satisfy approach (recipe based). The deem-to-satisfy method implies, that design criteria are met, which guarantee a certain specified service life. These criteria like minimum cover, or concrete composition to limit permeability for the governing environmental class, are mostly based on experience. Up to now optimization is not possible, because the rules do not offer the possibility for exchange (for instance smaller cover for improved concrete impermeability).

Contractual specifications for the service life to be regarded in design can be met in two ways (Fig. 3). On the one hand it is possible to reach the specified service life by avoiding the deterioration mechanisms. Possibilities are: − the application of membranes or coatings (extension of reaction) − the application of materials with low sensitivity to deterioration (like stainless steel) − preventing the reaction (by for instance cathodic protection)

The method of probabilistic modelling. According to this method the environmental load is compared with the resistance of the structure, taking account of the influence of time. On this basis the probability is calculated that damage will occur to the structure. With this method optimization is possible: the option for exchange is built in and new materials can be used as soon as their properties are known. It should be noted that the option “avoiding damaging reactions” should not be regarded with blind confidence. Fig. 4 shows the condition of a protective membrane that was applied on a structure in order to stop the penetration of chlorides: on the one hand the membrane has become brittle due to ultra violet radiation, on the other hand behind the membrane a moisture pressure has been developed, which leads to peeling-off of the membrane. The application of stainless steel is an interesting possibility too for durable design, especially since less expensive types of stainless steel are available on the market. Savings on the quality of the concrete can be accepted with confidence and economic solutions on another basis are possible. Also the possibility of combining stainless steel with normal “black” steel is a possibility. Most actual design recommendations are based on a deem-to-satisfy approach. EN 206 distinguishes 5 main environmental classes, including 18 subclasses. Design parameters to satisfy the requirements are

On the other hand materials and design methods can be chosen with the aim to delay or extend the chemical reaction that leads to corrosion. This method of design means that the sensitivity of the structure is reduced by appropriate choices for material and structural detailing. By the selection of a material with suitable properties in combination with the minimum structural dimensions the deterioration

Figure 2. Probabilistic determination of service life, Rostam (2001).

4

the environmental conditions, the conceptual design, the material choice, the dimensions, detailing, execution, maintenance and quality of archiving the data. The best available models for design for service life consider predominantly environment, material and geometrical dimensions: hardly any attention is given to detailing, execution and maintenance. Especially the quality of execution is a “blind spot”. Here a number of aspects play a role, such as placing the reinforcement, the casting procedure, compaction, curing, storage of materials, formwork, demoulding, labour conditions, quality control, organisation at the site, education and training of the executing labourers. Formulating requirements with regard to the quality of execution and the introduction of quality control procedures are no absolute guarantee that the criteria for service life are met. So, at handing over the structure to the client it should be controlled that the specified quality has indeed been achieved. At the start of its life the structure should be investigated with regard to the most important influencing properties governing durability. These findings should be laid down in reports. This initial control fits very well in the total plan of intermediate controls. On the basis of this inspection it is possible to determine whether the maintenance plan and the quality of the structure are in agreement with each other. Not reaching the specified quality should have consequences for the contractor, like for instance the obligation of repair or the payment of capitalized additional maintenance cost. In Japan the flow chart shown in fig. 5 is used. At handing-over the initial inspection is carried out. On the basis of the results of this inspection a maintenance category is defined and the deterioration of the structure is predicted. Intervals of control testing are programmed in advance. On the basis of these control tests the need for repair is determined. For small-scale

Figure 4. An unsuccessful attempt to stop the deterioration process with a coating on a quay wall.

the thickness of the cover and the concrete strength. The combination of the two has to satisfy certain criteria. Meanwhile considerable progress was made with regard to modelling the deterioration processes. Theoretically it is possible to describe the deterioration process on the basis of physical models. For the practical application of those models in design a number of basic elements are necessary: − − − − − −

limit state criteria a defined service life deterioration models compliance tests a strategy for maintenance and repair quality control systems

The method of probabilistic modelling implies that a concrete composition is chosen with regard to its capacity to control the speed of penetration of for instance chloride or carbon dioxide. By optimizing the concrete composition and applying an appropriate minimum concrete cover the structure can then be designed for a specified service life. The probabilistic approach is now only applied for structures with large relevance, such as tunnels and important bridges. For “all-days” structures the deemto-satisfy-method is sufficient. However, it is important to make sure that the results of both methods are compatible. In EN 206 a minimum strength class is given without any relation to the type of cement, and/ or the use of fillers. Moreover the quality of curing and the length of the curing period have a significant influence. Here parameter studies and comparison with experiments are necessary to increase the understanding. 2.3

Dealing with the quality of execution

Every design method should take into account the large number of influencing factors which influence service life. The most important factors, in sequence with their occurrence in the building and construction process are

Figure 5. Flow chart for inspection and maintenance of concrete structures.

5

The Maesland Storm Surge Barrier, was delivered in 1990. The design service life is 100 years. The design for durability was made on the basis of an extrapolation of the governing design code (which was based on a maintenance free service life of 50 years) with the √t formula. The calculation showed that for a maintenance free service life of 100 years a minimum cover of 35 √(100/50) = 50 mm is necessary. The Western Scheldt Tunnel and the High Speed Railway Amsterdam-Brussels were finished in 2000 and 2001 respectively. For the design probabilistic models according to the DuraCrete method were used. Further criteria were:

repairs the prediction of deterioration is updated. For large scale repairs the initial control is carried out again. The recording of the data is very essential. In stead of inspections, or additional to those, the condition of the structure can be followed by monitoring. 3

EXPERIENCE WITH DESIGN FOR DURABILITY UP TO NOW

3.1

Experience with infrastructural projects

In The Netherlands various large infrastructural projects have been realised in the last decades. Durability has been a point of growing attention. The sluices of Haringvliet, Fig. 6, were finished in 1960. With regard to the durability the following requirements have been formulated in the design stage: − − − −

− A maintenance free service life of 100 years − The end of service life was defined as the start of corrosion (end of initiation period) due to penetration of sea water (outside) or frost thaw salt (inside) − An accepted failure probability of 3,6% − Effect of exposure and curing to be regarded in calculations.

use of blast furnace cement water-cement ratio ≤ 0,45 concrete cover ≥ 70 mm prestressing where possible

3.2

Until now no problems with regard to durability have been noted: after nearly 50 years, only limited chloride penetration was measured. The probability that the chloride content at the reinforcement reaches a critical value during the first 100 years was calculated to be smaller than 3,5%. The Eastern Scheldt Barrier was finished in 1980. The specified design service life was 200 years. Calculations for determining the durability have been carried out until an age of 80 years on the basis of mean values. After 25 years no visual damage was observed. There was some chloride penetration. According to probabilistic calculations, based on physical models and inspection the probability of reaching a critical chloride content at the reinforcement was calculated to be less than 6,6% after 50 years and less than 14% after 200 years.

Figure 6.

Verification of the DuraCrete models

The Duracrete models contain a number of parameters. For those parameters a number of indicative values have been given. Whether these values are sufficiently representative has not yet been verified. A research project was carried out by CUR Committee B82. The project was entitled “Durability of Marine Structures”, abbreviated as DuMaCon. The most important topic of research was the chloride penetration as a function of time. The equation to be investigated on suitability was: C ( x, t ) = Cs [1 − erf ( x / 2 {KD0 (t0 / t ) nCl t})]

(1)

where Cs is the chloride content at the surface, x is the distance from the surface, K is a coefficient taking account of various influences (environment, curing), D0 is the diffusion coefficient for chloride at the reference time t0, nCl is the material dependant “reduction exponent” and t is the time. Also the chloride content which is initially in the concrete should be regarded. With the aid of this equation it can be predicted when the chloride content at the reinforcement reaches the critical content Ccr and corrosion is initiated (limit state G(t) = Ccr – C(x,t)). This happens at time ti. Field research was carried out at six structures, in which the thickness of the concrete cover, the steel potential and the electric resistance were measured. Drilled cores were used in order to establish the chloride penetration (profiles), the microstructure (by microscopy), the resistance and the strength. Measurements were carried out at and specimens taken from 17 places with an area of 1 × 1 m2. Subjects of

Haringvliet Sluices (1960).

6

4

investigation were i.e. the Pier of Scheveningen, the Haringvliet Sluices, the Eastern Scheldt Storm Surge Barrier and three quay walls in Rotterdam. These structures are now between 25 and 45 years old. It was remarkable that in most cases a considerable scatter was found in chloride penetration, even between 6 cores taken from one m2, see fig. 7. Also along the length of the Eastern Scheldt Storm Surge Barrier a considerable scatter was found. With regard to the backgrounds of this variation no clear explanation was found yet. In this respect more data are necessary in order to take full profit of new calculation models, like described in the PhD thesis of Li (2004). The chloride profiles as determined from the tests were compared with those from the theoretical model. In most cases the agreement was good. On the basis of those observations, in combination with laboratory tests, it was proposed to modify a number of parameters for blast furnace cement in a marine environment. This referred to the value of the exponent nCl and another calculation of the environment factor K: in this calculation the temperature is a decisive factor and not anymore the curing. Moreover a value for the surface chloride content was found which describes well the chloride load from about 10 years exposure for concrete from the water level to 7 meters above sea level: an average chloride content of 2,9% of the cement weight with a standard deviation of 0,8% was measured. Above 7 m for Cs strongly diverging results were found: between 1% and 5%. Apparently at a larger height the surface chloride content is influenced by the contrary effects of wetting by sea water and raining. Rain-protected areas (which are only in contact with fine seawater vapour) are therefore exposed to significantly higher chloride concentrations.

4.1

FURTHER NECESSARY STEPS FOR ENHANCING THE RELIABILITY OF SERVICE LIFE DESIGN Improvement of physical models like the DuraCrete models

When verifying transport models like the DuraCrete models on existing concrete structures, like described in 3.1, it turned out that in certain respects knowledge was lacking and understanding was not sufficiently developed: • There is no reliable value for the critical chloride content in practice (this holds true as well for Portland cement with fly ash). The B82 research project did not give sufficient evidence. This is felt as a major lack in knowledge. The significance of this parameter for taking the right decisions with regard to maintenance is large. Further research is therefore necessary. • The existing model for chloride transport contains simplifications and uncertainties. Improved models are needed, which are able to simulate the effect of moisture variations better. Also a new generation of models should be verified. Moreover it is useful to follow structures in time. The penetration of chlorides and carbon dioxides can then be estimated with better accuracy. In The Netherlands it was therefore advised to investigate the same structures another time after 10 years. In this respect use can be made from improved formulations for the penetration of chlorides, like for instance developed by Meijers, (2003). In his model account is taken of variable boundary conditions with regard to moisture, temperature and chloride, Fig. 8. • Exposure to chlorides at an early age of the concrete (<28 days) leads to increased penetration. A high speed of construction can therefore negatively influence the service life. This effect should be studied more in depth. • The background of spatial variation in chloride penetration is not yet well known. This should be further investigated, both with regard to the variation in exposure and the variation in resistance (macro scale and meso scale). This information is necessary to be able to evaluate results of local inspection measurements. • Maintaining concrete structures can be improved by better recording relevant data from the construction period. Here it should be formulated which data are necessary, to which degree they are measurable and to which extent they should be additionally determined. Relevant information concerns design criteria, information on the materials used, the way of producing and curing and further construction aspects including deviations and eventual control tests. The total collection of data should be the basis of the “Birth certificate”.

Figure 7. Scatter in chloride profiles within an area of 1 m2, see CUR B82 (2004).

7

• The “maintenance limit state” needs to be further specified. Mostly here the age at initiation of corrosion is taken, although for the realisation of corrosion more time is required (propagation period). An open question is which reliability index belongs to that limit, what the consequences for maintenance decisions are and how anticipating maintenance versus reacting maintenance works out on the cost along the service life.

4.2

cover and the water-cement ratio have an important effect. For projects of high relevance the quality of the structure should be determined at the “initial inspection”, upon delivery of the structure. Only then a definite maintenance plan can be made. The initial inspection, see also Fig. 5, is an important instrument for the determination of the initial state of the structure. On the basis of these data the development of the deterioration can be predicted and a maintenance plan can be made in combination with an inspection plan. The initial inspection is a significant element in the quality control. It can give an indication whether the execution has been carried out with sufficient quality. It should be further specified what is controlled at this initial inspection, what should be measured and how measurements are carried out. Further to that the consequences should be clear for the case that the specified quality has not been reached.

Simulation of the behaviour after depassivation of reinforcing steel

Until now it has always been assumed that the instant of depassivation of the steel defines the end of the damage-free service life. For this limit state a reliability index in the range 1,5–1,8 is defined. It is, however, important to find out what happens in reality after depassivation. Now qualitatively the following stages in the further deterioration process are distinguished: the occurrence of splitting cracks, spalling of the cover, and loosening of parts of the concrete. With regard to the length of those stages only limited information exists. This is important because it is not sure that depassivation is synonymous to the start of an accelerated corrosion process which finally leads to unacceptable damage. There is uncertainty with regard to the question after how much time damage will develop in relation to internal and external factors. 4.3

4.4

The significance of crack width criteria for the durability of structures and their economic impact

In the past a governing role for durability was attributed to the widths of the cracks. Later on it was demonstrated that the thickness and the quality of the cover play a much more important role. Nevertheless in many codes still maximum crack width limits are given in relation to the environmental class. These demands are not very well supported by evidence. Especially when the crack widths should be small the influence of this requirement on economy (required reinforcement) can be large. Whether the specified maximum crack width is 0,2 mm or 0,1 mm can mean an increase of the steel volume with 100%. Another important aspect is that the crack width control mainly pertains to the effect of loads and imposed deformations. These are the most harmless cracks, because they mainly cross the reinforcement perpendicularly. Cracks, however, can have their origin as well in many different causes which are more relevant. Especially cracks occurring before hardening should be considered.

Better accounting for the quality of execution

The quality of execution has a large influence on the durability of concrete structures. A better understanding is necessary into the role of the quality of compaction and curing. In the DuraCrete models those influences are only treated with a simple factor. A better solution is hardly possible because the quality of execution can only be determined after finishing the structure. Deviations from the specified concrete

4.5

Taking better account of probabilistic aspects

With the PhD thesis of Li (2004) first impetus was given to the introduction of probabilistics into inspection and maintenance. In this thesis the significance of a good probabilistic basis for the choice of the optimum repair strategy was well demonstrated. Deviations from the specified values should be taken into account and their effect on durability should be regarded. Measurements at the building site showed, that the deviations are often significant. At a Dutch building site it was found, that for a specified value of the water-cement ratio of 0,50 mm an average value

Figure 8. Comparison of experimentally and calculated chloride profiles after 2, 8 and 16 years according to Meijers (2003).

8

5

of 0,53 was measured, with an upper limit of 0,60 and a lower limit of 0,48 mm. Also for the concrete cover deviations were noted. At a site where the specified minimum cover thickness was 35 mm, an average thickness of 40,4 mm was measured, with a standard deviation of 9,5 mm. The measured minimum thickness was 17 mm and the maximum was 75 mm. Such imperfections can have a large influence on the durability. 4.6

In Fig. 2 it was shown that the bearing resistance of structures is reduced in time due to deterioration processes. On the other hand the traffic increases. This means that the reliability of the structure decreases. The decision to upgrade the structure, however, has considerable financial consequences and should therefore be taken on the basis of sound considerations. A very important aspect that has not been treated yet is that many structures have a residual bearing capacity. If it can be proven convincingly that the curve representing the real bearing capacity is higher than the curve representing the design-capacity, this means that strengthening can be extended. Fig. 9 shows this schematically. Residual bearing capacity can occur as a result of several reasons. A number of those reasons are given in the following survey:

Improvements in design methodology

For a number of important criteria no generally accepted values or definitions are available, such as: − Definition of the end of service life. − Determination of acceptable probability of exceeding a limit state. − Critical chloride content for various cement types used in practice, as well as blends. − Better definition of the exposure conditions (chloride, moisture variation). 4.7

DETERMINING THE REAL BEARING CAPACITY OF EXISTING STRUCTURES

a. Compressive membrane action. If a bridge deck is loaded by a heavy wheel load the punching shear capacity may be expected to be governing. In many bridge decks the real bearing capacity is larger than the theoretical punching load. The reason is the development of compressive membrane action. This occurs since deflection under the wheel load goes along with lateral extension of the concrete around the loaded area. b. Design calculations are mostly carried out according to lower bound models. An example is the strip method, assuming simplified load bearing paths. This method is transparent for the designer and leads generally to simple reinforcement geometries, appropriate for construction. When, however, a structure designed in this way is recalculated by e.g. the theory of elasticity, it is often found that the real bearing capacity is larger than determined on the basis of the design rules. c. After having calculated the necessary reinforcement on the basis of ULS calculations, often practical reinforcement is added. Examples are horizontal web reinforcement or reinforcement along free edges. This reinforcement adds to the real bearing capacity. d. At the time that the bearing capacity of a structure becomes subject of discussion the age is mostly in the range of 30–60 years. The concrete strength is then, as a result of continued hydration, much higher that the 28-days strength on which the design has been based. Many bridges in The Netherlands showed concrete compressive strengths of 60 MPA, whereas the design strengths was not higher than 25 MPa in the past. This may count especially in the case that the shear capacity is concerned. e. Solid slab bridges are mostly designed with shear formulas which are a result of the evaluation of

Robust criteria in stead of “advanced” calculations?

Many mechanisms that cause deterioration are based on penetration of chemicals in concrete. The microstructure of the material plays therefore an important role. However, the variation in the concentration of chemicals (like Cl− ions and CO2) and the variation in environmental conditions (temperature, relative humidity, wind, and solar radiation) are uncertainties. Moreover the properties of the material are determined by the way of casting in relation to the workability of material and quality of curing. Summarizing, there is a multitude of processes and boundary conditions, which cannot always be defined very accurately. In cope with this situation different approaches are possible. On the one hand one can try to describe all processes as accurately as possible and carry out parameter studies in order to get an idea about types of deterioration, their speed and their probability of occurrence. On the other hand one can argue that, because of the availability of so many unknown aspects, it may be better to distinguish a limited number of durability classes for the concrete. In a cooperation program between a number of South African universities, see Beushausen (2003), a proposal was formulated to classify the concrete in durability classes, on the basis of penetration tests for three different media, being oxygen, water and chlorides. On the basis of the results of those tests the concrete ends in one of the durability classes “excellent”, “good”, “moderate” or “bad”. This method is as well suitable for the classification of existing structures. It seems worthwhile to further investigate and evaluate such methods.

9

2.

3.

4. Figure 9. Extended service life by virtue of residual bearing capacity.

5.

deem-to-satisfy-rules. Probabilistic models will be available for structures of large relevance. An important task will be to assess the state of deterioration of existing structures, in order to define their remaining service life and plan the corresponding maintenance. To take full profit of advanced deterioration models the aspect of scatter deserves due consideration. Under many conditions an accurate definition of the input values is not possible. More “robust” models therefore should also be taken serious. The influence of executions on durability is a major influencing factor. It is therefore useful to test the condition of a new structure and record the data in a “birth certificate”. The aspect of residual bearing resistance of structures should be regarded when estimating the service life of existing structures.

REFERENCES

Figure 10.

Beushausen, H.D., Alexander, M.G., Mackechnie, J., “Concrete durability specifications in an international context”, Betonwerk + Fertigteiltechnik, 7, 2003, pp. 22–32. CUR Committee B82, “Durability of concrete structures in a marine environment”, Gouda, The Netherlands, 2004. Gaal, G.C.M., “Prediction of deterioration of concrete bridges”, PhD-thesis, TU Delft, June 2004. Li, Y., “Effect of spatial variability on maintenance and repair decisions for concrete structures”, PhD thesis, TU Delft, June 2004. Li, Y., Vrouwenvelder, T., Wijnants, G., Walraven, J.C. “Spatial variability of concrete deterioration and repair strategies”, Structural Concrete, Journal of the fib, Vol. 5, Nr. 3, September 2004, pp. 121–129. Meijers, S., “Computational modeling of chloride ingress in concrete”, PhD thesis, TU Delft, March 2003. Rostam, S., “Performance based design of concrete structures: the challenge of integrating strength, durability and sustainability”, Proceedings fib-Symposium “Concrete and the Environment”, Berlin, 3–5 Oct. 2001. Sitter, W.R. de, “Costs for Service Life Optimization: the Law of Fives”, Durability of Concrete Structures”, Workshop Report, Ed. Steen Rostam, 18–20 May, 1984, Copenhagen, Denmark, pp. 131–134.

Compressive membrane action.

beam-shear tests. In a beam a weak spot may have a considerable influence on the shear capacity, whereas in a slab a weak sport is compensated by regions with a higher strength. It can therefore be expected that slabs have a higher shear bearing resistance as a result of their larger redistribution capacity. f. Non-linear finite element programs can be very useful for the determination of the real bearing capacity. It is, however, necessary to calibrate those programs for the typical structure considered, to take full advantage of these numerical tools.

6

CONCLUSIONS

1. In future codes rules for the determination of the service life of structures will be given on various levels. Deterministic criteria will be given by

10

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Importance of microstructural understanding for durable and sustainable concrete Karen L. Scrivener Laboratory for Construction Materials, EPFL, Lausanne, Switzerland

ABSTRACT: This paper presents the importance of microstructural studies for understanding concrete durability. Three examples—sulfate attack; alkali silica reaction and bacteriogenic corrosion in sewers are used to illustrate the pitfalls which can arise from conventional empirical testing, particularly when blended or nonPortland cements are used. The pressure to provide more sustainable cements with lower CO2 emissions will inevitably lead to major changes in the chemistry of concrete. Therefore it is important to properly understand durability and the microstructural mechanisms involved to design appropriate performance based criteria to using the increasing diversity of concrete which will become available. (NB a similar version of this paper will also be presented at the Conference on “Microstructure Related Durability of Cementitious Composites”, Nanjing, China 13–15 October 2008).

1

INTRODUCTION

attack” plays an important role in the durability of certain structures. Damage due to sulfate interaction can result in the cracking and softening, with loss of strength of the concrete. Eminent scientists, including Vicat, Michaelis and Le Chatelier, recognised over a century ago that adverse reactions occurred between cement paste and aqueous solutions containing sulfate. The work of Thorvaldson established the role of C3A leading to ettringite formation in degradation and this led to the development in the 1920’s of “sulfate resisting cements” with low C3A content (ASTM type V). In general the use of low C3A cements has been successful in avoiding sulfate attack—such that Harboe in the Verbeck symposium of 1982 wrote:

Cementitious materials are the most used materials in the world, with more than one cubic metre of concrete being produced per person per year. It is a material of low environmental impact; the raw materials for fabrication are widely available; it is very flexible for fabrication on site and provides a strong durable material contributing to lower energy consumption of buildings in use. However, due to the huge volumes produced every year, sustainability is a big and growing issue. Since the turn of the century the production of cement has increased by about 50%, driven mainly by the growth in China. Forecasts indicate that the demand for cement will double or treble by 2050. It is estimated that cement production currently accounts for about 5% of man-made CO2 emissions. Therefore even to maintain current levels of GHG emissions it would be necessary to reduce the CO2 emission per tonne by half to two thirds. About half the CO2 emissions from cement production are due to the decomposition of CaCO3 to CaO as the first step in formation of the clinker phases. Therefore improving sustainability will almost certainly lead to changes in chemistry, for which the implications on durability are unknown. This challenge to provide durable sustainable concretes will require much better understanding of the mechanisms behind concrete degradation. This paper presents three examples of how microstructural studies are critical to understanding concrete durability. 2

“When I am asked to show someone a structure that has deteriorated due to sulfate attack, I am hard pressed to find good examples for them to see. For 40 years we have been building structure to resist sulfate attack, and the older structures that did deteriorate have now been replaced. When you add this to the fact that well over 90% of the concrete in the United States is not exposed to sulfate attack, it is not hard to understand why most people today have never seen sulfate attack [1].”

However now we need to know if blended cements can be used in conditions of sulfate exposure and it becomes clear that the test methods established for Portland cement cannot be easily generalised to determine criteria for sulfate resistance of blended cements. Worse, when sulfate resisting cements with an excellent track record of performance in the face of exposure to sulfates are tested according to common tests, they often fail the acceptance criteria. Some other questions which arise are:

SULFATE ATTACK

• High levels of slag addition are known to give good sulfate resisting blends, but what is the threshold

The degradation of concrete due to ingress of sulfate ions from the environment—so called “sulfate

11

level of slag needed, and how does this depend on the composition of the slag and cement • How is the formation of ettringite related to the expansions observed and can gypsum formation also produce expansion • What is the role of thaumasite formation in degradation due to sulfate ingress • How do additions of limestone affect resistance to sulfate attack. A series of research programmes co-ordinated by the Nanocem network [2], based on microstructural understanding, is starting to provide answers to these questions. The essential reaction implicated in sulfate related degradation is the reaction of calcium alumina monosulfate (monosulfate) to ettringite: C3A i CS i H12 + 2Ca ⇒ C3A i 3CS i H 32

2+

Figure 1. Microstructure of mortar immersed in sulfate solution in the laboratory.

2− 4

+ 2SO + 20 H (1)

around the aggregates is apparent. In this study it was observed that the rate of expansion started to increase dramatically, when the width of the outer expanded zone reached around 20–30% of the specimen width (1–1.5 mm on each side of a 10 mm prism). A further complication in the understanding of sulfate related degradation comes when the results obtained in the laboratory are compared with samples exposed in field conditions. It seems that there may be extensive decomposition of the original cement hydrates to leave a zone of decalcified C-S-H and gypsum at the surface, without any evidence of expansion having occurred. (Figure 2) Despite the complications in relating phase changes to mechanical degradation, it is clear that aluminate containing phases play a key role in determining resistance of sulfate attack. Therefore to understand the role of supplementary materials (e.g. slag, fly ash, etc) it is important to be able to understand how these change the distribution of alumina containing hydrates. With modern analytical methods—SEM, X-ray diffraction, Thermal analysis, etc. we are now able to do this. Figure 3 shows the distribution of alumina in the microstructure of a 100% CEM 1 and the same cement blended two different slags at 40 and 70%. In the 100% CEM1 most of the alumina ends up in AFm phases (monosulfate and monocarbonate), where in the presence of sulfate ions it can transform to ettringite and produce expansion. However nearly 20% of the total alumina is present in the C-S-H phase where Al can substitute for Si in the structure—in this phase it is unlikely to be able to react with sulfate to give ettringite. With the addition of 40% slag although the overall amount of alumina in the system increases, the amount in AFm phases stays about the same, with the extra alumina coming

When it is considered that the sulfate and water come from external sources, there is a volume increase of around 90% associated with this reaction. However, studies on temperature induced internal sulfate attack (commonly referred to as delayed ettringite formation—DEF) have shown that considerable ettringite formation can occur in a system without macroscopic expansion if the ettringite forms in pre-existing pores and voids. The formation of ettringite only leads to expansion when it forms from monosulfate finely intermixed with C-S-H, where the C-S-H provides the confinement needed to generate macroscopic expansion [3]. Investigations of mortar prisms in sulfate solution [4] indicate that the same mechanism operates in samples exposed to NaSO4 solutions in the laboratory. In this study, for samples exposed at both low (4 g/l) and high (44 g/l) concentrations of sulfate, three characteristic zones developed in the microstructure: − a core region characterised by a dense and compact microstructure and a good bonding between cement paste and aggregates. − a transition zone (1.0–3.0 mm depth from surface) showed first signs of changes in the microstructure of the cement paste. − a surface region (0–1.0 mm depth from surface) where the microstructure was characterised by crack formation between or around the aggregates (Figure 1). The cracks were either parallel or perpendicular to the sample surface and generally empty, expect in the samples exposed at the high concentration where some cracks were filled with gypsum. Figure 1 shows the microstructure of the outer zone, where the prevalence of cracks from gaps opening up

12

Figure 2. Microstructure of concrete immersed in sulfate groundwater for 7 years. Despite considerable penetration of sulfate and transformation of products in outer zone to gypsum and decalcified C-S-H, there is little cracking of the sample.

affect durability and design tests which reflect field performance.

3 ALKALI SILICA REACTION The second example looks at a very different mechanism of degradation—alkali silica reaction (ASR). Worldwide this is probably the second most prevalent form of concrete degradation after chloride induced corrosion of reinforcement. As time goes on more and more structures are diagnosed as suffering from ASR and it is becoming more and more difficult to find local sources of non-reactive aggregate. In the area of ASR we are currently undertaking two research projects—one to study the effect of reaction in dams and one to better identify the mechanism by which SCMs can prevent damage. Figure 3.

Distribution of alumina in Portland slag blends.

3.1

Modelling effect of ASR at a materials level

Despite the common view that alkali silicate gel forms at the edge of aggregates; in the vast majority of cases it in fact forms within the aggregates. The expansion of gel inside the aggregate causes the aggregates to crack and this leads to macroscopic expansion of the concrete. At the low levels of expansion, which may be enough to cause structural problems in unreinforced dams, the cement paste component is virtually undamaged (Figure 4). This is important information for the correct modelling of ASR in order to calculate its consequences on the properties of concrete. In the work of BenHaha [5] in was found that there was a unique relationship between the amount of reaction in the aggregates measured by image analysis and the macroscopic expansion (Figure 5). Based on this observation a model has been developed which explicitly models the evolution of expansive zones in the aggregates (Figure 6) using XFEM [6]. This model can be used to represent a whole mortar bar with an aggregate packing fraction

from the slag being combined either with magnesium in a hydrotalcite phase or in the C-S-H. The reaction of the slag lowers the calcium to silica ratio of the C-S-H and also increases the level of Al substitution in this phase. Finally in the material with 70% slag, the amount of alumina in AFm phase is drastically reduced compared to the previous two blends. This may explain the improved performance of the material with 70% slag substitution under sulfate exposure, compared to the other 100% Portland and 60/40 blends. The ongoing studies described above indicate how degradation in the presence of sulfate is a complex phenomena, for which we should not expect to obtain all the information needed from a simple empirical test, which imposes unrealistic conditions on the samples. Nevertheless, detailed study of the microstructure gives important insights into the mechanisms taking place. This will enable us soon to identify how additions of supplementary cementitious materials

13

Figure 4. Microstructure of ASR reaction–gel forms within aggregate.

of close to 70% and by adjusting the mechanical properties of the ASR gel the reaction-expansion curve of BenHaha is reproduced (Figure 5). 3.2

Avoiding ASR in new structures

There are 3 main ways recommended to avoid deleterious AAR: 1. Avoid the use of reactive aggregate. This depends on a reliable test to diagnose reactivity, which as mentioned above can be problematic with slowly reacting aggregate. However, more important is the fact, that non-reactive aggregates are becoming rare in many localities and transporting aggregate is expensive and ecologically unsound. Furthermore for many constructions, such as tunnels and dams it is an obligation to use the aggregate available on site for environmental reasons. 2. Use low alkali cement. Laboratory experience indicates that alkali contents of concrete below 4 kg/m3 should be sufficient to avoid ASR. However, this is becoming increasingly discredited as a way of avoiding ASR. Indeed, if this recommendation was valid dams should never experience ASR due to their low cement content (it is conservatively estimated that around 10–30% of dams worldwide suffer from ASR). Furthermore, low alkali cements are not environmentally friendly as they demand more energy and there are more waste by-products for their production. 3. Use supplementary cementitious materials (SCMs), to avoid deleterious expansion of concrete made with reactive aggregates. This is the most environmentally friendly options and if correctly done, probably the most reliable. Field experience in Canada and the UK indicates that concrete made with reactive aggregates and Portland cement blended with fly ash, performs well over more than 40 years. However there is very little knowledge available

Figure 5. Expansion-Reaction curve from Benhaha et al [5] left and model right.

Figure 6.

14

Model to study ASR reaction [6].

solution so as to keep the pH below the level at which deleterious AAR occurs. This hypothesis offers a new approach to establishing the combination of SCMs needed in a blended cement to suppress AAR—measure the amount and composition of the C-S-H formed. At first sight this may seem much more complicated and difficult to do, but in fact the C-S-H composition can be reliably measured in concretes (even with SCMs) form around 28 days—much sooner than 1 year. More importantly, generic behaviours can be established for given chemical combinations of cement and pozzolans in advance and then an appropriate blend chosen to give the combination producing an innocuous pore solution for a given aggregate. We are currently investigating the above hypotheses with the aim of providing better ways of formulating non reactive concretes.

of the right type and amount of SCMs needed to control the reaction of a given aggregate and also much debate on tests which can reliable predict safe combinations of aggregate and binder. Numerous laboratory studies and decades of field experience have shown SCMs to be effective in avoiding deleterious AAR in concrete made with reactive aggregates. Nevertheless the exact mechanism by which this is brought about is unclear. Furthermore not all SCMs are equally effective and the amount of substitution needed to suppress expansion with a given aggregate is unclear. Although the contribution of silica by the SCM is usually focussed on, there is some evidence that the role of alumina is also important. For example many studies have shown silica fume to have only a temporary effect on suppressing AAR, whereas pozzolans with a higher alumina content, e.g. metakaolin seem to be very effective. Several researchers have shown that the addition of SCMs brings down the alkalinity of the pore solution [e.g. 7]. The reasons for this are not totally clear. While pozzolans reduce the amount of calcium hydroxide, there are still significant quantities of this phase present in blends which do not suffer deleterious AAR. In addition, slags, which do not significantly change the amount of calcium hydroxide, are also found to be effective in suppressing deleterious ASR. Furthermore the alkalinity of the pore solution is not determined merely by the presence of calcium hydroxide which alone only gives a pH of 12.5 (much lower than that needed to provoke degradation of most AAR susceptible aggregates), but by the potassium and sodium ions present as impurities in the cement and sometimes released by the aggregates themselves. The picture is further complicated by the observation that in trials of large concrete blocks [8] it is found that deleterious expansion can occur even when low alkali cements (Na2Oeq < 0.5) are used. The most abundant hydrate phase is calcium silicate hydrate this has been shown to reversibly absorb alkalis [9]. Furthermore the amount of alkalis absorbed increases as the calcium to silica ratio of the C/S decreases. This would then appear to be a reasonable explanation for the action of pozzolans which both decrease the C/S ratio of the C-S-H and increase the amount of C-S-H, thereby increasing the capacity for absorption of alkalis. It is well known that aluminium can substitute for silicon in the structure of C-S-H in “so called” bridging sites, and it is known that this substitution of can further increase the adsorption of alkalis [10]. This would appear to explain the relative effects of silica fume and metakaolin mentioned above. Based on the above discussion, it is suggested that the suppression of deleterious AAR requires the formation of a sufficient amount of C-S-H of a composition which can absorb the alkalis ions in the pore

4

BACTERIOGENIC CORROSION IN SEWERS

Well functioning sewage network are critical to public health. Increasing urbanisation puts rising demands on existing networks and creates a need for new networks, especially in developing countries. Sewage networks provide a very aggressive environment for cementitious materials. At the same time their relative inaccessibility poses challenges for maintenance and repair. The main cause of degradation is the corrosion of concrete due to the in-situ production of sulphuric acid by bacteria. The production of sulphuric acid is the final step in a complex chain of processes which comprise bacteriogenic corrosion, illustrated in Figure 7. Sewage contains organic matter. When the flow of sewage is sluggish, and availability of oxygen limited; anaerobic bacteria breakdown this organic matter and in so doing produce H2S gas. This gas in turn serves as a food source for another group of aerobic bacteria, which colonise the walls and roof of the sewer above the water line. These bacteria breakdown the H2S to produce sulphuric acid, which dissolves the concrete. Damage is usually concentrated at the crown and water line of the sewer, Figure 7. The nature of the process means that the amount of acid produced is limited by the extent of activity of the bacteria so the phenomenon is very poorly modelled by tests in which mineral acid is used, especially where the quantity of acid is large in comparison the amount of cementitious materials. In fact the colonisation of the concrete surfaces is progressive. The bacteria responsible—Thiobacilli is a whole family of bacteria, the various strains of which thrive at different pH levels. Initially the surface of the concrete is highly alkaline, due to the nature of the hydrates, particularly, Portlandite, Ca(OH)2. The first colonisation is by the strains of bacteria capable of flourishing at the highest pH. Through the production

15

6

4

5 H2S

3

6

2

1 Figure 8. State of concrete pipe section in test sewer after 12 years. Portland/siliceous section (left) is heavily corroded, in parts the earth behind the pipe is visible. CAC/ siliceous section (right) is relatively unaffected; only black part around the water line shows up to 10 mm of erosion.

Figure 7. H2S Bacterial corrosion. 1 – Sulphates produced by bacteria; 2 – Reduction of sulphates to sulphide by anaerobic bacteria living in the slime layer; 3 – Turbulence ejects H2S dissolved in the effluent into the pipe atmosphere where it is carried by convection to the pipe crown; 4 – Oxidation of H2S to sulphur; 5 – Transformation of sulphur to sulphuric acid by aerobic bacteria; 6 – Destruction of the concrete.

of sulphuric acid they lower the pH level of the concrete surface and facilitate colonisation by new strains, adapted for lower pH conditions. The final stage of the process is colonisation by thiobacilli thioxidans, the most aggressive, these bacteria thrives best at a pH of 2 or less and is capable of generating enough sulphuric acid to bring the surface pH down to 1, which is highly aggressive to all cementitious materials. Even when colonies of thiobacilli thioxidans are well established the process of dissolution of the concrete tends to neutralise the acid created. Of the major oxides present in cementitious materials, CaO; SiO2; Al2O3 and Fe2O3 (C, S, A, F for short); S and F are virtually insoluble even at very low pH. Lime (C) is readily soluble at all acid pH and each mole of calcium which dissolves will neutralise 2 moles of acid: Ca(OH)2 + 2H+ ⇒ Ca2+ + 2H2O

(2) Figure 9. Test results from Chamber to study bacteriogenic corrosion[13]. Mass loss left and surface pH right. SC and CC, dotted lines are calcium aluminate cement mortars. The samples with solid lines are all Portland based systems, OPC = plain Psortland; srPC = sulfate resisting Portland cement and BFC = Portland/Blast Furnace slag blend.

(equivalent equations can be written for C-S-H which will also decalcify) Alumina is virtually insoluble above pH4, but below this pH dissolves to neutralise 3 moles of acid. Field experience shows outstanding performance of concretes made with calcium aluminate cements subject to bacteriogenic corrosion (Figure 8) [11]. However laboratory tests, using strong acids fail to show this improved performance due to the fact that they do not include the colonisation process of the bacteria. There is now overwhelming evidence, both from the field [12] and in laboratory studies [13], that

the pH on the surface of CAC subject to bacteriogenic corrosion only drop to just below 4 (as opposed to around 1 for Portland concrete) due to suppression the growth of bacteria (Figure 9). In such condition hydrous alumina has very low solubility and blocks the pore structure (Figure 10) [14].

16

REFERENCES [1]

[2] [3] [4]

[5]

[6]

Figure 10. Microstructure of Concrete from 30 year old sewer lining made of calcium aluminate cement (CAC).

5

[7]

CONCLUSIONS [8]

The examples presented here illustrate the importance of microstructural studies to understand the mechanisms leading to degradation of cementitious materials. Such studies are essential to be able to produce durable concrete. Furthermore the pressure to provide concrete with lower associated CO2 emissions means that concrete will be much more diverse in the future. The changes in microstructure brought about by mineral additions (or supplementary cementitious Materials, SCMs) or nonPortland cements mean that conventional empirical tests are unreliable or outright misleading to determine the performance of concrete in the field.

[9] [10] [11] [12]

ACKNOWLEDGEMENTS

[13]

I am very grateful to the students and associates who provided some of the figures for this paper from their on-going work and to the sponsors who have financed much of this research (names in brackets)—Thomas Schmidt (Cemsuisse), Aude Chaberlie (EU/Nanocem), Emmanuel Gallucci, Gwenn le Saout, Victor Fernandez (EU/Nanocem), Cyrille Dunant (Swiss Federal Office of Energy).

[14]

17

Harboe, E.M., “Long-Time Studies and Field Experiences with Sulfate Attack,” In P. Klieger (ed) George Verbeck Symposium on Sulfate Resistance of Concrete, SP-77, American Concrete Institute, Farmington Hills, MI, 1982, 1–20. http://www.nanocem.org Taylor, H.F.W., Famy, C., Scrivener, K.L., “Delayed Ettringite Formation”, Cem. Concr. Res, 31 (2001), 683–693. Schmidt, T., Lothenbach, B., Romer, M., Neuenschwander, J., Scrivener, K.L., “Physical and microstructural aspects of sulfate attack on ordinary and limestone blended Portland cements”, submitted to Cem. Concr. Res. Benhaha, M., Gallucci, E., Guidoum, A. Scrivener, K.L., “Relation of expansion due to alkali silica reaction to the degree of reaction measured by SEM image analysis” Cem. Concr. Res. 37 (8): 1206–1214. Dunant, C., Scrivener, K.L., Modelling ASR reaction, submitted to Cem. Concr. Res. Shehata, M.H., Thomas, M.D.A., Bleszynski, R.F. “The effects of fly ash composition on the chemistry of pore solution in hydrated cement pastes”, Cem. Conc. Res. 29 (1999) 1915–1920. Private communciation, K. Folliard and J. Ideker, U. Texas, for blocks exposed in field site. Hong, S.-Y., Glasser, F.P., “Alkali binding in cement pastes: Part I. The C-S-H phases”. Cem. Concr. Res. 29 (1999), pp. 1893–1903. Hong, S.-Y, Glasser, F.P., “Alkali sorption by C-S-H and C-A-S-H gels Part II. Role of alumina”, Cem. Conc. Res. 32 (2002) 1101–1111. Goyns, A.M., Calcium aluminate cement linings for cost-effective sewers : In Calcium Aluminate Cements 2001, 2001, 617–631. Lamberet, S., Guinot, D., Lempereur, E., Talley, J., Alt, C. “Field Investigations Of High Performance Calcium Aluminate Mortar For Wastewater Applications” In Calcium Aluminate Cements. Proceeding of the Centenary Conference, Avignon, France, 30 June—2 July 2008, IHS BRE Press, 2008, 269–277. Ehrich, S., Helard, L., Letourneux, R., Willocq, J., Bock, E., Biogenic and chemical sulfuric acid corrosion of mortars, Journal of Materials in Civil Engineering/November 1999, 340–344. Scrivener, K.L., Cabiron, J.L., Letourneux, R. “High Performance concretes from calcium aluminate cements”, Cem. Concr. Res. 29 (1999) 1215–1223.

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Advanced NDT methods for the assessment of concrete structures H. Wiggenhauser Federal Institute for Material Research and Testing (BAM), Berlin, Germany

ABSTRACT: The development of methods for the Non-Destructive Evaluation (NDE) of concrete and masonry structures has made considerable progress in the past two decades. Technology and knowledge transfer from other areas such as materials testing and medicine has been the main source for innovation, supported through multidisciplinary research by scientists, technicians and support personnel. A versatile toolbox of methods for the investigation of RC structures has emerged from this effort. In the area of simulation and 3D-reconstruction of ultrasonic and radar data, progress was driven by more powerful computers and development of state of the art software for 3D re-construction and visualization of data. Bringing these developments into the state of practice, into everyday use on construction sites by trained engineers rather than researchers still remains a major challenge. To earn the trust of industry, the reliability of these Non-Destructive Testing (NDT) methods should be established and adequately verified to prove that their application is worth the additional expense and effort. Validation of NDT methods has to prove that the customer requirements are met by the test. A focused effort is needed to implement a validation methodology for NDT methods in civil engineering, which is widely accepted and used. In an effort towards this goal, a number of reference specimens have been developed over the years at BAM and other locations, acting as a standard for specific testing tasks. Those specimens are available to the NDT community for comparing their methods and instruments under independent conditions. The development and application of NDT methods for concrete structures requires more than just using off-the-shelf testing equipment. Existing test methods must be improved for more reliable use. New methods borrowed from other areas of testing such as medicine or geophysics should be further developed. Using an array of methods and combining their results, preferably through data-fusion algorithms has a lot of potential. Over the years, the NDT group at BAM has developed and accumulated valuable expertise in some of the above-mentioned areas. The focus of the research and developments efforts at the NDT-CE division at BAM has been on methods used to evaluate the geometrical properties of structure such as thickness, location of components as well as detection of the presence of voids, delaminations, ungrouted tendon ducts or cracks.

1

ULTRASONIC PULSE ECHO (UPE)

systems developed at BAM, automated testing utilizing these transducers became a reality. Based on these experimental developments, reconstruction processes developed in other areas could be utilized

It became clear very early that ultrasonic pulse echo on concrete requires very low frequency transducers (Neisecke 1991) to overcome the scattering problem in coarse aggregate concrete. Multi position experiments and synthetic aperture data analysis led to reliable thickness measurements of concrete slabs. In order to reduce testing times and to ease experiments, arrays of ten transducers were used. In the mid 90s a new type of transducer was developed in Russia (www.acsys.ru 2006) which did not require special coupling to the concrete surface (Fig. 1). Based on transversal excited ultrasound, testing became much easier to perform with the availability of these commercial products. The individual transducers have ceramic pin contact to the concrete, 12 transducers each are combined and form a large aperture for transmitting and receiving the signals. Scanning of lines and areas was possible with the A1220 sensors in reasonable time, due to the fact that no special coupling was required. With the scanning

Figure 1. Ultrasonic device with point contact 55 kHz shear wave transducers. 12 point contacts used for transmission of ultrasound, 12 for receiving (ACSYS).

19

for ultrasonic data collected within a 2D aperture. The result of such tests is a 3D reconstruction of the volume below the test surface (Streicher et al. 2005). As an example of 3D imaging of concrete bridges a C-scan from a bridge is shown in Figure 2. It clearly shows tendon ducts in a depth of 10 cm over the entire length (10 m) of the tested web. The 3D reconstruction of ultrasonic echo data is based on the well known SAFT (Synthetic Aperture Focussing Technique) algorithm which calculates

Figure 4. Air coupled ultrasonic probes mounted in front of concrete block for pulse echo measurements. Aluminium sheets mounted for air wave suppression (BAM). Figure 2. Result from ultrasonic investigation of PT concrete bridge. Shown is the web with tendon ducts located in depth range around 10 cm. Darker colors indicate higher acoustical contrast.

the phase shift for each single measurement for each volume element. Calculation times have dropped dramatically due to faster processors and advanced programming techniques. Quasi real-time imaging is becoming a reality. Very recently a new ultrasonic device has become available which allows taking measurements and imaging a scan within seconds. Based upon the same point contact transducers, measurement time could be reduced to increase performance. With this unique device, quasi real-time imaging of concrete using a hard-held device is available now (Fig. 3). The coupling problem for ultrasound can be minimized using air coupled transducers. In transmission experiments this technique has shown its capabilities (Hillger 2004). However, in pulse echo mode, experimental obstacles are apparent. The separation of the echo signal from all other signals (surface echo, direct coupling between transmitter and receiver, surface wave) is quite difficult. Further research is needed to utilize this promising technique. The advantage of this technique is the superior testing speed which could be used for testing large areas in relatively short time. Figure 4 shows the experimental setup for testing concrete with air coupled ultrasound in pulse echo mode. 2

IMPACT-ECHO (IE)

Impact-Echo is an acoustical test method based on mechanical excitation of sound waves in concrete (Sansalone & Streett 1997, Pessikki & Olson 1997). Resonant vibrations of the surface near the point of contact are recorded and identified by transforming the measured time signal into the frequency domain (Fig. 5). Dominant frequencies are assigned to depth values by applying the well known IE formula

Figure 3. Ultrasonic array for real time imaging. Ten Elements are aligned to form a linear aperture. B-scan imaging within 4 seconds. Shear wave transducers, 55 kHz. (www.acsys.ru)

20

Figure 5. Impact-Echo signal: Time domain data (top) and frequency power spectrum (bottom). Thickness signal at 6.81 KHz, scanner resonance at 1.2 KHz.

d = vL / 2 f

Figure 6. Impact-Echo thickness gauge by Olson Instruments (right). Impact is generated by a solenoid, coupling to the surface through elastic material. Frequency range: 1–25 KHz. Rolling JE probe shown on the left. Impact is generated every inch (2.54 cm) for continuous measurements. (www.olsoninstruments.com)

(1)

with vL being the P-wave velocity and f the measured frequency. The wave speed must be determined at each concrete through calibration at a position of know thickness or by measuring on a core. Impact-Echo has been successfully applied to concrete thickness measurements and defect localization in many cases. For this purpose a number of commercially available Impact—Echo devices are available (Figs. 6, 7). These devices are for manual point testing and can be distinguished by their impactors and sensors. Sensor coupling is assured through lead caps or plastic material, impacts are generated either through hand held steel balls of different diameters or by an electrically driven solenoid. Latter has the advantage of being easily reproducible when the IE head is not moved between subsequent impacts. The diameter of the steel balls determines the frequency content of the impact, given by the formula (2)

Figure 7. Impact-Echo system of Germann. Impact generated manually with steel balls. Coupling to surface through thin lead sheets. Frequency range: 1–50 KHz. (www. germann.org)

with d being the diameter of the steel ball in mm and fmax being the maximum useable frequency in kHz (www.olsoninstruments.com 2006). The duration of the impact ultimately determines the maximum frequency of the impact. This equation may only be used for concrete with perfect surface. In practical cases, the surface must be checked for dirt and soft areas. The impact will generate different waves depending

upon the exact position where the hit takes place. If the impactor directly hits a hard aggregate, which may be hidden by a thin cement paste cover, the frequency range may be different from a hit into the softer cement paste. At BAM this point method was improved into a scanning test method to visualize test results as an Impact-Echogram, similar to a B-scan in ultrasonic

f max = 291 / d

21

pulse echo or a GPR radargram. The amplitude of the frequency power spectrum is displayed colour coded or in greyscale over the position of the measurements and frequency (Fig. 8). A handheld IE test device for scanning measurements was developed by Olson Instruments (www. olsoninstruments.com) as shown in Figure 6. A self driven scanner for horizontal surfaces was developed at BAM for both, Impact-Echo and Ultrasonic sensors (Fig. 9). With automated testing in dense measurement grids Impact-Echo measurements can show very detailed the thickness of a concrete slab. When scanning is extended to a 2-dimensional grid, IE-results can be shown as 3-D images as shown in Figure 8. It was successfully shown, that grouting defects in tendon ducts can be identified by the apparent increase in thickness of the concrete slab at the position of the duct. However, validation of this test is still

Figure 9. BAM NDT Stepper FOR Impact-Echo and Ultrasonic Pulse Echo.

pending to define the limits of this test method and making it a reliable tool in quality control (Fig. 10). In a recent study at University Karlsruhe the reliability of Impact-Echo for the localization of honeycombs in concrete slabs has been researched (Müller et al. in prep.). It could be shown, that Impact-Echo can directly and indirectly determine the presence of a honeycomb. The indirect method identifies a defect by the missing thickness signal. However, it should be clearly pointed out, that in this study the honeycombs could only be reliable located with scanning ImpactEcho. Point measurements were not showing reliable results (Fig. 11). Impact-Echo experiments on specimens with small dimensions may be interpreted with care. Since most of the impact energy is transformed into surface and shear waves, reflections of those from nearby edges cause interference patterns on the surface. Such signals can easily being mistaken for thickness signals or void indication if not validated through measurements at different points: Those must be separated well enough to exclude that the signal is triggered by geometry effects. Before testing, the operator has to assure that the system setup is meeting the requirements of the test. The sampling rate of the data acquisition system must be high enough to measure the highest envisaged frequency (Nyquist frequency). Also, the impactor must be able to generate the frequency for the given test problem. The sensor must work accordingly and the data acquisition system must record the signal properly. The dynamic range of the AD-converter becomes an issue when the initial signal after the impact is very high but the following resonances are highly attenuated. With an AD-converter of low dynamic range (e.g. 8 bit) the signal amplitudes may not be recorded properly and make the data less accurate. Data analysis of Impact-Echo signals includes time and frequency domain filtering. Typically a high pass filter is applied to remove the signal offset and low frequency resonances within the system. A time band

Figure 8. Thickness of concrete slab of 5 × 4 m2 (BAM reference specimen GBP. Top: plan of section with thickness variations. Middle: 3D plot of measurements in very dense grid (2 × 2 cm²). Bottom: B, C and D Scan of same measurements with signals of various thicknesses.

22

Due to the nature of electromagnetic waves it can be used for concrete and masonry structures but radar waves cannot penetrate metals. Any metallic layer, e.g. a metal sheet or a metallic tendon duct is an inpenetratable boundary (Fig. 12). However, reinforcement which leaves gaps between the rods can be passed by radar waves within limits. The propagation of electromagnetic waves in solids depends on the dielectric constant (relative permittivity) of the material. The permittivity of the material basically describes how the dielectric field in the material is following the applied electric field of the waves. The permittivity is a complex number. The real part determines the propagation velocity in the material

Figure 10. Impact-Echo testing of grouting defects in ducts. Top: drawing of defect location. Middle: B-scan along duct. Bottom: C-scan showing apparent increase of slab thickness at defect locations. (BAM, unpubl.)

vL = c0 / ε r

with c0 being the velocity of light in vacuum. The imaginary part is largely responsible for the frequency dependent attenuation of the radar waves. Radar waves are polarized, having the effect of a preferred direction. This is especially true for linear objects like reinforcing bars, which show up with highest intensity when the polarization of the antennae is parallel to the direction of the rebar. For this reason the localisation of rebars should always be performed using both polarization directions (Fig. 13). This is done by scanning a surface in two perpendicular directions and combining the two data sets. The availability of antennas with different frequencies allows choosing the penetration depth in accordance with the requirements of the test. Low frequency antennas penetrate deeper into the material, but have lower depth resolution. High frequency antennas have a higher depth resolution but less penetration depth. For application on structures mostly bow-tie antennae types are used. They can be slid over the surface for conducting the test. Horn antennas are mainly being used for road and train track measurements. They have a distance of several decimeters from the surface and can be mounted on vehicles for fast inspection.

Figure 11. Honeycombs in concrete slabs measured with scanning Impact-Echo. Left: Pre-fabricated honeycombs of different sizes. Top: Drawing of specimens with different reinforcing ratios. Middle: Pictures of specimens before casting. Bottom: Impact-Echo results, darker colour indicating larger thickness. Thickness is apparently increased over honeycombs.

pass excludes signals from times where the ImpactEcho signals should have attenuated. If the time band pass is set too small, frequency resolution is lost. Wavelet analysis has been used for data analysis as well and can serve very well for time-frequency analysis. Recently, the Hilbert-Huang transform has also been utilized for the successful analysis of ImpactEcho signals (Algernon 2006).

3

(3)

GROUND PENETRATING RADAR (GPR)

Ground Penetrating Radar has emerged from geophysical applications and has become a versatile and powerful instrument for civil engineering applications. GPR can be applied in general without surface preparation by sliding the antenna over the surface. The fast data collection allows covering a large area in reasonable time.

Figure 12. reflector.

23

Radar parabola signature when passing over

29.4 4

infrared radiator

Temperature in °C

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22 0

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Figure 14. Thermogram and temperature decay curves of carbon fiber reinforcements after impulse heating (1 min, 7.2 kW/2.25 m2.

been applied to carbon fiber reinforcements and shallow defects in concrete with good success. Figure 14 shows a thermogram of a concrete specimen with carbon fiber reinforcements which has some artificial defects. The temperature decay curves show a distinct difference between sound and voided areas (Helmerich et al. 2006). The quantitative analysis of the temperature after heating can either be done in the time domain by curve fitting or in the frequency domain. Later analysis is referred to as Pulse-Phase-Thermography and has certain advantages. Surface effects are largely eliminated in the PPT images, from the phase angle the depth of defects can be derived (Arndt 2006).

Figure 13. Influence of polarization on the detectability of tendons.

Typical applications for radar on concrete structures are (Maierhofer et al. 1995): − − − − − − −

Location and concrete cover of reinforcement Location and concrete cover of tendon ducts Thickness measurements of beams and slabs Moisture detection areas Localization of voids Localization of delaminations Measurement of the thickness of layers

5 TOMOGRAPHY

Conducting radar tests and the interpretation of the results is supported by advanced instrumentation which is commercially available. More important are powerful computer systems which can handle large amounts of data in reasonable time. Software tools for different purposes have been developed and are available for service providers or research. GPR applications for non-destructive testing in civil engineering have been a success story for many years now. More and more information has become available and known about the capabilities and limitations of such systems. New systems emerging from research will bring array systems into the market, allowing to scan a wide path with one sweep. An early system of this type has been developed by the Federal Highway Administration in the nineties (Scott et al. 2000).

Tomographic investigation of concrete and masonry structures is time consuming and complicated. X-ray tomography is not very common for large objects, acoustic tomography has been tried and results have been reported (Colla et al. 1997, Patitz & Illich 2000). New sensors (ACSYS ultrasonic device A1220 with dry coupling) and the availability of automated measurements with the scanner systems allow new approaches to tomography of large civil engineering structures. A more general approach of tomography is envisaged for this research, where large sensor arrays record signals which are transmitted or reflected. The combination of methods (radar, ultrasonic and sonic) allows a reconstruction of the volume under investigation on a much larger data basis than from manual testing with a single method. Figure 15 shows the results of tomographic measurements and inversion of the data on a masonry specimen.

4 THERMOGRAPHY

6

Impulse Thermography is widely used in nondestructive testing of composite materials, such as aircraft components. In civil engineering, thermography is mostly used in the quasi static mode, where the ambient temperature distribution is utilized to test an object. Impulse heating of a surface, followed by a recording and analysis of the temperature development over time and position leads to new applications. In recent studies, impulse thermography has

LIBS

The application of the Laser-induced Breakdown Spectroscopy (LIBS) for the quantitative evaluation of chlorine and sulphur content in building materials is being researched at BAM (Fig. 16). LIBS is a method which has the ability to evaluate the elements of the concrete matrix to get information of the bulk material and at the same measurement

24

R1 0

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Figure 15. Result of acoustic tomography of a masonry specimen with two voids (missing bricks). Forward modeling: acoustic waves. Transmitter T and Receiver R every 5 cm. Bricks/cement: v = 3000 m/s, ρ = 1.7 g/cm³, Voids: v = 300 m/s, ρ = “1” g/cm³, Inversion: 5 × 5 cm grid size, modified velocity model, five iterations, curved rays. Figure 17. Top: Typical spectrum of cement measured under helium containing chlorine, the spectral lines are labelled. Bottom: Comparison of spectra of cement and aggregate.

921.6 nm. For every measurement the ratio of a calcium to an oxygen spectral line is evaluated to take care for the heterogeneity of the concrete and relate the results directly to the cement content (Fig. 17). 7

ULTRASONIC ECHO AND IMAGING TECHNIQUES

Foundations have only one-sided access and require echo methods like ultrasonic echo (US-echo) for testing. The investigation of the above-mentioned slab has been carried out with an array of 24 dry point contact transducers working with transverse waves excited with a centre frequency of 15 kHz (ACSYS A1220). For a reliable imaging of the complex geometry an automated transducer positioning system (scanner) has been used. Data (8000 measuring points) have been recorded and processed with the help of a reconstruction calculation. The so-called SAFT (Synthetic Aperture Focusing Technique) focuses signals received at many aperture points by coherent superposition, yielding a high-resolution image of the region of interest. The following images are results of SAFT-reconstruction. Various sections through the

Figure 16. Experimental LIBS set-up in the BAM laboratory.

also information about trace elements. The method can be applied on surfaces without any preparation and the results are available directly after the measurement. LIBS is currently developed as a tool for on-line evaluation and for quality assurance during repair works. LIBS uses a pulsed infrared laser for plasma generation and a spectroscopic system for the analysis of the emitted plasma radiation. For the quantitative evaluation of chloride ingress the chlorine spectral line at 837.6 nm is used. The sulphur spectral line is for sulphate ingress at

25

reconstructed data volume can be processed and layers with significant reflections become obvious and visualise internal objects and geometry. Figure 18 shows the section parallel to the surface at a depth of 75 cm. The expected reflection of the back wall at that depth is clearly visible. In upper left and upper right corner (x/y 960/860 mm, x/y 4000/850 mm) two small areas show no back wall reflection. This is where the pile heads are located because the signals propagate from the slab further into the piles and are not reflected at the depth of the back wall. The limits of the thickness measurement depending on the reinforcement ratio become clearer looking at cross-sections shown in Figure 19. For the high reinforcement ratio shown at the top of Figure 20 (section b-b) only in the non-reinforced sections a clear back wall signal at depth of 75 cm and 125 cm appears. In the sections with upper and lower reinforcement a strong reflection in the surface near depth is visible. A weak signal at the 75 cm section allows depth estimation up to that depth. If the 28 mm diameter reinforcement is placed only in the lower level, as shown in section c-c at the bottom of Figure 19, the reinforcement bars produce a reflection in addition to the back wall reflection a few centimetres above. Depth measurement up to 125 cm is possible with this reinforcement ratio. Cross-section d-d shown in Figure 21 also reveals the geometry of the slab and the location of the piles. The back wall reflection at 75 cm and the bottom of the strip foundation at 125 cm are clearly visible. Also the detected width of the strip foundation of 50 cm and its location agree well with reality. The interrupted back wall echoes at the depth of 75 cm between x = 800 mm and 1100 mm and x = 3850 mm and 4150 mm mark the location of the pile heads.

Figure 19. Drawing of the designed foundation slab with varying thickness and reinforcement ratio.

Figure 20. Section b-b, ∅ = 28, s = 10 cm (upper and lower reinforcement).

Figure 21. Cross section d-d revealing the slab geometry and location of the pile heads.

Reliable thickness measurements have been possible up to 75 cm for areas with high reinforcement ratio (28 mm diameter at 100 mm spacing, crosswise, upper and lower reinforcement layer) and up to 125 cm with lower reinforcement ratio.

8 AUTOMATION

Figure 18. 75 cm.

Automation of non-destructive testing methods helps to overcome the disadvantages of manual point testing. Point measurements must be repeated several times for statistical proof of the results. Reproducibility is not always given in rough test environments and difficult surfaces. Furthermore test personnel get tired during long testing periods.

Section parallel to the surface at depth of

26

At BAM several automation devices have been developed for various applications: automation of point testing methods with physical contact during testing (impact-echo, ultrasonic echo) and those with non-contact testing (radar, rebar detection). The test devices are mounted on a two-axial frame which moves to a pre-defined position and triggers the measurement. Pneumatic installations are used to couple the sensors to the surface until the test is finished. The entire test procedure is computer controlled. All data about the test procedure, the test object and the applied methods is stored for documentation. Interface with test equipment is done by utilizing existing interfaces or in cooperation with manufacturers. Typical test times required for ultrasonic testing is 1.5 h per square meter at a spacing of 2 cm × 2 cm between test positions. Radar is much faster due to the non-contact nature of the test, at a spacing of 5 cm between lines the required time is 5 min per square meter. Scanners have been developed for both horizontal (bottom and ceiling) and vertical flat surfaces. The large scanner covers an area of 10 m × 4 m, smaller versions exist for 1.8 m × 1.8 m areas. Recent successful tests have supported the concept of vacuum mounting of a lightweight frame. This version does not require fixed mounting to the concrete surface (Fig. 22). Mounting and test times have been successfully reduced using lightweight frames and fast, network based programs which can be easily adapted to testing scenarios. Fully automated testing of a 10 m² area concrete area within a day is within reach. So far the biggest problem is accessibility, which should be eased by the vacuum attached lightweight scanner which can be mounted from a cherry picker of snooper truck. For horizontal testing the “BAM NDT-Stepper” was constructed (US Patent 2006/0191358 A1). A small vehicle moves with constant speed along a straight line. The sensor is pneumatically pressed to the surface in equally spaced positions along this line. The test position is maintained exactly during the test while the vehicle is moving. Spacing can be as long as 10 cm, the speed is defined by the spacing and the test time required per position (typical speed is 1–3 min/m). Air pressure is generated by an internal miniature pump. An internal battery allows testing time of more than four hours for typical test scenarios. (Fig. 23). New versions of the “BAM NDT-Stepper” include mounting of two sensors on either side of the vehicle. Array systems can be mounted on an extension of the mechanics which assures the positioning of the sensors. 9

BRIDGE TESTING Figure 22. BAM NDT-Scanners for concrete surfaces. Top: Vacuum scanner 1.2 × 1.2 m². Middle: Horizontal surfaces 10 × 4 m². Bottom: Vertical surface 10 × 16 m².

Non-destructive assessment of existing bridges is the most important application of the NDT developments at BAM. Bridges are a valuable infrastructure asset.

27

Figure 24. Visualization of the outer reinforcement layer of a box girder web from radar data. Measurements were taken in two perpendicular polarization directions, data was fused and reconstructed.

Figure 23. BAM NDT-Stepper with mounted ultrasonic sensor (ACSYS A1220).

Germany alone has app. 37,000 rc-bridges within in the highway system. Interruptions to the flow of traffic should be minimized for inspections and maintenance. In addition to the visual inspection based regular inspection, special testing is required in some cases where damages are possible but not identifiable visually. For prestressed concrete structures the investigation of tendon ducts is one of the most essential test problems. The localisation of tendon ducts, the determination of the thickness of the concrete cover and especially the detection and quantification of ungrouted areas inside the ducts are relevant questions, which can be solved by the combined application of the imaging echo methods radar, impact-echo and ultrasonic echo. Other points of investigations are detecting the reinforcement and determining the thickness of structures if accessibility is given from one side only. Investigations with the different test methods utilizing the scanning systems were carried out on several post-tensioned concrete bridges. The improvement of the measurement conditions and the more comprehensive data acquisition allowed a much simpler interpretation of the results. Reinforcement bars, which are arranged perpendicular could be visualised with a high resolution (Fig. 24). Tendons in the investigated bridge structures could be localised with the 1.5 GHz-antenna at measurement depths up to 16 cm and with ultrasonic echo at measurement depths up to 55 cm (Fig. 25). The position of couplings between tendons could be determined and hints to not completely filled tendon ducts could be given. The backside of structures were detected with ultrasonic and impact-echo up to a depth of 83 cm. The scanning systems show a reliable performance under field conditions. The time and effort could be considerably reduced by automation in comparison with manual measurements (Streicher et al. 2005).

Figure 25. Arrangement of tendon ducts in the cross section of a box girder web. Left: According to construction plan. Right: Located at a SAFT-B-projection by ultrasonic echo.

10 VALIDATION Validation is used in many technical fields, e.g. pharmacy, aviation, to furnish the technical proof that the customer’s requirements are fulfilled. Subject of the validation is the whole process of a successful solution for a testing problem. Well-documented validation improves the application safety for a service provider and makes his service transparent for the client. The client knows exactly what he buys and that he has full benefit from the results. According to DIN EN ISO 17025 (DIN EN ISO/IEC 17025, 2004) Validation is the confirmation by testing to furnish the proof that the requirements for a certain intended use can be fulfilled. The client—in some cases together with the service provider—expresses his/her requirements considering all necessary boundary conditions. If a testing method— suggested by the service provider—succeeds to satisfy these requirements the validation process is completed. According to Figure 26 the validation process consists of three steps: − Characterization of a testing methods − Requirements of the client − Proof that the client’s requirements have been fulfilled

28

Figure 26. Validation process.

Figure 26 also shows this process: characterization is done by regular testing in research and development and evaluating the results. Calculating the uncertainty of measurement, determining the precision and accuracy and limit of detection etc. are possibilities to characterize a method under certain boundary conditions (method, device, environment etc.). The customer requirements are also expressed by the properties mentioned before. They are influenced by time, costs, accessibility etc. All this together forms the individual testing problem. A testing method under the characterized methods that is suitable for the testing problem is marked by the intersection of characterization and customer requirement. The validation is at the end of the process if the proof is furnished that customer’s requirements are satisfied. This does not mean that a validation always has to be carried out at the customers testing object because this will often be impossible. The idea of the process is to provide a close characterization of the method considering a wide range of boundary conditions that fit for the customer’s testing problem. In one case it might be possible to predict precisely that the customer needs will be satisfied. In another case the results, e.g. the expected uncertainty, might be too close to the customer’s defined limit and the success of a test cannot be guaranteed. In case that the characterization of a method is not sufficient, further—but few—measurements considering the unknown boundary conditions have to be carried out. Validation has to be carried out individually for a specific testing problem because there is obviously no general validation for a testing method. The Guide to the Expression of Uncertainty in Measurement (GUM) (Guide to the Expression of Uncertainty in Measurement 1995) provides a good methodology that can be applied in the NDT-CE field. According to the GUM the uncertainty of thickness measurement of 70 to 120 cm thickness and having areas of different reinforcement ratio was measured. The testing was performed with an automated testing device using dry point contact sensors (ACSYS A1220). The measured data was first assessed as raw data and later assessed after improving it by creating the envelope of the transit time curve. In Figure 27 the effect of improved data analysis for different reinforcements is shown. We assess the data according to the GUM mentioned in the previous chapter the total

Figure 27. Total standard deviation of thickness measurement with ultrasonic-echo (transverse waves) in relation to thickness, reinforcement and data assessment.

standard deviation for the thickness measurement in relation to thickness, reinforcement and data assessment can be calculated (Gaussian error propagation) as given in Figure 27. The results shown in Figure 27 are the results of a systematic characterization. It helps the engineer to estimate up to which depth and to which reinforcement he may expect reliable results. If the engineer just needs a quick overview to decide if thickness is 70 cm or 120 cm he/she may be satisfied by assessing raw data. If he/she reliably wants to know the slab thickness, e.g. in case of a static calculation, he/she will get results with a standard deviation less than 5% only by using “improved data” (in this case creating the envelope) and up to a reinforcement of ∅ 12 mm with 15 cm spacing (upper and lower layer). 11

NDT-CE COMPENDIUM

Information about the availability and performance of NDT-CE methods is not readily available. The BAM in Germany has compiled a compendium on such methods (ZFPBau-Kompendium) and published it later in the internet: www.bam.de/zfpbau-kompendium.htm. The English translation is under way and will then also be freely available worldwide on the same address. 12

REFERENCE SPECIMEN

Information Improvements of NDT methods can best be shown through reference tests at suitable specimens. For this purpose, a number of large scale reference test specimens have been constructed at BAM. Well defined test problems are incorporated into those, well separated from other defects. Special care is taken of size effects, e.g. the geometry effects in impact echo. The specimens typically have a size of at least 2 m × 1.5 m × 0.5 m. The size is mainly limited by the ability to move the specimens with a fork lift.

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Arndt, R. 2006. Rechteckimpuls-Thermografie im Frequenzbereich. Berlin: Dissertation eingereicht bei der TU. Colla, C., Das, P., McCann, D. & Forde, M. 1997: Sonic, electromagnetic and impulse radar investigation of stone masonry bridges. in: NDT&E International Sonderheft Vol. 30 4: 249–254. DIN EN ISO/IEC 17025: Allgemeine Anforderungen an die Kompetenz von Prüf- und Kalibrierlaboratorien (ISO/IEC 17025:1999); Ausgabe: 2000-04, Beuth Verlag, Berlin. Guide to the Expression of Uncertainty in Measurement: deutsche Übersetzung 1995. Leitfaden zur Angabe der Unsicherheit beim Messen, Beuth-Verlag, Berlin. Hasenstab, A., Krause, M., Rieck, C. & Hillemeier, B. 2004. Materialuntersuchung an Holz mit niederfrequenter Ultraschall-Echotechnik, Berichtsband 89-CD, DACH— Jahrestagung 2004, 17–19. Mai 2004, Salzburg. Helmerich, R., Niederleithinger, E. & Wiggenhauser, H. 2006. Toolbox with Nondestructive Testing Methods for Condition Assessment of Railway Bridges; Transportation Research Record: Journal of the Transportation Research Board, no. 1943, Transportation Research Board: 65–73. Hillger, W. 2004. Imaging of Defects in Concrete Components with Non-Contact Ultrasonic Testing, 16th World Conference on NDT, 30. August–03. September 2004, Montréal, Canada. Maierhofer, C., Borchardt, K. & Henschen J. 1995. Application and Optimization of Impulse Radar for Nondestructive Testing in Civil Engineering and further articles therein, Proceedings of the International Symposium on Non-Destructive Testing in Civil Engineering, Deutsche Gesellschaft für Zerstörungsfreie Prüfung e.V., Eds: Schickert, G. & Wiggenhauser, H., Berlin: 663–672. Müller, H.S., Fenchel, M., Wiggenhauser, H., Maierhofer, C. & Krause, M. in prep.: Zerstörungsfreie Ortung von Gefügestörungen im Beton, DAfStB Forschungsbericht. Neisecke, J. 1991. Ultraschallprüfung im Bauwesen—eine kritische Wertung, in: Schickert, G. (Hrsg.); “Zerstörungsfreie Prüfung im Bauwesen”, Tagungsbericht Int. ZfPBau-Symposium 27. February—1. Macrch 1991 in Berlin, BAM, DGZfP, Berlin: 182–191. Patitz, G. & Illich, B. 2000. Blick ins Innere: Zerstörungsarme Untersuchungen: Voraussetzung für ein substanzschonendes Sanierungskonzept/Teil 2, Bautenschutz + Bausanierung, Vol. 23, no. 4: 35–39. Pessikki, S. & Olson, L. 1997. Innovations in NonDestructive Testing of Conncrete, Americam Concrete Instituite, Farmington Hills, SP-168. Sansalone, M. & Streett, B. 1997: Impact-Echo. Jersey Shore, Bullbrier Press. Scott, M., Rezai, A., Moore, M. & Washer, G. 2000. The HERMES Bridge Inspector Project, Reference Manual, Geophysics. Streicher, D., Wiggenhauser, H., Holst, R. & Haardt, P. 2005. Zerstörungsfreie Prüfung im Bauwesen: Automatisierte Messungen mit Radar, Ultraschallecho und Impact-Echo an der Fuldatalbrücke. Beton- und Stahlbetonbau 100, 3: 216–224. US Patent 2006/0191358 A1. www.acsys.ru 2006. www.olsoninstruments.com 2006. ZFPBau-Kompendium der Bundesanstalt für Materialforschung und -prüfung; http://www.bam.de/service/ publikationen/zfp_kompendium/welcome.html.

Figure 28. Large concrete slab on BAM campus (10 m × 4 m, thickness 30 cm).

Figure 29. Large concrete slab during construction: rear section with tendon ducts, front section with thickness variations and honeycombs.

In addition to the laboratory specimens, even larger specimens have been established for special tasks: − The large concrete slab (LCS) with test problems typical for bridge decks (Figs. 28, 29) − Foundation slab with different reinforcements, thickness sections and test piles underneath All test specimens are available for round robin tests to other interested groups upon request. ACKNOWLEDGEMENTS The research described in this contribution is the primary result of ongoing teamwork in division VIII.2 of BAM. The dedication and enthusiasm of all members is greatly appreciated. Special thanks to all research partners within national and international projects. REFERENCES Algernon, D. 2006. Analyse akustischer Wellen in Beton. Berlin: Dissertation Technische Universität Berlin.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Efficiency control of electrochemical repair techniques C. Andrade, I. Martínez, M. Castellote & P. García de Viedma Institute Eduardo Torroja of Construction Science, Madrid, Spain

ABSTRACT: The main electrochemical techniques used for reducing corrosion in reinforced structures are Cathodic Protection (CP), Electrochemical Chloride Extraction (ECE) and Realkalisation (ER). Traditionally, for controlling the efficiency of CP, standard methods based on the depolarization of rebar are used, which have the main inconvenience of requiring the interruption of the protection current even for several hours. Concerning ECE and ER, the usual methods involve extraction and chemical analysis (chloride and hydroxyl ions respectively) of cores from the structure. In this paper, non-destructive methods, for monitoring the performance of the electrochemical repair techniques, are presented and analyzed. In the case of CP, a new developed methodology, called passivity verification technique, PVT, that makes use of the Electrochemical Impedance Spectroscopy technique, is described. As far as ECE an ER are concerned, the density of electrical charge passed, potential of corrosion and rate of corrosion measured by the Polarization Resistance technique are analyzed as non-destructive indicators. Additionally, for RE, the detection of a sudden increase in the intensity of electrical current passing, and a parallel establishment of electro-osmotic flux, is also postulated as a key parameter for practical control of the efficiency of the realkalisation. Finally, a new parameter “standardized by the resistance charge, SRC” is also presented, that would be a more indicative parameter than the simple recording of coulombs to check the efficiency during ECE treatment. Keywords: Concrete, Cathodic Protection, Electrochemical Chloride Extraction, Realkalisation, Electroosmotic flux.

1

INTRODUCTION

techniques. This document describes the most important aspects of the techniques, including methods of monitoring their efficiency. The control of the efficiency of CP, ECE and ER in controlling or eliminating corrosion is precisely one of the aspects still lacking for full satisfaction of the users. Thus, in the case of CP, due to the permanent cathodic polarization, the methods used for checking absence of corrosion (Instant-off potential or depolarized potential) give only indirect information on the real corrosion state of the reinforcement. In the case of ECE and ER, it is necessary to drill cores after the treatment or to wait weeks to verify the absence of corrosion. It seems a desirable goal to develop nondestructive methods to verify the efficiency of these treatments during their application and at the end, in order to have a contractual parameter to be easily checked by the owner of the structure. In the present paper, new methods to control the efficiency during and after the treatment are presented for the three techniques.

The oldest electrochemical technique used for repair/ protection of reinforced concrete structures damaged by corrosion is Cathodic protection (CP) (Wagner, 1952), while Electrochemical chloride extraction (ECE) and Electrochemical realkalisation (ER) are more recently applied. CP has been applied to submerged or buried concrete structures from the 50’s (Hausmann, 1969) and from the 70’s to concrete aerial structures, either in existing or in new structures (Berto1ini et al, 2002). Cathodic protection is currently standardised in different countries (NACE, 1990; Concrete Society, 1989; CUR, 1996; CEN, 2000 a,b). Electrochemical chloride extraction ECE, and Realkalisation, ER, are relatively recent methods, and they are still being developed today. The application of these techniques is based on the concept of removing the cause that produces the corrosion of the steel (chlorides or pH decrease around the steel). ECE has not been standardized yet, but CEN is producing a draft of a European Standard (CEN, 2000b). The European Federation of Corrosion published in 1998 a complete state of the art (Mietz, 1998) and it is also remarkable the work than can be found in (Bennet and Schue, 1990). On the other hand, the European COST Action 521 (COST, 521, 2002), has produced a document on these three protection/repair electrochemical

2

EXPERIMENTAL

The experimental section considers the three electrochemical repair techniques, in order to present the trials to describe the proposed methods that better control their efficiency.

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2.1

the PVT were made at different distances from the current injection point. Instant- on and -off potentials were also simultaneously measured in all the points.

Cathodic Protection (CP)

For a better non-destructive control of CP it is necessary to try techniques able to be applied without switching off the cathodic polarization. This is feasible by EIS and a method has been developed and patented (Invention patent No. ES2151410) as PVT: Passivation Verification Technique (PVT). This technique is based on the measurement of the Electrochemical Impedance Spectroscopy in the polarized structure. For its application where the bars have a semi-infinite area, it is necessary to apply a modulated confinement of the current by means of a controlled guard ring by two “electrical field followers” (Andrade et al, 2004). For the PVT, the instrument applies five different predetermined frequencies 100, 10, 1, 0.1, 0.01 Hz by using a central counter ring surrounded by a guard ring which modulates the confinement of the current by means of two “electrical field followers” and measures the phase angle between applied and recorded signals (Andrade et al, 2004). In the present paper, results are described of a real structure located in the South of Spain, the Algeciras Market (Figure 1). The structure was built in 1935, E. Torroja and Sanchez Arcas being the designers. The structure is constituted by a spherical dome of 47.76 m outside diameter supported by a set of eight columns. Due to chloride attack from the marine wind, in recent years the structural elements were showing some signs of deterioration in some parts of the roof and in the “belt” that links the roof with the columns. The corrosion of the octagonal belt represents a serious danger for the structural stability, especially because the steel bars cannot be replaced. For the repair, a reinforced concrete overlay was added, complementing the belt’s traction resistance, and cathodic protection was installed in all its reinforcement. A continuous Titanium wire was used for the octagonal belt cathodic protection, injecting the current by a unique rectifier through an electrical transformer placed in one point of the whole perimeter. The measurements of the CP efficiency by means of

2.2

Electrochemical Chloride Extraction (ECE)

For ECE there are two possibilities of non-destructive control of the progression of the treatment: a) monitoring of the electrical charge (Coulombs) applied (Castellote et al, 1999), or b) measurement of the corrosion rate or the potential at the end of the treatment. The tests presented here were performed on cubic 7 × 7 × 7 cm paste specimens cast with three rebars embedded in their central part, which acted as negative electrodes. Four different binders were tested: plain OPC (CEM I-42.5 R/SR); cement type CEM I42.5 R/SR with a 9% Silica Fume substitution; cement CEM III B/32.5 SR/NR (between 66% and 80% of blast furnace slag substitution) and cement IV-B-32.5 SR/BC, with about 65% Fly ash substitution. In all the mixes, NaCl % by mass of cement, ranging from 1.2 and 1.5 was added when casting. The specimens were held in a chamber at 100% RH for about 150 days in order to assure full maturity of every sample and the establishment of stable potential and corrosion rate values. Then, ECR treatment was performed, applying an electrical field between an external titanium mesh anode and the rebars embedded in the specimens. The current density applied was 1.4 A/m2 during 1 and 2 months, which corresponds to charge densities of 1000 and 2000 Ah/m2 respectively. Once the experiment was finished, Corrosion Potential and Polarization Resistance (Rp) measurements were performed and compared with nontreated specimens. The Rp was measured using a polarization from –10 mV to +10 mV at a sweep rate of 10 mV/min. The Rp was calculated from the slope of ∆E/∆I, and the corrosion rate through the expression: Icorr = 26 mV/Rp. 2.3

Electrochemical Realkalisation (ER)

To control the efficiency of ER the most used methods are: a) Control of the electrical charge applied, maintaining it between the recommended limits: 0.8 to 2 A/m2 during 1 to 2 weeks orb) Phenolpinthalein test that can be performed at the end of the experiment after drilling cores from the structure. More occasionally, electrochemical techniques have been used after treatment, as measurements of corrosion potential polarization resistance Rp, and potentiostatic anodic polarisation curves. In addition, the authors have used observation of the electro-osmotic flux by means of monitoring the current increase (Andrade et al, 1999; Castellote et al, 2006) to detect the end point of the treatment.

Figure 1. Algeciras Market.

32

3.1

In order to illustrate the advances in ER for monitoring the efficiency during and after treatment, data from several trials on carbonated (65% RH, 20 ± 2ºC and 100% CO2 concentration) OPC matrixes with and without embedded rebar, are reported here for the specimens without rebar, two different types of samples have been tested:

Figure 2 depicts the results obtained in the real structure (Algeciras Market). It can be deduced that there are big differences in the polarized potential shown in different belt perimeter points. This is the consequence of the current being applied only from one point, so that the difference of potential-on from one point close to the transformer to the opposite one is about 400 mV due to the high resistance of the circuit. This deficiency is well detected by means of the PVT because it identifies that the level of efficiency is higher the shorter the distance between the transformer (current injection point) and the measurement point (Figure 3). So, only point 0 and, in some cases point 1 have good protection levels, while points 2 and 3 do not reach the accurately level of protection in any case. Finally, it was confirmed as represented in Figure 4, that the phase angle at low frequencies is lower when good protection is detected.

1. On the one hand, small cylinders of 8 mm diameter and 10 mm length of normalised OPC mortar on tubular cells (Castellote, 1995), in which both compartments were made of glass tubes of 4 mm internal diameter. Two activated titanium wires were used as electrodes. Distilled water was introduced into the compartment where the cathode was located (catholyte) and 0.1 and 0.2 M of Na2CO3 solutions were used as the anolyte for the different experiments. The voltage drop applied between the electrodes was 30 and 12 V volts for the cases of 0.1 and 0.2 M respectively. 2. On the other hand, a disk of 50 mm thick carbonated OPC concrete was submitted to ER by using a “classical migration cell”, as reported in detail in (Castellote et al, 1999). In this case two corrugated rebars of 10 cm in length and 0.6 cm in diameter were used as electrodes, with a voltage drop of 40 V applied. In order to better reproduce the real conditions, the rebar acting as cathode was allowed to freely corrode in distilled water for two weeks before the treatment. This pre-corrosion induced the formation of sufficient rust on the rebar surface. Distilled water was introduced into the catholyte and 1 M of Na2CO3 solution was used as the anolyte. The corrosion rate and the corrosion potential of the rebars, acting as cathode, were measured before and after the trial by means of the Polarisation Resistance Technique.

3.2

Electrochemical Chloride Extraction (ECE)

With respect to non-destructive control of the treatment by means of monitoring the electrical charge density (A-h/m2) applied, in Figure 5a, the increase in the percentage of extraction of chlorides, averaged for the whole of the sample, is presented. From this figure it can be deduced that the efficiency in the removal of chlorides is dependent on the cement binder, the efficiencies being encountered in the following order: Plain OPC ≥ Silica Fume > Slag substituted binder ≥ Fly ash The amount of total chlorides remaining in the cathodic part of the samples after the maximum amount of charge passed, 2000 A-h/m2 is applied, is

The last trial consists of the realkalisation of a cylindrical concrete specimen with an embedded rebar that acts as cathode. In order to do so, a ponding cell (Mietz, 1995) was used, by gluing a rectangular pool on one side of the specimen parallel to the rebar containing a 1 M Na2CO3 solution as anolyte, and an external Ti mesh acting as anode. A voltage drop of 50 V was applied between the electrodes. The corrosion rate and the corrosion potential of the rebar were measured before and after the trial by means of the Polarisation Resistance Technique. During all these experiments, the establishment of the electroosmotic flux was studied by means of monitoring the abrupt increase in the current intensity during the treatment (Yeih and Chang, 2005). 3

Cathodic Protection (CP)

E (pillar) = –235 mV E (ring) = –124 mV

E (pillar) = –225 mV E (ring) = –140 mV E (pilar) = –209 mV E (ring) = –290 mV

E (pilar) = –317 mV E (ring) = –128 mV

MARKET

E (pilar) = –180 mV E (ring) = –153 mV

E (pilar) = –132 mV E (ring) = –139 mV

30 m

E (Pilar) = –600 mV E (ring) = –272 mV

E (pilar) = –562 mV E (ring) = –276 mV

Figure 2. Potential-on measurements taken through embedded reference electrodes along the belt perimeter. The current injection point is in the right lower part in the figure (E piller = –600). The values indicate that the potentials are less cathodic the further zones are from the injection of the cathodic current.

RESULTS

For the three electrochemical repair techniques, results are presented related to the control of the efficiency, as described.

33

100

Mean percentage of extraction (%)

CATHODIC PROTECTION IN ALGECIRAS MARKET

PVT Protection level (%)

100

PROTECTED 90 Point 1 -A point 1* -A Point 2* -A Point 3 -A Point 1 -B Point 2 -B Point 0 -B

80

NON PROTECTED 70

OPC SF SLAG FA

75

a)

50

25

0 0

500

1000

1500

2000

2500

2

charge density passed (A-h/m )

60 -1900

-1400

-900

E- ON (mV)

-400

100

*Measurement over the anode

Figure 3. Percentage of protection obtained in the different measurement points of the belt of Algeciras Market.

1,2

0,8

0,4

0

ALGECIRAS MARKET

phase

70 60 50 40 30

OPC

SF

SLAG

FA

Point 3: Non protected area Point 1: protected area

Figure 5. a) Evolution of the mean percentage of extraction in the samples, as a function of the charge density passed b) Initial and final amount of chlorides in the cathodic part of the samples (% by mass of cement) after the maximum amount of charge passed, 2000 A-h/m2.

20 10 0 0.01

Initial After 2000 A-h/m2

b)

1,6

%Cl by mass of binder

-2400

0.1

1

Frequency (Hz)

10

100 Ecorr (mVSCE) vs time (days)

Figure 4. Phase angles measured using PVT for the determination of the cathodic protection effectiveness in Algeciras Market.

200 0 -200 -400 -600

depicted in Figure 5b, where the initial amount is also given. After 2 months of treatment, smaller values than the limit of 0.4% of chloride by cement mass remain in the OPC and SF, while for Slag and FA (see Figure 5b) cements, the current density passed has not been enough for reaching that level of chloride extraction. Due to the higher amounts of bound chlorides in slag and FA cements, they could not be released during the treatment and they remain without being removed. With regard to the efficiency of the treatment in relation to the passivation of the rebars, the Corrosion Potential and the Rp were measured in order to determine the corrosion current, and the results obtained are given in Figure 6 (a,b). Before the electrochemical treatment, all the specimens were actively corroding, with values of corrosion potential around –600 mV SCE and corrosion current higher than 0.2 µA/cm2. After treatment, and with time (the specimens have been monitored for more than three years after finishing the trials), all the values of Corrosion Potential remain at values more positive than –200 mV, typical of passive steel. However, according to the values of corrosion rate, the rebars are passive only for the samples OPC and SF. The values

-800 0

T R E A T M E N T

a)

200

400

600

800

1000

1200

1400

1600

-1000 -1200 SF

SLAG

FA

OPC

Icorr (uA/cm2) vs time (days) 100 SF 10

SLAG

T R E A T M E N T

1

0.1

FA

OPC

b)

0.01 0

200

400

600

800

1000

1200

1400

1600

Figure 6. Evolution of a) Ecorr and b) Icorr before and after the ECE treatment for all the specimens studied.

for the FA specimens are in the range between 0.1 and 0.2 µA/cm2, which is considered in the border of active corrosion; in the case of slag cement specimens, at a certain time after the treatment, the bars showed Icorr values characteristic of active corrosion. Therefore, the Icorr value is the parameter that better correlates with the remaining chloride content.

34

Before the electrochemical treatment, both specimens were actively corroding, with values of corrosion potential around –600 mV SCE for the rebar corroding in distilled water and around –250 mV SCE for the rebar embedded in the carbonated concrete, and corrosion current much higher than 0.2 µA/cm2. After the ER treatments, the steels present very high Icorr values for a short period of hours and they decrease afterwards. In the case of the steel in the cathodic solution of the migration cell, it remains active during the two months recorded while in the case of the embedded rebar the steel repassivates.

100

Mortar-0.1M

300

b)

Mortar-0.2M Concrete-1M

current intensity (mA)

75

I/I0 (mA)

200

100

50

25

a) 0 0

1000

2000

3000

0 0

1000

2000

3000

charge density (A-h/m2)

charge density (A-h/m2)

Figure 7. Evolution of the current during the ER treatment as function of the charge density passed (A-h/m2) for the different experiments. a) Specimens without embedded rebar. b) Specimen with embedded rebar. The limits for the recommended values of charge density passed are marked with vertical dotted lines.

4 4.1

before realkalisation -15

after realkalisation

-5

5

15

35

45

The level of efficiency indicated by the PVT has been shown based on the appearance or not of a Faradaic response in the low frequency region of the EIS spectra. The choice made seems to empirically agree with the indication of the traditional 24 hour depolarisation decay leading to the same conclusions about the CP efficiency. Thus, the PVT provides an alternative for monitoring the cathodic protection efficiency. As the measurement is fast and does not need the disconnection of the protection current, it is possible to use it in existing structures needing periodic supervision.

55

0

Ecorr (mV)

-600 -900 -1200

Concrete-rebar in catholyte

a)

-1500

Concrete-rebar embedded

b)

100 10

Icorr (uA/cm2)

Cathodic protection

time (days) 25

-300

1 0.1

4.2

0.01 0.001

-15

0.0001 -5

before realkalisation

5

15

25

35

45

after realkalisation

Realkalization and chloride removal

In the case of ECE and ER, there are two stages at which it is necessary to measure the efficiency of the treatments: during their application in order to decide when the treatment should cease, and after the treatment in order to monitor the maintenance of the repassivation achieved.

55 time (days)

Figure 8. Evolution of a) the Corrosion Potential and b) Corrosion rate of the rebars before and after the ER experiments.

3.3

DISCUSSION

4.2.1 During the treatment application For ECE, enough efficiency was assumed if a certain total electrical charge density, Q (Ah/m2) has been passed (Castellote et al, 2006a). However, present results show that due to the different micro-structural characteristics of the different cement types (indicated by the resistivity) and their different binding ability, this parameter is not reliable enough to inform on the re-installing of passivity. In Figure 9a it can be seen that the mixes for which the repassivation has not been reached exhibit values of Q/R below 1800 Ah/m2KΩ. Thus, a new parameter is proposed: The “standardized by the resistance charges, SRC”, that would be a more indicative parameter than the simple recording of coulombs to check the efficiency during ECE treatment. A threshold of 1800 Ah/m2KΩ could be a suitable value. For measuring the progress of ER, present results have shown that the total charge is not only insuffi-

Electrochemical Realkalisation (ER)

Figure 7 (a,b) shows the evolution of the current passing during the treatment as a function of the charge density passed (A-h/m2) for the different experiments. Figure 7a) and b) reports the data corresponding to the specimens without and with embedded rebar respectively. Since the voltage applied in the different experiments was not the same, for the sake of clarity, in Figure 7-a, the values of current have been normalised to the initial value passing through the specimens. It can be deduced that in most cases, the recommended values of charge densities are not enough in order to reach the development of the electro-osmotic flux. In Figure 8, the Corrosion Potential and the Corrosion Rate of the rebars having been realkalized here depicted.

35

5400

3. The realkalization is only efficient if an electroosmotic flux of carbonates is induced to penetrate to the bar, which can be noticed during the treatment by means of sudden dramatic increase of the current monitored. Only in these conditions can realkalization be fully effective.

3000

0.41

Q/R=It/(V/I)=I2t/V (A-h/m2K)

2500 3600

2000 A-h/m2

0.66

1.18

1800

1.51

1500 1000 500 0

0

a)

OPC

SF

SLAG

FA

b)

OPC

SF

SLAG

The recommended values of charge density passed the realkalisation treatment (14 days at 2 A/m2) seems to be insufficient for getting the electro-osmotic flux in most cases. An amount above 1700 Ah/m2 is recommended. The measurement of Rp after ECR or ER is a reliable indication of the degree of steel repassivation, and enables monitoring during aging. When the Icorr values remain at levels above 0.1 µA/cm2 after some days or weeks, the treatment has not being efficient enough and new treatment periods should be considered.

FA

Figure 9. a) Q/R calculated at the end of the experiment. b) Charge density passed that would have been enough for reaching whole passivation, taking 1800 Ah/m2KΩ, as the threshold value for SRC (standarized by the resistance).

cient in the value recommended until now, but it is also not enough if a sudden increase (the peaks of Figure 7) is not recorded, because these peaks are the only clear indication of the developing of the electroosmotic flux. Thus, it is always necessary to first monitor the appearance of the current peaks.

ACKNOWLEDGMENTS The authors are grateful to funding of the Ministry of Education and Science of Spain through the project of the program SEDUREC CONSOLIDER-2006 INGENIO 2010.

4.2.2 After the treatment The measurement of the corrosion potential and definitively, the measurement of the Rp, have been the best techniques to monitor the maintenance of the repassivation, until even 3 years after, in the case of ECE. Regarding realkalization, the phenolphthalein test is not a sufficient tool to prove the state of the steel surface (Castellote et al, 2003), as has been detected after electrochemical measurements on realkalised specimens that turned pink after applying the acid-basic indicator. The corrosion rate obtained from the Rp is the most reliable indication of negligable corrosion (Yeih and Chang, 2005; Elsener, 2001) and it has enabled to verify that efficient repassivation is feasible if the correct treatment conditions are applied.

5

REFERENCES Andrade C., Castellote M., Sarría J., Alonso C., 1999. Evolution of pore solution chemistry, electro-osmosis and rebar corrosion rate induced by realkalisation. Materials and Structures, 32, 427–436. Andrade C.I., Martínez I., Lasa I., Tronconis de Rincón O., Torres-Acosta A.A., Martínez M., 2004. Cathodic protection efficiency measurements made on concrete bridges using a new technique called passivity verification technique (PVT) without disconnecting the protection current, in: Proc. NACE CORROSION 2004. paper nº. 04329. Bennet J.E., Schue T.S., 1990. Electrochemical chloride removal from concrete. A SHRP Contract Status Report. Corrosion 90, Las Vegas, Nevada, Paper nº. 316, April. Berto1ini, L., Gastaldi M., Pedeferri M.P., Pedeferri P., Redaelli E., 2002. Cathodic protection of steel in concrete and Cathodic prevention, Proceedings COST 521 Workshop, J. Mattila, ed. Tampere, Finland. Castellote M., Andrade C., Alonso C., 1999. Electrochemical chloride extraction: int1uence of testing conditions and mathematical mode1ling. Advances in Cement Research Vol. 11, nº. 2, Apr., pp. 63–80. Castellote M., Llorente I., Andrade C. 2003. Influence of the external solution in the electroosmotic flux induced by realkalisation; Mater Construc, Vol. 53, nº. 271–272, pp. 101–111. Castellote M., Llorente I., Andrade C., Turrillas X., Alonso C., Campo J. 2006a. In-situ monitoring the realkalisation process by neutron diffraction: electroosmotic flux and portlandite formation. Cement and Concrete Research 36, 791–800.

CONCLUSIONS

Present results enable the following conclusions to be drawn up regarding the three repair techniques evaluated. 1. The method called passivity verification technique, PVT, based on the analysis of Impedance Spectra has shown to qualitatively inform on the efficiency of CP through the measurement of the angle phase change, with the advantage that PVT does not need to disconnect the system, as traditional methods require. 2. Chloride removal should be measured during the treatment through the amount of effective electrical charge made to pass, standardized by the electrical resistance measured at the beginning of the treatment, SCR. This parameter should be at least 1800 Ah/m2KΩ.

36

Castellote M., Llorente I., Andrade C., 2006b. Influence of the composition of the binder and the carbonation in the zeta potential values of hardened cementitious materials, Cement and Concrete Research 6, 1915–1921. CEN, 2000a, Cathodic protection of steel in concrete, EN 12696. CEN, 2000b, Electrochemical rea1kalisation and ch1oride extraction treatments for reinforced concrete—Part 1 Realkalisation, prEN 14038-1. Concrete Society, 1989, Cathodic protection of reinforced concrete, Tech Report no. 36, Concrete Society & Corrosion Engineering Association. COST 521. 2002. Corrosion of steel in reinforced concrete structures. Final Report on “Electrochemical maintenance methods” R. Polder, Ed. By Romain Wydert 123–164. CUR, 1996, Kathodische bescherming van wapening in betonconstru’ (Cathodic protection of reinforcement in concrete structures), CUR Tecb Recommendation 45, Gouda, in Dutch. Elsener B. 2001. Half cell potencial mapping to assess repair work on RC structures. Contr. Build. Mater, 15 (2–3), pp. 133–139. Hausmann DA. 1969. Criteria For Cathodic Protection Of Steel In Concrete Structures, Materials Protection 8 (10): 23-&.

Invention patent nº. ES2151410. Publication date: 1 de february 2003. Inventors: C. Andrade, J. Fullea, J.A. Bolaño, F. Jiménez, A. Navarro and I. Martínez “Procedimiento y dispositivo para la detección de corrosión en acero enterrado y protegido catódicamente, especialmente en armaduras de hormigón, o determinar si esta pasivo”. Mietz J. 1995. Electrochemical realkalisation for rehabilitation of reinforced concrete structures, Werkstoffe und Korrosion , Volume 46, Issue 9, Pages 527–533. Mietz J. 1998. Electrochemica1 rehabilitation methods for reinforced concrete structures, a state of the art report, European Federation of Corrosion Publications No. 24, IOM Communications, London. NACE, 1990, STANDARD RPO290-90 Cathodic protection of Reinforcing Steel in Atmospherically Exposed Structures. Wagner C., 1952. Contribution To The Theory Of Cathodic Protection, Journal Of The Electrochemical Society 99 (1): 1–12. Yeih W., Chang J.J. 2005. A study on the efficiency of electrochemical realkalisation of carbonated concrete, Construction and Building Materials, Volume 19, issue 7, pages 516–524.

37

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

External strengthening of continuous beams with CFRP L. Taerwe, L. Vasseur & S. Matthys Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Ghent, Belgium

ABSTRACT: This paper discusses the use of FRP EBR for the flexural strengthening of continuous beams. With respect to this type of beams, few research has been reported. By means of an experimental and analytical study the non-linear behaviour of 2-span reinforced concrete beams, strengthened with externally bonded FRP laminates, has been analyzed. It is demonstrated that FRP strengthened cross-sections restrict the rotation of a plastic hinge at that location, and allow additional plasting hinge formation in unstrengthened cross-sections. In addition, the bond behaviour between the FRP and the concrete in the case of flexural strengthening of continuous beams has been studied, it is shown that the debonding mechanisms are also governed by the shear force and moment redistribution occurring in multi-span beams.

1

between the moment and the curvature is obtained, namely

INTRODUCTION

Structures may need to be strengthened for different reasons, among which a change in function, implementation of additional services or to repair damage. Different strengthening techniques exist. Often applied is externally bonded reinforcement (EBR), based on fibre reinforced polymers (FRP), the so-called FRP EBR. FRP EBR can be applied for the strengthening of existing structures, enhancing the flexural and shear capacity or to strengthen columns by means of confinement. This paper discusses flexural strengthening of 2 span reinforced concrete beams. CFRP (Carbon FRP) laminates are glued on the soffit of the spans and/or on the top of the mid-support (Ashour, et al. 2004, El-Refaie, et al. 2003). The efficiency of the FRP EBR strengthening technique is often limited by the capability to transfer stresses in the bond interface. Hereby bond failure between the laminate and the concrete may occur. For unstrengthened continuous beams a moment redistribution can be observed especially after yielding of one of the critical sections. As a consequence, a plastic hinge will be formed. For strengthened continuous beams, after reaching the yield moment, the FRP strengthened cross-section is still able to carry additional load and the formation of a plastic hinge will be restricted. The aim of this study is to have a better insight into the behaviour of reinforced concrete structures strengthened in flexure in a multi-span situation.

1 M = r EI

(1)

with 1/r the curvature, M the bending moment and K = EI the flexural stiffness. This stiffness is assumed to be constant and therefore independent of the value of the bending moment. However, for the cross-section of a concrete beam the moment-curvature diagram is non-linear. This non-linear character is caused by the variable flexural stiffness, as shown in Figure 1. Two cases are shown in this graph : a cross-section with externally bonded FRP (strengthened) and a cross-section without FRP (unstrengthened). An important difference between these cases is the flexural stiffness (slope of lines K0, K1 and K2). With FRP higher values for K are obtained than without FRP. This different behaviour will influence the moment redistribution of a continuous beam. If Figure 1 is applied to a continuous beam, we start with the uncracked phase along the whole length of the beam, corresponding to the use of K0 as flexural stiffness. By increasing the load, the beam is

M'u

K'2

M'120 y Mu My

2

K2 K'1

CALCULATION MODEL FOR CONTINUOUS BEAMS

K1 with FRP

2.1

Mc

Non-linear moment-curvature diagram

1/r 0

Performing an analysis of a construction according to the linear elasticity theory, a linear relationship

Figure 1.

39

without FRP

K0 1/r'y1/r'y

35

1/r'u

Moment-curvature diagram.

1/ru

characterized by cracked and uncracked zones, each with the related value of flexural stiffness. This change of stiffness causes a first redistribution of moments. At the yield load Fy, one or more cross-sections reach the yield moment (My). In yield zones without FRP EBR, the bending stiffness K2 is so small that plastic deformations appear in the critical cross-section and in a restricted area near to it. This is the formation of a so-called plastic hinge. The increasing load is mainly carried by the non plastic zones, during which the bending moment in the plastic hinge remains almost constant (Mu ~ My) or is slowly increasing. In zones with FRP EBR, the value of the bending stiffness is higher (K 2' ). Also plastic deformations appear, but in a more limited way. The yielding zone still carries a significant part of the increasing load and the formation of the plastic hinge is restricted. 2.2

Figure 2. Continuous beam with variable bending stiffness (simplified to 2 stiffness zones).

1,0

0,8

General behaviour of continuous beams

0,6

a b

m=

M support M span

k=

K support K span

m/1+m

Consider a continuous beam with two identical spans and symmetrically loaded by two point loads (Figure 2). In one span, two zones can be distinguished : a zone with negative moments (above the mid-support) and another with positive moments (in the spans). It is assumed that in each zone the bending stiffness is constant. So the mid-support zone and the span zone have stiffness Ksupport and Kspan, respectively. Further, we define:

λ=

0,0 0,0

0,2

0,4

0,6

0,8

1,0

k/1+k

(2)

Figure 3. Relation between the bending moment ratio m and the ratio of bending stiffnesses k.

internal moment is linear, as in the case of isostatic beams. By further increasing the load, the changing bending stiffnesses in different cross-sections modifies k thus the relation between the internal moments m. As a result, the moment distribution deviates from the classic theory to the so-called non-linear moment redistribution.

(3)

With Eq. (3) the internal forces in the continuous beam can be calculated. In what follows, calculations are performed for a = 2 m and b = 3 m. Hence with λ = 2/3 Eq. (3) changes into. 36m3 + ( 45 − 8k ) m2 − 34 km − 35k = 0

0,4

0,2

By considering that the angle of rotation above the mid-support equals zero, the following equation can be obtained (Taerwe, et al. 1989): (2 + 3λ )m3 + (3 + 3λ − 2λ 2 )m2 − k λ(3 + 4 λ)m − (1 + λ )(1 + 2λ ) k = 0

k=1

3

3.1 (4)

DEBONDING MECHANISMS ON CONTINUOUS BEAMS Different debonding mechanisms

Bond failure in case of FRP EBR implies the loss of composite action between the concrete and the FRP reinforcement. This type of failure is often very sudden and brittle. According to Matthys [3] different bond failure aspects can be distinguished.

This equation is shown in Figure 3. For loads below the cracking moment, the mid-support zone and field span zone are uncracked and the two zones nearly have the same bending stiffness. This condition correspond with k = 1. From Eq. (4) we obtain then m = 0.9722 = mel. This value of m corresponds to the moment distribution following the classic theory. Hereby, the relationship between acting load and

3.1.1 Crack bridging The externally bonded FRP will need to bridge cracks. In regions with significant shear forces, shear or flexural cracks have a vertical (v) and a horizontal

40

(w) displacement. The vertical displacement of the concrete causes tensile stresses perpendicular to the FRP EBR, which initiates debonding of the laminate (Figure 4). 3.1.2 Force transfer The variation of tensile force in the FRP, due to the composite action between the FRP EBR and the concrete beam initiates bond shear stresses at the interface. The bond shear stress considered between two sections at a distance ∆x equals (Figure 5):

τb =

∆N fd

(5)

b f ∆x

L Figure 6.

These shear stresses have to be smaller than the bond strength between the concrete and the FRP reinforcement.

Concrete rip-off also can occur due to a stress concentration at the laminate end.

3.1.3 Curtailment and anchorage length Theoretically the FRP reinforcement can be curtailed when the axial tensile force can be carried by the internal steel only. The remaining force in the FRP at this point needs to be anchored. The anchorage capacity of the interface is however limited, and hence the FRP may be extended to zones corresponding to low FRP tensile stresses.

3.2

v peeling action w

Peeling-off caused at shear cracks.

Figure 5.

Peeling-off caused at shear cracks.

Debonding mechanisms in continuous beams

To predict the debonding load, the available calculation models [4] are based on formulas which have been derived from experiments on isostatic reinforced beam and pure shear bond tests. The difference between isostatic beams and continuous beams, which is critical for these debonding mechanisms in continuous reinforced concrete beams, is the moment line with opposite signs. While the moment in the span is positive, the moment at the mid-support is negative. As a result, the compression zones in the spans are situated at the top of the beam, at the support the compression zone is situated at the soffit of the beam (shaded zones in Figure 7). In contrast to reinforced isostatic beams, this allows to anchor the CFRP laminates in the compression zones (except for the end supports) (Figure 8). By extending a laminate into these compression zones, two out of the four different debonding mechanisms will be avoided: debonding by a limited anchorage length and debonding by end shear failure (concrete rip-off). Debonding by limited anchorage length is prevented by extending the laminate into the compression zone because in this situation the tensile stress in the laminate is gradually reduced to zero, and anchored in a zone with small compressive stresses (no significant risk for buckling). Debonding by end shear failure occurs at a shear crack at the end of the laminate. By extending the laminate into the compression zone, the plate-end reaches a zone where no shear cracks will be formed and neither concrete rip-off will appear. Debonding mechanisms can be avoided by extending the laminate into the compression zone, hence beyond the point of contraflexure, which is the location where the internal moment equals zero. For

3.1.4 Concrete rip-off If a shear crack appears at the plate-end, this crack may propagate as a debonding failure at the level of the internal steel reinforcement. In this case the laminate as well as a thick layer of concrete will rip off (Figure 6).

Figure 4.

Concrete rip-off by shear at the laminate end.

41

Figure 7. beams.

Moments with opposite signs in continuous Figure 9. Internal and external reinforcement configuration. Table 1. Properties of concrete and reinforcement materials. Concrete

Steel

CFRP

36.0

–

–

– – 3.3 0.35* 32000

570 0.28 670 12.40 210000

– – 2768 1.46 189900

Figure 8. Anchoring laminates into compression zones. Compres. strength [N/mm2] Yielding strength [N/mm2] Yielding strain [%] Tensile strength [N/mm2] Failure strain [%] E-modulus [N/mm2]

calculating the exact location of this point, it has to be noticed that the point of contraflexure moves with increasing load, due to the non-linear moment redistribution. 3.3

Specific debonding aspects related to continuous beams

*in compression.

In the case of strengthened continuous beams, some particular aspects can be noted, which may also influence the moment of debonding. This is illustrated in the following by means of an analytical study for the beam and strengthening configuration shown in Figure 9. The applied internal reinforcement is kept constant during the analytical study and is based on the elastic theory. In this case almost the same amount of internal reinforcement is used in the spans and in the mid-support (reinforcement ratio’s ρs,span = 0.68 % and ρs,support = 0.61 %). The properties assumed in the analysis are given in Table 1, whereas the amount of FRP strengthening in the spans and mid-support zone is varied (FRP widths of 60 mm, 100 mm, 150 mm and 200 mm are used in this study). The length of the FRP is chosen in such a way that all four debonding mechanisms can occur. Herewith the laminates are not anchored into the compression zones as described in section 3.2. The length of the laminate at the soffit of the span equals 2000 mm and is applied in such a way that the center of the laminate is just beneath the point load. The laminate at the top of the beam above the mid-support equals 1600 mm (see Figure 9). The influence of the amount of FRP strengthening on the acting shear forces is illustrated in Figure 10, for a point load F of 100 kN. Herewith, V1 (solid lines) is the shear force acting between the outer support and the point load. V2 = F−V1 (dashed lines) is the shear force acting between the point load and the midsupport (Figure 2). As can be noted, the value of V1 (and hence V2) is influenced by the FRP reinforcement

ratio’s of both the span (ρf,span) and the mid-support (ρf,support). By increasing the width of the laminate above the mid-support (increasing wf,support or ρf,support), for wf,span = constante, V1 decreases and V2 increases. This is due to the moment redistribution which is dependent on both the external and internal reinforcement ratio used over the length of the beam. Owing to this, also the distribution of the reactive forces in the supports is dependent on the reinforcement ratio’s. As a result, the part of the applied load which is carried by the mid-support increases with an increasing amount of FRP at the mid-support. If wf,span (or ρf,span) is increased as well, the decrease of V1 will be less pronounced and V1 may even increase (compared to the unstrengthened beam). As debonding phenomena are often related to the acting shear force, this means that possible debonding of a FRP laminate not only depends on the FRP configuration at that location, but also on the amount of FRP in the zone with opposite moment sign. Another significant aspect with respect to the values of ρf,span and ρf,support relative to each other, is their influence on the point of contraflexure. Indeed, by increasing ρf,support (increasing the width of the laminate above the mid-support), Msupport will increase and Mspan will decrease. Herewith, the point of contraflexure moves towards the mid-support. On the opposite, by increasing ρf,span at the soffit of the span, Msupport will decrease and Mspan will increase. As a result, the point of contraflexure moves away from the mid-support. Because of this, a change of

42

anisms will be investigated in function of both ρf,span and ρf,support. Hereby, a differentiation is made between debonding of the top laminate (case A), debonding of the laminate at the soffit of the span between the point load and the mid-support (case B) and debonding of the laminate at the soffit of the span between the point load and the outer support (case C) (Figure 9). The calculations are performed according to section 2 and fib-bulletin 14 (fib 2001). Results of the debonding load calculations are given as far as they do not exceed the ultimate load of the strengthened beam assuming full composite action. Herewith it is assumed that debonding of the FRP does not occur, and that the construction only can fail by concrete crushing or by exceeding the tensile strength of steel or FRP reinforcement. In Table 2 a summary is given of the effect of the external reinforcement on the debonding load.

65

V 1, V 2 [kN]

60 wf,span = 0 mm wf,span = 60 mm wf,span = 100 mm wf,span = 150 mm wf,span = 200 mm

55 50 45 40 0

50

100 w f,support [mm]

150

200

Figure 10. Shear force V1 in function of width of laminates.

4 Figure 11. Table 2. load.

4.1

Differentiation between places of debonding.

Amount of top Amount of laminate above laminate at the the mid-support ? soffit of the span ? Crack bridging debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

debonding load ? debonding load ? debonding load ?

General overview of test program

4.1.1 Configuration For the experimental study, the test set-up of Figure 7 is used. The total depth of the continuous concrete beam equals 400 mm and the width 200 mm. The continuous beam exists of two spans, each with a length of 5 m. The beam is loaded with one point load in each span. The locations of the point loads are at a distance of 3 meter of the mid-support and 2 meter from the end supports. Hence, a equals 2 m, b equals 3 m and λ = 2/3 (referring to Eq. 3). In the experimental program three full-scale continuous beams are tested with the same crosssection but different configurations of the internal and external reinforcement. The reinforcement configuration is shown in Figures 12, 16 and 21. Beam CB1 is reinforced with a small amount of internal reinforcement in the spans and a large amount at the support. To compensate the small amount at the spans, externally bonded reinforcement (EBR) is applied only in the spans. The next beam (CB2) has internal reinforcement based on the linear elastic theory. In this case almost the same amount of internal reinforcement is used in the spans as at the mid support. As external reinforcement, laminates are glued on top of the beam above the mid-support as well as at the soffit of the beam in the spans. Finally a third beam is tested (CB3) with a large amount of internal reinforcement in the spans and a small amount at the support. As external reinforcement, EBR is only applied at the top of the beam above the mid-support. During the tests both manual and electronic measurements are performed. Strain gauges are glued on top of the laminates. Load cells are placed beneath each support, by which the moment redistribution

Effect of external reinforcement on debonding

A B C Force transfer A B C Anchorage failure A B C Concrete rip-off A B C

EXPERIMENTAL STUDY

the distance between the laminate end and the place where the internal moment equals zero (L) can be observed. Indirectly also the anchorage length (lt) will change. Due to this, the debonding mechanisms anchorage failure and concrete rip-off once more will be dependent on the amount of external reinforcement along the beam. In the following paragraphs, the load at which debonding occurs, for the different debonding mech-

43

Table 3.

CB1 CB2 CB3

Table 4.

Overview of ultimate loads. Fcollaps,calc [kN]

Fcollaps,exp [kN]

Ratio [%]

157 197 124

153 172 115

97.5 87.3 92.7

can be made. A first observation is the good correspondence between the predicted and the experimentally obtained moment redistribution. Secondly, it can be concluded that the obtained debonding failure load is somewhat lower than predicted. This is especially the case for beams CB2 and CB3, for which debonding of the top laminate occurred. This can also be noted from tables 1 and 2, which give an overview of the ultimate loads and the debonding mechanisms. In Table 5, a comparison is made between the ultimate load of the tested beams and the calculated ultimate load of these beams if they would not have been strengthened. Whereas the failure aspect of the strengthened beams is characterized by debonding, the failure aspect of the unstrengthened beams (as obtained from the calculation model) is characterized by yielding of the steel followed by concrete crushing.

Overview of debonding mechanisms. Debonding mechanism

CB1 CB2 CB3

By crack bridging of laminate at soffit By crack bridging of laminate at top By crack bridging of laminate at top

Table 5. Comparison between reinforced and unreinforced continuous beams.

CB1 CB2 CB3

Fstrengthened [kN]

Funstrengthened [kN]

Ratio

4.2

Continuous Beam 1 (CB1)

157 197 124

122 118 102

1.29 1.67 1.22

4.2.1 Configuration The first beam tested has internal reinforcement as shown in Figure 12. The beam has a low internal reinforcement ratio in the spans (ρs,span = 0.48%)

can be calculated for the reaction measures. Also the deflection in the spans is measured continuously by the use of LVDT’s. Finally, the strain of the internal steel and the concrete, especially in the compression zones, is measured manually. 4.1.2 Moment distribution The moment redistribution is illustrated in Figures 13, 17 and 22. These graphs give the span moment Mspan and the central-support moment Msupport at the critical section (where the moment is maximum), in function of the acting point load F (see Figure 2). In each graph four different curves concerning the moment distribution are observed. First there is the linear curve which is the moment distribution calculated following the elastic theory. Hereby, the relationship between the acting load and the internal moment is linear. Following, there is a non-linear dashed curve. This curve illustrates the non-linear moment distribution of the unstrengthened beam calculated according to the above mentioned non-linear theory. In addition, there are two non-linear curves, which represent the calculated and experimental non-linear moment distribution. Finally a dashed horizontal line is drawn in the graphs. This curve illustrates the calculated load value where debonding is expected (calculations based on fib bulletin 14).

Figure 12. Table 6.

Internal steel configuration of CB1. Properties of concrete and CFRP.

Compr. strength Tensile strength Failure strain E-modulus

Table 7.

44

CFRP

38.0 N/mm² 3.4 N/mm² 0.35% 35500 N/mm²

2768 N/mm² 1.46% 189900 N/mm²

Properties of steel reinforcement.

Yielding strength Yielding strain Tensile strength Failure strain E-modulus

4.1.3 Overview of test results Based on the graphs of the moment redistribution (Figures 13, 17 and 22), two important conclusions

Concrete

Reinforcement in span

Reinforcement at support

601 N/mm² 0.28% 677 N/mm² 12.40% 218000 N/mm²

530 N/mm² 0.25% 701 N/mm² 12.40% 216000 N/mm²

and a high concentration of reinforcement above the mid-support (ρs,support = 1.29 %). As external reinforcement, two CFRP laminates with a length of 3750 mm are applied in the spans. The section of the CFRP laminate is 100 mm × 1.2 mm (ρf,span = 0.17 %). The characteristics of the materials are given in Tables 6 and 7. These values result from standard tensile and compression tests. 4.2.2 Moment redistribution In Figure 13, the moment redistribution of CB1 is illustrated as obtained from analytical calculations. For the unstrengthened beam, the formation of a plastic hinge can be noticed (vertical part of the dashed moment distribution curve). Whereas by the strengthened beam, although the strengthened spans still start to yield first, the FRP allows the spans to continue resisting the additional load. At increasing load when the support starts to yield, a plastic hinge will be formed at this mid-support. Debonding of the FRP EBR is predicted following fib bulletin 14 at 157 kN. Concerning the experimental data, a good agreement is observed with the calculated curve.

Figure 14.

Debonding in the span by crack bridging.

10

%

0,01

20 30

0,005

40

mm

0 200

400

600

800

-0,005

F [kN]

240

100

120

span yields

140

Experimental data

120

Linear elastic moment distribution

100

130 140 150

Figure 15.

Strain in compression zone of laminate.

Figure 16.

Internal steel configuration of CB2.

measured by six strain gauges, is linear over the laminate end. By visual inspection of the laminate ends, anchored in the compression zones, during the test, no buckling of these laminate ends could be noticed. In Figure 15 a (small) shift of the point of contraflexure, caused by the non-linear moment redistribution can be observed.

80 60 Non-linear moment distribution (without FRP EBR)

40 span cracks 20 support cracks 0 -80

Mspan [kNm] - span moment

Figure 13.

125

-0,02

160

-120

110

-0,015

180

support yields

-160

105

200

support collapses

-200

80 90

-0,01

Non-linear moment distribution (with FRP EBR)

220

60 70

0

4.2.3 Debonding mechanism The strengthened continuous beam fails by debonding of one of the CFRP laminates in the span. The mechanism which occurs is debonding by crack bridging. Debonding starts at a crack, located near the right point load, and propagates towards the mid support (Figure 14). By testing the beam, the laminate debonds at a load of 153 kN. This is 2.5% lower than the calculated value. With its length of 3750 mm, the end of the laminate, near to the mid-support, extends about 500 mm in the compression zone. As mentioned above, this is done to avoid some debonding mechanisms. On the contrary the laminate has to resist to compressive strain in this zone. Figure 15 gives a visual representation of the compression strains. As shown in the graph, the strain,

50

-40

4.3 0

40

80

120

160

Continuous Beam 2 (CB2)

200

4.3.1 Configuration The second tested continuous beam has internal reinforcement as shown in Figure 16. The beam has an

Msupport [kNm] - support moment

Moment redistribution of CB1.

45

internal reinforcement ratio calculated according the linear elastic theory (ρs,span = 0.68 %) and (ρs,support = 0.61 %). As EBR, external reinforcement is used in the spans as well as at the mid support. Two CFRP laminates with a length of 3750 mm are applied in the spans (ρf,span = 0.17 %), while one CFRP laminate with a length of 5000 mm is applied at the central support (ρf,support = 0.17 %). The section of the CFRP laminates is 100 mm × 1.2 mm. The characteristics of the materials are given in Tables 8 and 9. These values result from standard tensile and compression tests.

Following the non-linear theory, the support and the span yield at nearly the same load, both in the strengthened and the unstrengthened beam. For the unstrengthened beam this results in a mechanism (formation of 3 plastic hinges at the same time). For the strengthened beam, due to the FRP EBR, the yielding sections are still able to carry additional load and at the same time plastic hinge formation is restricted. Concerning the experimental data, a good agreement is observed with the calculated curve. 4.3.3 Debonding mechanism In this case debonding occurs at the top laminate, above the mid support, by crack bridging (Figure 18). Following the calculations, a debonding load of 197 kN is expected. Experimentally the laminate debonds at 172 kN. This is a difference of 25 kN or 12.7% with the calculated value following fib bulletin 14. The debonding starts at a crack, located at the mid support, and debonds towards the left point load in Figure 18. The strain distribution at the FRP ends anchored in the compression zone is given in Figures 19 and 20. As can be seen in Figure 19, for the end of the soffit laminate near to the mid support, and in Figure 20, for both ends of the top laminate, the strain caused by com-

4.3.2 Moment redistribution The moment redistribution of CB2 is illustrated in Fig. 17. Because the used amount of internal and external reinforcement is chosen nearly according to the linear elastic moment distribution, hardly any moment redistribution is observed. Table 8.

Properties of Concrete and CFRP.

Compr. strength Tensile strength Failure strain E-modulus

Table 9.

Concrete

CFRP

36.0 N/mm² 3.3 N/mm² 0.35% 32000 N/mm²

2768 N/mm² 1.46% 189900 N/mm²

Properties of steel reinforcement. Reinforcement Reinforcement in span* at support

Yielding strength Yielding strain Tensile strength Failure strain E-modulus

570 N/mm² 0.28% 670 N/mm² 12.40% 210000 N/mm²

570 N/mm² 0.28% 670 N/mm² 12.40% 210000 N/mm²

* not tested, same values assumed as span.

240 span collapses

Non-linear moment distribution (with FRP EBR)

F [kN]

220

Figure 18. bridging.

200

Debonding at the mid-support by crack

180

140 120

Non-linear moment distribution (without FRP EBR)

100

60

-200

-160

-120

MM - span moment span [kNm] [kNm] - span moment span

Figure 17.

10 20 30 40

0,005 mm

Linear elastic moment distribution

80

40 span cracks 20 support cracks 0 -80 -40 0

0,01

%

160 span yields support yields

0 -0,005

Experimental data

50 60

0

200

400

600

800

70 80 90 100

-0,01

110 120 130

-0,015 40

80

120

160

200

139 150

Msupport [kNm] - supportmoment moment Msupport [kNm] - support

-0,02

Figure 19.

Moment redistribution of CB2.

46

160

Strain in compression zone of soffit laminate.

Table 10.

0 1200 -0,01

Properties of Concrete and CFRP.

10

%

0,01

mm

20

Concrete

CFRP

35.3 N/mm² 3.2 N/mm² 0.35% 32000 N/mm²

– 2768 N/mm² 1.46% 189900 N/mm²

30

1400

1600

1800

2000

2200

2400

40

Compr. strength Tensile strength Failure strain E-modulus

50 60 70

-0,02

80 90

-0,03

100 110

-0,04

120 130

Table 11.

139

-0,05

150

Properties of steel reinforcement.

160

-0,06

Figure 20.

Strain in compression zone of top laminate.

Yielding strength Yielding strain Tensile strength Failure strain E-modulus

Reinforcement in span*

Reinforcement at support

589 N/mm² 0.26% 674 N/mm² 12.40% 22300 N/mm²

589 N/mm² 0.26% 674 N/mm² 12.40% 223000 N/mm²

*not tested, same values assumed as span. 240

F [kN]

220

Figure 21.

200

Internal steel configuration of CB3.

Non-linear moment distribution (with FRP EBR)

180 160 span collapses 140

pression is quasi linear over the length of the laminate end. In both cases the strain is measured by four strain gauges. By visual inspection of the laminate ends, anchored in the compression zones, during the test no buckling of the laminate ends could be noticed.

span yields

120

100 support yields 80

Experimentaldata Linear elastic moment distribution

60

Non-linear moment distribution (without FRP EBR)

40

4.4

Continuous Beam 3 (CB3) -200

-160

-120

M span [kNm] - span moment

4.4.1 Configuration The last tested beam has internal reinforcement as shown in Figure 21. The beam is designed with high internal reinforcement ratio in the spans (ρs,span = 0.90%) and low amount of reinforcement above the mid-support (ρs,support = 0.29%). As external reinforcement, one CFRP laminate with the length of 5000 mm is applied at the mid support. The section of the CFRP laminate is 100 mm × 1.2 mm (ρf,support = 0.17%). The characteristics of the materials are given in Tables 10 and 11. These values result from standard tensile and compression tests.

Figure 22.

span cracks 20 support cracks 0 -80 -40 0

40

80

120

160

200

Msupport [kNm] - support moment

Moment redistribution of CB3.

formed in the (unstrengthened) spans. The calculated debonding load following fib bulletin 14 is equal to 124 kN. Concerning the experimental data, a good agreement is observed with the calculated curve. 4.4.3 Debonding mechanism For this beam again debonding of the top laminate at the mid support occurs. Following the calculations, a debonding load of 124 kN is expected. Experimentally the laminate debonds at 115 kN. This is a difference of 9 kN or 7.3% of the calculated value. The debonding starts at a crack, located at the mid support, and debonds towards the right point load in Figure 23. Fig. 24 gives the measured compression strains along the length of the laminate end, which is ancored in the compression zone. The measurements are carried out by the use of six strain gauges. As shown in

4.4.2 Moment redistribution In Figure 22, the moment redistribution of CB3 is illustrated. Following the non-linear theory, the central support yields first. For the unstrengthened beam, after yielding of the mid support, a plastic hinge is formed (vertical part of the dashed moment distribution curve). For the strengthened beam, although the strengthened central support still starts to yield first, the FRP allows the mid support to continue resisting the additional load. At increasing load when the spans start to yield, plastic hinges will be

47

0,01

Debonding at the mid support by crack

%

Figure 23. bridging.

on the amount of internal steel reinforcement in the span and mid-support and the strengthening configuration. After reaching the yield moment, the FRP strengthened cross-section is still able to contribute in carrying the additional load. Hence, such FRP strengthened cross-sections restrict the rotation of a plastic hinge at that location (and the related moment redistribution), but allow to transfer plastic hinge formation to unstrengthened cross-sections with high internal steel reinforcement ratio. Concerning the debonding of the FRP-EBR, continuous beams compression zones are available at which FRP laminates can be anchored. Here two debonding mechanisms (concrete rip-off and anchorage failure) can be avoided. By means of an analytical study, it has been demonstrated that the debonding loads are also governed by the shear force and moment redistribution. This redistribution is occurring in FRP strengthened continuous beams and depends on the amount of FRP in the spans and at the mid-support, relative to each other. Because of the specific influence on the debonding load, redistribution of internal forces should be considered when verifying the debonding load of strengthened continuous beams. Depending on the situation (amount of FRP, type and location of the debonding phenomenon) both an increased or decreased value of the debonding load may be obtained.

0 1200 -0,01

mm 10

1400

1600

1800

2000

2200

2400

20 30 40 50 60

-0,02

70

-0,03

80 90

-0,04

100 110

-0,05

116

-0,06

Figure 24.

ACKNOWLEDGEMENTS Strain in compression zone of top laminate.

The authors acknowledge the financial support by FWO-Vlaanderen. the graph, the strain has a roughly linear character over the length of the laminate end. By visual inspection of the laminate ends, anchored in the compression zones, during the test, no buckling of the laminate ends could be noticed. In Figure 24 a (small) shift of the point of contraflexure caused by the non-linear moment redistribution can be observed.

5

REFERENCES Ashour, A.F., El-Refaie, S.A. and Garrity, S.W. 2004 Flexural strengthening of RC continuous beams using CFRP laminates. Cement and Concrete Composites 26(7): pp. 765–775. El-Refaie, S.A., Ashour, A.F. and Garrity, S.W. 2003 Sagging and Hogging Strengthening of Continuous Reinforced Concrete Beams Using Carbon Fiber-Reinforced polymer Sheets. ACI Structural Journal. Vol. 100: pp. 446–453. fib 2001 fib bulletin 14, Externally bonded FRP reinforcement for RC structures. International federation for structural concrete, Lausanne. Matthys, S. 2000 Structural behaviour and design of concrete members strengthened with externally bonded FRP reinforcement. Ghent University, Department of structural engineering. Taerwe, L. and Espion, B. 1989 Serviceability and the Nonlinear Design of Concrete Structures. IABSE PERIODICA 2/1989.

CONCLUSIONS

For unstrengthened continuous beams a considerable moment redistribution can be observed, especially after plastic hinge formation. The latter occurs after reaching the yield moment in the critical crosssection (where the moment is maximum). Almost no moment redistribution is however observed if the yield moment is reached at the same time in both the spans and the mid-support. In the case of FRP EBR strengthened continuous beams the observed behaviour largely depends

48

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Diagnostic analysis and therapy of severely cracked bridges Hans-Peter Andrä & Markus Maier Leonhardt, Andrä und Partner, Stuttgart, Germany

ABSTRACT: Severe cracking in reinforced and/or prestressed concrete structures usually results from the accumulation of a number of causes, such as errors in structural design, detailing of reinforcement, concrete composition, concrete curing, overloading under construction, early lowering of falsework, underestimated thermal loadings, fatigue, and lack of maintenance etc. The paper presents examples of a consistent strategy of diagnosis and therapy.

1

INTRODUCTION

The spans were erected with an overhead erection gantry where the segments are dry jointed and the spans post tensioned with external prestressing cables within the void of the box. The 33 piers consist of two types of reinforced concrete substructures. Pier #19 and Pier #23 are twin column portal frames supported on pile caps founded on bored cast in place piles. All other piers are T-shaped piers (see figure 1). The column cross section is orthogonal, the width across the flats amounts to 3.62 m. The columns range from 1,5 m to 16 m in height and are supported on 8 m × 8 m square and 3 m deep pile caps which are founded on 4 bored cast in place piles with 1,8 m diameter. Each column supports a 18.65 m long cross head which is cantilevering out on both sides about 7.50 m. The cross heads depth varies from 2 m at the tip to 3,5 m at the column face, with an overall width of 4 m. These cantilever crossheads are the focus of the case study, which is presented within this paper.

Severe structural defects in buildings are generally due to the accumulation of poor design, poor construction and construction material and poor maintenance. The predominant cause is human error, which can only be reduced by means of thorough failure mode and effect analyses (FMEA). Human error often results from carelessness, disregard, insufficient knowledge, underestimating of decisive influences, blind faith in computer analysis, lack of plausibility checks and inadequate time and cost pressures. Preemptive risk analyses by independent reviewers and checkers as well as site supervision and continuous monitoring can significantly reduce the risk of error accumulation. Severe cracks in the cross heads of the T-shaped piers of the Kepong Flyover in Kuala Lumpur are a typical result of the accumulation of such errors. Lessons learned from these errors are described in detail as follows.

2

DESCRIPTION OF THE STRUCTURE

The Kepong Flyover is a dual three-lane elevated carriageway as a part of the exterior highway system of Middle Ring Road II (MRRII) in Kuala Lumpur. The overall length of the superstructure is 1537 m between abutments. The cross section consists of twin precast match cast single cell box girders which are 11,5 m wide between barriers with an overall depth of 2,50 m. The bridge has 34 spans constructed by a span by span construction method. The 45,2 m long spans are simply supported with a continuous link slab at the piers to provide continuity of the running surface. Movement joints are provided at every fifth pier. At each pier each box is supported on two bearings, one underneath each web.

Figure 1. Kepong Flyover Viaduct at Middle Ring Road 2 in Kuala Lumpur, Malaysia.

49

Expert reports from various parties with various interpretations finally resulted in inconsistent conclusions about the causes of the cracks and the cracks influence on the structure load bearing capacity. A major concern of the reports was the sensibility for delayed ettringite formation (DEF) in the concrete, where crack growth results from ettringite expansion. The observed typical crack patterns, however, could not be explained by DEF. Different expert opinions about the causes of measured test results are a well known phenomenon both in research and in surveys on defects of existing structures. This results from the fact that everybody draws conclusions from the basis of his own experience and will only find what he is searching for. In 2005 key management within JKR changed and new Director General, an excellent structural expert himself, realized the potential structural risk resulting out of this bridge and the structural findings. In October 2005, LAP was appointed for final and detail design for remedial works of the piers on the basis of a complete critical analysis of design, construction, material properties and maintenance. At the same time the bridge was closed for traffic. The focus of the remedial design concept was to eliminate potential structural deficiencies, to relief the structures stresses and to provide appropriate response on potential DEF occurrence. The contract for remedial design was awarded to a contractor on March 2006. Execution works began after only a one month mobilization period in April 2006. During the following 6 months all 33 T-shaped piers were structurally rehabilitated, with maximum of 18 piers being remedied simultaneously. Site logistics and strengthening sequence was most challenging, both for contractor, LAP site supervision and client, while traffic on highway underneath the bridge was fully ongoing. The strengthening principles will be illustrated in chapter 7 in more detail. In November 2006 the 100% strengthened and rehabilitated bridge could be reopened to unlimited traffic.

3 TRACK RECORD OF THE STRUCTURE The contract for the flyover structure was awarded in May 1999 on a design & built basis. Design was done between July 1999 and August 2001. The construction started in February 2000 and was completed in April 2002. The flyover was opened to traffic in May 2002. First cracks on the crossheads were observed already during construction. Shortly after opening of the bridge for traffic further cracks were developing and excessively widening. First reports about these cracks are dated back to January 2003 respectively November 2003. Detail crack mapping of all 33 piers including abutments was subsequently appointed to a specialist contractor in May 2004, in order to systematically have documentation of crack patterns including crack sizes (see figure 2). Various field investigations including cover meter tests, verification of rebar sizes and spacing, core sampling for strength tests and monitoring installations were initiated. Typical crack patterns consist of • two longitudinal (along the direction of the cross head) cracks in between the two adjacent bearings, more or less partitioning the cross head into 3 segments, • a onionskin shaped crack mesh on the upper flanks at the cross head centre, partly interconnected in transverse direction by cracks on the top surface, • horizontal cracks over the piers underneath the onionskin mesh, • inclined cracks running down the face of the cross beams, and • inclined cracks underneath and in the vicinity of the bearings.

side face pier #9, max. crack width 3.2 mm

4

LOADING HISTORY OF THE CROSS HEADS

The cross heads carry the superstructure dead load and final stage live loads (vertical and horizontal). During construction, the gantry loads in addition to the superstructure dead load are a major, sometimes even governing load case. 4.1 top and bottom faces pier #32, max crack width 3.0 mm

Thermal loading due to curing of (mass) concrete

The pouring in one shot of the massive cross-section of the crossheads with 4.00 m in width and 3.50 m in depth and the subsequent curing of the concrete

Figure 2. detailed scaled crack mapping of all crosshead surfaces.

50

causes a heat development and distribution within the member, arising from cement hydration. The applied concrete mixture proved a high content of Portland cement which promotes heat development. Temperature gradients from cooling down outer surfaces exposed to formwork or air while the inner core of the structure is still hot causes internal tensile stresses, which may exceed tensile strength of the immature concrete. Special caution must therefore be undertaken to control early thermal cracking by providing sufficient reinforcement in size and arrangement to control regular crack spacing as well as limit individual cracks width. On the other hand concrete mixture must be carefully selected to avoid excessive heat development, e.g. by using cement with lower hydration heat development. 4.2

Loading during erection of the superstructure

The superstructure was erected using an overhead erection gantry. The gantry was supported on the cross head and carried the load of the precast superstructure elements which were attached and winched up into position. Once all 15 segments with a weight of 80.8 tons each were in position and adjusted for fit, line and level the external post-tensioning was applied. Upon completion of stressing the completed superstructure was lowered onto the piers transferring the superstructure load from the gantry supports to the permanent bearings. After completion of the first superstructure, the gantry was shifted in transverse direction to the adjacent carriageway for the erection of the second superstructure in the same span using the same erection sequence. Upon completion of the second superstructure the gantry was launched forward to the next pier for the next span erection. This cycle was then repeated starting from abutment A in the north and then proceeded until the entire viaduct was completed. The crossheads are more or less symmetrically loaded due to superstructure dead loads in the final stage. In the erection stage however they are loaded by high out of balance loads as well as Torsion resulting from the rear and font gantry nose beam supports. (see figure 3). The overall weight of 15 segments per span sums up to 1212 tons, the overall erection gantry tonnage sums up to 617 tons which is about half of the superstructures weight. The maximum loading coming from rear and front gantry support occurs just prior to placement of posttensioned box girder onto the final bearings, i.e. when the completed box girder is still fully supported by the gantry. The total weight which is resting on the crossheads gantry support locations short before lowering down the assembled superstructure to the final bearings sums up to 1829 tons. Both gantry supports are eccentric to centerlines of the piers, therefore the gantry loadings generate high

Figure 3. principle sketch of erection stage with rear and front gantry support.

torsion and high out of balance bending moments especially at the front gantry support during the erection of the first superstructure. However since the gantry progressed span by span, each pier suffers clockwise torsion from front gantry support during erection of this span, while suffering anticlockwise torsion from rear gantry support during erection of next span. 4.3

Loading transferred from the superstructure while “in-service”

British Standard BS 5400 was applied for the bridge design. The design loads for the in-service condition include dead load from superstructure and superimposed dead load comprising from asphalt surface and parapets as well as live loads resulting from HA, and HB 45 units load design vehicles both for serviceability state (SLS) and ultimate limit state (ULS) with its load combinations as specified in BS 5400. The bridge is 11.50 m wide between the kerbs, which acc. BS5400 corresponds to four notial design lanes, although each deck of the bridge will actually only be used as a three lane deck. Figure 4 shows the governing HA/HB45 load arrangement on a single span, which generates worst case supporting forces to be transferred onto the bearing pads, which in turn generates maximum crossheads in service stresses (bending, out of balance moments, shear and torsion).

51

Maximum crosshead cantilever bending moment e.g. results to: due to dead loads due to live loads

MDead MLive

= 76.50 MNm = 31.50 MNm

total

Mtot

= 108.00 MNm

Horizontal loadings depend on and result from the bearing lay out of the superstructure. Thermal elongation/shrinkage creating bearing friction forces as well as other horizontal loadings due to acceleration/braking and wind are minor load effects, but these effects must not be neglected. The resulting stresses are superimposed by other horizontal local stresses such splitting or spalling stresses underneath the bearings. The accumulated stresses are of critical magnitude if no adequate reinforcement is provided. This is even more important for cyclic loadings. Both vertical loads coming from live load design vehicles as well as horizontal loads resulting from bearing pattern are considered to be cyclic. 4.4

Effects of the load history on the observed cracks

Both thermal effects and construction loadings had not been sufficiently addressed in the design. Thermal effects result in a mesh of micro cracks, weakening the complete crossbeam envelope and in DEF vulnerability. The inclined shear cracks at the cross beam faces are due to underestimated torsion loadings during construction. Ignorance of construction loadings and final material composition is a general problem, because concrete composition according to local availability and construction equipment are not fully known in the design phase or because there is a discontinuity in the interface between design and construction. This discontinuity between design and construction generally results from contractual liability reasons and not from engineering reasons. Contracts and attitudes between the parties involved which disregard the success of the project as a whole imply significant risk potentials.

5

DESIGN AND DETAILING OF THE CROSS HEADS

The structural system of the T shaped pier seems to be quite simple. Bending moments and shear forces at governing sections can be determined and designed for by hand calculation. The determination of internal stresses resulting from the local application and introduction of the bearing forces into the cross head section however is rather complex.

Figure 4. Governing vehicle load arrangement for 4 notial design lane analysis and resulting supporting forces on bearings.

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5.1

A comparison between the results of a finite element analysis and a strut and tie model is shown in Figs. 5 and 6. Even though both approaches provide equal results in terms of equilibrium, the finite element analysis requires a significant amount of interpretation and is much more susceptible to human error than the strut and tie model. The finite element analysis should rather be used as a tool to verify the alignment of the strut and tie model. Detailing according to the strut and tie model however is quite simple, and the necessity of reinforcement in three dimensions is obvious. An erroneous interpretation of the finite analysis model could suggest that transverse reinforcement between the bearings and splitting reinforcement underneath the bearings would not be needed because the concrete tensile stresses are low. The integral of these stresses however results in large forces. FE analysis determines the governing tensile stresses in a concrete zone which is pre-cracked due to thermal reasons and which is considerably intersected by the reinforcement, Fig. 7. Due to this intersection of rebars there is virtually no concrete section left which could transfer tensile stresses. This does not comply with the suppositions of the FE analysis. The superposition of individually negligible horizontal loadings from bearing friction, braking, wind, horizontal temperature gradients etc. add up to significant stresses and result in fatigue failure of non reinforced concrete sections. The sudden and brittle collapse of the bridge seat of the Reichsbruecke in Vienna 1976, Fig. 8, demonstrates that this may happen even after years of operation. The lack of sufficient transverse reinforcement and of proper splitting reinforcement is responsible for the two longitudinal separating cross head cracks and the diagonal cracks underneath the bearings. Blind faith in even highly sophisticated FE computer analyses implies significant risk potentials.

Strut and tie model versus FE analysis

Due to the fact that the loads are not applied as line loads across the section but in terms of single loads along the edges of the cantilever, the internal load path actually corresponds to a strut and tie space frame with compression and tensile forces in three directions. This space frame reproduces the load path from the bearings through the cross head to the pier junction and through the pier and the raft into the piles. The space frame model includes both the global force actions and the internal thrust actions due to the local deviation of the load path. Both compression struts and tensile ties need to be analyzed and all tensile forces need to be covered by properly detailed and anchored reinforcement. “The Art of Detailing”, published by the late Professor Fritz Leonhardt [1] more than 40 years ago describes the engineering concept of strut and tie modeling in detail. Structural design on the basis of strut and tie modeling has now been incorporated in the relevant provisions of Eurocode 2.

Figure 5.

Finite Element Analysis of the cross head.

Figure 6. Strut and Tie Space Frame Model of the cross head.

Figure 7. Tensile zone in concrete intersected by rebars.

53

Figure 9.

Figure 8.

5.2

Reinforcement of the cross beam.

Collapse of the Reichsbruecke Vienna.

Detailing of reinforcement

The reinforcement of the crossbeam is shown in Fig. 9. The top two layers of the crossbeam T40 bars at a spacing of 120 mm were lapped over the column, i.e. at the location of maximum bending moments. The lap length was chosen at its minima allowed acc. to BS 5400. The necessary lap length according to the German DIN Code and US AASHTO codes in an area with “unfavourable bond condition”, i.e. “top reinforcement so placed that more than 30.5 cm of concrete are cast below” would be about 1.8 times longer and would require transverse stirrups to anchor splitting forces from the transmission of the forces between the lapped bars. The transmission of the forces from each bar of approximately 500 kN to the surrounding concrete induces significant tensile splitting stresses normal to the bar axes. Due to the extremely high concentration of lap splices, the remaining concrete strips between the bars are to small to transfer the high transmission forces, Fig. 10. Bond failure of this lap splice was anavoidable. Locating the laps outside the maximum bending moment zone and staggering hooked laps accordingly would have complied with the art of detailing rules [1]. The column bars were not lapped with the crossbeam bars, so that there is no sufficient reinforcement to resist the transfer of the cantilever moment between crossbeam and column due to eccentric crossbeam loadings during construction. Overall stability is still possible as long as the eccentricity of the normal force in the column produces a bending moment of equal magnitude. Horizontal cracks above the pier underneath the onionskin mesh however are

Figure 10. Transmission of forces between lapped bars [2].

due to this unreinforced section between column and crossbeam. 6

CONCRETE COMPOSITION AND MAINTENANCE

The 3,5 m by 4,0 m cross-section of the cross head is massive concrete section which would have required particular concrete composition with low cement contents and cooling of the fresh concrete to avoid temperatures of more than 70°C due to the liberated hydration heat. High hydration temperatures do not only result in internal stresses and cause early cracking but also create optimal conditions for ettringite crystallization. Both primary ettringite formation at the start of hydration and ettringite formation in the hardened concrete texture may occur. The latter is referred to as delayed ettringite formation, DEF. There is a controversial discussion in literature on whether the expansion caused by ettringite was the primary cause of texture damage or whether it was a subsequent manifestation of prior damage. Regarding the examined concrete composition of the relevant piers (maximum cement content 560 kg/m³, SO2 content about 3%), the ettringite density of 1,77 g/cm³

54

5. Add lacking vertical reinforcement legs, adhesively bonded within concrete member to take the tensile component forces, resulting from lapsplice action according Figure 10. 6. Placing of self compacting concrete to close the highly reinforced lap-splice area in order to achieve best bond conditions for load transfer within lap splice. 7. Apply moderate permanent post-tensioning force for crack control by 18 externally bonded CFRP tendons, each stressed to 400 kN. 8. Seal and finish the surface for waterproofing reasons to take necessary actions in terms of suspected DEF problems.

corresponds to a volume of some 5% ettringite per cubic metre of concrete on the average. Contrasted with an average pore volume of 13%, the maximum amount of ettringite that can be formed is significantly smaller than the total pore volume. Damages of the concrete texture due to DEF are therefore very unlikely. There is no doubt however that the ingress of moisture and water into concrete cracks due to tensile stresses promotes the local accumulation of newly formed ettringite on the flanks of these cracks. Recrystallized ettringite is thus seen not as the cause of damage but rather as its consequence. Due to defective piping of the drainage of the bridge superstructure over the piers, the crossbeams were regularly flooded and wetted in the rainy season. In addition, the local humidity is very high. Water and moisture induced ettringite formation thus expanded the original excessive cracking from structural deficiencies. Ettringite formation can be stopped after elastic grouting of the existing cracks and waterproofing of the pier head surface.

7 7.1

7.2

Installation of tie-frame

Since no vertical support was accepted, to relieve the structure during strengthening, an external tie-frame was installed, which could be stressed up to 18 MN. This external horizontal force introduction on the end faces of the crossbeam allowed a stepwise relieve of the internal reinforcement stresses, without changing the statical system. The total resultant force of the external tie-frame was at same level than the top tension force within the reinforcement layers resulting out of cantilever bending moment. The tie-frame together with end buttress including working platform with a total weight of 80 tons were lifted in one piece during night (Fig. 11). Immediately after the tie-frame and the buttresses were installed in final position, the tie-frame stressing works commenced. Strands aligned within the main chords of the tie-frame were stressed to 65% UTS to secure the structure.

REHABILITATION General

Since the identified main crack patterns were addressed to have systematic structural causes, and since every pier was built according the same drawing, it was required to apply strengthening and rehabilitation procedures to every pier, regardless of the present degree of cracking excess. It would have been only a matter of time, once cracking excess would have propagated, since the stability could only be explained by assuming concrete tensile strength, which is not known as a reliable concretes property while bearing cyclic loadings. Following was the sequence of strengthening procedure for each and every pier:

7.3

Grouting of detected cracks

All cracks were grouted with epoxy resin. The grouting works were carried out sequentially and directionally from the bottom of the cross-head to the top. At some piers the epoxy resin could easily be pumped at the onion shell-like cracks over a length of 4 m through the complete pier section. The maximum quantity which was pumped into one packer was about 60 litres which was an indication of –beside cracks-poorly compacted concrete as well as presence of large honey combs.

1. Securing the crossbeam structure with a temporary, horizontal and external tie-frame to release the bending stresses by temporary post-tensioning up to 18 MN stressing force. The tie-frame was at same time providing the working platform. Temporary vertical supports were not accepted, since the highway below the bridge must be in full operation. 2. Grouting of all detected cracks with epoxy resin to fill the voids. 3. Transverse post-tensioning of crossheads with prestressing bars to compensate the lack of transverse reinforcement and to stress cracks faces of longitudinal cracks against grout. 4. Replacing and removal of highly and undefined cracked concrete within lap-splice area down to 3rd layer of lap-splice reinforcement by means of hydro-water-jetting.

7.4

Transverse post-tensioning

A total of 15 nos. of prestressing bars per pier diameter 32 and 40 mm respectively were to be installed to cater for insufficient transverse reinforcement below the bearings. The location of the existing reinforcement was checked to identify the optimized coring location in order to avoid, cutting or damaging of existing vertical

55

For structural safety reasons the lap splice was opened in two steps: First opening of the centre section with a width of 2 m and repair of that section (Area I). Then opening of the adjacent areas followed, each with an approximate width of 1m (Area II). A high pressure water jetting unit was set up below the flyover serving two demolition areas simultaneously. The area of the lap splice was completely covered to prevent parts from falling onto the traffic below the flyover structure. Specially designed temporary drain channels were installed to collect the waste water and divert it to a collection tank for subsequent cleaning purposes. Initially, the concrete cover was removed by water jetting to expose the top reinforcement bars. 250 mm deep pilot holes each with a spacing of 250 mm were drilled around the perimeter of Area 1. These pilot holes helped to speed up and control the consistency of the water jetting process. Finally all bars with diameter 40 mm were fully exposed with a minimum clearance of 20 mm to all sides. For proper bonding purposes all rebar was cleaned of cement slurry and loose particles. A full lap-splice mapping followed. The main purpose of the remedial exercise was to repair the deficiencies in the lap splice area. All of the T40 rebar were generally lapped in a vertical manner at the maximum bending moment area. To prevent the concrete in the splice zone from spalling off, additional transverse (vertical) reinforcement (U-Bars) needed to be installed as the existing stirrup reinforcement was insufficient. To install the additional U-bar reinforcment, holes were drilled into the concrete between the T40 bars and bond-anchored with epoxy resin.

Figure 11a. External tie-frame stressed up to 18MN, to relieve the structure during strengthening, allowed to open the lap splice.

Figure 11b. Lifting procedure of 80 tons tie-frame incl. working platform.

7.6

reinforcement bars. The anchorages of the transverse stressing bars had to be placed inbetween the gap of the 3rd to 4th CFRP tendon layer, later running longitudinally on both sides of the cross-head. Therefore it was extremely important to limit the drilling deviation over the length of 4 m through the crosshead to only 20 mm. However this was successfully achieved by the competent coring subcontractor. The bars were stressed to 65% UTS, grouted and final anchor head corrosion protection was applied. 7.5

Pouring and closing lap splice with self compacting concrete (SCC)

After the U-bars were installed the lap splice area was closed by self compacting concrete (SCC). SCC was chosen for the following 5 reasons: − The spacing between the spliced bars was expected to vary substantially. − The early concrete strength had to be high to allow re-opening of the flyover soon after concrete placement. − The bonding quality of the concrete has to be extremely good in the lap splice area. − Since relatively small concrete quantities were required and for quality assurance reasons the aggregates for the SCC had to be supplied in sacks and mixed on site. − Requirement of low shrinkage properties.

Lap splice strengthening

The most important remedial work procedure was the rehabilitation of the lap-splice. To allow to open the lap-splice without any vertical support of the cantilever ends the temporary prestressing force, horizontally introduced by tie-frame jacking was increased to 18MN, which represents the ultimate tensile force of the top two reinforcement layers.

All lap splices had to be concreted during weekends when the flyover could be temporarily closed for traffic. to give the SCC sufficient hardening time until reopening of the flyover for traffic.

56

The SCC was mixed on top of scaffolding situated on top of the flyover. The concrete was discharged directly from the scaffolding through a PVC-pipe into the lap-splice area. Because of the required shifting of the scaffolding the mixer was later on installed on a truck which could easily service up to 6 piers in one night. To achieve the required mix parameters (e.g. fresh concrete temperature, viscosity, flow spread etc.) it was important to implement the following additional measures: − Cooling of the pre-packed mixture in an airconditioned container for about 24 hours. − Cooling of the water to a predetermined temperature. A strict QC/QA procedure was established to garantee satisfactory SCC properties. After curing time of the SCC the temporary tieframe post-tensioning force was relieved from 18MN down to 9MN to get the lap-splice activated. 7.7

Figure 13. Permanent steel buttresses stressed against end faces of crossbeam by 18 Nos. of CFRP tendons.

Permanent longitudinal post-tensioning by CFRP tendons

The cross-heads were permanently prestressed with 18 Nos. of CFRP tendons, each prestressed to 400kN. CFRP-tendons were chosen for the following reasons:

stressing force was simultaneously relieved by deinstallation strand by strand. Finally the tie-frame was lowered down. Fig. 13 shows the anchorage detail and the final situation, where the steel buttress installed on both end faces of the crossbeam are permanently interconnected by 18 post-tensioned CFRP tendons, properly covered and protected against mechanical impact.

− Easy handling of the tendons due to its light weight. − High durability in the hot and humid climate of Malaysia. − Easy bonding of the tendons with the surface of the concrete, which increases the factor of safety during the ULS and provides crack control and cracks width limitation.

8

CONCLUSIONS

Structural design, construction process, selection of materials and maintenance are susceptible to human error. Thorough risk analysis is required to prevent severe defects or failures due to the accumulation of these errors. The paper presents the focuses of a chain of errors which could have been prevented by means of thorough failure mode and effect analyses (FMEA). Dominant errors were

The tension force was applied in incremental steps, to allow filling the gap behind the CFRP-tendons with adhesive. Starting with each 6 CFRP tendons on side faces the top 6 tendons were installed. During this stressing process, while increasing the permanent CFRP post-tensioning force, the temporary tie-frame

• • • • •

Imperfect loading assumptions Blind application of code provisions Blind faith in FE analysis Ignorance of the art of detailing Inconsistency of the interface between design and the construction process • Inadequate material selection • Inadequate contractual relationships • Inadequate maintenance.

Even the first steps of the rehabilitation process were again governed by human error, especially due to legal and liability reasons.

Figure 12. Opened lap splice additionally reinforced by vertical U-Bars.

57

[3] Fritz Leonhardt, “Vorlesungen über Massivbau, Teil 1, 2, 3”, Springer Verlag, Berlin 1975. [4] Leonhardt, Andrä und Partner, “Summary on Investigations, Findings and Conclusions Concerning Cracks in Crossheads at MRR2 in Kuala Lumpur”, Berlin, Stuttgart, Nov. 2004. [5] Leonhardt, Andrä und Partner, “Report on Remedial Detail Design for Main Viaduct Piers P1 to P33 and Abutment B”, Berlin/Stuttgart, Dec. 2005. [6] Lee Coon Siang, Ismail Mohamad Taib, Dato’ Mohamad Razali b Othman, Causes and Retrofitting for Failure at Kuala Lumpur MRR II Highway Viaduct Malaysia, IABSE Symposium Lisbon 2005.

The prevention of the errors described requires a fundamental change in economical competition standards. Applied science has to dominate short dated profit considerations, not vice versa.

REFERENCES [1] Fritz Leonhardt, “über die Kunst des Bewehrens von Stahlbetontragwerken”, Beton- und Stahlbetonbau, Vol 60, 1965, No. 8, pp. 181–192; No. 9, pp. 212–220. [2] Fritz Leonhardt, “Das Bewehren von Stahlbetontragwerken”, Betonkalender 1971, Wilhelm Ernst & Sohn, Berlin, Part II, pp. 303–330

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Invited papers

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Research in the field of repair – Actual approaches and future needs M. Raupach Institute of Building Materials Research (ibac), RWTH Aachen University, Germany

ABSTRACT: Repair and protection of concrete structures are generally complex tasks requiring special knowledge in different fields. During recent years extensive scientific research has been carried out to study the corrosion mechanisms (chloride induced corrosion, hydrogen embrittlement, etc.), but only limited systematic research has been conducted in the area of the effectiveness, durability and sustainability of materials and systems for repair and protection. Since the 1980th the main focus of research in the area of concrete repair have been special methods, mainly electrochemical methods to protect the reinforcement against corrosion like cathodic protection, electrochemical desalination, electrochemical realcalisation or the use of inhibitors. This paper deals with selected actual approaches, ideas and needs for future research in the area of repair and protection of concrete structures.

1

GENERAL

During recent years protection and repair of concrete structures has become an economically important and scientifically interesting field. World-wide the investments for maintenance, repair and restoration of the infrastructure are increasing. As buildings generally are unique with own histories individual solutions are required for maintenance and repair. This has lead to the development of lots of different basic strategies and solutions for the repair of concrete structures. However, these strategies and solutions can be classified: The new European Standard Series EN 1504 lists e.g. 43 methods to protect and repair concrete structures according to altogether 11 different principles. It is the task of the designer of repair or protection works to select appropriate solutions taking the relevant technical and economical facts into account. Repair and protection of concrete structures are complex tasks requiring special knowledge in different fields (see Fig. 1):

Figure 1. Relevant aspects for protection and repair of concrete structures.

about 1975 the amount of damages is increasing considerably resulting in the fact, that in the meantime reinforcement corrosion has become one of the major problems of maintaining the infrastructure. Most critical are structures exposed to chloride environment like offshore of coastal structures exposed to seawater or bridges or parking garages exposed to de-icing salts. However, also carbonation of the concrete can lead to severe problems, especially when the concrete cover is to low or in cases of insufficient compaction of the concrete. Since the 1980th the main focus of research in the area of concrete repair has been special methods, mainly electrochemical methods to protect the reinforcement against corrosion, e.g.

• • • •

Static and dynamic behaviour of buildings Damage processes and corrosion behaviour Repair options, principles and methods Properties and behaviour of materials and systems for repair • Quality control and maintenance. Extensive scientific research has been carried out to study the corrosion mechanisms (chloride induced corrosion, hydrogen embitterment, etc.), but only limited research has been conducted in the area of the effectiveness, durability and sustainability of materials and systems for repair and protection. In the 1960s first major damages on concrete buildings induced by reinforcement corrosion have been documented. Since

• Cathodic protection of the reinforcement • Electrochemical removal of chlorides from the concrete • Electrochemical realcalisation of the concrete • Corrosion inhibitors for concrete to protect the reinforcement.

61

Repair materials like repair mortars, materials for crack injection, hydrophobic agents, impregnations, coatings, glues, etc. are typically developed by the industry, sometimes in cooperation with universities, but usually not as part of scientific research programs. Therefore the knowledge regarding the properties of repair materials is limited and it can be assumed that there is a large potential in improving the materials and to develop totally new materials with a taylor-made profile of properties needed for repair and protection. This paper deals with selected actual approaches, ideas and needs for future research in the area of repair and protection of concrete structures.

2

DIAGNOSIS AND PROGNOSIS

Diagnosis of the actual status of a structure and prognosis of the expected progress of damages are important first steps in repair: The better the quality of diagnosis and prognosis the better the repair and protection works can be designed and carried out. Especially in the case of chloride induced corrosion a diagnosis of the whole concrete surface is much more effective than taking single samples from selected points. Figure 2 shows potential mapping of a parking deck in lines with a distance of 50 cm which give information of the probability of corrosion at the whole investigated surface. Figure 3 gives an example of such a potential map. The potential maps are a basis to decide which repair methods should be used in which areas: In areas with low chloride contents protective coating of the concrete may be sufficient while in areas with high chloride contents at the reinforcement replacement of the concrete or electrochemical methods like cathodic protection may be required. However, usually selected additional tests (chloride analysis, determination of the remaining cross section of the reinforcement, etc.) are carried out in areas with typical potentials. Figure 4 shows as an example the classification of a large concrete surface into sections with different repair needs. To promote the idea of a full-surface diagnosis in practise, systems need to be developed which allow a determination of the relevant parameters in a short time for a low cost. One idea is the Betoscan-System, which is presented in another paper of this conference (Figure 5). As the Betoscan is designed only for horizontal surfaces, also other systems for vertical walls or overhead are required. In the future furthermore evaluation tools should be developed to support the designer to select the best repair and protection method for each section of a structure. Another approach to improve the quality of diagnosis is the use of new techniques of chemical analysis of the concrete. Figure 6 shows as an example

Figure 2.

Potential mapping on a parking deck.

Figure 3. Potential map showing areas with different probabilities of corrosion.

Figure 4. Schematic representation of the classification of a concrete surface into sections with different repair needs.

an instrument using the laser induced breakdown spectroscopy (LIBS) for a two-dimensional analysis of selected chemical elements of a concrete surface. This method is also described in an own paper of this conference.

62

Figure 5. Example for an instrument allowing a fullsurface diagnosis.

Figure 7. Area with insufficient adhesion of a polymer coating on concrete.

Figure 6. Example for an instrument allowing a twodimensional quick analysis of relevant chemical elements in concrete using the LIBS-technology (ILCOM).

3

Figure 8. SEM view of a zone with a delaminated coating on concrete.

UNDERSTANDING OF THE ADHESION MECHANISM OF REPAIR MATERIALS ON CONCRETE

repair materials on concrete are desired. These investigations must take the influence of ageing and physical, chemical or mechanical actions on the bond into account.

Bad adhesion is still one of the major problems and sources of damages in the area of repair. Figures 7 and 8 show examples of areas with insufficient adhesion. For more than 20 years research has been conducted to understand the mechanisms of adhesion between repair materials and concrete surfaces. However, as here is a huge amount of influencing factors (properties of concrete and repair material, preparation of the concrete, application, environmental conditions, etc.) and the adhesion mechanisms seem to be quite complex, actually there are only general rules how to apply repair materials on concrete, but there is no detailed understanding of the mechanisms behind or the quantitative effects of the relevant parameters. As quality control usually only the pull-off strength is compared with a limit value, which gives in many cases no reliable information on the quality of adhesion, especially regarding the long-term performance. To improve this situation fundamental research including tests on the nano scale on the adhesion of

4

EFFECTIVENESS AND DURABILITY OF THE METHODS FOR REPAIR ACCORDING TO EN 1504

As already mentioned the European Standard Series EN 1504 lists altogether 43 methods to repair or protect concrete structures according to altogether 11 principles. To support the designer selecting the optimal method for each project the possible fields of application and special questions regarding durability, required maintenance and sustainability of the methods should be investigated in detail. Such research is especially required for quite new applications. Scientific research on special questions regarding the application of cathodic protection

63

Figure 10. Glass systems for the protection of concrete against aggressive environment.

Figure 9. Cathodic protection on a parking deck exposed to deicing salts.

(see Figure 9) is e.g. given in a separate paper of this Conference. 5

DEVELOPMENT OF MATERIALS FOR REPAIR AND PROTECTION ON CONCRETE STRUCTURES

In the area of concrete repair and protection often materials with limited performance are used. Often the durability of repair materials is only in the range of about 10 years, especially in aggressive environment or under traffic. On the other hand the desired lifetime of a building is usually in the range of 50 years. Consequently there is a high potential for developments of durable and sustainable repair and protection systems. Figures 10 and 11 show the development of glass systems for the use in sewage plants or wastewater pipes with a high risk to biogenous sulfuric acid corrosion. Special glasses are available which are resistant against such acid attacks and highly durable. Using special glues glass sheets or panels can be mounted on concrete surfaces and protect them from aggressive substances. Looking to the long-term performance, it can be expected that repairs or protection measures using glass systems could last for decades or even longer. Besides the glass systems there is a need to develop also durable repair and protection systems for other environments. Looking into the future nanotechnology is expected to be the key to develop a new generation of repair materials with exiting properties and to improve existing repair materials considerably. The development is running from passive nanostructures (e.g. nano particles in coatings) via active nanostructures (e.g. amplifiers, adaptive structures) towards 3D nano systems with heterogeneous nano components and molecular nano systems with heterogeneous molecules.

Figure 11. Glass panels installed into a wastewater pipe as protection against biogenous sulfuric acid attack.

Figure 12. Leaves of a wet flower with drops of water as example for a self cleaning surface, which can be achieved using nano-technology.

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As an example actually the thickness of a polymer coating on a concrete surface can only be detected destructively by taking cores. As the owners usually do not wish to partially destroy the new coating often the coating thickness is not checked, but only the wet film thickness during application. On the other hand is the coating thickness one key parameter for the effectiveness and durability of a coating. One new approach to measure the coating thickness non-destructively on site is the NMR-mouse technology, which is presented in another paper of this conference, see Figure 13. Besides the thickness of coating layers also the humidity distribution or the ingress profiles of hydrophobic agents or impregnations should be detectable at the building economically and non-destructively.

Actually fundamental research is proceeding to create ultra-high performance materials with the following properties interesting for repair materials: • Self cleaning surfaces (superhydrophobic, super hydrophilic, “Lotus-effect”, see Figure 12, etc.), • Self healing coating systems (self-repairing, e.g. using nano tubes with filling material, shape memory effect), • Photo catalytic (e.g. TiO2, reduction of air pollution), • Antibacterial (bio-active, e.g. using nano-scale silver particles), • Corrosion resistant (e.g. nano coatings for metals), • Photovoltaic (e.g. surfaces of roofs), • Energy active materials (e.g. phase change materials PCM for heat control), • High strength surfaces (e.g. using carbon nano tubes CNT with extreme properties). Furthermore in future nano sensors integrated within the repair materials might show critical conditions e.g. by changing their color.

6

7

MAINTENANCE: SERVICE LIFE DESIGN AND MONITORING SYSTEMS

Service life design of new and repaired concrete structures and especially service life design based on monitoring data are key tools for the maintenance of buildings. An extensive German research project on corrosion models for the propagation phase of reinforcement corrosion is actually progressing (www.dfg-for537. de.vu). The status of the work is presented in an own session of this conference. Figure 14 shows the exposure of concrete specimens in seawater environment on the island Helgoland, where systematic data for the models are gathered. For design verification purposes and to update the service life design from time to time corrosion monitoring systems are an important tool. Figures 15 and 16 show examples of sensors monitoring the ingress of the critical chloride front with respect to reinforcement corrosion into the concrete.

QUALITY CONTROL

To achieve a high quality of repair and protection works quality control plays an important task. However, for some essential properties there are still no suitable test methods available.

Figure 13. Schematic representation of the laboratory set-up to investigate the suitability of the NMR mouse technology to measure coating thickness.

Figure 14. Exposure test in seawater of the island Helgoland supervised by the University of Stuttgart, Germany.

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Figure 15. Corrosion Sensor to monitor the critical depth and time-to-corrosion. Figure 16. The piles of the 36 km long Hangzhou Bay Bridge in China equipped with corrosion sensor shown in Figure 15.

It would be important to develop also sensor systems for other relevant deterioration mechanisms to continue the idea of “smart” structures.

8

Future research is expected to develop new systems and materials for repair based on nano-technology. Based on fundamental research in this field high performance solutions with new properties will be created. However, as shown in this paper there are also interesting approaches in the fields of diagnosis, maintenance and monitoring.

CONCLUSIONS

Actually interesting research projects are running in the field of repair. Lots of them are presented at this conference.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Reflections on future needs in concrete durability research and development Y. Ballim School of Civil & Environmental Engineering, University of the Witwatersrand, Johannesburg, South Africa

M.G. Alexander, H.D. Beushausen & P. Moyo Department of Civil Engineering, University of Cape Town, Cape Town, South Africa

ABSTRACT: There is no doubt that, over the past two decades, we have made enormous advances in the understanding and practice of concrete durability. Spurred by the often experienced early deterioration of reinforced concrete structures, with high capital investment for repair and rehabilitation, conceptions of design for durability have gained an increasingly higher level of importance in recent years. Durability design is beginning to be considered of equal importance as design for safety and serviceability. Nevertheless, durability concerns remain and early deterioration still attracts much expenditure. This paper is aimed at identifying important developments made in the field of concrete durability during the past two decades. Based on current design practice and current knowledge, future research and development needs are discussed, focusing on the influences of constituent materials, deterioration prevention methods, service life modelling of reinforced concrete structures, and performance-based test methods.

1

INTRODUCTION

fundamental questions of concrete durability, influenced the practice of concrete technology, or attended to needs of technology transfer in this field. While the issues raised are intended to have general application, the reader will find something of a South African bias.

It is not difficult to recognise or to acknowledge the significant changes that have taken place in the field of concrete durability over the last 25 years. This period has seen an enormous growth in our understanding of the mechanisms and processes of deterioration, the ways in which to protect concrete against early deterioration and the effective utilisation of concrete-making materials that will enhance durability. Equally important has been the extent to which the advocacy of durability—as a fundamental property of concrete—has resulted in a conceptual change in the minds of those involved in the design and construction of concrete structures. For practitioners, concepts of durability are beginning to be woven into the everyday discourse of concrete technology. In large measure, this change has been driven by the efforts of researchers throughout the world, both in developing our understanding of the theoretical issues and in making the case for greater attention to durability in the cement and concrete sector in general. In the best tradition of the path of knowledge development, the debates in the field of concrete durability have certainly been vigorous and many misconceptions or flawed theories lie on the sides of this path. However, in many areas, the issues and concerns are far from settled and the need for continuing research and development remains. This paper therefore presents some reflections on the future research needs in concrete durability. Much of this discussion relies on observations of the extent to which researchers have addressed some of the

2 2.1

MATERIALS Influence of material properties on reinforcement corrosion

Significant advances have been made in our understanding of the influence of concrete mixture

Figure 1. Reinforcement corrosion remains the most significant threats to the durability of reinforced concrete structures.

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area, other deterioration mechanisms are also in need of research attention. Some examples of such issues are:

constituents on the likelihood and rate of reinforcement corrosion. Much of this work has focused at understanding the role of cement extenders and admixtures in modifying the ionic and fluid transport properties of the concretes that we use in chlorideladen environments. Greater understanding of the notion of the ‘threshold’ level of chloride content in relation to the binder type has also helped to clarify much of the complexity of empirical data. Our sense is that continued work is necessary in developing a more fundamental understanding of the physico-chemical process related to the interactions between the products of hydration and corrosioninducing species such as chlorides or carbon dioxide. These studies should be focused on ingress, binding and flow of these materials in concrete. This is necessary in light of the rapidly changing nature of cementitious materials that are being used in concrete. Our reliance on models developed from empirical data, without fundamental understanding, limits our ability to predict the long-term performance of new materials used in concrete. Furthermore, the increasing use of combinations of supplementary cementitious materials—so called ternary blends—presents particular challenges in this context. The past five years have also seen enormous growth in the materials and technology of concrete admixtures. However, there is little information on the effects of these new materials on the general durability and deterioration of concrete. In particular, there is a serious lack of understanding of the effects of these materials on concrete exposed to the potential of damage due to reinforcement corrosion. In the areas of limiting the effects of corrosion of reinforcing steel, the following are some of the issues that will require attention:

− Our understanding of the mechanisms and process of sulphate attack remains hazy and there is much confusion and sometimes contradiction on this phenomenon (Neville, 2006). − In the case of alkali silica reaction, our knowledge of the strategies for prevention is fairly advanced. However, more work is clearly needed in the repair and rehabilitation of concrete damaged by this mechanism of attack. − Soft water attack and other dissolution processes also requires further research, particularly to develop our understanding of the strategies for protecting concrete against such deterioration. Much of the comments related to the materials research needs for reinforcement corrosion apply here in equal measure.

3

3.1

General

For reinforced concrete structures, the most important durability considerations concern reinforcement corrosion. As a result, the modeling of the ingress of aggressive agents such as chlorides and carbon dioxides has received considerable attention on recent years. However, many questions yet have to be answered in determining reliable and practical design procedures for reinforced concrete members subjected to chloride ingress or carbonation. Design approaches for durability can be divided into prescriptive concepts, also termed deemed to satisfy concepts, and performance-based concepts. Prescriptive concepts are based on material specification from given parameters such as exposure classes and life span of the structure. Following this approach, durability specifications in most existing codes and standards are based primarily on establishing constraints to the mix proportions of the concrete as a function of the severity of the exposure. Durability specifications in the South African standards (SANS 2005) and the new European standards (BS EN 2004), for example, follow the prescriptive concept and are of the ‘recipe’ type, setting limits on w/c ratios, cement contents, and compressive strength for different exposure classes. The design for durability includes the correct choice of exposure class and compliance with material requirements, concrete cover specifications, and curing procedures. However, durability is a concept that incorporates material properties, processing technology and environmental exposure conditions and, as such, it cannot easily be assessed through intrinsic material properties.

− Corrosion inhibitors: future research is needed to establish long-term performance and applicability − Service life of protective coatings − The development of corrosion-resistant reinforcement. There is certainly a need for an international shared database on long-term performance of concrete structures in relation to concrete mixture constituents. This will act as a touchstone for researchers to assess the suitability of proposed models against a historical knowledgebase of recorded performance. There is a special concern here regarding the potential problems with new materials such as high-performance concrete, for which long-term behaviour is unrecorded. In these materials, the likely effects of large amounts of unhydrated cement on durability (as well as longterm deformations) remain a concern. 2.2

SERVICE LIFE MODELING OF REINFORCED CONCRETE STRUCTURES

Other deterioration mechanisms

While deterioration due to reinforcement corrosion remains the most significant global concern in this

68

based on environmental classes. The difference between the deemed to satisfy approach envisaged by the fib Model Code and traditional service life design rules is that the latter are commonly not based on physically and chemically correct models but largely on practical experience, whereas the fib method is intended to be calibrated against the full probabilistic approach. The fourth level of service life design (avoidance of deterioration) is based on the use of non-reactive materials such as stainless steel, or protection systems, such as coatings. In the presentation of the above design options, fib has taken a major step towards designing concrete structures for durability. Fundamental principles and design options for concrete durability have been clearly laid out. For the successful application of the various design options, however, further work is necessary in the following fields:

The prescriptive approach ignores, to a large extent, the different performance of the different cement types and of the mineral components added to the cements or to the concrete itself, as well as the influences of onsite practice during the construction process. Performance concepts, on the other hand, are based on quantitative predictions for durability from exposure conditions and measured material parameters. The resistance of the structure, measured through durability parameters of the actual concrete used, is compared against the environmental load, taking the influences of time into account. On this basis, the probability of damage occurring to the structure during its lifetime is calculated using appropriate deterioration models. Different levels of sophistication may be applied to performance-based design for durability, including the use of durability indexes, the application of analytical deterioration models, and full probabilistic methods. At the moment, various approaches are being developed worldwide, however yet with very limited application in real projects. Of course, the added challenge of assessing such models against long-term structural performance remains as an important future research need. 3.2

− Defining rational limit state criteria, − Testing actual material performance against relevant material deterioration models, − Calibrating service life models against uncertainties in material properties and environmental influences (probabilistic methods), − Identifying suitable test methods for the characterization of relevant concrete properties. For existing test methods, clear guidance needs to be provided on how to interpret test results and on how to incorporate them into service life models, − Identifying efficiency and durability of protection systems, such as coatings.

Current international developments

The recently published fib Model Code for Service Life Design (fib, 2006) proposes a design approach to avoid deterioration caused by environmental action comparable to load design. Based on quantifiable models for the load side (environmental actions) and the resistance side (resistance of the concrete against the considered environmental action), the following design options are presented:

Many international research efforts are and have been underway in resolving the above listed issues. However, most countries struggle with the implementation of durability design guidelines, due to the many questions yet to be answered. In contrast, in South Africa, durability design guidelines have been implemented in the past decade and are nowadays frequently applied in concrete construction. The approach has been developed to a point where it can be used with some confidence; however the development of the approach is ongoing as remaining uncertainties still need to be addressed. Based on the South African durability design approach, some of the most common shortcomings associated with durability design are discussed in the following.

− Option 1: Full probabilistic approach − Option 2: Semi probabilistic approach (partial safety factor design) − Option 3: Deemed to satisfy rules − Option 4: Avoidance of deterioration. The full probabilistic approach, which is intended to be used for exceptional structures only, should be based on probabilistic models that are sufficiently validated to give realistic and representative results of deterioration mechanisms and material resistance. The basis of this approach is formed by appropriate test methods and statistical evaluation models. In the partial safety factor approach, the probabilistic nature of the problem (scatter of material resistance and load) is considered through partial safety factors. It is based on the same models as for the full probabilistic approach and intends to present a practical, yet statistically reliable design tool. The deemed to satisfy approach is comparable to the durability specifications given in most current codes and standards, i.e. specifications based on a selection of certain design values (dimensioning, material and product selection, execution procedures)

3.3

The South African approach for durability design

The philosophy of the South African durability design approach involves the understanding that durability will be improved only when unambiguous measurements of appropriate cover concrete properties can

69

In view of the various options for service life design presented by fib (2006), the South African Durability Index approach principally adopted the strategy of providing deemed to satisfy rules, which limit durability index values and cover depths for given environmental classes and selected binder types. Limiting durability index values are based on service life models, which in turn relate to partial factor design and partly to full probabilistic modelling of relevant parameters. The South African approach therefore aims at combining Options 1, 2, and 3, as presented by fib, to provide a practical tool for durability design and quality control. The durability index approach is currently being applied in a number of large scale construction projects in South Africa. This is considered a large step towards improvement of concrete quality and structural durability. However, much work remains to be done, in particular generating correlations between indexes and actual structural performance. Future research needs in the field of durability index design and service life modelling are discussed in the following.

be made. Such measurements must reflect the in situ properties of concrete, influenced by the dual aspects of material potential and construction quality. Key stages in formulating this approach were developing suitable test methods, characterizing a range of concretes using these tests, studying in-situ concrete performance, and applying the results to practical construction. The approach links durability index parameters, service life prediction models, and performance specifications. Concrete quality is characterized insitu and/or on laboratory specimens by use of durability index tests, covering oxygen permeation, water absorption, and chloride conduction (Alexander et al 2001, Beushausen et al 2003, Alexander and Stanish 2005). The service life models in turn are based on the relevant DI parameter, depending on whether the design accounts for carbonation-induced or chlorideinduced corrosion. Designers and constructors can use the approach to optimize the balance between required concrete quality and cover thickness for a given environment and binder system. A framework for the development and application of performance-based specification methods for concrete durability is illustrated schematically in Figure 2.

MATERIAL INDEXING Characterization of concrete (surface layer) using easily measured physical properties, such as permeability, sorptivity and possibly strength.

QUALITY CONTROL

Correlations DIRECT DURABILITY TESTING

Suite of accelerated tests (lab): Modelling of environment and mechanisms of deterioration

Correlations

Correlations

Long-term tests (lab or site-based). As close to ‘real’ conditions as possible

STRUCTURAL PERFORMANCE Evaluation of structural performance; Consequences of deterioration; Management of economic strategies

FUNDAMENTAL MECHANISTIC STUDIES

Figure 2.

Framework for carrying out durability studies in South Africa.

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PREDICTION

E N V I R O N M E N T

Micro- and macro- environment. Presence of deteriorating agents; measurement of aggressiveness of environment

3.4

Future needs in durability design and service life modelling

Table 1. Environmental Classes adopted in South Africa (Natural environments only) (after EN206).

3.4.1

Calibration of service life models and test methods against field performance of concrete structures The design against chloride ingress, using chloride conductivity indexes, is based on service life models that were developed at the University of Cape Town (Mackechnie, 2001). The relationship between conductivity index values and the potential field performance of reinforced concrete structures was established from 2 different techniques:

Carbonation-Induced Corrosion Designation

Description

XC1 XC2 XC3

Permanently dry or permanently wet Wet, rarely dry Moderate humidity (60–80%) (Ext. concrete sheltered from rain) Cyclic wet and dry

XC4

Corrosion Induced by Chlorides from Seawater

− Correlation between 28-day conductivity index values and chloride ingress in structures in the Western Cape Province. − Laboratory-based experimental correlations between 28-day conductivity index values and chloride diffusion coefficients.

Designation

Description

XS1

Exposed to airborne salt but not in direct contact with seawater Permanently submerged XS2a + exposed to abrasion Tidal, splash and spray zones Buried elements in desert areas exposed to salt spray XS3a + exposed to abrasion

XS2a* XS2b* XS3a*

Further work is necessary to test 28-day conductivity index values against chloride ingress in various marine environments in South Africa, taking into account that chloride ingress is dependent on environmental conditions such as water and air temperature and relative humidity. In a current study at the University of Cape Town, chloride ingress into various different types of concrete is being investigated, based on site exposure in the Cape Town and Durban harbours, which, from a temperature and relative humidity point of view, represent very different environmental conditions. Measurements taken on site-exposed samples are correlated to laboratorybased measurements of chloride conductivity index values and diffusion coefficients, in order to refine existing service life models for application in various regions in South Africa. This project is described in more detail in another paper presented at ICCRRR 2008 (Heiyantuduwa, 2008). The design against carbonation using oxygen permeability indexes is based on empirical relationships between 28-day OPI values and carbonation depth measurements on actual structures and laboratorycured samples (under accelerated carbonating conditions). Carbonation depth measurements, particularly on real structures, show large variations, making statistical evaluations of test results difficult. It is therefore important to collect more data to calibrate the service life models used for the prediction of carbonation. This needs to be done taking various climatic conditions into account and researching fundamental influences on carbonation of concrete.

XS3b*

*These sub clauses have been added for South African coastal conditions.

established whether the current environmental classes are sufficient in describing potential deterioration mechanisms, or whether a more refined approach needs to be developed that classifies environmental exposure based on the prevailing climate conditions. 3.4.3 Development of probabilistic models The natural variability in the concrete material makes it inevitable to use probability theory in formally including the uncertainties in the service life prediction model parameters. A framework for the application of probabilistic models in durability design of reinforced concrete (RC) structures in South African marine conditions has been developed in a research project at the University of Cape Town. This involved incorporating the steady state diffusion coefficient derived theoretically from the chloride conductivity test results in a probabilistic model as discussed in Muigai et al (2008). Statistical information for each parameter in the model was applied in providing improved estimates of the predicted service life. The research study also demonstrated the use of the probabilistic model in specifying limiting values for chloride conductivity based on initiation limit state target probability values given in fib Model Code for Service Life Design (fib, 2006). However, additional data sets still need to be acquired for each of the quantified model parameters to improve on the accuracy of the model. Further research in this field also requires the application of

3.4.2 Definition of environmental classes The environmental classes used in durability design in South Africa are related to the EN 206 classes as modified for South African conditions (Table 1). As discussed in the previous section, it needs to be

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the oxygen permeability test in the carbonation service life prediction model based on the same probabilistic framework.

addressing the above 2 aspects. Aspects relating to the structural capacity of corrosion-damaged reinforced beams are discussed by Malumbela et al (2008).

3.4.4 Definition of limit state criteria A number of service life models for reinforced concrete structures exist. Many of these adopt the twostage service life model first proposed by Tuutti (1982), in which the deterioration is split into two distinct phases, namely the initiation period and the propagation period, as shown in Figure 3. Most service life models assume that the end of the initiation period denotes the end of service life. However, since a concrete structure does not immediately lose its strength or functionality at the onset of the propagation period, a more sensible approach would make use of a detailed maintenance strategy for corrosion-damaged structures. The definition of a suitable maintenance strategy depends on knowledge of the following aspects:

3.4.5

Investigating the effects of cracking on concrete durability Most service life prediction models cover the ingress of harmful substances into uncracked concrete only. As a result, transport mechanisms for chlorides and carbon dioxide are primarily assumed to relate to established diffusion models. However, cracks do frequently occur in concrete structures, especially under the influence of load-induced stresses. Transport mechanisms and corrosion cell development in cracked concrete may be very different, in comparison to uncracked concrete. Further research needs to be done to establish the influence of cracks on the corrosion of steel reinforcement. It appears ineffective to model the ingress of harmful substances into uncracked concrete when existing cracks may accelerate the deterioration process. In particular, the influences of various crack widths on the ingress of harmful substances into concrete needs to be established and considered in the modeling of service life. In another paper presented at ICCRRR 2008, Otieno et al (2008) discuss the influence of crack widths on reinforcement corrosion in more detail.

− influence of steel reinforcement corrosion on the load-bearing capacity of structural members − performance (and durability) of materials and systems for repair and protection of corrosiondamaged structures. Both of the above aspects still need further research to be fully understood in the context of service-life modeling. Two research projects are currently being undertaken at the University of Cape Town,

3.4.6

Remedial measures for structures that do not meet durability specifications The South African Durability Index approach enables engineers to specify certain durability parameters (indexes) in relation to the anticipated service life, environmental conditions, binder types and cover depth requirements. Durability specifications commonly comprise limiting values for the thickness and penetrability of the concrete cover. When limiting values, obtained on the as-built structure, meet the specified requirements, the structure is considered to be inherently durable. However, a clear design methodology for concrete structures that do not meet the specified requirements needs to be established. If limiting durability index values have not been achieved, the owner of the structure principally has the following options:

Level of deterioration Initiation period

4

Propagation period

3 2 1

Time of exposure [years] 1

Depassivation of the reinforcement Initiation period

2

Formation of cracks

3

Spalling of concrete cover

4

Failure of the structure through bond failure or reduction in cross section of the load bearing reinforcement

− Demolish and rebuild the structure − Accept that the anticipated service life duration may not be reached − Protect the structure against the ingress of harmful substances, such as carbon dioxide and chlorides − Accept that harmful substances can reach the reinforcement, but protect the reinforcement against corrosion.

Propagation period

For most projects, the first of the above options will be undesirable, for obvious reasons. The second option

Figure 3. Deterioration process of reinforcement corrosion: 2-phase model for service life (Tuuti 1982, fib 2006).

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experience of the knowledgebase of concrete deterioration and durability research. It is important to acknowledge that the global concrete research community has made remarkable contributions to our understanding of concrete durability and deterioration. In the same breath, this paper has tried to emphasise that there is much understanding that has yet to be developed through research. The particular problem of deterioration due to reinforcement corrosion will continue to demand much research time and effort before we can say that international concerns in this area have largely been allayed. This paper has also made a strong case for a more rigorous approach to service life modelling, durability design and specification. In this regard, the proposals presented in this paper are intended as a basis for the development of a framework towards addressing this concern. Finally, the important challenge of technology transfer should not be neglected. Our success in these areas of research endeavour will be measured by the extent to which we positively influence the practice of concrete technology to produce more durable concrete structures.

involves a re-evaluation of the original design parameters and may in many cases also not be acceptable. Probably in most cases, the third or fourth option will be aimed at, i.e. protecting the structure against deterioration to ensure that the design service life can be reached. Such methodology may for example include the application of protective surface coatings or corrosion inhibitors. Depending on the quality of the structure (by how much did it not reach the limiting durability design parameters?), a once-off application may be sufficient, whereas in other cases a detailed maintenance plan may need to be established, taking repeated application of protective measures into account. The decision of appropriate repair and maintenance strategies needs to be based on an evaluation of the expected service life. For this, the measured durability index value needs to be used as an input parameter in the service life model, with which the original design parameter was established. This will allow an estimation of the actual service life duration that can be expected. This, in turn, will give the information of how many years of additional service life the protective measure needs to provide. Based on this, it can for example be argued that a coating, which prevents the ingress of harmful substances over that required duration, presents a suitable protective measure, bringing the structure back to its original service life. However, a clear philosophy needs to be developed, based on which the design of appropriate protective measures can be carried out. The design engineer and the owner of the structure need to be given clear guidance on what steps to follow and on what options are available. From a technical point of view, it needs to be established, which coatings can be used to either prevent or slow down the ingress of chlorides or carbon dioxide sufficiently. The performance of protective coatings commercially available can commonly be shown to be promising in short-term tests. However, there is a lack of data available on the durability of such coatings. Future research is needed to fill this gap of knowledge. A current project at the University of Cape Town is dealing with these issues. Another promising protection method for reinforced concrete structures is the application of corrosion inhibitors. However, also for these materials the long-term efficiency still needs to be established.

4

REFERENCES Alexander, M.G., Mackechnie, J.R. and Ballim, Y. Guide to the use of durability indexes for achieving durability in concrete structures, Research Monogram No. 2, Department of Civil Engineering, University of Cape Town, March 1999, 35 pp. Alexander, M.G. and Stanish, K. (2005), ‘Durability design and specification of reinforced concrete structures using a multi-factor approach’. Mindess Symposium, Third Int. Conference on Construction Materials, Vancouver, August 2005, CD ROM, University of British Columbia, Vancouver, 2005, 10 pp. Beushausen, H., Alexander, M.G. and Mackechnie, J. Concrete durability aspects in an international context, Concrete Plant and Precast Technology BFT, vol. 7, 2003, Germany, pp. 22–32. European Committee for Standardization (2004), BS EN 1992-1-1 Eurocode 2: Design of Concrete Structures— Part 1-1: General rules and rules for buildings (European standard prEN1992-1-1). CEN, Brussels, Belgium. fib bulletin 34: Model Code for Service Life Design, Switzerland, 2006, 110 pp. Heiyantuduwa, R. and Alexander, M.G., Studies on prediction models for concrete durability, International Conference on Concrete Repair, Rehabilitation and Retrofitting, Proceedings ICCRRR 2008, Cape Town, 24–26 November 2008. Mackechnie, J.R. Predictions of Reinforced Concrete Durability in the Marine Environment, Research Monograph No. 1, Department of Civil Engineering, University of Cape Town, 2001, 28 pp. Malumbela, G., Alexander, M.G. and Moyo, P. Structural behaviour of beams under simultaneous load and steel

CLOSURE AND OUTLOOK

In this paper, we have tried to undertake a particularly difficult task: to define what future users of concrete will need from the research that we are currently undertaking. Nevertheless, we feel that the issues we have raised are practical and draw directly from our

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corrosion, International Conference on Concrete Repair, Rehabilitation and Retrofitting, Proceedings ICCRRR 2008, Cape Town, 24–26 November 2008. Neville, A. Concrete: Neville’s Insights and Issues, Thomas Telford, London, 2006. 314 pp. Otieno, M., Alexander, M.G. and Beushausen, H. Corrosion Propagation in Cracked and Uncracked Concrete, International Conference on Concrete Repair, Rehabilitation

and Retrofitting, Proceedings ICCRRR 2008, Cape Town, 24–26 November 2008. SANS 10100-2 (2005), Structural use of concrete Part 2— materials and execution work, 3rd Edition. Tuutti, K. Corrosion of steel in concrete, Stockholm: Swedish Cement and Concrete Research Institute. In: CBI Research, Report No. 4:82, 1982.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Reinforcement corrosion: Research needs C. Andrade Institute Eduardo Torroja of Construction Science, Spain

ABSTRACT: Corrosion of reinforcements has extensively being studied during the last 3 decades in spite of which many questions remain unsolved. Steel corrosion embedded in concrete can be suppressed or slown down but at a cost that still has not been correctly accounted from the design phase. The paper reviews some of the subjects, although not all related to the fundamentals of corrosion, the management of the service life and the repair techniques. The task to summarize all important aspects is too wide and much further basic work is needed in the area.

1

INTRODUCTION

initiated by a symposium in 1969, set up the TC-60 chaired by P. Schiessl (Schiessl, 1988). From then to now, there have been 3 decades of exponential production of papers, organization of congresses and publications of specialized committees. To summarize in few papers the state of the art after these decades of high production is a difficult task. In present paper only some highlighting of the subject will be approached, making emphasis in the parts related to assessment and repair. The main aspects related to corrosion of reinforcement that will be mentioned are:

Reinforced concrete as building material represents development and opportunities for infrastructures. After about one century of its industrial use, the material has demonstrated to be versatile, able of multiple applications and has evolved towards a much higher range of mechanical strengths and types of reinforcing as the very high performance concretes are enabling. However, not all are advantages, the material is sensitive to its process of fabrication and curing and properties as creep or shrinkage are still subjects needing long term results. Being a material which aspires to offer a long service life without maintenance, corrosion of reinforcements is still, after 3–4 decades of intense research around the matter, a problem that subsists in its basic consequences. In aggressive environments as are marine ambient, to reach 100 years of time life without damages, seems to be feasible only if special preventive measures are adopted from the design. Thus, it has to be recognized that the material can offer long life in many environments but for very aggressive ones, special measures have to be prescribed: concretes of very high mechanical strengths above the needed from resistance demands in order to lower the porosity and with it the chloride ingress, or reinforcements of stainless steel or monitored cathodic protection. To try to summarize in few pages a state of the art on reinforcement corrosion is not an easy task or perhaps it is a too ambitious attempt. Let us then underline some few aspects that, in authors perspective are of interest at present for the sake of repair and rehabilitation. The intensive study of corrosion of reinforcement started in the decade of 70’s (Alexeyev, 1969; Page, 1982; Verbeck, 1975; Gouda, 1965; Hausmann, 1964) when the corrosive effect of chlorides was noticed more evidently. Only at the initiation of the 80’s, specialized workshops started to be organized (Arup, 1993). RILEM, following the work on durability

1. 2. 3. 4. 5. 6. 7. 8.

Fundamental mechanisms and corrosion initiation Corrosion propagation Modeling of service life Corrosion measurement techniques Prevention methods Repair techniques Management of technical life and repair techniques Structural performance.

2

FUNDAMENTAL MECHANISMS

Although all publications repeat that the steel embedded in concrete is covered by a film of oxides that passives the bar providing the surrounding remain alkaline or free of Cl–, it has to be recognized that the studies on the nature of the passive layer are very scarse (MaCDougall, 1995; Benzekri, 1990; Haupt, 1987) and many questions remain unexplored on such a fundamental property. Similar situation of lack of basic studies exists on two basic mechanisms of depassivation and the initial states of the corrosion. Figure 1 shows for instance, AFM views (Sánchez, 2007) on the initial attack of a prestressing steel by chlorides which make evident that only ferrite dissolves remaining unaltered the cementite which is interpreted to act as cathode.

75

Pol Potential (mV, SCE)

300 low C3A

200

low C3A+35%FA(1)

100

high C3A(1)

0

high C3A+O1+35%FA(1)

-100

Aluminous

-200

filler

-300 -400 -500 -600 0

0.5

1

1.5 2 % Free CL-

2.5

3

3.5

Figure 2. Dependence of the amount of water soluble chloride content which induce corrosion with the potential applied.

Figure 1. Three-dimensional AFM images of the sample surface after a) 0 h 0 min, b) 0 h 30 min c) 0 h 50 min. and d) 2 h 15 min of exposure to a 0.05 M NaCl solution.

The aspects like how the corrosion nucleates and progresses at the initial stages and how the pit progress in extension and depth need much more attention not devoted until now in spite of the huge amount of literature published. Related to the initiation of corrosion, the functioning of the electrochemical cell is too often mistaken assuming that the corroding zone is always a pure anode, when the corrosion progress by microcells and not only by macrocells. This fundamental error of considering the whole corroding zone as anodic has many practical misleading consequences, one of which is the fact that as this zone acidifies due to the generation of protons during the corrosion reaction (Galvele, 1976) the progress of the pits may not need oxygen as the protons provide the cathodic reagent. The reduction of the ferric oxide in the redox reaction ferrous-ferric provides the reactant needed to maintain the corrosion active. With respect to the chloride threshold that, on the opposite has attracted lot of authors, has generated more heat that light and still is valid the consideration of a limit of 0.4% of Cl− based in a simple calibration of the Cl− bound with the aluminate phases of an OPC (Treadaway, 1979). It seems now clear that the threshold: 1) is not unique and depends on many variables (Bamforth, 1994: Whiting, 1981; Buenfeld, 1995), 2) it depends on the steel electrical potential (figure 2) (Izquierdo, 2004) that is not constant but evolves with time and depends on the concrete characteristics. The observation is the nucleation of pits in the voids or pores at the steel-concrete interface does not makes a special difference (Glass, 2000) because there the local potential is likely different than in the zones of a perfect contact of the solid phase with the steel.

Figure 3. Brittle surface of fracture of Cold drawn steel. Load obtained in the type c specimen at –275 mVAg/AgCl potential.

That is at microscopical level a void means a different resistivity and free chloride concentration than in the gel pores and therefore a different potential. A final comment should be devoted to the fundamentals of the SCC mechanisms in the prestressing steels that is also very scarcely studied (Nurnbeger, 2005; Galvele, 1987; Sánchez, 2007). In spite of the recent functioning of a European COST ACTION534 specially devoted to prestressed concrete, very few are the papers on the basic mechanisms. This important phenomenon can lead into brittle failure and then should be studied from a multidisciplinar approach, merging electrochemical knowledge to the fracture mechanic fundamentals (figure 3).

3

CORROSION PROPAGATION

Numerous studies have been made on the observation of the level of corrosion and its consequences, but the majority are related to parametric or comparative

76

Corrosion rate (Icorr-µA/cm²)

10 1

Explotation and Service Life

0.1

Owner

Demolition Design

Construction Maintenance

Repair

PostRepair

0.01 Initial Safety

0.001 Jan-99

Apr-99

Jun-99 Sep-99

Dec-99

Mar-00

Jun-00 Sep-00

Dec-00

Mar-01

Minimum Safety

Figure 4. Evolution of corrosion rate with the atmospheric climatic events.

Figure 5. Steps in the life of reinforced concrete structures.

studies on the influence of the different local materials to fabricate concrete. Comparatively very few study the mechanisms of progression of the corrosion. Also it has to be mentioned how often the corrosion rate, that is a differential or instantaneous value, is mistaken with the accumulated corrosion (Andrade, 1978) that appears an integral value. Another remark to make is the fact that most of the studies are made in laboratory conditions when most of the structures are exposed to the climatic actions. This fact makes a great difference, as the natural events happen at changing temperature and the structures are submitted to rain, snow and wind. In these conditions all the corrosion parameters evolve with time as figure 4 shows and then it is necessary to define an annually averaged or representative corrosion rate, Icorr,REP , which could facilitate the prediction of the corrosion progression (Andrade 1999), in order to calculate the remaining time with a certain limit state.

4

Better calibration has been achieved in the case of the modeling of the propagation period because, on-site techniques have been available (Rodríguez, 1994) and calibration in real conditions has been made (Figure 4). Thus, providing the corrosion rate is known, annually averaged of the progression of corrosion is lineal with respect to time Px = Icorr t, where Px is the accumulated penetration of corrosion, Icoor is the instantaneous corrosion rate and t is the time. Another remark to be have made is the good relation found between concrete resistivity and corrosion rate due to the resistivity is an indication of the water content of the concrete (Millard, 1992; Andrade, 2002). The service life has been established as a limit state that can be handling from a technical economical and points of view. This, from a technical perspective, to reach a certain unacceptable corrosion degree is a limit state, either of service, SLS, or ultimate, ULS, depending upon the consequences (figure 6). This limit state can be treated in a deterministic or in a probabilistic manner (Siemes, 1985; DuraCrete, 1998; Frangopol, 1997) which has open the door to the proposal to introduce the calculation of durability in the codes and standards. As table 1 show, four levels of calculation are being developed: level 1 is the use of deemed-tosatisfy rules that is the traditional prescription of concrete mix proportioning. Level 2 is based in the use of “durability indicators” and performance tests that are related to durability but without the explicit use of the time. That is, durability indicators only implicitly are related to the time to corrosion. Level 3 consists in the calculation of service life by means of a model that explicitly considers the time to the corrosion onset. Finally level 4 makes use of reliability analysis. The methods for calculating service life of reinforcements At this respect, in spite that the fib and the JCSS (Fib, 2007) have published probabilistic model codes or procedures applied to reinforcement corrosion, it has to be emphasized, that the level of optimum safety is needed to be established by comparison to the consequences and the economical costs. That is, the reliability index β has to be optimized by comparison of the different technical solutions with their

MODELLING OF SERVICE LIFE

Figure 5 shows with an arrow the different steps in the life of a concrete structure. Service life management is one of the subjects studied in which the progress made is really remarkable from the publication of Tuutti’s model defining the initiation and propagation periods (Tuutti, 1982). The first aspect largely studied after Tuutti proposal is the modeling of the initiation period, that is that related to the carbonation and the chloride ingress rates. The need to approach it of computational tools has attracted many engineers to propose models (Andrade, 1978; Andrade, 1999; Tuutti, 1982; Martín-Pérez, 2000; Petre-Lazar, 2003; Guliker, 2004; Mangat, 1994; Bamfort, 1993) that however lack of proper calibration as they are relatively new. Based in the theory of diffusion these models until now, neither can appraise all the expected life of more than 50 years nor can model the local environment with the climatic changes. Thus, the prediction differs significantly among them (Andrade, 2006). This lack of calibration should however not stop the research in the area.

77

t5

Damage level

the basic phenomena (Andrade 1978; Sagües, 1991; Sagües, 1992). The application of these techniques to the measurement in large real structures was made at the beginning of the 90’s (Feliú, 1990; Feliú, 1996; Vennesland, 1997) after solving the calculation of the steel area polarized by the current applied from a small counter electrode or “critical length”. This length is not a constant and therefore requires an intelligent mode of confining the cement applied to a predetermined area. This can be achieved by the so called “modulated confinement” by means of a second annual counter that surrounds the principal one and maintains the current confined between both: central disc and grand ring. An alternative to the modulated confinement is the direct measurement of the critical length by means of placing influence electrodes in the surface of the concrete at precise distances and follow the alternation of the potential with the distance. An important development on the techniques is the use of embedded sensors (Andrade, 2006) which present the advantage of the possibility of monitoring with the undetected sensors, only few measure the Rp being more common the recording of galvanic currents between two sensors. At this respects it has to emphasize that the galvanic current is always only a part of the corrosion current (see figure 1) and cannot be used for prediction of the residual life but simply to detect despassivation. A final comment is the need to development of wireless communication because this avoids the cabling although the technology is still far to be developed because of the requirement of long life (more than 50 year) and the high current demand, which cannot still be provided by the present commercial wireless technology. The monitoring of data in real structures reports important information but, on the other hand, demands the management of a huge amount of data and their processing and incorporation into engineering models. The filtering of erroneous data and the interpretation of them still will need development.

Loss of structural safety

t4 t3

Section loss

Cracking

t1

t2 Corrosion Despasivation

initiation

propagation

time

Figure 6. Limit states of corrosion: depassivation, cracking, loss in cross section or bond and the structural consequences of reinforcement corrosion.

Table 1.

Levels for design of durability.

Design level

Calculation method for durability

I II III IV

Concrete mix Performance tests Models Probabilistic treatment

associated costs in order to establish the optimum safety level (Rackwitz, 2000). Until present, there is a lack of real data which could help to advance more the state of the art in this subject. It is necessary to collect data in real structures. A life cycle analysis or safety calculations should be more than an interesting computational exercise.

5

6

PROTECTION METHODS

The most efficient methods to extend the corrosion onset of the reinforcement are the cathodic protection (Pedeferri, 1989) or the use of stainless steel, galvanized (Yeomans, 1994) or composite rebars (Mailvaganam, 1992). That is, to change normal steel by a corrosion resistant material or to apply a current to avoid anodic dissolution of the iron. Some inhibitors added (Berke, 1989) in the mix are also efficient in extending the life while most of epoxy coatings (De Rincón, 1996) have shown to faire in less than 20 years. All these methods are efficient but render more

CORROSION MEASUREMENT TECHNIQUES

Proper techniques were used from very early studies (Gouda, 1965; ASTM C876-91) in the laboratory and so, corrosion potential and corrosion rate (through the Polarization Resistance technique) measurements have enabled to understand many of

78

expensive the concrete. They should be analyzed in a life cycle cost approach.

e. Execution of the repair work. f. Monitoring of the efficiency of the repair.

7

Figure 7 shows the main steps of the repair proces with two examples of optimization methods for the best selection.

REPAIR TECHNIQUES

The most common technique of repairing is patching and cosmetic repair of the concrete. However the most efficient is cathodic protection. Other two techniques have advanced very remarkably in the last decade: the electrochemical realkalization and the chloride extraction. Both suppose the need of the application of high electrical current and they demand particular conditions to efficiently stop corrosion and restore passivity. In the case of realkalization, it is necessary to notice the soaking of the external carbonate solution into the concrete to be sure the technique will maintain the alkalinity (Castellote, 2006). The restoration of geometrical and strength values above the threshold levels on the design requirements(Izquierdo, 1998; REHABCON IPS2000-00063), is an essential contribution to the updating of the economical value of the structure. The selection of the best repair option consists meanly in:

8

The building of concrete structures represents an investment for the owners. The value of the structures should be maintained during their service life in order to avoid undesirable losses of the initial investment. They need then to try to update the investment from an economical point of view and to maintain the safety level in that defined during the design phase. Concrete structures are designed and built complying with the design requirements of safety, serviceability and aesthetic regulated in Eurocodes. Apart from that of mechanical stability and resistance to fire covered by the Eurocodes, concrete structures must comply with the other essential requirements. They also need a maintenance policy in order to comply with the essential requirements during the period of their service life. Maintenance policy of concrete structures is an essential part of their management strategy in order to maintain the values of the initial investment. The management is at present made by means of empirical methods and using a performance approach not well developed. The main aspects of a Management System (Castellote, 2006,) are: Inventory of the structures, periodical inspections, analysis of collected data and proposal of interventions and investments (REHABCON IPS-2000-00063). The requirements considered are of technical and non-technical nature. They can be grouped in four main headings:

a. Identification of available repair methods suiting in the needed restoration of initial condition of the structure. b. Benchmarking or comparison of best repair methods. c. Selection of the optimum repair method. d. Definition of specifications of the repair work in order the repair structure fulfils again the design requirements by executing the selected repair method.

PRINCIPAL ASSESSMENT

REPAIR SELECT

EXECUT REPAIR

MANAGEMENT OF SERVICE LIFE

• • • • •

MONITOR REPAIR

Economical and financial Social and cultural Environmental and Functional Social and political. General goals for buildings and structures are:

EN1504

Figure 7.

• To have a reasonable low risk of accidents and fire. • They could withstand reasonable loads associated with the normal use and exceptional climatic events. • They are made by a design which provide reasonable means of access and egress. • They have adequate ventilation and sanitation facilities to maintain the health and comfort of occupants.

PBA

CHOOSE OPTION

IDENTIFY INDICATORS

CHOOSE PRINCIPLE

DEFINE THRESHOLD VALUES

CHOOSE METHOD

TESTING VERIFICATIONS

A final step in the Management is the monitoring of the performance. Feedback from monitoring gives input for need of corrective actions (repair,

Main steps in the repair process.

79

STRUCTURE LIFE

DESIGN CONSTRUCT.

The scientific challenges identified to approach, at least by simplified models (Rodriguez, 1995; Molina, 1993), in order to integrate the durability with the criteria of structural safety are very numerous, because they imply the establishment of parameters of materials, as new or deteriorated, or when they are associated to a repair material,. In addition, a data base should be built for the probabilistic treatment. Some of the challenges are:

POST-REPAIR LIFE

SERVICE LIFE

PT EI

PERFORMANCE TARGETS, PT

PERFORMANCE THRESHOLD

ECONOMICAL INVEST, EI

ACTORS IN THE PROCESS

MANAGEMENT OF STRUCTURES PROCESS

REQUIREMENTS

CLIENT

CLIENT

OWNER

OWNER

ASSET INVENTORY

ROUTINE INSPECTION

CONSULTANT

MATERIAL SUPPLIER

PRINCIPAL ASSESSMENT

REPAIR SELECT

EXECUT REPAIR

CLIENT CONSULTANT OWNER

MONITOR REPAIR

PRINCIPAL ASSET

ESSENTIAL •ECONOMICAL AND FINANCIAL •SOCIAL AND CULTURAL •ENVIRONMENTAL •FUNCTIONAL EN1504

METHODS TO SELECT BEST REPAIR OPTION

REPAIR SPECIALIST CONSTRUC

• Characterization of environmental actions that induce damage. Probabilistic treatment. • Calibration of existing mathematical models with the behavior in real conditions and development of models that integrate real the environmental aggressiveness with the predictions. • Simplified mathematical models of physicochemical actions other than those inducing metallic corrosion. • Establishment of material parameters for their integration in the numerical modeling: stress-strain curves which indicate corrosion and cracking for instance. • Behavior of repaired structures: changes in the performance of the structural element.

PBA

CHOOSE OPTION

IDENTIFY INDICATORS

CHOOSE PRINCIPLE

DEFINE THRESHOLD VALUES

CHOOSE METHOD

TESTING VERIFICATIONS

Figure 8. General process of service life and repair/post repair components of the management of concrete structures. CONREP-NET(54).

rehabilitation, remedial) or preventive actions. A process map for making a Performance Based Intervention (PBI) is presented in Figure 8 taken from Conrepnet (Conrepnet, 2007). 9

ACKNOWLEDGEMENTS

ON THE INTEGRATION OF MECHANICAL AND ENVIRONMENTAL ACTIONS IN THE LIFE CYCLE OF THE STRUCTURE

The author is grateful to all colleagues that have dis cussed and comment these many aspects contributing to the solution of a distress so economically important.

A final trend scientific challenge is the necessity to design the durability with the same principles undertaken by the mechanical behavior (Reinhardt, 1985; Mazars, 1989). That is to say, it is necessary to design the structures with a service life technically guaranteed and to evaluate existing structures being able to predict their residual life (Rodriguez, 1995). This capacity of calculation has to take into account the atmosphere where the structure will be erected for which is necessary to characterize and to rank its aggressivity in form of “environmental actions”. The methodology is parallel to which it is used in the traditional calculation. This means that the methodology that is based on the definition of durability Limit States, with regard to aesthetic, economic criteria and of the state of knowledge. The figure shows to diverse limit states related to the corrosion of the reinforcement. It is necessary the collection of data and its probabilistic treatment in order to apply these concepts in a rigorous manner to the structures in the risk or being deteriorated. Moreover, and given to the enormous effort and cost that the experimentation of deteriorated structures entails, it is vital the characterization of the parameters that enable structural numerical modeling either of the environmental actions as of the performance of the structural and material deterioration.

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Andrade, C. and Gónzalez, J.A., “Quantitative measurements of corrosion rate of reinforcing steels embedded in concrete using polarization resistance measurements”, Werkst. Korros., 29, 515 (1978). Arup, H. Sorensen, B. Frederiksen, J. and Thaulow, N. “The Rapid Chloride Permeation Test- An Assessment”, CORROSION/93, paper nº 334, March 7–12, New Orleans, Louisiana (1993). ASTM C876-91. “Standard Test Method for Half Cell Potentials of Uncoated Reinforcing Steel n Concrete”. Bamforth, P. and Chapman-Andrews, J. “Long term performance of RC elements under U.K. coastal exposure conditions”, International Conference on Corrosion and Corrosion Protection of Steel in Concrete, Sheffield Academic Press, Ed. N. Swamy Ed. (1994) p. 139–156. Bamforth, P.B. and Price, W.F. Concrete 2000, Economic and durable construction through excellence, (E&F Spon, Dundee), Vol 2, September (1993), 1105–1118. Benzerkri N., Keddam M. and Takenouti H. “a.c. Response of RRDE during the Passivation of Iron”. Corros. Sci. 31, 1 (1990). Berke, N.S. and Rosenberg, A., “Technical Review of Calcium Nitrite Corrosion lnhibitor in Concrete”, Transportation Research Record 1211, Transportation Research Board, National Research Council, Washington, D.C. (1989) 18–27. Buenfeld, N.R., Shurafa-Daoudi M.-T. and McLoughlin I.M. (1995): Chloride transport due to wick action in concrete. Paper presented at the RILEM International Workshop on Chloride Penetration into Concrete, October 15–18, Saint-Rémy-les-Chevreuse, France. Castellote M., Llorente I., Andrade C. and Turrillas X. Alonso C., Campo J. In situ monitoring the realkalization process by neutron diffrection. electrosmotic flux and portlandite formation. Cement and Concrete Research 36 (2006)791–800. Conrrepnet. Thematic Network on performance based remediation of reinforced concrete structures. BRE press. London 2007. Dura Crete (1998) Probabilistic Performance based durability design of concrete structures Brite EuRam Project 95–1347. Feliú S., Gonzalez J.A. and Andrade C. “Multiple-electrode method for estimatinfg the polarization resistance in large structures”. Journal of applied electrochemistry 26. pp. 305–309. (1996). Feliú, S. , González, J.A., Feliú, S. Jr., and Andrade, C., “Confinement of the electrical signal or in-situ measurement of Polarization Resistance in Reinforced concrete,” ACI Mater. J., 87, pp 457. (1990). Fib. Model code for service life design. Task group 5.6–2007. Frangopol, D.M. Application of life cycle reliability-based criteria to bridge assessment and design. Safety of Bridges, Institution of Civil Engineers, Highway Agency, Thomas Thelford Ed. (1997). Galvele, J.R. “Transport processes and the mechanism of pitting of Metals”, J. Electrochem. Soc., 123 (4), (1976) 464–474. Galvele, R.S. “A stress corrosion cracking mechanism based on surface mobility”, Corrosion Science, vol. 27 no. 1, (1987) 1–33. Glass G.K., Reddy B. and Blenfeld N.R. The participation of bound chloride in passive film breakdown on steel in concrete. Corrosion Science 42(2000).

Gouda, V.K. and Monfore, G.E. “A rapid method for studying corrosion inhibiton of steel in concrete”, Journal PCA, Septiembre, no. 3, (1965) 24. Gouda, V.K. and Monfore, G.E., “A rapid method for studying corrosion inhibiton of steel in concrete”, Journal PCA, Septiembre, no. 3, (1965) 24. Gulikers, J. (2004), Critical evaluation of service life models applied on an existing marine concrete structure. NORECON Seminar 2004: Repair and Maintenance of Concrete Structures, Copenhagen April 19–20, 2004. H. Kaerche-Zcment-Kalk-Gips 12(1959) No.7, 911. Haupt S. and Strehblow H-H.. “Corrosion, Layer Formation, and Oxide Reduction of Passive Iron in Alkaline Solut.: A Combined Electrochemical and Surface Analytical Study”. Langmuir. 3, 6, p. 873 (1987). Hausmann, D.A. “Electrochemical behaviour of steel in concrete”, Journal A.C.I. Febrero, (1964) 171. Izquierdo D., Alonso C., Andrade C. and Castellote M. Potentiostatic determination of chloride threshold values for rebar depassivation. Experimental and statistical study. Electrochimica Acta 49 (2004) 2731–2739. Izquierdo, D. and Andrade, C. (2002) Strategy for Maintenance and rehabilitation of concrete structures Innovation REHABCON Lewis, R.W. Schrefler, B.A., The Finite Element Method in the Static and Dynamic Deformation and Consolidation of Porous Media, 2nd ed., Wiley & Sons, Chichester, 1998. MaCDougall B. and Graham M.J. “Growth and Stability of Passive films”. En “Corrosion Mechanisms in Theory and Practice”, P. Marcus, J. Oudar Eds. Marcel Dekker, Inc. pp. 143, (1995). Mailvaganam N.P. “Repair and Protection of Concrete Structuresl”. CRC Press, Boca Raton, USA (1992). Mangat, P.S. and Molloy, B.T. (1994): Predicting of long term chloride concentration in concrete. Materials and Structures, Vol. 27, pp. 338–346. Martín-Pérez, B, Zibara, H., Hooton R.D. and Thomas, M.D. A,” A study of the effect of chloride binding on service life predictions, Cement and Concrete Research, 30 (2000) 1215–1223. Mazars, J. and Pijaudier-Cabot, J. Continuum damage theory—application to concrete, J. of Eng. Mechanics ASCE 115 (2) (1989) 345–365. Millard, S.G. and Gowers, K.R. “Resistivity assessment of in-situ concrete: the influence of conductive and resistive surface layers”, Proc. Inst. Civil Engrs. Struct. & Bldgs, 94, paper 9876, pp.389–396. (1992). Molina, F.J., Andrade, C. and Alonso, C. “Cover cracking as a function of bar corrosion: Part II—Numerical model”, Materials and Structures, 26, (1993), pp. 532–548. Nurnbeger U. Stainless steel in concrete structures. In H. Böhni. Corrosion in reinforced concrete structureswoodhead publishing Ltd. CCondridge 2005, 135–162. Page C.L. and Treadaway K.W.J. Aspects of the electrochemistry of steel in concrete. Narute 297 (1982) No. 5862, 109–115. Pedeferri, P. “Corrosione e protezione delle strutture metalliche e in cemento armato negli ambienti naturali”, ClupMilano. Piazza Leonardo da Vinci, 32, Milano; 1989. Petre-Lazar, I., L. Abdou, C. Franco, and I. Sadri (2003) THI—A Physical Model for Estimating the Coupled Transport of Heat, Moisture and Chloride Ions in Concrete. 2nd International RILEM Workshop on Life Prediction and Aging Management of Concrete Structures, Paris, 5–6 May 2003, France.

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Rackwitz, R. Optimization, the basis of code, ranking and reliability simplification, Structural safety, vol. 22, no.1, 2000, 27–60. Rehabcon (2000) Strategy for Maintenance and rehabilitation of concrete structures Innovation Project IPS 2000-00063. REHABCON IPS-2000-00063, “Strategy for Maintenance and Rehabilitation in Concrete Structures”, DG Enterprise. Reinhardt, H. and Cornelissen, H.K.W. “Dauerstand-zugfestigkeit von Beton baustoff 85”, Bauverlag, Wiesbaden, 1985. Rincón O., Reincón A., Morón O., Villasmil M. and Fernández. R. “Use of Epoxy Portland Cement Coatings to Repair Reinforced Concrete Structures”. Mater. Perform., 35, 10, p. 55 (1996). Rodriguez, J., Ortega, L.M. and Casal, J. “Load carrying capacity of concrete structures with corroded reinforcement”, International conference on structural faults & repairs, London, UK (Ed. M.C. Forde), Vol. 2, pp. 189– 198, 1995. Rodriguez, J., Ortega, L.M. García, A.M., Johansson, L. and Petterson, K. “On-site corrosion rate measurements in concrete structures using a device developed under the Eureka Project EU-401-Int. Conference on Concrete Across Borders, Odense, Denmark, vol.I, pp.215–226. Sagües A.A. and Kranc S.C. “Low-Frequency Electrochemical Impedance for Measuring Corrosion of Epoxi-Coated Reinforcing Steel in concrete”. Corrosion (NACE), 47, 11, p. 852 (1991). Sagües, A.A. and Kranc, S.C. “Computational modeling of the electrochemical impedance response of macroscopic reinforcing steel in concrete systems”, 15th

Panoamerican Congress on Corrosion (Mar del Plata), Argentina, Oct. (1992). Sánchez J., Fullea J. and Andrade C. Stress corrosion cracking mechanism of prestressing steels in bicarbonate solutions. Corrosion Science 49 (2007) 4069–4080. Sánchez J., Fullea J., Andrade C., Gaitero J. and Porro A. AFM Study of the corrosion of a high strength steel in a diluted sodium chloride solution. SPM’07 Workshop. San Sebastian-Spain. Siemes T., Vrouwenvelder T. and Van del Beukel T. (1985) “Durability of buildings: a reliability analysis HERON Vol. 30, no. 3. Treadaway K.W.J. Durability of steel in concrete-SCILondon-Corrosion of steel reinforcement in concrte construction (1979) 1–14. Tuutti, K. “Corrosion of steel in concrete”, CBI Research Report 4: 82, Swedish Cement and Concrete Research Institute, Stockholm, (1982). Vennesland, ∅. 'Electrochemical parameters of repaired and non repaired concrete at Gimsoystaumen Bridge— International Conference on Repair of Concrete Structures from theory to practice in a marine environment, A. Blankvoll ed. Svolvaer Norway May (1997) pp 253–262. Verbeck G.J. Mechanism of corrosion. Corrosion of metal in concrte. ACI-Publication SP-49-Detroit (1975) 21–38. Whiting, D., “Rapid determination of the chloride permeability of concrete”, Report no. FHWA-RD-81-119, NTIS DB no. 82140724, Federal Highway Administration, Washington, DC, (1981) 174. Yeomans S.R. Performance of black, galvanized and epoxicoated. Corrosion Engineering, January (1994) 72–81.

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Theme 1: Concrete durability aspects Innovative materials and influences of material composition

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

High performance concrete for extreme applications B. Hillemeier & R. Wens Technische Universität, Berlin, Germany

L. Hänisch Deutsches Elektronen Synchrotron, Hamburg, Germany

ABSTRACT: High Performance Concrete (HPC) does not only mean high strength! A large field of applications for HPCs is the development of materials with specific properties, for example high durability in extreme environments. Another application field for HPCs is in steel fiber concrete. Between 1 and 10 V-% steel fibers are used to create constructions for high-tech applications that are stable, crack free and with even surfaces, or crack free constructions for waterproofing horizontal and sloped areas. One example shown for using steel fiber reinforced concrete is a 7.000 m³ concrete slab as foundation and floor for a new experimental hall in the DESY in Hamburg—the longest slab in the world, made by concrete without joints.

1

INTRODUCTION

The German electron synchrotron DESY in the Helmholtz community in Hamburg, Germany, is one of the leading accelerator centres in the world. The accelerator Petra is being improved to be the most brilliant worldwide storage ring-based X-ray radiation sources worldwide. This new light source of the superlative offers excellent conditions for top research with particularly intensive and sharp joint X-ray radiation. The decisive advantage is the so capillary X-ray radiation of an especially high brilliance: Also, tiny tested materials can be examined and their pictures highly resolved in the order of their atoms. Almost 300 m of the 2.3 km long Petra ring must be completely modified and a new experimentation hall set up for the retrofitting. The basis of the hall is a 1 m thick concrete slab which carries the accelerator and the experiments. The slab is 24 m wide and 280 m long. It follows the circular arc of the accelerator ring. The concrete slab protects the highly precise measurement equipment from mechanical vibrations. The slab is decoupled from the building. It had to be built without joints and cracks. Therefore the slab had to be concreted without interruption. The evenness of the final floor slab was better than 4 mm/10 m. At a concentrated load of 1 kN the floor slab may deform vertically only by 1 µm. The unusually high requirements call for correct planning, faultless production, highly sophisticated quality assurance, and an experienced contractor.

2

Figure 1. Aerial photo DESY with Petra ring. In the foreground the planned new experiment hall. (Photo: DESY)

hindrance despite the bituminous sliding layer must be taken into consideration. If the slab is shortening by cooling off after its maximum temperature gained by hydration heat, it has to suffer a maximum tensile stress in its central cross section through each of the approximately 150 m long ends. If these tensile stresses remain below the tensile strength of the concrete, reinforcement could be foregone. The shortening of the slab starts three days after the beginning of concreting. For this circumstance a concrete mix had to be designed which produced a hydration temperature as low as possible while reaching a tensile strength of at least 2 N/mm2. The mixture had to be proven in internal performance tests by the construction firm. Because all conditions could not be known for this special construction task one had to take care that a crack does not lead to a gaping joint in the concrete slab. Because of this the planned tensile strength of the young concrete had to be achieved for certain. The achievable tensile strength can only be guaranteed

PHILOSOPHY OF THE ENGINEERING DESIGN

The dimensioning of the floor slab was planned with the following considerations: A minimal deformation

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when steel fibres are mixed into the concrete. Steel fibres do not increase the tensile strength of the concrete, however, they guarantee it. This guarantee is reached only by a fibre content of approximately one volume per cent. For this reason we suggested 75 kg of steel fibres per cubic metre of concrete. Micro cracks arise during the hydration process in the very young concrete. Late cracks arise by shrinkage at a higher age, by restraint and other internal tensions. Micro cracks caused by autogenous shrinkage or by tensions through hydration heat act later as crack starters if loads are effecting the structure. These fine micro cracks do not further enlarge if they are held small and locked by well bonding scraped fibres. To cope with the shrinkage crack formation in the very early age we chose scraped fibres with the best bonding performance in young concrete. The amount was 40 kg scraped fibres per cubic metre of concrete. For cracks arising later we chose crack bridging fibres with double bended ends, a length of 50 mm, diameter 0.8 mm, and 35 kg per cubic metre of concrete. Cracks have to be avoided primarily in the surface. Near the surface of a concrete the influences are more various and more intense than inside the structure. Therefore we planned for the steel fibre concrete to be on the top 50 centimetres, one half of the slab thickness. The lower area was conventionally reinforced with a single reinforcing layer to prevent crack widths bigger than 0.3 mm. (Fig. 2.) The middle of the 300 m long concrete slab was chosen as the fixed point. A 1 m deep trench was exe-

Figure 2.

Figure 3. The glass like surface of the finished concrete slab.

cuted monolithically in the central section of the slab. Concreting must start from here in the two directions simultaneously.

3

RESULTS

The 7000 m3 of concrete for the floor slab of the hall were placed within approx. 60 hours. The curing time was 31 days. The concrete was left in the side formwork for the complete curing time. Immediately after the completion of work the monitoring of the strength and deformation behaviour of the slab began. The floor slab behaved as calculated. The ends were shortened with a speed of 30–40 nm/s after three days. Each end shortened by about 40 mm. Compared with the values determined in the laboratory the temperature rise inside the concrete slab increased at most 32 K. The reason was the low fresh concrete temperature and the cool weather during the execution. After the length deformation stopped the floor slab was measured in a grid of 2.0 m × 2, 0 m. Such a high level of evenness was achieved that a planned additional levelling screed can most likely be foregone. Partners: Züblin AG, Direction North, Hamburg Holcim Concrete and Aggregates GmbH, Hamburg GuD Consult, Berlin IFDB, Berlin.

Layers of the floor slab.

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Durability and microstructural development during hydration in ultra-high performance concrete B. Möser, C. Pfeifer & J. Stark F.A. Finger-Institute for Building Materials Science, Bauhaus-University Weimar, Germany

ABSTRACT: The investigations on Ultra-High Performance Concrete (UHPC) with different high resolution electron microscopy imaging techniques and differential calorimetric analysis show a strong retardation of the hydration process caused by a high amount of superplasticizer. The C-S-H phases with a maximum length of 200 nm are visible after 18 hours of hydration time. During the hardening process an extremely dense and compact UHPC microstructure is formed. For this purpose a FE-SEM with high resolution low vacuum imaging capabilities in the low voltage mode (using a Helix detector) was deployed. The existing results of pretreated UHPC stored in water or heat treated up to 90°C show no significant deterioration in the microstructure and only marginal elongation after the storage in a climate chamber. In intentionally pre-damaged samples with microcracks a “secondary hydration” of unreacted clinker particles could be observed. In heat treated samples also a secondary ettringite formation was found. Nevertheless, the elongation damage threshold value of 0.4 mm/m was not exceeded.

2 1

RESULTS AND DISCUSSION

INTRODUCTION 2.1

So far only a few investigations (Fehling et al. 2008, Reda et al. 1999) on the hydration process, the formation of the microstructure and the durability under defined climate conditions of UHPC have been published. In this paper, the hydration of UHPC and formation of its microstructure were investigated on samples which had been stored in water or had undergone heat treatment. High resolution electron microscopy imaging techniques (ESEM-FEG and FE-SEM) were used for imaging of the hydration process and for characterization of the complex microstructure of UHPC. In addition, differential calorimetric analysis and quantitative X-ray phase analysis (Rietveld method) were used. These results are not presented here but can be found in the literature (Möser et al. 2007a, Pfeifer et al. 2007). Furthermore, durability of UHPC was evaluated by means of cyclic climate storage, a procedure developed in Weimar at Bauhaus-University. This program simulates the climate conditions in Central Europe in an accelerated manner for constructions exposed to ambient weathering (Seyfarth et al. 2006). All investigations were carried out on a standardized fine grain UHPC mixture (M2Q). Furthermore, durability studies were performed on a standardized coarse grain UHPC mixture (B4Q).

Hydration process and microstructure of UHPC

The aspect of extremely dense microstructures is particularly important during the investigation of UHPC by means of electron microscopy. Because of the spatial limitation caused by the dense mixture it comes to an extreme obstruction of growth of the hydration products occurring. Only by optimizing the contrast by means of Nova NanoSEM (using a Helix detector) at low accelerating voltages in a low water vapour atmosphere, it becomes possible to image such extremely dense microstructures in detail at high resolution and without charge buildup artifacts, see Figure 1. The hydration and reaction products in the microstructure can now clearly be imaged at a high resolution. A further option to characterize ultra dense microstructures is the FESEM analysis using backscattered electron imaging in combination with X-ray microanalysis. 2.2

Durability

In the following paragraph the UHPC reference mixtures M2Q and B4Q are evaluated with regard to their durability. In Figure 2 expansion of the reference mixture M2Q is shown. It is obvious that the expansion of the UHPC reference mixture is below the threshold

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29 cycles (603 days) no increase in expansion could be measured. By means of electron microscopy investigations it was found that in intentionally pre-damaged samples ‘second hydration’ takes place namely in the crack areas. After the 4th cycle of cyclic climate storage it can be observed that in heat treated and pre-damaged samples a secondary ettringite formation occurs in cracks having a width smaller than 10 µm. Cracks which exceeded this value showed a strong carbonation effect. 3

Different HR-SEM imaging techniques are necessary for the characterization of the hydration process and studying the development of microstructure of UHPC: ESEM in WET mode for the early hydration process, NanoSEM for the examination of the extremely dense microstructures of hardened samples and BSE imaging analysis for polished microsections. A strong retardation of the hydration process caused by a high amount of superplasticizer can be observed. Furthermore, the growth of ettringite is influenced. The length of ettringite crystals is smaller than 600 nm whereas under “normal” cement hydration conditions a length of up to 3 µm is visible. Clinker grains smaller than 2 µm had often been completely dissolved, resulting in hollow shell grains in the UHPC microstructure. The bonding between the aggregates and the UHPC matrix is very strong. The interfacial region shows no gaps between these components. Individual silica fume agglomerates up to a diameter of 250 µm are visible in the UHPC microstructure irrespective of the storage conditions of the samples. Locally concentrated initiation of ASR can be observed but the typical crack formation occurs only around the silica fume aggregates. No damages due to ASR occur. Investigations on the durability in a climate simulation chamber show that the expansion of UHPC reference mixtures M2Q is below the threshold value of 0.4 mm/m regardless of water storage, heat treatment of intentional pre-damaging. In heat treated and predamaged samples a secondary ettringite formation in micro-cracks with a crack width smaller than 10 µm can be observed whereas in wider cracks a signifcant carbonation occurs. Nevertheless the expansion threshold value is not exceeded.

Figure 1. 2d hydration time: by optimizing the contrast even very dense microstructures can be imaged in detail.

WS-pre-dam WS-undam Concrete /Giebson et al. 2006/

0,50

HT-pre-dam HT-undam

0,35 0,30

8. Cycle

6. Cycle

Expansion [mm/m]

0,45 0,40

0,25 0,20 0,15 0,10 0,05 0,00 8

50

CONCLUSIONS

92 134 176 218 260 302 344 386 428 470 512 554 596 638 Sample age [d]

Figure 2. Expansion measurements of reference mixture M2Q: water storage with intentional pre-damage [WS-predam], heat treatment with intentional pre-damage [HT-predam], water storage undamaged [WS-undam], heat treatment undamaged [HT-undam].

of 0.4 mm/m regardless of water storage, heat treatment or pre-damaging. During the first two cycles of cyclic climate storage, a minor increase in the expansion caused by a hygric expansion can be observed. For comparison pur poses, normal concrete of CEM I 32,5R (European nomenclature), greywacke, quartz sand with a water-to-cement ratio of 0.45 has been included in Figure 2. It shows a continuous increase of expansion. After 11 cycles of storage the threshold is trespassed. The expansion progression of the heat treated reference mixture M2Q shows double the amount of expansion of water stored samples. A value of 0.2 mm/m was not exceeded. Only a minor difference can be observed between the undamaged and pre-damaged series. After

ACKNOWLEDGEMENTS The authors would like to thank the German Research Foundation (DFG) for the funding within the framework of the priority program SPP1182 “Sustainable Building with UHPC”.

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Concrete ability to resist chloride ions using ternary blended cement I. Stipanovic´ Oslakovic´ & R. Roskovic´ Civil Engineering Institute of Croatia, Zagreb, Croatia

D. Bjegovic´ Civil Engineering Institute of Croatia, Zagreb, Croatia Faculty of Civil Engineering, University of Zagreb, Zagreb, Croatia

ABSTRACT: Concrete resistance to chloride ions is of crucial importance for reinforced concrete structures exposed to aggressive marine environment. This study shows one of the possibilities to increase concrete resistance to chloride penetration by use of ternary blended cement. Combination of three mineral additions in cement such as slag, fly ash and limestone, which were investigated in this study, is insuring lesser penetration of chloride ions into concrete. It was shown that diffusion of chloride ions into the concrete prepared with different types of ternary blended cements significantly decreased in comparison to the referent concrete. Concrete prepared by cement with 50% of mineral additions has shown to be more compact and less porous concrete and with almost the same other mechanical properties such as compressive and flexural strengths, comparing to the referent concrete.

1

INTRODUCTION

There are numerous reinforced concrete structures on Croatian coast exposed to very aggressive marine environment. After 20 years of exploitation, majority of the reinforced concrete structures are demonstrating severe damages. Those damages are caused by the sea chlorides where construction parts are directly exposed to the sea water or only to the temporary splashes (Fig. 1). Major factor influencing durability of the reinforced concrete structures is microstructure of the concrete cover to the reinforcement, i.e. cohesion of the pores in the cement paste and transfer zone. Porosity of the concrete and structure of the pores has an extensive influence on the concrete permeability, which enables penetration of the fluids, and speeds up corrosion process and damage of the concrete cover. This paper emphasizes the influence of the mineral additives in concrete through the usage of ternary blended cements, on the chloride penetration properties and finally on overall structural durability.

2

Figure 1. Corrosion of reinforced concrete elements due to chlorides from the sea.

process. Within the time there is dissolving or chemical dissolution along with rising of cement paste structure, which increases porosity and penetration. Often there are new products built in chemical reactions, which are causing concrete shattering, pH value decrease and decrease in reinforcement protection. As a consequence there is a decrease in concrete strength and undesirable changes in volume. Depending on destruction mechanism and major typical changes in concrete, cement influence on concrete corrosion could be divided on following processes: • Dissolution of hydration products; • Solid components transformation; • Swelling of solid phase.

CEMENT INFLUENCE

Durability of reinforced concrete depends on possibility of penetration of aggressive agents into the concrete, which may penetrate through absorption, diffusion and fluid flow under pressure through cement paste structure. Penetration of chlorides into the concrete is a most frequent cause of corrosion of reinforcing steel.

Durability of concrete structure in non-aggressive environment is increased due to the hydration progress. Nevertheless, if such structure is exposed to the aggressive environment its durability is decreased. Basically there are two factors of concrete corrosion: type of hydration product and cement matrix structure which is formed over the period of hydration

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3

5 4,5 4 3,5 3 2,5 2 1,5 1 0,5 0

400 350 300 250 200 150

1,5% 1,0%

0,8%

100 50

0

35 Mineral addition (%)

50

kg cem./m3 concrete

Coulumb x 103

The type of cement has a considerable effect on the concrete resistance to chloride ions penetration. Ordinary portland cement in transition zone (zone of contact of aggregate and cement stone) is increased by content of lime crystal which increases porosity. In case of blended cement the improvement of durability of concrete is described in their particular attributes and impacts, generally characterized by pozzolans reaction (fly ash, silica fume and blast furnace slag) and volume filling of pores (silica fume and limestone filler).

0 Coulumb x 1000 Air contents (%) kg cem./m3 concrete

Figure 2. Comparison of air content, amount of cement clincer and concrete permeability to chloride penetration.

EXPERIMENTAL TESTING AND RESULTS

For experimental testing 3 types of cement mixtures were used, one produced only with cement clinker, and two with ternary blended cements, with 30% and 50% of mineral additions, such as slag, fly ash and limestone. Experimental testing was performed on three types of concrete which were:

depend on the pore system (size and connectivity of the pores) and the quantity of water in the pores. From Figure 4 it can be seen that with the elevation of mineral additions in ternary blended cement, the chloride penetration into the concrete is significantly reduced, which will have the effect of increased durability of reinforced concrete structures. By increasing of mineral admixture in cement, quantity of clinker in 1 m3 of concrete will be decrease, what in cement production significantly contribute to better economic efficiency and sustainable development throw:

• RC = reference concrete, • C1 = concrete, with ternary blended cement with 35% mineral additions, • C2 = concrete, with ternary blended cement with 50% mineral additions. The first concrete was made with referential cement—ordinary Portland cement and it was marked as RC. The second concrete was made with the usage of ternary blended cement which consisted of 65% of referential cement, 20% of slag, 10% of fly ash and 5% of limestone and it was marked as C1. The third concrete was made of ternary blended cement which consisted of 50% of referential cement, 20% of slag, 20% of fly ash and 10% of limestone, and it was marked as C2. The testing of chloride diffusion have been conducted according to ASTM C 1202:1997. This method represents laboratory testing of electric conductivity of material sample thus enabling accelerated determining of resistance to the diffusion of chloride ions. The samples for the determining of chloride diffusion are cylinders Ø 100 mm, drilled from cubes and cut to be 50 mm high. Concretes in which mineral admixture are used have lower air content in fresh concrete and lower concrete ability to resist chloride ions relative to the increase of mineral admixture quantity, which is represented in Figure 2. Durability of concrete structures is primarily dependent on the environmental influences, i.e. the penetration of aggressive substances in the structural element from the environment. Processes of absorption, diffusion and the flow of substances through the concrete

• Natural resources and fossil fuel consumption decreases • CO2 emission decreases. 4

CONCLUSION

By using slag and pozzolan, with pozzolanic reaction the amount of Ca(OH)2 quantity is decreased. Decrease in Ca(OH)2 quantity is enabling reaction with chlorides from aggressive environment and prevents formation of C-S-H gel. By using fine lime stone filler there is a synergy effect with additional pore decrease. Research results presented in this study show that using mineral admixtures to cement, therefore in concrete occur: • Decrease in pore quantity; • Decrease of penetration chloride ions by using ternary blended cements. Number of previous researches pointed out the possibility of increasing concrete resistance to fluid penetration by using mineral admixtures in cement. This research on ternary blended cement is ongoing with the goal to improve overall concrete properties, by the implementation of other chemical and mineral additives.

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Sulphate resistance of high volume fly ash cement paste composites E. Aydin Cyprus International University, Mersin 10 Turkey, North Cyprus

ABSTRACT: This research was carried out to evaluate the effects of using high volume Class C fly ash on strength and sulphate resistance of construction materials. Physical, mechanical and sulphate resistance tests were conducted on the Φ50 mm/100 mm specimens. Physical tests considered were apparent specific gravity, water absorption and dry unit weight. Mechanical properties considered were compressive strength, and flexural strength. The durability properties considered was: sulphate resistance. In general strength and sulphate resistance of High Volume Fly Ash (HVFA) were considerably affected by amount of fly ash. Also, the strength properties for the 20 % fly ash mixtures were either comparable or superior to the no-fly ash concrete. The sulphate resistance of HVFA composites was either comparable to or better than the no-fly ash composites. All the mixtures, with and without fly ash, tested in this investigation conformed to the strength and durability requirements for excellent quality structural grade concretes. Based on the sulphate test results of this study indicates that the engineering performance of the final product can be adequate for using them in the manufacturing of construction materials (brick, ceramic tile, paving stone and briquette) and various civil engineering applications such as construction of structural fills, embankments, grouting injection, road bases and sub-bases.

Based on the Electric Power Research Institute (EPRI) report, more than 600 million tons of coal ashes are produced annually in the world, with 105 million tons per year produced in the United States. The disposal cost of fly ash is $10–20 per ton. Therefore, $6 billion is needed per annum. It is necessary to utilize a large volume of fly ash for the future activities. Utilization of high volume fly ash as a resource has been studied for decades in many areas such as cement/ concrete applications, brick, ceramic tile, lightweight aggregate, highway pavements. Based on the durability tests of this study indicates that the engineering performance of the final product can be adequate for using them in the manufacturing of construction materials (brick, ceramic tile, paving stone and briquette) and various civil engineering applications such as construction of structural fills, embankments, grouting injection, road bases and sub-bases. It is necessary to utilize larger volumes of fly ash in the construction industry. End-products made with HVFA have superior engineering properties, as well as economic benefits. The objective of this study is:

groups were selected to manufacture the building materials. Design mix groups were prepared at 100 mm slump for the manufacturing of briquettes, paving stones, bricks, tiles. Briquettes were prepared by using briquette molds and the others were by using 50 mm cubic, 40 mm*40 mm* 160 mm prismatic and ‘50/100 mm molds. The workability of the mixture combinations were measured by using slump test according to the ASTM C143–90a and flow table test according to ASTM C230–90. The numerical analyses results indicate that the relationship between the consistency (w/b, slump, flow), the physical (apparent specific gravity, dry unit weight) and mechanical properties (unconfined compressive and flexural strength) of fly ash mix groups of different physicochemical properties are highly correlated with each other. Based on the UCS values of HCP (hardened cement paste) composite; the final composites are adequate for manufacturing precast/prestressed elements, construction of structural fill, base and sub-base course, construction of catch basins and manholes and both non-load bearing and load bearing elements. The final composites can also be used in manufacturing Class MX and Class NX paving bricks, standard and special type tile production and manhole brick Grade MS and Grade MM; those bricks are intended to be used in manholes and catch basins not requiring high degrees of abrasive resistance. The final composites are not suitable for heavy vehicular brick applications. Based on the results of sodium sulphate solution indicate that a moderate to high performance in terms of the durability of the composite is expected.

• To investigate the physical, mechanical, sulphate resistance properties of HVFA fly ash (>75 %) mainly composed of silica fume, lime, cement and water reducing admixture (WRA). • To produce cost-effective and environmental friendly building products. These HVFA cement paste composites can satisfactorily be used in manufacturing building materials such as bricks, briquettes, tiles and paving stones. Based on the experimental results obtained in cold bonded engineering properties, five of the mix

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slump and strength based mix design. However; combination of WRA (water reducing admixture), silica fume together with HVFA shows better engineering properties than HVFA cement paste mixtures. Higher ash replacements are become acceptable and some ashes which normally do not meet activity standards is become acceptable. Published research on mixtures containing both a fly ash and WRA is limited. The importance of making trial mixes with each source of fly ash is emphasized. The fly ash used in this study is a Class C (high lime, Soma) fly ash, and the mix proportions shown are for this specific fly ash. The engineering properties of the material both in fresh and hardened state are highly influenced by the physical (fineness, grain size distribution, particle shape) and chemical (pozzolanic activity/rate and degree of hydration) properties of the mix ingredients, mainly by the properties of fly ash being the main constituent. The design water content in fresh state, thereby the porosity in hardened state are highly influenced by the physical properties of the mix ingredients; the Soma fly ash, having a well graded grain size distribution is more sensitive to the change in w/b in terms of variation in the slump and porosity. A fundamental approach, in this manner, is to use neat high volume fly ash cement paste in the construction/production of semi—structural/isolating materials. The strength and the durability of the fly ash can be improved to a great extent by addition of considerably low amounts of cement and/or lime and other mineral admixtures, letting the final product still be within economical limits in comparison to the same kind conventional construction materials.

Based on the sodium sulphate test results; the final composites are classified as medium to high sulfate resistant (standard specification of weight loss should be in between % 6–16). The originality of this study is to manufacturing of cost-effective environmental friendly building products by using HVFA cement paste. Based on the report published by the Governmental Development of Turkish Republic, for 1 tone of brick production 1.3 tones, and for 1 brick/tile production, 4 kilogram raw material (clay, shale, etc.) is needed. The replacement of fireclay and shale material by fly ash could save about $ 10 per tone for materials, the cost of the energy, and the time required to complete burnout of the clay component is replaced by fly ash. The clay minerals in coals are fired during coal combustion, so the energy consumption from firing during brick manufacture is not needed, resulting in energy savings. Few published literature is available on this subject and neither of them is included HVFA cement paste composites. HVFA composites show adequate sulphate resistance and the trend is exponentially increasing by fly ash content for all slump classes. Use of silica fume together with WRA shows better sulphate resistance than ordinary Portland cement-fly ash system. Removal of calcium hydrates (CH) and decrease of porosity by C-S-H formation causes an improvement in sulphate resistance. The numerical analyses results indicate that the relationship between the consistency, the physical and mechanical properties of mix groups are highly correlated with each other, good enough to construct a nomograph enabling target

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Durability of light-weight concrete with expanded clay aggregate M. Hubertova R&D in Lias Vintirov, LSM (Light-weight Building Material) k.s., Czech Republic

R. Hela & R. Stavinoha Brno University of Technology, Faculty of Civil Engineering, Institute of Technology of Building Materials and Components, Brno, Czech Republic

ABSTRACT: The paper describes development of light-weight concrete using light-weight expanded clay aggregate. The objective of this research was to evaluate durability of light-weight concrete exposed to corrosive liquids and gases (high concentrations of sulphate, chloride ions, Diesel oil and gaseous CO2 and SO2 environments).

1

INSTRUCTION

action of different corrosive environments, in particular gaseous CO2, gaseous SO2, solution of NaCl and Diesel-oil. After 28 days, basic physico mechanical properties were tested. Other samples were placed in corrosive environments for the period of 12 months. Then physico mechanical and physico chemical properties of hardened concrete were tested and changes of properties of light weight concrete as a consequence of corrosive environment were observed. The influence of fine additives on durability of concrete was observed. Gaseous corrosive environment was created in hermetically sealed corrosion boxes with constant levels of concentration and relative humidity of air. Corrosion atmosphere was renewed every two days. Liquid corrosive environments were created with the solution of NaCl into which the test samples were submerged. Constant value of concentration of corrosive agent in solution was maintained by means of regular renewal every 7 days. To simulate real conditions of corrosion of concrete as exactly as possible we tested corrosion of concrete by cyclic action of chlorides or Diesel Oil. In one cycle, the test samples were placed for 24 hours in test solution and then for 24 hours in standard laboratory conditions. Temperature of liquid environment was 20 ± 2 oC. Characteristics of individual environments are stated in Table 2. Determination of resistance of concrete surface to action of water and chemical de-icers was carried out in accordance with CSN 73 1326 (Czech standard). The principle of this standard consists in the fact that the surface subjected to freezing is covered with 3% solution of NaCl. The test was carried out in accordance with version A of above mentioned Standard.

As a part of the research of light-weight concrete we test it’s resistance to corrosive environment (chlorides, gaseous CO2, gaseous SO2, Diesel-oil) as a part of research work. This paper describes evaluation of a part of mix designs after 1 year of exposition to corrosive environments and compares influence of selected additives (including meta kaolin) on resistance of light-weight concrete. 2

EXPERIMENTAL PART

We developed a set of 5 mix-designs with difference only in type of admixture used. The amount of Portland cement (370 kg/m3), additives (40% by volume of cement) and admixtures (poly carboxylates based super plasticizers) was equal for all mix-designs. The composition of aggregate was also the same for all mix-designs, we used a combination of light weight expanded clay based aggregate and natural dense stone. The amount of effective water to keep constant consistence was between 160 and 170 kg/m3 depending on additive used. Basic reference mix-design (MIX I-A) with coal fly ash from Detmarovice in the proportion of 40% by volume of cement was mixed. This mix-design was modified with meta kaolin in the proportion of 5% by volume of cement (MIX I-B), powder micro silica in the proportion of 5% by volume of cement (MIX I-C). The mix-design (MIX I-D) contained micronized lime stone in the proportion of 40% by volume of cement. Durability of concrete was experimentally tested with these mix designs. Test samples made from above mentioned mix-designs were subjected to

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their microstructure, which would mean degradation of concrete matrix due to chlorides. We have to highlight that none of tested mix-designs showed after 360 days in chlorides occurrence of Fridels salt or other minerals, which could cause expansion pressures in microstructure of the material and degrade matrix to the extent of decreasing strengths of tested concrete. Modified MIX I-B showed increase of strength, which indicates positive effect of meta kaolin. Even though we expected higher level of degradation through cyclical action of chlorides, we did not find any major changes compared to constant exposition of test samples. Samples exposed to Diesel oil for 360 days did not show any major changes of microstructure of cement matrix. The most evident proof contamination by oil products (Diesel oil) is the loss by firing. Results of this analysis show slight contamination of surface layers of tested mix-designs. Comparison of samples from 200 mm depth of specimen placed in Diesel oil and in external environment show, that only the surface of tested concrete was contaminated. Contamination of surface of samples with cyclical exposition to Diesel Oil is lower than that of samples in constant exposition. Difference in strength compared to reference values are negligible, only within 3%. Comparison of contamination of different mix-designs shows that mix-design I-A (with fly ash) and I-B (with meta kaolin) unambiguously lead as regards resistance to penetration with oil products (Diesel oil). As regards resistance of surfaces to water and chemical de-icers, the observed samples (including the mix-designs with meta kaolin) show second degree of failure after 100 frost cycles (up to 500 g/m2), only I-D mix-design (with micronized lime stone) shows fifth degree of failure (over 3000 g/m2). Mix designs are in strength classes LC 30/33 to LC 35/38 and volume classes D 1.6 to D 2.0. This paper focused on influence of corrosive environment on selected light-weight concrete. Based on the results found in the research we can state that using coal fly ash is unambiguously positive as regards resistance and durability of light-weight concrete in corrosive environment, in particular CO2 and SO2. The analyses imply that mix-design I-A (with fly ash) and in particular mix-design with fly ash modified with meta kaolin (I-B) are much more resistant to corrosive agents than other mix-designs. This outcome has been achieved with the financial support of the Ministry of Industry and Trade of the Czech Reublic, MPO FI-IM5/016 “Development of light-weight high performance concrete for monolithic constructions and for precast elements” and with the financial support of project GA 103/07/076.

OBSERVATIONS AND COMMENTS

Based on the results of physico mechanical analyses we can state following: Samples exposed to CO2 in particular surface layers (0–20 mm deep from the surface) are in the second stage of carbonation with the exception of MIX I-A with fly ash and MIX I-B with meta kaolin. The degree of carbonation is also confirmed by occurrence of carbonation products (calcite, aragonite, vaterite) in the micro structure of concrete matrix of these mix-designs. In the second stage of carbonation other hydration products of cement are altered. Newly formed modification of CaCO3 together with amorphous gel of silicic acid form a crystalline neoformations of CaCO3 with very fine grain. Properties of concrete do not change much, which explains only little differences in compressive strengths and volume weights of mix-designs. MIX I-A (with fly ash) and I-B (with meta kaolin) are in the first stage of carbonation after 360 days of exposition to 98% CO2 and 75% relative humidity of air, which is the same condition as that of samples placed in exterior environment as regards the level of carbonation. In the first stage of carbonation, calcium hydroxide in the microstructure of the cement matrix (both crystalline—portlandit and from the space between grains) is attacked by carbon dioxide. The product of these chemical reactions is calcium carbonate crystallizing in the form of calcite. In the second stage of carbonation, carbon dioxide reacts with calcium-hydro-silicates forming fine-grain calcium carbonate in the microstructure of concrete in particular in the form of aragonite and vaterite. All mix designs were classed in the stage 1 of sulphation after 360 days of exposition to 98% SO2 with relative humidity of air 75%. After comparison of SO2 content in samples located in corrosive gas and external environment we have to state considerable increase of the content of SO2. Increased degradation of surface layers was also confirmed by marked coloring of samples. However, results of X-ray diffraction analysis did not confirm occurrence of products of sulphation (gypstone, monosulphate, trisulphate) indicating increased degradation of cement matrix. In the first stage, Ca(OH)2 (or its solution) in the spaced between grains is altered to hemi-hydrate of calcium sulphate, which partly fills pores. Strengths of concrete rise, but the value of pH decreases. Comparison of results of mineralogical composition of mix-design placed in chlorides and in external environment showed that action of chlorides on mix-designs modified with fly ash in the time period of 360 days did not cause formation of new phases in

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On the potential of rubber aggregates obtained by grinding end-of-life tyres to improve the strain capacity of concrete A.C. Ho & A. Turatsinze Université de Toulouse, UPS, INSA, LMDC (Laboratoire Matériaux et Durabilité des Constructions), Toulouse cedex 4, France

D.C. Vu Laboratoire Principal Routier, ITST (Institut de Transport Science et Technologie), Ha Noi, Vietnam

ABSTRACT: Cement-based materials suffer from low tensile strength and poor strain capacity. They are brittle and highly sensitive to cracking, notably to cracking due to length changes whatever the original cause of the length change. While being inspired on a well-known technique in metals by which the drilling of a hole at the crack tip induces stress relaxation and slows down its propagation, this work shows that a partial replacement of natural aggregates by rubber aggregates obtained by grinding end-of-life tyres is a suitable solution to improve the strain capacity before the crack localization. Experimental results confirm that the ideal, namely low modulus and high strength are two characteristics mutually exclusive. Despite its low strength, the strain capacity of rubberised concrete is significantly improved and the potential for cracking is reduced. They are some conditions to carry out durable applications in particular for concrete large area such as pavements and slabs and their repair. 1

3

INTRODUCTION

3.1

This paper reports the impact of rubber aggregates in concrete formulation. As expected, the study confirms that rubber aggregate incorporation improves the strain capacity of concrete before macrocrack localization. As a consequence, when resistance to the cracking due to imposed deformation is a priority, the technique should be considered as a suitable solution to improve durability in order to reduce maintenance expenses and an opportunity for a material recovery. It also appears that rubberized concrete can advantageously answer a special application: the Controlled Modulus Columns (CMC). 2 2.1

RESULTS AND DISCUSSIONS Compressive and splitting tensile tests

The compressive (fc) and splitting tensile (ft) strengths of the concrete mixtures were determined at 28 days. The results are presented in figure 1. The variation in the modulus of elasticity with rubber aggregate content is illustrated in figure 2. 3.2

Four-point flexure tests

The curves in figure 3, typical examples of the results obtained, enable each mix to be characterized by the load bearing capacity Fmax and by the corresponding deflection named the strain capacity (δFmax).

EXPERIMENTAL PROGRAMME Materials

3.3

CXR designates the mix where the volumetric ratio of rubber aggregates to natural sand is X%. Table 1.

One of the advantages expected from the enhanced strain capacity of rubberized concrete is an improved resistance to cracking due to restrained length change. Tests were performed according to a standard test method for determining age at cracking and induced tensile stress characteristics of concrete under restrained shrinkage (ASTM standard C 1581-04). This method consists of casting concrete in a circular mould around a steel ring having two strain gages as indicated in figure 4. Steel ring strain versus specimen age is illustrated in figure 5.

Concrete mix proportions (value in kg/m3). C0R

Cement Sand (0–4 mm) Rubber (0–4 mm) Gravel (4–10 mm) Water Superplasticizer Stabilizer

323 872 0 967 153 3.03 0.91

C20R

C30R

C40R

698 79

611 118

524 157

3.29

3.61

Effect of rubber aggregate incorporation on the concrete resistance to the cracking due to restrained shrinkage

3.99

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Figure 1. Effect of rubber aggregate content on compressive and splitting tensile strengths.

Figure 5.

Steel ring strain versus specimen age.

Table 2. Effect of rubber aggregates on the flexure load bearing capacity and on the strain capacity.

Mix

Load bearing capacity Fmaxr kN

Strain capacity δFmax mm

C0R C20RR C40R

16.6 12.1 11.5

0.05 0.08 0.10

Table 3. Effect of rubber aggregate incorporation on the concrete potential for cracking according to ASTM standard C1581–04.

Figure 2. Modulus of elasticity vs rubber aggregates content.

Net time-tocracking, tcr

Average stress rate, S

Mix

Days

MPa/day

Potential for cracking

C0R C20R C40R

9.25 15.50 33.25 (no cracking)

0.39 0.16 0.05

High Moderate low Low

4

With regard to mechanical properties, rubber aggregate incorporation is highly detrimental to the compressive and splitting tensile strengths of the material. The modulus of elasticity is also affected but it appears that its variation does not follow the classical empirical relationship with the compressive strength. In addition, rubber aggregate incorporation is a suitable solution to design a concrete fulfilling the required properties for CMC application. A drop in flexural strength is observed when rubber aggregates are used but the strain capacity is significantly increased. Tests clearly demonstrate that rubber aggregate incorporation is a suitable solution to reduce the potential for cracking due to restrained shrinkage. It can be concluded that if resistance to cracking due to imposed deformation is a priority, rubberized concrete is of great interest to achieve durable applications. In particular, an ongoing work focuses on performance of rubberized concrete as thin bonded repair material.

Figure 3. Flexure load—deflection curves—effect of rubber aggregates on the load bearing capacity Fmax and on the strain capacity.

Figure 4. mould.

CONCLUSION

Ring test specimen after extraction of the outer

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Investigations on the PCC-Microstructure after Mechanical Load A. Flohr, A. Dimmig-Osburg & K.A. Bode F.A. Finger-Institute of Building Material Science, Bauhaus-University Weimar, Weimar, Germany

ABSTRACT: For PCC as a constructional material two parameters are of particular importance: the deformation behaviour and the stiffness evolution at mechanical loading. The microstructure and the composition are the main differences between normal concrete and polymer modified concrete. To understand the deformation behaviour of PCC it is necessary to survey the behaviour of the polymer matrix under mechanical load. There are two ways how concrete is influenced by polymers under mechanical load. Firstly, strong polymer films demand a higher load to open the micro-crack edges so that the tensile strength increases. Secondly, a small particle size raises the specific adhesion between the matrix and the sand grains, which also leads to an increasing tensile strength. In fact, it is a co-action of both ways, depending on the kind of polymers, their particle sizes, their film tensile strengths, and their film formation temperatures.

1

INTRODUCTION

In principle the load regime of the tensile and compressive tests were all about the same.

A polymer modification primarily affects the bond phase of the concrete. It develops a polymer network, which is more or less strongly involved in the load transfer together with the hardened cement paste matrix. This raises the question on the influence of loading on the microstructure of PCC and how this affects the statical behaviour of modified concrete. In the development and testing as well as in the application of new constructional materials it is necessary to know the material parameters which have a decisive influence on the deformation behaviour. In this way the dimensioning of the structural part is ensured. 2

3.1

In the first part of the investigations free polymer films (foils) were produced and tested under different thermal conditions. Polymer 4 with a film formation temperature of approximately 30 °C, filmed in the drying chamber at 40 °C. In temperature range from 0 °C up to 20 °C the tensile strength of the polymer foils decreased and the strain increased, caused by the increase of the molecule mobility at rising temperature. During further rises in temperature the tensile strength of the polymer foils increased again.

MATERIALS

3.2

Compressive test on concrete cylinders

The compressive tests have been executed on concrete cylinders. The reference specimens were stressed up to 90 % and the PCC up to 80 % of their compressive strength. The maximum load was retained for 3 minutes, after that reduced to 1 kN and again retained for another 3 minutes. The examined specimens show clear differences in the deformation behaviour (Fig. 1). It can be stated that the deformations increase, i.e. the elastic and plastic deformation, crack strain but also creep.

For the investigations an epoxy resin (polymer 1), a re-dispersible powder (polymer 2) and a dispersion (polymer 4) on the basis of styrene/acrylate were used. With given processing and storage conditions (20 °C, 65 % rel. humidity) polymer 1 and 2 form continuous films. Polymer 4 does not form any films under these conditions. The polymer particles do not merge, but they have an adhesive bond to the aggregates and to the hardened cement matrix. The researches on concrete were made with Portland cement with an effective water-cement-ratio of 0.5. The polymer-cement-ratio was 0.15. 3

Invetigations of free polymer films

3.3

Tensile test on reinforced concrete cylinders

Concrete cylinders with an axial reinforcing steel were tested. The load was increased up to 50 kN and retained for 3 min, thereafter the load was reduced to 1 kN and was retained for another 3 min. The stress-deformation curves (Fig. 2) illustrate that PCC shows smaller deformations, higher

METHODS AND RESULTS

The load tests were accomplished with the same polymers. Thus, it is possible to compare the reactions of the polymer matrix in the structure of concrete.

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Figure 1. Transversal strain and longitudinal strain during the compressive test.

Figure 3. Matrix between grains with stretched and torn polymer films (PCC 2).

Figure 2.

Stress-strain relationship during the tensile test.

strength to produce cracks and an increase of slip of the reinforcement. 3.4

Detecting microstructure

After the load tests the microstructure of the loaded specimens was analyzed. The unmodified specimen has a typical relatively closed structure. Mostly the grains are pulled out of the matrix. The grains of PCC 2 and PCC 4 show adhesions of the polymers. There is a good bond between matrix, uniformly distributed polymer films and grains. In the compressive test tensile stress occurs, where structure movement caused by micro cracks is possible and crack surfaces move away from each other, i. e. in the transverse tension area. The result is a uniformly distributed system of cracks, which generally formed along the grain surfaces. If there is a good bond between the polymer films, the matrix and the grains, higher load is necessary to destroy this bond. In the tensile stressed areas a system of stretched polymer films occurs (Fig. 3) which inhibit a further drifting apart of the crack surfaces. That not only the formation of tension proof films is necessary to increase the tensile strength of concrete shows the tensile strength of PCC 4. The stressed PCC 4 (Fig. 4) shows only small stretched polymer films without strength. Essential is here the increasing of the adhesive bond of matrix and grains. The specific adhesion increases, so that the tensile strength increases.

Figure 4. Only small stretched films without tensile strength (PCC 4).

4

CONCLUSIONS

Tests have been made to determine the properties of free polymer films, and the deformation behaviour of statically stressed concrete samples. Coherences between the tensile strength of free polymer films and also the particle sizes of the polymers and the statical behaviour of PCC were found. There are two simultaneous ways how mortar and concrete are influenced by polymers under mechanical load. Firstly, strong polymer films demand a higher load to open the micro-crack edges so that the tensile strength increases. Secondly, a small particle size raises the specific adhesion between the cement matrix and the sand grains, which also leads to an increased tensile strength.

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Composites in structures – Use of special strength theories based on the form of anisotropy V. Lackovic´ Civil Engineering Institute of Croatia, Zagreb, Croatia

J. Krolo & M. Rak Faculty of Civil Engineering, University of Zagreb, Zagreb, Croatia

ABSTRACT: Almost all materials currently used in construction industry belong to the group of composite materials. These materials may be traditional ones, usually made of wood, stone, concrete or brick or modern ones, which are increasingly represented by polymer composites. When studying resistance of such materials to mechanical load, it is important to define the form of anisotropy, and an acceptable strength theory—based on analytical and experimental methods—that will assist us in the analysis of the limit state of materials when subjected to complex mechanical load. The limit state analysis for an orthotropic fibre-based composite was conducted by experiment, and the results obtained were numerically verified based on the second order strength theory. The validity of the selected strength theory was confirmed by complex experimental study, involving multi-axial loading of appropriate material components.

1

INTRODUCTION

properties other than original materials composing them. The structure of a composite material has the disperse phase and the base-matrix. Depending on the type and arrangement (geometry) of the disperse phase there are two forms of composite materials in terms of anisotropy. Composite materials reinforced with continuous fibres are commonly used in construction industry. In terms of form of anisotropy those materials are mostly classified as orthotropic composites.

The designing and execution of structures made of composite materials subjected to loading require a detailed analysis of their behaviour regarding mechanical load, which applies primarily to the complex static load. The first step is to define the form of anisotropy of the material to be used, or has been used, for construction of the structure. The next step is determination of mechanical properties of material at uniaxial load (compressive, tensile and shear strength and elasticity constants). Behaviour of material under mechanical load is determined based on the results of uniaxial testing and then the selection of a strength theory is required, for determination of material behaviour when subjected to complex—multi-axial static load. The Paper summarises behaviour of orthotropic polymer reinforced with continuous fibres subjected to complex mechanical load; the results of experimental investigation are separately presented. Finally, upon analysis of results of personal past investigations, certain suggestions are given for future investigations of composite materials in terms of mechanical load.

2 2.1

2.2

Special strength theories

The criterion of strength at anisotropic materials comprises several characteristics of material; one of the most important ones is the form of anisotropy. The applicability of acceptable theory of the first or second class of special strength theories is determined based on the form of anisotropy. When analysing the relation between deformities and stress it is important to bear in mind that in isotropic materials the deviator of stress represents a part of stress that does not result in change of volume; however, in anisotropic material the change of volume results from subjecting the material to normal and shear stresses. Goldenblat-Kopnov criterion is one of polynomial criteria consolidating anisotropic properties of material strength, which can be simply presented in the form of polynomial out of components of strength tensors of different orders and components of stress tensors for anisotropic materials. This is an appropriate criterion for evaluation of stress limits for anisotropic materials of different strengths at stretching

THEORETICAL ASSUMPTIONS Forms of anisotropy of composite materials

Composite materials are materials composed of two or more different types of materials, which display

99

and pressure in each direction, as well as different shear strength depending on the direction of tangential stresses in each defined plane. Elaborated form of Goldenblat-Kopnov criterion of strength in the layer in a state of plane stress is as follows: In case of a state of plane stress, when the axes of the main stresses concur with the anisotropic axes of material: σ1 = σx, σ2 = σy, τ12 = 0, the elaborated strength criterion is as follows: Π11σ 1 + Π22 σ 2 +

3 3.1

(Π

)

σ 12 + Π2222 σ 22 + 2Π1122 σ 1σ 2 ≤ 1

1111

EXPERIMENTAL TESTING

Figure 1. Pipes which make part of equipment for testing of stress and deformities at biaxial load.

About experiment

Experimental testing was conducted in the Laboratory of the Technical Mechanics Department of the Faculty of Civil Engineering in Zagreb and the Construction Laboratory in the Civil Engineering Institute of Croatia Zagreb. The testing was comprised of four parts. The first three parts referred to testing of samples to tensile load, compressive load and shear load and the fourth part referred to the control of estimated strength criterion for two combinations of biaxial state of stress in pipe element. Testing of samples at uniaxial load Mechanical properties of samples were tested according to theoretical postulates required for calculation of components of strength tensors. The following testing results were obtained: σyv = 46.87 MPa σyt = 120.9 MPa τ–45 = 14.89 MPa

σxv = 77.84 MPa σxt = 123.6 MPa τ+45 = 25.33 MPa

Testing of orthotropic composite sample subjected to biaxial load. The experiment was conceived to test stress and critical strength in two instances, on two models, at simultaneous compressive load applied along the pipe generatrix and tensile load obtained by water compression in model elements, at water pressure of: a) 0.1 MPa and b) 0.2 MPa. Figure 1 presents testing of elements at biaxial state of stress, performed in the Laboratory of the Department for Technical Mechanics. F = σ2(1) ⋅ Aring surface = 1.62 MPa ⋅ 4160.5 mm2 = 6.740 kN σ1 = 3.24 MPa = const.

Table 1. Value of stresses at material failure at biaxial state of stress.

σ1 (MPa) σ2 (MPa) ε (%0)

Experiment 1

Experiment 2

75.02 67.40 0.96

2.70 3.24 1.46

The critical forces at material failure are read from the diagrams. Table 1 presents the values of critical stresses and deformities in the direction of the main anisotropic axis L, read from the diagrams, for two performed experiments.

4

CONCLUSION

The analysis of theoretical and experimental investigations results in conclusion that mechanical properties of composite materials significantly depend on the form of anisotropy of those materials. This especially applies to the selection of one of special strength theories to be used for evaluation of mechanical properties (critical stresses and strength) at complex mechanical load. The experiment detailed in the Paper indicates the complexity of the procedure for the selected form of anisotropy, specific for polymer reinforced with continuous fibres.

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Concrete containers for containment of vitrified high-level radioactive waste: The Belgian approach B. Craeye & G. De Schutter Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Belgium

H. Van Humbeeck ONDRAF/NIRAS, Belgian Agency for Radioactive Waste and Enriched Fissile Materials, Belgium

A. Van Cotthem Tractebel Development Engineering, Consulting Company, Belgium

ABSTRACT: ONDRAF/NIRAS has selected the supercontainer as the preferred Belgian design for disposal of vitrified high-level radioactive waste. This supercontainer is based on the use of an integrated waste package composed of a carbon steel overpack surrounded by a concrete buffer based on Ordinary Portland Cement. An outer stainless steel liner encloses the concrete buffer. This buffer provides a favourable chemical environment for the overpack and radiological protection for the workers and should remain intact for on the order of a thousand years. For the choice of the cementitious buffer, two different compositions are being considered: a Self Compacting Concrete (SCC) and a traditional vibrated concrete (RPC). An intensive laboratory characterization program with large scale tests has been carried out. The obtained mechanical and thermo physical data are used to simulate the behaviour of the concrete buffer during the different stages of manufacturing by using a 3D thermal and crack modelling program. Through-going cracks in the concrete buffer should, at all times, be avoided.

1

INTRODUCTION

For the final disposal of long-lived, heat-emitting vitrified waste and spent fuel (category C waste) in the deep Boom Clay layer, ONDRAF/NIRAS has developed a design, called supercontainer, based on the use of an integrated waste package composed of a carbon steel overpack surrounded by a concrete buffer based on Ordinary Portland Cement. For the choice of the cementitious buffer, two compositions are being considered: a traditional vibrated concrete (RPC) and a self-compacting concrete (SCC). Several thermal, physical and mechanical tests are performed to predict the difference between these two compositions. This is part of an intensive laboratory characterization program in order to obtain concrete parameters used in a 3D thermal and crack modelling program. 2

of years for the spent fuels assemblies. For corrosion protection purposes, the overpack is enveloped by a high pH concrete buffer (high alkaline concrete). This buffer, with a thickness of about 70 cm, also performs as a well-defined radiological protection buffer for the workers and simplifies underground waste transportation operations. This buffer is surrounded by a stainless steel cylindrical envelope (also called liner). The outside radius of the supercontainer is about 1.9 m and it has a total length of 4.2 m (about 6 m for spent fuel assembly).

THE SUPERCONTAINER CONCEPT

The supercontainer is intended for the disposal of (vitrified) high level heat-emitting waste and for the disposal of spent fuel assemblies. In this concept (Fig. 1), the vitrified waste canisters or spent fuel assemblies are enclosed in a carbon steel overpack of about 30 mm thick. This overpack has to prevent contact of the waste with the water coming from the host formation during the thermal phase i.e. several 100’s of years for vitrified waste and 1000’s

Figure 1a. Cross-section in a disposal gallery for vitrified HLW. b. Axisymmetrical cross-section of the supercontainer.

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CONCRETE BUFFER COMPOSITION

There are certain restrictions to the different components of the buffer. It is recommended that CEM I, Portland Cement, with limited hydration heat production to avoid or limit thermal cracking, is used to prevent portlandite consumption (with additional pH drop), with additional restriction that the cement has a low SO3 and C3A content. This avoids formation of dense hydrogarnet with resulting increases of porosity and permeability. The concrete must have high sulphate resistance to better resist to sulphur species present in Boom Clay pore water. It is also recommended that both fine and coarse aggregates should be limestone (calcium, calcite). This limits the formation of CSH phases (alkali-aggregate reaction) resulting in expansion and cracking. No other organic additives are acceptable except, as low quantity as possible, a small amount of superplasticizer. 4

Point 3

Figure 2. Contourplot of Syy (left) after 36 h and Szz (right) after 36 h, casting temperature 20°C, environmental temperature 20°C (SCC). 4

3 fct 3 3

SZZ (MPa)

3

0,7 fct 3

2

1

THE NUMERICAL SIMULATION TOOL 0 0

The obtained concrete properties of SCC and RPC will be implemented in the material database of the finite elements programme HEAT. This material database can be subdivided into three categories: thermal properties, maturity related properties and mechanical properties. For the numerical simulation of the first stage of manufacturing of the supercontainer, the finite element programme HEAT has been used. HEAT calculates the stresses (due to temperature and humidity effects) and the strength in the concrete structure using a state parameter approach linked by a material database. First, the actual state parameters are calculated (like degree of hydration or maturity, temperature, humidity), and afterwards, stress calculations are realized. 5 5.1

First simulation: 20°C

In a first simulation, the casting temperature of the concrete and the outside temperature is equal to 20°C. For SCC the maximum temperature is 54.24°C at 41 h. The temperature peak of RPC is slightly lower (52.76°C) and occurs 4 hours earlier (at 37 h). It shows that early age cracking of the buffer will not occur. At all times, the radial (Sxx), axial (Syy) and tangential (Szz) stresses remain smaller than 0.7 fct. Furthermore, the stresses inside the SCC are slightly higher than inside the RPC. The region with the highest stress build up is located near the outer surface of the buffer (Fig. 2). 5.2

Influence of temperature

If we change the casting temperature of the concrete and the outside temperature to a value of 10°C, we notice that the temperature peak is lower (about 15°C difference) than in the case of 20°C and occurs about

48

72

96

-1

Time (h)

Figure 3. Time graph of Syy development for SCC at point 3, casting temperature 30°C, environmental temperature 30°C.

12 hours later. The developed stresses generate no early age cracking danger. In case of a temperature increase, up to 30°C, the temperature peak occurs 10 hours earlier and is about 15°C higher than in the case of 20°C. The axial and the tangential stresses become slightly higher than the 0.7 fct limit (Fig. 3). In reality, to prevent early age cracking, measures has to be taken. 6

POST-PROCESSING

24

CONCLUSIONS

Massive hardening concrete elements are very prone to early age thermal cracking due to the heat of hydration. Finite element simulations are performed for the first production step of the supercontainer, where the concrete buffer is cast into the outer steel liner, in order to investigate the problem of early age cracking. This supercontainer is used for the final disposal of long-lived, heat-emitting vitrified waste and spent fuel. From the simulations, the following conclusions can be obtained. • If the environmental temperature does not exceed a value of 20°C, no early age cracking is expected. The axial, radial and tangential stresses remain, at all times, smaller than 0.7 fct. • The temperature development, due to the hydration heat, is smaller in case of the traditional concrete (RPC) in comparison with self-compacting concrete (SCC). • Measures have to be taken in case of an environmental temperature of 30°C, in order to avoid thermal cracking.

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Performance of different types of Pakistani cements exposed to aggressive environments A.R. Khan NED University of Engineering & Technology, Karachi, Pakistan

ABSTRACT: This paper reports the results of a study conducted to evaluate the performance of different types of Pakistani cements exposed to various aggressive sulfate and chloride environments. Mortar and concrete (plain and embedded with reinforcing steel) specimens were cast using three types of cements, namely Type I (OPC), Type V (SRC) and Type I plus granulated blast furnace slag (BFSC). Mortar specimens were exposed to three sodium and magnesium sulfate solutions to study the deterioration due to sulfate attack. Plain concrete specimens were exposed to 4% NaCl, MgSO4 and Na2SO4 solutions. Concrete specimens embedded with reinforcing steel were exposed to wetting and drying cycles in 4% MgSO4 solution to determine the corrosion resistance by comparing the weight loss measurements. Maximum deterioration was noted in BFSC cement followed by OPC and SRC cements. The performance of OPC and BFSC was not significantly different from each other.

1

INTRODUCTION

drying cycles in 4% MgSO4 to determine the corrosion resistance of embedded rebars.

Performance of concrete is generally judged by strength and durability properties. Probably the most important durability issues with reinforced concrete are deterioration due to reinforcement corrosion and degradation of concrete exposed to sulfate-bearing environments. Problems of concrete durability are a cause of major concern all over the country especially in the coastal areas where the structures have undergone deterioration well before their expected life and need proper attention and care. This paper reports the results of a study conducted to evaluate the performance of different types of Pakistani cements namely Type I (OPC), Type V (SRC) and Type I plus granulated blast furnace slag (BFSC), exposed to various aggressive sulfate and chloride environments. The performance was evaluated by evaluating the reduction in compressive strength and assessing the resistance to salt attack by visual inspection and weight loss measurements.

2

EXPERIMENTAL PROGRAM

Mortar and concrete (plain and embedded with reinforcing steel) specimens were cast using three types of cement available in market namely OPC, SRC, BFSC. Mortar specimens were exposed to three sodium and magnesium sulfate solutions with sulfate concentrations of 1%, 2%, and 4% and to study the deterioration due to sulfate attack. Plain concrete specimens were exposed to 4% NaCl, 4% MgSO4 and 4% Na2SO4 solutions. Concrete specimens embedded with reinforcing steel were exposed to wetting and

2.1

Mortar specimens

Cement mortar specimens measuring 50 × 50 × 50 mm were cast with sand to cement ratio of 2.5 and the water to cement ratio of 0.5. After 28 days of curing, the mortar specimens were divided into two groups. One group of specimens was continuously cured under water while the second group was placed in 1%, 2%, and 4% sodium and magnesium sulfate solutions. These conditions represent very severe sulfate exposure conditions according to ACI 318–99. The effect of sulfate concentration on the performance of selected cements was evaluated by visual examination and measuring the reduction in compressive strength. 2.2

Plain and reinforced concrete specimens

Two types of concrete specimens (plain and embedded with reinforcing steel were used in this study. The concrete had a water to cement ratio of 0.65 and a cement content of 300 kg/m3. Cylinders having dimensions of 100 × 200 mm were cast for the compressive strength. After 28 days of curing, specimens were divided into two groups. One group of specimens was continuously cured under water while the second group was placed in three tanks with 4% NaCl, 4% MgSO4 and 4% Na2SO4 solutions. The performance of plain concrete specimens was evaluated by measuring the reduction in compressive strength. The reinforced concrete specimens for the corrosion resistance of embedded bars were 100 × 200 mm

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concrete cylinders in which a 16 mm diameter steel bar was centrally embedded. After 28 days of curing, all the specimens were completely immersed in 4% MgSO4 solution. The specimens were maintained in the same condition for 15 days and then subjected to drying in open air at room temperature for another 15 days. Each wetting and drying cycle thus consisted of 30 days. All the specimens were subjected to 6 complete cycles (180 days) of test period. 3 3.1

RESULTS AND DISCUSSION

3.3

Weight loss measurements

The corrosion resistance of OPC, SRC and BFSC reinforced concrete specimens was related to weight loss in embedded rebars. It was observed that the weight loss was minimum (2.5% per 180 days) in SRC specimens followed by BFSC and OPC specimens (6.1% and 10.8% per 180 days respectively).

Visual Inspection

No wear and tear was observed in the OPC and SRC cement specimens in all sulfate concentrations while marginal deterioration was noted in the BFSC specimens exposed to 2% and 4% Na2SO4 solutions. The deterioration increased with the period of exposure in BFSC specimens only. For cement mortar specimens exposed to 1%, 2%, and 4% MgSO4 solutions for 180 days, disintegration of the edges and breakup was noted in the OPC and SRC cement specimens, while cracking of the surface skin, localized at the edges, was noted in BFSC specimens. The severity of deterioration in all specimens exposed to 4% sulfate solution was, however, more than that in similar specimens exposed to solutions with lower sulfate concentration. Rate of deterioration generally increased with increasing sulfate concentration in all cements. 3.2

specimens it was 8%. Highest strength reduction (12%, 13% and 17%) was observed in SRC, BFSC and OPC specimens exposed to 4% MgSO4 solutions respectively.

Reduction in compressive strength

The compressive strength of the BFSC specimens decreased from 6% to 27% for 180 days of exposure to 1%, 2% and 4% Na2SO4 solutions. No strength reduction was observed in OPC and SRC cement specimens within 180 days of exposure. A higher strength reduction, ranging from 18% to 71%, was noted in BFSC specimens exposed to 1%, 2% and 4% MgSO4 solution. Strength reduction in OPC cement specimens, ranging from 17% to 26%, was more than that in SRC cement specimens (2% to 3%). After 180 days of exposure to the sulfate solutions the reduction in strength, due to sulfate attack, was the highest in BFSC specimens followed by OPC and SRC specimens. No strength reduction was observed in all concrete specimens within 180 days of exposure to 4% NaCl solution. For plain concrete specimens exposed to 4% Na2SO4 solutions, strength reduction in SRC concrete specimens was 5% while for OPC and BFSC

4

CONCLUSIONS

1. For the cement mortar specimens exposed to 1%, 2% and 4% Na2SO4 solutions, maximum strength reduction, due to sulfate attack, was noted in BFSC mortar specimens, while no reduction in strength was observed in OPC and SRC cement mortar specimens. The strength reduction in BFSC specimens increased with the increase in sulfate concentration in the exposure solution. 2. For the cement mortar specimens exposed to 1%, 2% and 4% MgSO4 solutions, maximum strength reduction, due to sulfate attack, was noted in BFSC mortar specimens followed by OPC and SRC mortar specimens. The strength reduction in all specimens increased with the increase in sulfate concentration in the exposure solution. 3. For the plain concrete specimens exposed to 4% NaCl, 4% Na2SO4 and 4% MgSO4 solutions maximum strength reduction, was noted in OPC and BFSC specimens followed by SRC cement specimens. The strength reduction in all specimens was maximum in MgSO4 solution followed by Na2SO4 solution. No re3duction in strength was observed in NaCl Solution. 4. The weight loss method for determining the corrosion resistance indicated that the SRC reinforced concrete specimens exposed to severe MgSO4 environment had superior performance as compared to OPC and BFSC specimens. 5. Poor performance of BFSC mortar and plain and reinforced concrete specimens can be attributed towards the fact that the amount of BFS blended with OPC was not available from the manufacturer and quantity of replacement material may play a vital role in the sulfate resistance of concrete.

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Durability requirements in self-compacting concrete mix design A. Ioani & J. Domsa Technical University of Cluj-Napoca, Romania

C. Mircea & H. Szilagyi National Institute for Research in Construction (INCERC), Cluj-Napoca Branch, Romania

ABSTRACT: The paper presents results furnished by a national research program developed at the Technical University of Cluj-Napoca in cooperation with specialists from INCERC Cluj-Napoca Branch, in order to implement self-compacting concrete in the Romanian precast concrete industry. Mixes for C50/60, C40/50 and C30/37 strength classes with cement CEM I 52.5R, cement and silica fume or cement and limestone filler have been designed and tested, and properties in fresh and hardened state have been evaluated. Mix constituents (sand, gravel) are materials currently in use in a local precast concrete plant and admixtures (HRWR) and additions (silica fume) are provided by Sika Romania Ltd. Mixture proportions and parameters such as cement type, cement content, w/c ratio and concrete strength class are discussed with respect to the limit values specified in Romanian Standard SR EN 206-1: 2006, for durability reasons.

1

MIXTURE PROPORTIONING

Similar to the case of conventional vibrated concrete, the primary specification for the hardened SCC is to meet the exposure class (EN 206-1) and the characteristic compressive strength at 28 days. When SCC is used in precast application and particularly for prestressed members, the dominant requirement is a high early strength of concrete which enable the workers to demould, to prestress or to move the elements within a short production cycle. Taking into account the characteristics of concrete members cast in the factory (shape, size, reinforcement, etc.), in the experimental program the following requirements for the fresh SCC were established: − − − −

− determine the cement amount in order to reach the required compressive strength at transfer and at 28 days, and to satisfy the limit values prescribed for the selected exposure (durability) classes in the National Annex of SR EN 206-1: 2006. 2

EXPERIMENTAL PROGRAM

The mix design procedure was successfully tested on more than 50 mixes for the precast concrete industry.

Slump-Flow class SF2, where 660 < SF < 750 mm; Viscosity class VF2, where 9 < VF < 25 s; Passing ability class PL2, where PL ≥ 0.8; Segregation resistance class SR2, SR < 15%.

The authors proposed a mixture proportioning procedure, aiming to fulfil the SCC requirements in fresh and hardened state: − maximise the total aggregate volume in order to obtain less paste; − reduce as much as possible the powder content and use the possible lowest water/powder ratio; − select the optimal sand/aggregate ratio (S/Agg) which leads to a minimum volume of voids; in experiments best results have been obtained using S/Agg = 0.56; − select the proportions between the aggregate ingredients in order to maintain the aggregate grading curve in “the recommended range” (Fig. 1); such range covers all important grading curves proposed as suitable for SCC in the technical literature, curves which represent grading with high packing densities, favourable in SCC mixes;

Figure 1. Aggregate grading ranges for SCC (Ioani & Szilagyi 2008).

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Representative SCC mix compositions are presented in Table 1, where C50/60 and C40/50 mixes have been designed for prestressed applications and C30/37 mix for R/C precast elements. The principal characteristics of SCC made with admixtures and additions produced by Sika, are shown in Table 2.

3

DURABILITY REQUIREMENTS

According to the Romanian Standard SR EN 206-1: 2006, the design of durability is based on 7 exposure classes (X0, XC, XS, XD, XF, XA and XM). For each class, in the National Annex a set of limit values for the concrete composition and concrete properties are given (w/c, cement content, cement type, concrete strength class, etc.). SCC mixes for prestressed applications are characterized by low water/cement ratios (0.38–0.39), high dosages of cement (481–510 kg/m3) and high concrete strength classes (C50/60 and C40/50) and consequently, all SCC compositions satisfy even the most restrictive limit values given in SR EN 206-1: 2006 (w/c ≤ 0.45, cement content ≥ 360 kg/m3 and a strength class Table 1.

SCC mix composition.

Strength classes Mix index SCC Cement (kg) Limestone filler (kg) SikaFume (kg) Sand 0–4 mm (kg) Gravel 4–8 mm (kg) Gravel 8–16 mm (kg) Water (l) water/cement water/powder Sand/Agg Vagg (l) Vpaste (l) ViscoCrete 20 Gold (kg) % of powder

C50/60

C40/50

C30/37

18

19”

31’

50

510 – – 920 230 493 199 0.39 0.39 0.56 620 380 6.12

481 – 25 920 230 493 202 0.38 0.40 0.56 620 380 6.32

410 95 – 920 230 493 192 0.47 0.38 0.56 620 380 6.5

350 155 – 920 230 493 190 0.54 0.37 0.56 620 380 5.5

1.20

1.25

1.30

1.58

≥ C35/45). For certain exposure classes, special measures should be taken: − for elements exposed to chlorides, a type of cement resistant to sea water actions should be used; − for elements exposed to freeze/thaw attack in the XF4 exposure class, supplementary air entraining admixture should be used in order to reach an air content greater than 5.5% (for SCC made with aggregate having a maximum size of 16 mm); − for elements exposed to chemical attack XA2 and XA3, the use of sulfate resisting cement are highly recommended (CEM II/B-S, CEM III/A). SCC mixtures developed in the research program for R/C precast elements are characterized by a moderate water/cement ratios (w/c = 0.54), moderate dosages of cement (350 kg/m3) and a concrete strength class C30/37. Having these characteristics, the proposed mixes meet the durability requirements corresponding to the following exposure classes: − XC1, XC2 and XC3, elements subjected to risk of corrosion induced by carbonation; − XS1 and XD1, elements subjected to risk of corrosion induced by chlorides from sea water or from other sources; − XF2 and XF3, elements exposed to freeze/thaw attack when air entraining admixtures are used in compositions to produce air content in concrete greater than 5.5%; − XA1, concrete subjected to slightly aggressive chemical environment (soils and ground water); − XM1, concrete subjected to mechanical attack (abrasion) which produces a moderate wear on industrial floors, slabs and platforms due to pneumatic wheel vehicular traffic. The cement type (CEM I 52.5R) used in experiments for SCC mixes is accepted in the National Annex of SR EN 206-1: 2006 for all exposure classes; this cement was primarily selected in precast/prestressed applications for its capacity to produce concrete with high early strength. 4

CONCLUSIONS

Table 2. Test results on fresh and hardened state. C50/60 Strength classes Mix index SCC Slump-flow (mm) T500 Slump-flow (s) V-funnel time (s) L-box passing ratio Segregation resistance (%) Cube compressive strength. (MPa) at – 1 day – 7 days – 28 days

C40/50

C30/37

18

19’’

31’

50

680 3 9.2 0.87 6.0

635 3.5 10.5 0.82 5.2

683 3.5 11 0.80 7.2

670 3 12.0 0.80 4.75

44.0 53.5 64.4

45.2 58.8 66.6

40.3 49.6 55.8

34.0 43.6 52.8

The results obtained in the experimental program confirm the possibility to produce with local materials, SCC with remarkable properties in fresh state and hardened state and an impeccable surface after removal from the mould. The proposed mixes for prestressed application (C50/60 and C40/50 strength classes) satisfy entirely the limit values specified in the National Annex SR EN 206-1: 2006 for concrete composition and concrete properties for whole exposure classes. The mix designed for C30/37 strength class satisfy only partial the demands of exposure classes and future tests will be need to accomplish a SCC composition with lower water/cement ratio.

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Chemical attack of SCC: Immersion tests and X-ray CT V. Boel Department of Construction, Faculty of Applied Engineering Sciences, University College Ghent, Ghent, Belgium

K. Audenaert & G. De Schutter Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Ghent, Belgium

ABSTRACT: Due to its alkaline nature, concrete in general is susceptible to acid attack causing the components of the cement paste to decompose. Especially concrete building components in agricultural constructions are subject to acid attack for which lactic, acetic and sulphuric acid seem to be the most important acids. In this paper destructive tests as well as non-destructive tests on self compacting concrete or cement paste will be discussed.

1

computed tomography illustrate the deterioration process caused by acid attack.

INTRODUCTION

Self-compacting concrete (SCC) and traditional concrete (TC) are based on different mix designs. As such, a different pore and microstructure of the cement matrix might occur causing a different durability. A research project has been set up in order to investigate the behaviour of SCC concerning chemical attack. This project includes the study of several SCC mixtures and TC mixtures. The concrete specimens are immersed in a solution of sulphuric acid in water and in a solution of acetic and lactic acid. The mass loss as well as the strength loss of the specimens is measured. The following parameters are studied: SCC versus TC, type of filler (fly ash and two types of limestone filler with a different grading curve), type of coarse aggregate (river gravel and calcareous rubble), the type of cement, cement/powder ratio, water/cement ratio, water/powder ratio and the amount of powder. A more direct, visual investigation of the pore system and the microstructure of the cement matrix can be realized by means of X-ray computed tomography (X-ray CT). X-ray CT provides non-destructive threedimensional visualization, creating images that map the variation of X-ray attenuation within objects. The X-ray transmission through an object is function of the material composition (effective atomic number), density and thickness. X-ray computed tomography produces a stack of 2D shadow images of complete internal 3D structures by reconstructing a matrix of X-ray attenuation coefficients. These images can then be volume-rendered to provide 3D information. As such this technique is interesting regarding the investigation of deterioration in concrete/hardened cement paste. Some tests performed by X-ray

2 2.1

EXPERIMENTAL PROGRAMME Immersion tests on concrete

These tests have been performed on samples immersed in a solution of lactic and acetic acid in water and in a solution of sulphuric acid in water. In this paper only the results obtained from the immersion tests in a solution of lactic and acetic acid in water will be discussed. Three cores of the specimens were placed in water and three cores were placed in a solution of 30 g/l acetic acid (CH3COOH) and 30 g/l lactic acid (C3H6O3) in water (pH = 2.5), all at a temperature of 20°C. After this test, the samples are taken out from the solutions and are put for 2 weeks in a climate room at 20°C ± 2°C and more than 90% R.H. They are prepared for a compression test in order to be able to determine the remaining compressive strength due to deterioration. The rate of deterioration is strongly depending on the transport properties which are depending on the microstructure. The link of water transport parameters (capillary suction, water permeability, water immersion) with the mass loss due to submersion in the solution of acetic and lactic acid has been studied. It has been found that the best correlation was found when the parameters derived from capillary suction were used. As those parameters are a function of the calculated capillary porosity, according to Powers’ model, the mass loss has been plotted against the capillary porosity as well (Figure 1).

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0 -2

0

2

4

6

8

10

12

14

mass loss [%]

-4 y = -0.98x - 2.90 r2 = 0.72

-6 -8 -10 -12 -14

SCC TC

(a)

-16

(b)

calculated capillary porosity [%]

Figure 1. Mass loss as a function of the calculated capillary porosity.

Figure 2. Scan after 25 h (a) and 21 d (b) of exposure to lactic and acetic acid.

From this figure it can be noticed that the capillary porosity has its important influence on the deterioration of TC and SCC mixtures due to acid attack as a combination of acetic and lactic acid. Figure 1 also illustrates that the acid attack of SCC is similar to that of traditional concrete, as far as the capillary porosity is similar. 2.2

X-ray computed tomography

The possibilities to use X-ray computed tomography as a technique to investigate chemical attack were explored in a small experimental program exposing samples to different solutions. The samples were scanned after about one day of exposure and after three weeks of exposure. The samples were also scanned before the attack, as a reference. In each test, a sample of hardened cement paste with diameter 5.5 mm was scanned (resolution of 8–10 µm). The composition of the samples was equivalent to a concrete mixture with 300 kg/m³ CEM III 42.5 LA, 300 kg/m³ limestone powder, 165 kg/m³ water and 1.8 l/m³ super plasticizer. 2.2.1 Attack by lactic and acetic acid The sample of hardened cement paste has been exposed to a solution of 30 g/l acetic acid (CH3COOH) and 30 g/l lactic acid (C3H6O3) in water (pH = 2.5) at a temperature of 20°C, leading to a disintegration of the hardened cement paste. The sample was scanned after 25 h and after 21 days of exposure. Reconstructed cross-sections of the samples are shown in Figure 2. After 25 h of exposure, an outer zone with lower density has been formed around the inner undamaged core. After 21 days the exposure had proceeded in such way that the outer part had partially been cracked and broken off. The front of attack is clearly visible in Figure 2(b). After 21 days, only 2.90 mm of the initial 5.50 mm were not damaged yet. 2.2.2 Attack by sulphuric acid The sample of hardened cement paste, exposed to sulphuric acid (H2SO4; pH = 1), has been scanned after 22 h and after 21 days (Figure 3).

(a)

(b)

Figure 3. Scan after 22 h (a) and 21 d (b) of exposure to sulphuric acid.

After 22 h only the outer zone (of 0.15 mm thick) is visibly attacked and pushed off. Within the rest of the sample no effect was noticeable. After 21 days, the damaged outer zone had a thickness of 0.45 mm, but it seemed like it consisted out of two layers with an air layer in between (with a maximum thickness of 0.20 mm).

3

CONCLUSION

The deterioration due to acetic and lactic acid is strongly depending on the concrete composition and the penetrability of the cement matrix. Some measures which could be taken to decrease the deterioration are: the use of fly ash, the use of filler enhancing a dense structure, a high C/P, low powder content, a low W/C and avoidance of calcareous rubble. Acid attack of SCC seems to be similar to that of traditional concrete, as far as the capillary porosity is similar. The chemical attack of hardened cement paste, due to immersion in a solution of acetic and lactic acid in water and in a solution of sulphuric acid in water was visualized by means of X-ray computed tomography. Different deterioration mechanisms can be observed.

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Study of chloride penetration in self-compacting concrete by simulation of tidal zone K. Audenaert & G. De Schutter Magnel Laboratory for Concrete Research, Ghent University, Ghent, Belgium

ABSTRACT: By cyclic wetting and drying of concrete in the tidal zone, chloride ions are penetrating rapidly under the combined action of capillary suction of seawater followed by diffusion of chloride ions. In order to study this behaviour, a test set-up was used immersing specimens in a chloride containing solution, followed by a drying period. After distinct test durations the penetration depth is determined. This test method was applied in the framework of a project studying the durability of Self-Compacting Concrete (SCC), including an experimental program on 16 SCC and 4 traditional mixes. Four types of cement and three types of filler are implemented and the influence of the amount of powder, the water/cement ratio and the cement/powder ratio is studied. The influence of the combined action of capillary absorption and chloride diffusion on the penetration is investigated and a simplified model is proposed, leading to an accurate prediction.

1

INTRODUCTION

Chloride induced reinforcement corrosion is the main durability problem for concrete structures in a marine environment. If the chlorides reach the reinforcement steel, it will depassivate and start to corrode in presence of air and water. Since the corrosion products have a larger volume than the initial components, concrete stresses are induced, leading to spalling and degradation of the concrete structures. The most common transport mechanisms determining the chloride penetration velocity are diffusion (chloride gradients in standing water), capillary suction (chlorides transported with moving water) and permeation (chlorides transported with water under pressure). All three transport mechanisms may occur simultaneously. The chloride penetration in concrete is largest for the part of a structure in the tidal zone. The chloride containing sea water is penetrating the concrete by capillary suction in the wetting period. Because capillary suction is a fast transport mechanism, the chloride ions are penetrating to a relatively high penetration depth. In the drying period, the water is evaporating from the concrete, leaving the chloride ions in the concrete. From there, the ions will diffuse into the concrete. In the following wetting period, new chlorides will be penetrating into the concrete together with the sea water by capillary suction. The continuously increasing amount of chloride ions at the penetration depth by capillary suction is creating a high concentration gradient over the remaining concrete. From this a high diffusion velocity is created. In this way, the penetration velocity of the chloride ions in the concrete is higher for the tidal zone than for example for the permanently immersed zone.

To simulate real chloride penetration conditions in a tidal zone, a test method was developed, immersing specimens in a solution containing chlorides and exposing the specimens to air. The specimens are cylindrical with a diameter of 230 mm and a height of 70 mm. Each cycle takes approximately 1 hour, 1/3 wetting and 2/3 drying. In this test a complex combination of diffusion, capillary suction, wetting and drying is created. After 6, 12, 18, 24, 30 and 36 weeks the specimens are broken and the penetration depth is determined with silver nitrate solution. Self-compacting concrete (SCC) combines a high flowability and a high segregation resistance, obtained by a large amount of fine particles and the use of superplasticizers. As the concept of self-compacting concrete is different from traditional concrete, changes in durability are to be expected. As part of a larger project studying the durability of self-compacting concrete, the chloride penetration in self-compacting concrete was investigated. An experimental program was carried out on 16 self-compacting concrete mixes and 4 traditional concrete mixes. Four types of cement and three types of filler (fly ash and two types of limestone filler with a different grading curve) are implemented and the influence of the amount of powder, the water/cement ratio and the cement/powder ratio is studied. The obtained chloride penetration profiles are analysed and the influence of different parameters in the concrete composition are investigated and discussed. In order to analyse and model the test results, the following approach was applied. In each wetting cycle the water, containing chlorides, will bring the chloride into the concrete. During the drying cycle, the water will evaporate and the chlorides will remain in the concrete. From the penetration depth of the

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capillary absorbed water, xcap, the chloride ions will start to diffuse further into the concrete. The diffusion of chloride in concrete is following Ficks second law and can be written as: ⎞⎞ ⎟⎟ ⎠⎠

H [mm/Vweek]

⎛ ⎛ x C x = Co ⎜1 − erf ⎜ ⎝ 2 Dt ⎝

4.5

(1)

4.0

experimental

3.5

calculated

3.0 2.5 2.0 1.5 1.0 0.5

with Cx = chloride concentration in the concrete at a distance x from the surface; Co = surface chloride concentration; D = the diffusion coefficient, in this formula considered constant, and t = the time. This equation can also be written as: ⎛ C ⎞ x dif = 2 D erf −1 ⎜ 1 − x ⎟ Co ⎠ ⎝

t

=A t

(2)

If the penetration depth is determined with silver nitrate, Cx will be a constant (0.07 N), and thus A will be a constant. Also water absorption by capillary action during short time periods can be modelled by this kind of equation: x cap = S t

(3)

with x = the penetration depth of the water, t = time and S = a capillary suction coefficient. A first rough approximation of combined diffusion and capillary absorption is to take both coefficients A (in Equation 2) and S (in Equation 3) into one coefficient H: x tot = x cap + x dif = H t

(4)

Using this test method, it is impossible to determine A and S separately. Therefore H is proposed to model the test results and is determined based on regression. For all concrete mixes, this regression value was determined and is given in Figure 1. For the modelling of the results, the height of capillary absorption was calculated and the penetration depth

0.0 1

2

3

4

5

6

7

8

9

SCC

Figure 1.

10 11 12 13 14 15 16 1

2

3

4

TC

Experimental and calculated values for H.

of the chloride ions driven by diffusion was formulated in function of time. From this input, the penetration depth in function of time was determined and the regression constant H deduced. This value is also given in Figure 1.

2

CONCLUSIONS

An experimental program on 16 self-compacting concrete mixes and 4 traditional concrete mixes is carried out concerning the behaviour with respect to chlorides in a test simulating real conditions (wetting and drying, capillary suction and diffusion). Four types of cement and three types of filler (fly ash and two types of limestone filler with a different grading curve) are used and the influence of the amount of powder and water, the water/cement ratio and the cement/powder ratio is studied. Based on these tests, the conclusion is that the penetration depth in real conditions is strongly influenced by the water/cement and water/(cement + filler) ratios. Decreasing one of these ratios or both is leading to a decreasing penetration depth. Another important conclusion is that the chloride penetration depth in SCC by cyclic immersion is lower than the penetration depth in TC. The test results were modelled based on the capillary absorption behaviour and the diffusion process. A good agreement is obtained, certainly for selfcompacting concrete.

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Transport of water and gases in crack-free and cracked textile reinforced concrete Rabea Barhum, Matthias Lieboldt & Viktor Mechtcherine Institute of Construction Materials, TU Dresden, Germany

ABSTRACT: Textile Reinforced Concrete (TRC) exhibits a very favorable stress-strain behavior but is accompanied by the formation of a considerable number of fine cracks. This paper addresses the effect of such multiple cracking as well as of the textile reinforcement on the transport of fluids and gases into and through this novel material. Water absorption, gas permeability, and water permeability were investigated on both crack-free and cracked specimens. Very low gas permeability perpendicular to the textile layers was observed in crackfree specimens. However, a pronounced increase in water and gas permeability was observed under conditions of multiple cracking. The transport values obtained correlated to the residual strains (deformations remaining after unloading) as well as to the cumulative crack length. Furthermore, the capillary suction by multifilament yarns influenced significantly the water absorption of TRC in the direction parallel to the textile reinforcement. Finally, the effect of self-healing phenomena on the transport properties of TRC has been considered. 1

Table 1.

INTRODUCTION

In the case of the utilisation of textile reinforced concrete (TRC) as a repair material, the protection of the old structural members against corrosion in concrete and its steel reinforcement is an important function of the TRC layer. Therefore, knowledge and understanding of mechanisms which govern the transport of water and gases in TRC are required for cracked and crack-free material. In this study the water absorption as well as the gas and water permeability of TRC were investigated. Special attention was given to the effects of the reinforcing yarns and typical fine cracks on the transport properties of TRC. Furthermore, the influence of the self-healing of cracks was considered. 2

PREPARATION OF THE SPECIMENS

Table 1 gives the compositions of the used matrix. Table 2 sets out the fineness of the weft and warp threads and the spacing between the yarns. Four different types of biaxial textile made of alkali-resistant glass (AR-glass) were used as reinforcement. Rectangular plates 600 mm long and 120 mm wide were produced using a lamination technique. Depending on the test methods to be applied, the thickness of the plates varied from 6 mm to 12 mm and 20 mm. Quadratic samples with a side length of 12 mm were cut out of the plates to test the capillary water absorption. Round specimens with a diameter of 100 mm were cored out of the plates for the permeability tests with water and gases. Uniaxial tension tests were performed on the TRC plates to produce crack patterns similar to those in practical applications. In order to obtain crack widths similar to those under service conditions, the TRC

Composition of the matrix [kg/m³].

CEM III B 32,5 Fly ash Micro silica suspension (solid:water = 50:50) Fine sand 0/1 Water Super plasticizer

550 248 55 1101 248 8

Table 2. Textiles used as reinforcement. Warp Textile structure Fineness used [tex]

Spacing [mm]

Fineness [tex]

Spacing [mm]

008 010 016 019

7.2 7.2 7.2 7.2

2*640 1200 1200 2*640

7.2 7.2 7.2 7.2

2*640 2*640 640 2400

Weft

plates were purposely “overstretched”, i.e. loaded above usual loading levels, and then unloaded. During the unloading the cracks become narrower but still possessed representative widths, estimated by dividing the remaining, non-elastic deformation by the number of visible cracks counted. 3

WATER ABSORPTION TESTS

Samples of both crack-free and cracked plates were used for the water absorption tests. Before testing the specimens were dried at a moderate temperature of 50°C until no further significant loss in weight could be observed. These tests were performed on two types of crack- free specimens (three specimens of each type), one with and

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TRC-Specimen (d = 100 mm, h = 12 mm): 4 layers AR-glass, 1280 x 2400 tex Cumulative crack length: 355 mm Crack width: 5…20 µm

3.0

1.6

Water flow rate [cm³/(s·m²)]

Capillary water absorption [g/cm²]

1.8

1.4 Without textile

1.2 1.0

With 4 layers of textile (008) weft thread direction

0.8 0.6 0.4 0.2

2.5

Cycle: 2 d in water saturated 1 d under 150 kPa

2.0 1.5

1st Cycle 2nd Cycle

1.0

3rd Cycle 0.5

0.0 0

10

20

30

40

50

0.0

60

0

Time [h]

Figure 1. Results of water absorption tests on the crackfree specimens with and without textile reinforcement (parallel to the yarns).

2

4

6

8

10 12 14 Time [h]

16

18

20

22

24

Figure 3. Water flow through TRC under pressure of 150 kPa.

20 Specimen 1 before water test

Cracked specimens with 4 layers of textile (008) Crack widths: 0-40 µm

0.30 0.25 0.20 0.15 0.10

176

225

8

0

0

20

40

60 Pressure [kPa]

80

100

0

120

Figure 2. Effect of applied pressure and cumulative crack length on the oxygen volume flow rate.

one without textile reinforcement. The reinforced specimens possessed 4 layers of textile 008 as in Table 2. The results of the tests showed that due to the capillary action of the multifilament yarns, the specimens with layers of textile absorbed water much faster than the specimens without textile reinforcement, cf. Figure 1. GAS PERMEABILITY TESTS

Round plates 100 mm in diameter and 12 mm thick were used as specimens in the gas permeability tests. The plates contained 4 layers of textile 008 (see Table 2). The specimens were placed into a metal test cell and pressures from 10 to 100 kPa were applied to one side of the cracked specimens with the opposite side at ambient pressure. The pressure levels in the tests on crack-free specimens were up to 250 kPa. Measurement began only after a stationary flow condition with constant pressures on both sides was attained. Figure 2 shows a linear increase of the oxygen volume flow rate with increasing pressure. Higher values of the cumulative crack length lead to a significant increase of the flow rate as well. 5

Specimen 2 after water test

235

0.00

4

Specimen 2 before water test 12

4

Cumulative crack length 0.05

Specimen 1 after water test

16 Volume flow rate [cm³/s]

Volume flow rate [cm³/(s·cm²)]

0.35

WATER PERMEABILITY TESTS

After the tests of gas permeability the same specimens were used for measuring water permeability. Before

20

40

60

80 100 Pressure [kPa]

120

140

160

Figure 4. The results of gas permeability tests before and after 3 cycles of water permeability experiments.

the water permeation tests, vacuuming and saturation were performed for each specimen (48 hours). Figure 3 shows the development of the water flow rate in time for all three cycles. While a considerable decrease of the water flow rate could be observed during the first cycle, only minor changes occurred during the second and the third cycles. Comparison of the water flow curves shows that the volume flow rate decreases from cycle to cycle. This can be attributed most probably to the effects related to the swelling of the cementitious matrix and to the self-healing of cracks. 6

SELF-HEALING OF CRACKS

Oxygen permeability was measured under four pressure levels (25, 50, 100 and 150 kPa) before the water permeability tests and after three cycles of exposure to the water pressure. The tests were performed on two different cracked specimens, each of them reinforced with 4 layers of textile 019 (see Table 2). Figure 4 shows a dramatic decrease in the volume flow rate of oxygen through the specimens after they were subjected to the water permeability tests. Since the specimens were dried after the water treatment, the effect of the swelling of the cement matrix can be excluded. Hence the only obvious explanation for the measured decrease in the gas volume flow rate is the self-healing of fine cracks.

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Application of FRC and FRSC in structural elements of St Rok tunnel E. Seferovic & E. Barisic Civil Engineering Institute of Croatia, Zagreb, Croatia

ABSTRACT: This paper describes the application of Fiber Reinforced Concrete (FRC) and Fiber Reinforced Sprayed Concrete (FRSC) in structural tunnel elements. FRC and FRSC, produced with polypropylene (PP) and steel fibers, were applied in primary and secondary tunnel lining, as well as in prefabricated concrete elements of draining system of St Rok Tunnel (5.661 m), one of the longest road tunnels in Croatia. A survey of preliminary laboratory testing of FRC and FRSC is given, with the review of the application in future tunnel structural elements. Durability properties of FRC and FRSC and fire resistance of FRSC reinforced with PP fibers are discussed in the paper.

1

INTRODUCTION

Concrete is the world’s most widely used material in construction of bridges, buildings, hydrotechnical, road and industrial structures as well as in fabrication of various types of precast elements. Concrete is a brittle material, which is sensitive to cracking and has low tensile strength. Accelerated degradation of infrastructure created the need to quickly develop innovative, long-lasting and cost effective methods for maintenance, repair and construction of new structures. One of the possible ways to deal with this problem is by applying fiber reinforced concrete. Fiber reinforced concrete is a composite material. Discontinuous or continuous fibers of high tensile strength are added to the ingredients of regular concrete during the process of mixing. Adding fibers to the matrix, improves specific deformation, matrix ductility, as well as energy absorption capacity, toughness and dynamic properties of concrete.

2

FRSC AND FRC WITH STEEL FIBERS

The properties of FRSC and FRC with steel fibers depend on fiber properties (volume ratio, strength, modulus of elasticity, shape and dimensions), and properties of cement matrix (strength, composition, volume ratio, modulus of elasticity). Extensive research of steel fiber reinforced concrete was carried out during the construction of St Rok tunnel in Croatia, and it was the first investigation of that type in Croatia. The advantages of FRSC with steel fibers over sprayed concrete reinforced with steel fabric reinforcement are: − a lesser amount of sprayed concrete is required (since smaller thickness is needed)

− a lesser amount of work is needed (scaffold is not needed) − it is easier to apply FRSC to unleveled parts of the excavation, due to its configuration − work is safer with FRSC − use of FRSC saves a lot of time − it is easier to organize works in a confined space. Apart from basic tests (compression and bending strength), the most important tests that were carried out were toughness and energy absorption tests. Toughness (post-cracking load bearing capacity) is the best indicator that differentiates FRSC from un-reinforced sprayed concrete. It is defined as the total energy of deformations, absorbed before a complete failure of the specimen, determined from the area under the load-deflection curve. It is obtained by bending a prism loaded at mid-span. The results of tests show that brittle failure occurred in regular concrete (without steel fibers), after the appearance of the first crack, so there is no postcracking load bearing capacity. On the other hand, steel fiber reinforced sprayed concrete possesses a significant post-cracking load bearing capacity (toughness). 60 × 60 × 10 cm plates were manufactured for the purposes of testing of load bearing capacity testing (energy absorption), plates made of standard sprayed concrete, plates made of standard steel FRSC and plates made of standard sprayed concrete reinforced with steel fabric reinforcement Q-139. The relations load-deformation (deflection) and energy absorption (plate capacity)—deformation were obtained by these tests. Based on test results, it can be concluded that sprayed concrete with added steel fibers is more suitable for securing underground excavation—of tunnels in soil classes II and III, than standard sprayed concrete or sprayed concrete with steel fabric reinforcement.

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3

FRSC AND FRC WITH POLYPROPYLENE FIBERS

Since the surface of PP fibres is hydrophobic, they do not absorb a part of the water from the matrix, and therefore do not reduce concrete workability. The main characteristics of PP fibers are the prevention of plastic cracks in young concrete and great resistance to fire action. A minimum amount of PP fibers that should be used to prevent plastic cracks in concrete is 0.1 vol%, i.e. 0.9 kg/m3. Addition of PP fibers to concrete does not have a significant effect on its compressive and tensile strength. However, during the determination of compressive strength in specimens of FRC with PP fibers, very ductile behavior was observed at failure. Toughness tested according to ASTM C 1018, increased with the increase in the amount of fibers, while bending strength increased insignificantly. Impact strength and absorbed energy in polypropylene FRC is significantly greater than in regular concrete, especially after the appearance of the first crack, and with the increase in the amount of fibers they increase as well. During impact load testing, the absorbed energy was from 33 to even 1000% greater in polypropylene fiber reinforced concrete than in regular concrete. Polypropylene fibers can be added to concrete in a smaller amount to increase its resistance to plastic cracking or in a greater amount to improve toughness and residual strength, impact strength, resistance to fire action and prevention of crack growth due to long-lasting shrinkage caused by drying or temperature changes. On the site of St Rok tunnel, specimens with PP fibers and without them were fabricated, of dimensions 60 × 60 and 30 cm in height, of the same thickness as the designed secondary tunnel lining. These specimens were tested for fire resistance at the temperature of 1350°C for two hours in a specially constructed furnace. During testing, the temperature was measured in specimens, at distances of 4, 10 and 15 cm from the specimen surface. The temperatures in the specimen without PP fibers are considerably higher than temperatures in the specimen with PP fibers. This phenomenon leads to great differences between the damage in specimens with PP fibers and the damage in specimens without PP fibers. Deep scaling and significant cracking in the specimens without fibers were observed, the reinforcement that was located at 4 cm from the surface melted, while specimens with fibers sustained only slight damage, 1–2 cm in depth.

Adding PP fibers increases the resistance to fire action and decreases scaling and structural failure. On visual inspection of the primary concrete lining of St Rok Tunnel made of FRSC with the addition of PP fibers, the following was found: − adhesion of FRSC with PP fibers to the rock mass is considerably better than the adherence of regular sprayed concrete − loose rebound material mass during placement of FRSC with PP fibers is significantly smaller than in standard sprayed concrete − falling of fresh sprayed concrete with PP fibers, from the arch part in particular, is insignificant due to the enhanced adhesion − damage to the hardened sprayed concrete close to mining location (front of the tunnel) is considerably smaller damage sustained by standard sprayed concrete − at the positions of accidental lateral impacts of machinery in the tunnel, it was noticed that damage was of a lesser degree when FRSC with PP fibers was used.

4

CONCLUSION

Based on the facts mentioned in this paper, it can be concluded that: − Sprayed concrete reinforced with steel fibers of required properties can be used instead of steel fabric reinforcement, for securing tunnel excavation in soil classes II and III. − Sprayed concrete reinforced with steel fibers of required properties can be used instead of standard reinforcement in the production of hollow concrete elements (gutters) for tunnel draining, as well as for other thin elements (concrete slabs, kerb units, walls of tunnel facilities etc.). − Sprayed concrete reinforced with polypropylene fibers of required properties can be used for securing tunnel excavation in soil class II. − Sprayed concrete reinforced with polypropylene fibers of required properties can be used in combination with standard reinforcement or without it, for secondary tunnel lining, thus achieving a great degree of resistance to fire action and reducing the number of micro cracks in young concrete and in hardened concrete later on and consequently an improvement of durability properties of concrete lining and the entire tunnel structure.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Durability of Strain-Hardening Cement Composites (SHCC) – An overview G.P.A.G. van Zijl Department of Civil Engineering, University of Stellenbosch, South Africa Faculty of Architecture, Delft University of Technology, Delft, The Netherlands

ABSTRACT: It has become possible to design fibre reinforced cement-based composites to desired mechanical performances. Among the various classes of High Performance Fibre-Reinforced Cement-based Composites (HPFRCC) that have been developed, a particular class of moderate tensile strength and pseudo strain-hardening to ultimate strain levels of 3% and beyond is of interest here. Fibre-reinforced Strain-Hardening Cement Composites (SHCC) have such superior tensile behaviour. The pseudo strain-hardening is achieved by multiple cracking through effective fibre bridging of the matrix cracks. This enables resistance to higher tensile loads without significant crack widening, but rather successive cracks arising at the weak points in the matrix. It is the crack control, generally to widths below 100 um despite large tensile deformation, which is of significance for durable repair, rehabilitation and retrofitting strategies. The largest source of damage in cement-based composites like steel Reinforced Concrete (RC) may be attributed to moisture, gas and chlorides ingress, whereby steel reinforcement is subjected to degradation processes. Crack width limitation or control is a well established durability concept in RC design. In RC crack width limitation is generally obtained through steel detailing. In SHCC this is an inherent material property. This is shown for SHCC on the material level and on the structural scale, as well as for steel reinforced SHCC elements (R/SHCC).

INTRODUCTION

High Performance Fibre-Reinforced Cement-based Composites (HPFRCC) are actively researched and developed internationally. In addition to enhanced structural requirements, improved durability of cement-based construction materials has been a continuous striving. This has lead to the development of various classes of HPFRCC. It has become possible to design HPFRCC to have either extreme tensile ductility, with moderate strength, or high tensile and especially flexural strength, but limited tensile deformability. Fibrereinforced Strain-Hardening Cement Composites (SHCC) do not significantly increase the matrix tensile strength, but are designed for increased tensile strength beyond first cracking and large tensile ductility. SHCC contain low to moderate amounts of fibres (1% ≤ Vf ≤ 3%). In contrast, Ultra-High Performance Fibre-Reinforced Concretes (UHPFRC) are designed to have high tensile and flexural strength, as well as extremely high compressive strength, but reach these strengths at moderate strain levels. Durability considerations are increasingly important. It is generally agreed that gas and moisture permeability is an important measure of concrete durability, as they may be media for ingress of deleterious materials which may lead to degradation of the cement-based material or steel reinforcement. In the near surface zone capillary sorption dominates moisture intake while moisture diffusion governs the longer term

migration of water in the material through the micropores. By matrix densification the capillary absorption is significantly reduced. In SHCC the principle is rather reduced permeability in the service condition, by inherent crack control. 2

CRACK CONTROL FOR DURABILITY

Crack width limitation is a well established concept in Reinforced Concrete (RC) design. The pseudo strain-hardening of SHCC involves the formation of multiple cracks, as shown in Figure 1. Crack widths are arrested and new cracks arise at increased deformation. Resistance to gas and moisture ingress and migration in cement-based composites is increased by matrix densification. However, upon the formation of cracks, permeability

4

3 Stress (MPa)

1

2

1

0 0

Figure 1.

115

1

2

3

4 Strain (%)

5

6

7

8

Direct tensile stress-strain response of SHCC.

is increased. A significantly increased coefficient of permeability is found beyond a crack width threshold of roughly 0,1 mm in cement composites (Wang et al. 1997, Lepech & Li 2005). This means that, by inherent crack-control to widths below this threshold, SHCC has a low permeability or high resistance to these degradation mechanisms. Increased crack width is related to higher chloride penetration rate in cement composites. The effective chloride diffusion coefficient is strongly dependent on crack width. Beyond the threshold crack width, this diffusion coefficient increases by several orders of magnitude. However, the crack width in SHCC is insensitive to the deformation level. This explains the diversion of chloride diffusivity of mortar specimens from that of SHCC with increased deformation level (Sahmaran et al. 2007). The controlled, small crack widths and associated resistance to water and chloride diffusion leads to significantly reduced corrosion rates in R/SHCC compared with R/C (Miyazato & Hiraishi 2005). 3

CRACK CONTROL IN SHCC

To exploit the inherent crack control as fundamental mechanism of structural durability, it is imperative that all modes of mechanical and environmental resistance are studied to ensure that crack control is maintained. Through balanced properties of the cement matrix, fibres and their interfaces, based on consideration of micro-mechanical mechanisms (Li et al. 1995), pseudo strain-hardening tensile behaviour can be achieved. Crack widths remain small (40–60 µm) throughout the pseudo strain-hardening part, ended by widening and eventual localisation at a single crack. This is within the threshold crack width for low diffusivity. Crack control is retained in flexure. The typical load-deflection response suggests that the multiple cracking phase may be utilized in structural design for the service condition (up to point II) in Figure 2. There is evidence that the multiple cracking characteristic of SHCC is retained in shear. It is postulated

III II

Py

II

Plastic hinge

Multiple cracking

Pe

I

III

LOADING HISTORY DEPENDENCE

If structures of (R/)SHCC are designed to operate in the multiple cracking region under service actions, it must be ensured that crack width control is retained under all loading histories, including cyclic loading, loading at varying rates and sustained load. Research results from monotonic tensile tests, as well as cyclic tensile tests of SHCC specimens under deformation control, as well as under force control (Yun & Mechtcherine 2007) indicate that the number of cracks that arise during the multiple cracking phase is relatively insensitive to these different loading histories. The total deformability is virtually unchanged, which indicates that crack widths may be insensitive to the loading history as well. From SHCC tensile creep test results of Boshoff & van Zijl (2007), it appears that a smaller number of cracks arise in specimens which have reached a particular tensile deformation level in creep, compared with tensile specimens loaded monotonically to the same level of tensile deformation. Although new cracks do arise during the sustained load phase, widening of existing cracks appear to contribute to the time-dependent increase in deformation. In applications of SHCC with high sustained loads, such timedependent widening must be considered for sustained durability. This is a current research focus. 5

CONCLUSIONS

Localisation

ACKNOWLEDGEMENTS

I

∆

∆e

4

Evidence of the inherent crack control of SHCC under various loading conditions and histories has been presented. Such crack control causes increased resistance to water and chloride diffusion in SHCC. Hereby, SHCC has significant potential as a repair material and general construction material which affords structural durability in service conditions, which inevitably include cracks.

P Pu

that the tensile ductility enables micro-crack arrest and apparent crack rotation. The behaviour of shear dominated beams of R/SHCC, and the resistance to combined bending and shear require further investigation to ascertain crack control under service conditions. Recent test results on SHCC, R/SHCC and RC beams indicate crack width control retention until close to peak flexural-shear response in R/SHCC and RC specimens.

∆y

∆u

Figure 2. Schematic representation of (R/)SHCC beam load-deflection phases.

The support of the South African Ministry of Trade and Industry through the Technology and Human Resources for Industry Programme (THRIP), as well as the industrial partners of the THRIP project SAPERCS is gratefully acknowledged.

116

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

A novel durability design approach for new cementitious materials: Modelling chloride ingress in strain-hardening cement-based composites F. Altmann & V. Mechtcherine Institute of Construction Materials, TU Dresden, Germany

U. Reuter Department of Civil Engineering, TU Dresden, Germany Formerly: Institute of Statics and Dynamics of Structures, TU Dresden, Germany

ABSTRACT: To fully utilise the advantageous durability properties of new cementitious materials such as fibre-reinforced Strain-Hardening Cement-based Composites (SHCC), a performance-based durability design approach is required. For crack-free ordinary concrete first promising concepts exist, e.g. the DuraCrete model. However, the parameters of these models, which were quantified for ordinary concrete, cannot be applied to SHCC due to its significantly different composition. As multiple cracking with limited crack width under tensile load is a characteristic material property of SHCC, any durability model for this material furthermore has to consider the cracked state. In this paper a novel durability design approach is presented exemplary for chloride ingress in SHCC. The developed approach takes into account the limited availability of relevant data for new materials, which prevents a reliable stochastic quantification of model parameters. The DuraCrete ingress model was adapted for SHCC with the help of fuzzy-probability theory, which allows the quantification of parameters using expert knowledge even if the data basis is very limited. 1

INTRODUCTION

3

SHCC has promising durability properties due to the formation of multiple, well-distributed cracks under tensile load. To design efficient and durable SHCC structures, a performance-based durability design concept similar to the probabilistic DuraCrete model for crack-free ordinary concrete is required. Such a design approach can be developed for SHCC by expanding the DuraCrete model to cover the cracked state. However, the composition of SHCC is very different from that of ordinary concrete and due to a lack of data the model parameters cannot be quantified stochastically for SHCC. Using fuzzy probability theory, it is possible to quantify parameters based on limited information and expert knowledge and to develop a fuzzy-probabilistic chloride ingress model for SHCC with the available information.

FUZZY-PROBABILISTIC MODEL FOR CHLORIDE INGRESS IN SHCC

Using tildes to denote fuzzy and fuzzy-stochastic variables, the DuraCrete model for chloride ingress can be expressed according to Equation 1: ⎛ ⎞ x C
SHCC ( x, t) = C
saSHCC erfc ⎜ ⎟ SHCC ⎝ 2 D
a t ⎠

where C˜SHCC(x, t) = the total chloride concentration in the concrete, C˜saSHCC = the (constant) surface chloride concentration, x = the distance from the concrete sur~ (m) µm 1

0

2

FUZZY-PROBABILITY THEORY Figure 1.

If insufficient data for a stochastic parameter description is available, this lack of knowledge may be interpreted as fuzzy uncertainty. Fuzziness can be quantified with a membership function µ (see Figure 1). Random and fuzzy uncertainties exist simultaneously, resulting in fuzzy-randomly distributed parameters. Fuzzy random parameters may be quantified by defining the variables of their probability functions as fuzzy variables. Figure 2 shows a fuzzy probability distribution function for a normally distributed parameter with a fuzzy mean value.

(1)

~ m

a

b

c

m

d

Membership-function of fuzzy mean value.

~ F(x) 1

~ F(x)

µ=1 µ=0

0.5

0 a

Figure 2.

117

b c

d

Fuzzy probability distribution function.

x

⎡ t ′ex ⎞ 1 ⎢⎛ D
aSHCC = D
0SHCC ⎜ 1+ ⎟ t ⎟⎠ 1− n
⎢ ⎜⎝ ⎣

1− n


⎛ t′ ⎞ − ⎜ ex ⎟ ⎜⎝ t ⎟⎠

1− n


⎤ ⎛ ⎞ ⎥ ⎜ t ′0 ⎟ ⎥ ⎜⎝ t ⎟⎠ ⎦

n


(2)

where D˜0SHCC = the actual diffusion coefficient under field conditions at the time t'0, t = duration of exposure, t'0 = reference time, t'ex = concrete age at first exposure, n˜ = age factor. Introducing a damage factor k˜cr to account for the influence of cracks, it is D


SHCC 0

= k
e k
t k
c k
cr D


SHCC ex ,0

experimental conditions were modelled, the chloride surface concentration C˜sa and the parameters k˜e and k˜c were defined as crisp values. The test method factor k˜t was defined as a purely fuzzy variable as insufficient information for a fuzzy-probabilistic definition was available. The results of the subsequent analysis are presented in Figures 3 and 4, with Figure 3 showing the influence of time on uncertainty. As can be seen in Figure 4 the non-stochastic uncertainty for quantiles relevant for design is much more significant than for the mean value. This highlights the importance of a thorough definition of membership functions. Cl- concentration [mass-% of binder] of binder] Cl - [mass-%

face, D˜aSHCC = the apparent diffusion coefficient, t = the time and erfc is the complement to the error function. ˜ SHCC Instead of the time-dependant function for D a prescribed in the DuraCrete model, a time-integrated average value is used:

(3)

with k˜e = environment parameter, k˜t = test method parameter, k˜c = curing parameter according to the DuraCrete model. D˜ex,0SHCC = the experimentally determined diffusion coefficient at the time t'0.

µ (100 years)

0.5 µ (1 year) 0 02550 0

In Table 1 the parameters of the proposed model are quantified exemplarily to demonstrate the possibilities of the model. These values do not represent a reliable parameter quantification. For the mean value of the diffusion coefficient in crack-free SHCC a triangular membership function was determined based on experimental results for two specimens. Its maximum µ(D) = 1.0 was defined at the average experimental diffusion coefficient. To estimate the upper boundary of the function, the regression line for the diffusion coefficient as a function of the number of cracks was used, which yielded a value of 7.75 × 10−12 m²/s for crack-free SHCC. The lower boundary was defined to have the same distance from the maximum. The same distances between the boundaries and the maximum of the membership function were used for cracked SHCC. The age factor n˜ was determined similarly. As well-controlled Table 1.

µ=1 µ=0

1

EXAMPLE

25

1.5

µ=1 µ=0

1

µ (mean)

0.5

µ (5%-quantile) 0 0

25

xDepth [mm] x [mm]

50

Figure 4. Fuzzy mean value and 5%-quantile of chloride profile after 10 years for cracked SHCC (µ = 1 and µ = 0).

Fuzzy-stochastic parameters for chloride ingress in SHCC. Mean

Parameter SHCC ˜ * D ex,uncr,0

Dimension *

–12

D˜ex,cr,0SHCC *

[m²/s 10 ] Random [m²/s* 10–12]

n˜ uncr

[–]

n˜ cr

[–]

k˜ t C˜sa ke kc

[–] [wt.-% binder] [–] [–]

*

xDepth [mm] x [mm]

50

Figure 3. Fuzzy mean value of chloride profile for cracked SHCC after one and 100 year (µ = 1 and µ = 0). Cl- concentration [mass-% of binder] of binder] Cl- [mass-%

4

1.5

Type of Uncertainty Fuzzy Distribution Fuzzy Random Fuzzy Random Fuzzy Random Fuzzy Crisp Crisp Crisp

Distribution Normal Normal Distribution Beta, a = 0, b=1 Beta, a = 0, b=1 – – – –

D˜ex,uncr,0SHCC = k˜cr × D˜ex,0SHCC F Distribution or b = c the membership function in Figure 1 is triangular.

**

118

a

Standard Deviation b = c**

d

a

b = c**

5.75

6.75

7.75

56.22

57.22

58.22

0.65

0.69

0.80

0.05

0.06

0.07

0.69

0.75

0.85

0.05

0.06

0.07

0.58

0.68 1.44 1 1

– – –

0.75 – – –

1.35 11.2

– – – –

1.92

d

16.0

– – – –

3.85 30.0

– – – –

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

A two component bacteria-based self-healing concrete Henk M. Jonkers & Erik Schlangen Delft University of Technology, Faculty of Civil Engineering and GeoSciences/Microlab, Delft, The Netherlands

ABSTRACT: The aim of this research project is the development of a new type of concrete in which integrated bacteria promote self-healing of cracks. Traditional concrete does usually show some self-healing capacity what is due to excess non-hydrated cement particles present in the material matrix. These particles can undergo secondary hydration by crack ingress water resulting in formation of fresh hydration products which can seal or heal smaller cracks. However, the integration of excess cement in concrete is unwanted from both an economical and environmental viewpoint. Cement is expensive and, moreover, its production contributes significantly to global atmospheric CO2 emissions. In this study we developed a two-component self-healing system which is composed of bacteria which catalyze the metabolic conversion of organic compounds (the second component) to calcite. Both components are mixed with the fresh cement paste, thus becoming an integral part of the concrete. Experimental results show that ingress water channeled through freshly formed cracks activate present bacteria which through metabolic conversion of organic mineral-precursor compounds produce copious amounts of calcite. This new bio-based two-component system may represent a new class of self-healing mechanisms which can be applied to cement-based systems. The self-healing capacity of this system is currently being quantified what should result in an estimate of the materials durability increase. A self-healing concrete may be beneficial for both economical and environmental reasons. The bacteria based concrete proposed here could substantially reduce maintenance, repair and premature structure degradation what not only saves money but also reduces atmospheric CO2 emissions considerably as less cement is needed for this type of self-healing concrete.

1

INTRODUCTION

Autogenous crack-healing capacity of concrete has been recognized in several recent studies (Neville 2002; Reinhardt & Jooss 2003; Li & Yang 2007; Edvardsen 1999). The main goal of the present research was to develop a type of sustainable self-healing concrete using a sustainable self-healing agent, i.e. limestoneproducing bacteria. Although bacteria, and particularly acid-producing bacteria, have been traditionally considered as harmful organisms for concrete, recent research has shown that specific species such as ureolytic and other bacteria can actually be useful as a tool to repair cracks or clean the surface of concrete (de Graef et al. 2005; de Muynck et al. 2005). In this research project we investigated the bio-mineral producing capacity of two systems. In system one only bacteria were incorporated in the cement stone matrix, while the biomineral precursor compound was provided externally. The second system comprised the cement stone matrix in which both components of the healing agent were incorporated.

2

MATERIALS AND METHODS

Spore-forming alkali-resistant bacteria were isolated from sediment originating from a natural soda lake (strain C2-C2-1 A) as well as obtained from the German Collection of Microorganisms and Cell Cultures (DSMZ), Braunschweig, Germany (strain Bacillus pseudofirmus DSM 8715).

Three different organic compounds, peptone, calcium lactate and calcium glutamate, which could be used by both bacterial strains as energy and carbon sources for grows, were tested for their potential applicability as bio-mineral precursor compound in concrete. To investigate bio-mineral producing potential of applied healing agents, test specimen comprising one healing agent component (bacteria only; strain C2-C2-1 A), both healing agent components (bacteria plus calcium lactate) or no healing agents (controls) were prepared. Specimen fragments were directly studied for mineral formation at crack surfaces by environmental scanning electron microscopy (Philips ESEM XL30 Series) equipped with an X-ray microanalysis (EDAX) system.

3 3.1

RESULTS Effect of healing agent additions on specimen strength

Incorporation of a high number of bacteria (5.8 × 108 cm−3 cement stone) appeared to have a mildly negative effect on compressive strength development as bacterial test specimen appeared almost 10% weaker then control specimen at all tested times. Effect of organic compound incorporation on development of strength appeared however strongly dependent on compound identity. Peptone and calcium glutamate additions resulted in significantly lower compressive strength values. However, calcium lactate did not significantly affect strength development as values were only slightly lower or higher then control specimen.

119

3.2

Mineral formation on specimen crack surface

Specimen with only incorporated bacteria but incubated in sodium glutamate plus yeast extract amended medium appeared to produce copious amounts of precipitates which formed plate-like aggregates on specimen crack surfaces (Figure 1). Major element composition of a control specimen incubated in methanol instead of water to inhibit further hydration or carbonation (Figure 2) is for comparison also given in Table 1.

4

DISCUSSION

Self-healing concrete should be able to heal or seal, by filler material formation, freshly formed cracks to inhibit ingress of water and other chemicals which could cause preliminary degradation of the material matrix or embedded reinforcement. In this study we investigated the bio-mineral production capacity of cement stone specimen in which bacteria were incorporated as healing agent. In practice, an integrated two-component healing system such as bacteria plus calcium lactate may not always be needed. Particularly concrete in wet environments such as marine or subsurface structures may be exposed to present bacteria and/or organic compounds dissolved in ingress water. However, as environmental

Figure 1. Plate-like aggregate formation on the crack surface of cement stone specimen containing 5.8 × 108 cm−3 bacteria of strain C2-C2-1 A; incubated in glutamate and yeast extract amended medium. ESEM photograph 100x magnification. Table 1. X-ray micro analysis of specimen surface. Elements expressed as percentage of atom number in sample.

Element: Calcium Silicon Carbon Ca/C Ca/Si

Control/ methanol

Control/ medium

C2-C2-1 A/ medium (Fig. 1A)

15.2 5.5 8.7 1.8 2.8

13.2 0.5 27.9 0.5 24.4

18.2 0.4 27.0 0.7 41.3

Figure 2. Reference cement stone specimen (control) kept in methanol. ESEM photograph 100x magnification.

conditions vary widely, types of locally present bacteria and organic compounds may not result in formation of crack-filling products, but instead exert a negative influence on the concrete matrix or embedded steel reinforcement. Particularly acid-producing bacteria such as sulfur- or ammonium-oxidizing strains are likely harmful as their metabolic waste products etch and dissolve the concrete matrix. To act as a fully self-healing mechanisms, however, both bacteria and filler precursor compounds need to be integrated in the material matrix. Incorporating high numbers of appropriate bacteria and filler material precursor compounds in the material already during construction lowers the risk of possible negative effects of harmful bacteria intruding the concrete matrix with ingress water at a later stage. As this practice seems straightforward for the incorporation of spore-forming bacteria it appears from this study that this may not apply for mineral precursor compounds. Here we have shown that incorporated bacteria can produce copious amounts of calcium carbonate enriched precipitates from sodium glutamate present in ingress water. However, substantially less mineral production was observed when both healing components, bacteria and calcium lactate, were incorporated. A reason for this phenomenon may be that calcium lactate added to the cement paste mixture may have become completely integrated in the material matrix and thus not accessible for bacterial conversion later on. Apparently only a minor amount of incorporated calcium lactate was present in the matrix capillary pore system resulting in only insignificant mineral formation. A potential solution to this problem could be encapsulation of the healing agent before addition to the cement paste mixture, a mechanism analogous to what was proposed by White et al. (2001) for self-healing in polymer composites. This would ensure the presence of a substantial reservoir in the material matrix containing original chemical healing agent which is also accessible for bacterial metabolic conversion. The conclusion of this study is that a bacteria-based self-healing system in concrete appears promising. However, before a functional two-component system needed for truly autogenous healing can be applied, further technical developments are required.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Self healing properties with various crack widths under continuous water leakage A. Hosoda & S. Komatsu Yokohama National University, Kanagawa, Japan

T. Ahn & T. Kishi Institute of Industrial Science, University of Tokyo, Japan

S. Ikeno & K. Kobayashi East Japan Railway Company, Japan

ABSTRACT: Crack healing properties of 3 mix proportions under continuous water leakage were investigated by observation with microscope and water permeability test. Self healing concrete under continuous water leakage showed better healing performances than cured in still water. An improved type self healing concrete showed remarkable decrease of crack width with water passing through crack all the time.

1

INTRODUCTION

Several researches have been reported about self healing concrete containing admixtures such as carbonate compounds or expansive additive whose water to binder ratio was around 25–60% (Hosoda, 2007). In all these researches, specimens were cured in still water after cracking. No research has been conducted to verify healing performances of self healing concrete with water passing through the crack. In this study, self healing performances of 3 mix proportions including the one which showed very good performance in our past research are verified under continuous water leakage. Crack width is set in 3 levels, such as 0.1 mm, 0.2 mm, and 0.4 mm. Half of the specimens are cured in still water, and the others are subjected to water head and continuous water leakage just after cracking. Healing performances are verified by the observation with a microscope and by water permeability test. 2

EXPERIMENTAL PROCEDURES

As a resource of Calcium ion for self healing, CalciumSulfo-Aluminate (Ettringite) type expansive additive was used. As other special additives for self healing, a cement crystal producing material developed by the third and the fourth authors and a commercial cement crystal breeding material which had been used in the previous study by the authors(Hosoda, 2007) were used. No.1 mix proportion is the control specimen, which includes no special additive. No.2 mix proportion (JR-mix, East Japan Railway Company mix) is a self healing concrete with low heat Portland cement

which showed good healing performance in the previous research. No.3 mix proportion (UT-mix, University of Tokyo mix) is an improved type self healing concrete which has been developed by the third and the fourth author and has exhibited good self healing performance in still water. 3

RESULTS AND DISCUSSION

Figure 1 to Figure 2 show the change of crack width of UT-mix. The improved mix (UT-mix) showed the largest healing. Better healing was observed under continuous water leakage condition. Between two planes of the specimen, on the upside plane, better healing was observed especially in the case of large crack width of 0.4 mm. Healing of UT-mix observed by microscope was shown in Figures 5 and 6. Between crack planes, brown looking products were observed. The healing mechanism of UT-mix contains the precipitation of CaCO3. Therefore in the case of the specimens cured in still water, due to high pH of surrounding water, the precipitation of CaCO3 might have been restrained compared to the case in the best condition. On the other hand, in the case of the specimens with continuous water leakage, pH of flowing water in the crack was nearly 7, which might have contributed to better precipitation of CaCO3. Changes of water flow rate were shown in Figures 3 and 4. Though the initial water flow rates varied widely, they became larger with the increase of crack width. These initial water flow rates were much larger than those calculated based on Edvardsen’s results (Edvardsen, 1999) which corresponded to limiting water flow rate for self healing.

121

Change of Crack Width (initial: 0.1mm) with Water Flow (Up)

with Water Flow (Down)

Still Water (Up)

Still Water (Down)

Crack Width (mm)

0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0

Figure 1.

14 28 Time after Cracking (days)

Crack width of UT-mix (0.1 mm). Change of Crack Width (initial: 0.4mm) with Water Flow (Up)

with Water Flow (Down)

Still Water (Up)

Still Water (Down)

Figure 5.

Crack before healing (UT-mix, 0.4 mm).

Figure 6.

Crack after healing (UT-mix, 0.4 mm).

Crack Width (mm)

0.5 0.4 0.3 0.2 0.1 0 0

Figure 2.

14 28 Time after Cracking (days)

Crack width of UT-mix (0.4 mm).

Water Flow (mm/s)

Still Water, 0.1 mm Still Water, 0.4 mm With Water Flow, 0.2 mm 450 400 350 300 250 200 150 100 50 0 0

Still Water, 0.2 mm with Water Flow, 0.1 mm with Water Flow, 0.4 mm

For all mix proportions, the decrease of water flow rate was larger in the case of the specimens with continuous water leakage than in the case of still water curing. 7

14 21 Time after Cracking (days)

28

Figure 3. Water flow of normal concrete.

Water Flow (mm/s)

Still Water, 0.1 mm Still Water, 0.4 mm With Water Flow, 0.2 mm 500 450 400 350 300 250 200 150 100 50 0 0

Still Water, 0.2 mm with Water Flow, 0.1 mm with Water Flow, 0.4 mm

4

Self healing concrete under continuous water leakage showed better healing performances than cured in still water. The improved type self healing concrete showed remarkable decrease of crack width with water passing through crack all the time. 5

7

14 21 Time after Cracking (days)

Figure 4. Water flow of UT-mix.

28

CONCLUSIONS

REFERENCES

Edvardsen, C. 1999. Water Permeability and Autogenous Healing of Cracks in Concrete. ACI Material Journal/ July–August, pp.448–454. Hosoda, A., Kishi, T., Arita, H., and Takakuwa, Y. 2007. Self healing of crack and water permeability of expansive concrete, First International Conference on Self Healing Materials, Noordwijk, The Netherlands. (CD-ROM).

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Using natural wood fibers to self heal concrete M.R. de Rooij, S. Qian, H. Liu, W.F. Gard & J.W.G. van de Kuilen Faculty of Civil Engineering and Geosciences, Delft University of Technology, The Netherlands

ABSTRACT: Changing from damage prevention to damage management, thereby allowing some cracks in a structure as long as the cracks self-repair over time, is the basic concept behind self-healing materials. Following nature, where wood fibers allow for both transport and bonding possibilities, this paper describes various options to apply wood fibers. Preliminary results on how to obtain single wood fibers and initial experiments on concrete crack width are presented. It is shown that the boundary conditions for a successful application of wood fibers as self-healing carriers can be met. However, it is also shown that many very practical obstacles in the manufacturing of self healing concrete still need to be cleared.

1 1.1

CONCEPT INSTRUCTIONS The concepts of self-healing

All strategies to improve the strength and reliability of materials developed over the past 20 centuries are ultimately based on the paradigm of damage prevention, i.e. the materials are designed and prepared in such a way that the formation and extension of damage as a function of load and/or time is postponed as much as possible (Van der Zwaag, 2007). Damage is defined here as the presence of micro- or macroscopic cracks not being present initially. However, in recent years it has been realized that an alternative strategy can be followed to make materials effectively stronger and more reliable through damage management, i.e. materials have a built in capability to repair the damage incurred during use. Cracks are allowed to form, but the material itself is capable of repairing the crack and restoring the functionality of the material. The material is self-healing. It is fair to say that concrete-like materials have a good track record for self-healing to start from. This paper is focusing on the transport properties of wood fibers. The paper describes various options to choose wood fibers, preliminary results on how to obtain single wood fibers and initial experiments on crack width. First however, some boundary conditions for concrete are discussed based on earlier research work. 2 2.1

EARLIER RESEARCH WORK Healing of early age cracks

The first research work on concrete material that was related to self-healing materials at Delft University of Technology, consisted of crack healing of early age cracks in concrete (Ter Heide, 2005; 2007). In this study, three point bending tests were performed on prismatic concrete specimens to fracture the specimens. These tests were performed between 20 to 72

hours after casting, producing crack (mouth) openings of the crack between 20–150 µm. After cracking the samples were stored in different relative humidity (RH) environments. After two weeks the samples were taken from their different conditions and tested again under three point bending. From this research, the following conclusions could be drawn: − Cracks do heal under the conditions that the cracks are made at an early age and the cracks are closed again (a compressive stress is applied) and the specimens are stored under water. − With increasing age of the specimen at the moment of cracking, a decrease in strength recovery is found. This seems to be related to the degree of hydration at the age of the specimen when the first crack is made. From this research it became clear that self-healing is possible, but the crack faces should be close together. In other words, the crack width should not be too large. 2.2

Keeping crack width small

The group of prof. Li (Li, 2003) has developed Engineered Cementitious Composites (ECC). When an ECC structural element is loaded (flexure or shear) to beyond the elastic range, the inelastic deformation is associated with microcracking. The microcrack width is dependent on the type of fiber and interface properties. However, it is generally less than 100 micron when PVA fiber is used.

3 3.1

CONCRETE AND WOOD Choice of wood fibers

Building form the knowledge of ECC material some ideal desirable fiber properties could be formulated

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to keep crack width small enough for self-healing to have a chance. Ideally the fiber should have a length around 8–12 mm with good processibility and adequate bonding capacities. The diameter should ideally be less than 100 µm. For aspect ratio we would be looking for something like 100 ∼ 200. Taking a first inventory on the physical and mechanical properties of a wide variety of fibers resulted in the overview presented in Table 1. For comparison also some synthetic fiber values are mentioned. It was concluded that the diameter of single fibers would not be a problem in theory to meet our ideal desirable value. It actually might even be a little on the small size. For the length the variety of values is rather wide, which means that the check whether or not our desired value would be obtained, depends on the actual choice of material. 3.2

Parallel preliminary experiments

While searching for the wood fiber type to be used in future research, preliminary experiments on producing concrete with natural fibers were under-

taken in a parallel research program. The first tests described here were set up with sisal fiber, because that was readily available in large enough quantities to produce mortar specimens. These tests should give insight in the workability problems that could occur, as well as distribution of fibers through the mortar. Furthermore, it would also give a first idea on crack distribution and crack width. Results showed that the number of cracks ranged from 3 to 7, with an average crack width ranging from 120 µm to 70–80 µm. Most noticeable was that the fibers were well distributed, but that they had not been broken. 3.3

As a future outlook on this very early research path, the next mark is getting extracted fibers from Oregon pine filled with water and then sealed with a coating to keep the water inside. Water is used as filling material as a first basis.

4 Table 1. Overview of first inventory on physical and mechanical properties of possible fibers.

Fibers Synthetic PE PVA PP Stem fibers Flax Hemp Jute Ramie Hibiscus Sugarcane Bamboo Deciduous wood Coniferous wood Seed-hair fibers Cotton Coir Leaf fibers Sisal Banana

Length mm

Diameter µm

Tensile strength MPa

12.7 8–12 6

38 39 12

2700 1620 770–880

27.4–36.1 8.3–14.1 1.9–3.2 60–250 160–1500 0.8–2.8 2.8 0.3–2.5 1.0–9.0

17.8–21.6 17.0–22.8 15.9–20.7 28.1–35.0 40–350 6.6–26 10–40 10–45 15–60

500–900 310–750 250–350 870

12–65 0.9–1.2

12–20 16.2–19.5

300–600 130–175

1.8–3.1 2.2–5.5

18.3–23.7 18–30

250–550 530–750

170–290 350–500 700

Future steps: filling fibers

CONCLUSIONS

In this article a very first inside look has been presented on the possible concepts of using natural fibers to self-heal concrete. Preliminary literature research has shown that the boundary conditions for a successful application of wood fibers in concrete can be met. However, it has also been observed that many and also very practical obstacles in the research still have to be cleared, like: − obtaining (bundles of) fibers in enough quantity to perform the research, − making a mixture composition in which the cracks remain small enough for self-healing to have a chance, − producing fibers that break even with these very small crack widths, − producing a coating that keeps the repair agent in the fiber during mixing of the concrete and during undamaged service life of the concrete, − producing a coating that is not so strong, that it will prevent the fiber to break. As we know it is very well possible to manipulate the properties of the natural fibers as much as it is possible to tune the properties of the future coating on the fiber, we are looking forward to the challenging research in the new field of self-healing materials.

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The effect of geo-materials on the autogenous healing behavior of cracked concrete T.H. Ahn & T. Kishi Department of Human and Social Systems, Institute of Industrial Science, The University of Tokyo, Japan

ABSTRACT: In this research, the self-healing phenomenon of cracked concrete incorporating geo-materials as a partial cement replacement was investigated in terms of recrystallization in the cracked portion by geopolymeric gel. Morphology and the shape and size of precipitated particles in the cracks were examined by microscopy and SEM-EDS. In order to apply this concept to the field, a self-healing concrete, designed by the author, was fabricated by ready-mixed car in a ready-mixed concrete factory, then used for the construction of artificial water-retaining structures. The results show that cracked cementitious composite concrete incorporating geomaterials exhibit much higher self-healing behavior than cracked normal concrete when cured in water after cracking. Geo-materials appear to be effective in inducing the aluminosilicate products in the cracks due to the diffusion and increase of Al and Si ions. From these results, it is considered that the utilization of appropriate dosages of geo-materials is desirable for autogenous healing of cracked concrete.

1

INTRODUCTION

The serviceability and durability of concrete structures has been extensively studied by various researchers. The serviceability limit of concrete structures is primarily governed by the extent of damage. Cracks, one of various types of damage, play an important role in the serviceability limit. However, if it were possible to know the reason for differing behavior of concrete structures exposed to largely similar conditions, we might have the key for designing high-durability structures with low or negligible maintenance and repair costs. Furthermore, the serviceability limit of concrete structures by cracking might be overcome by crack control methodologies; this enhanced service life of concrete structures would reduce the demand for crack maintenance and repair. The aim of this study is to develop autogenous healing concrete using geo-materials for practical industrial application. Research has been done on the healing of cracks in aged concrete. This study has focused on two primary issues: (1) experimental and analytical design of cementitious materials with self-healing capabilities, (2) development of a self-healing concrete using new cementitious materials at normal water/binder ratio [over W/B = 0.45]

2 2.1

EXPERIMENTAL Materials

(b) Geo-materials & Chemical agents In order to compare the fluidity and compressive strength of cementitious composite with various compositions, mineral admixtures such as expansive agent, geo-materials and chemical agents were used. The expansive agent, two geo-materials (A, B), chemical agent and various superplasticizers used were commercial products produced in Japan. 3 3.1

Self-healing capability of cementitious composite materials

Figure 1 shows the healing process of the cracked three-component system paste [OPC90% + CSA5% + Geo-materials5%] under water supply. In this case, the crack with an initial width of 0.2 mm was almost healed after 28 days. Rehydration products between cracks were clearly observed after 14 days, and the self-healed perfectly even though there were small cracks between the rehydration products after 200 days, as shown in Figure 1 (f). Figure 2 shows the re-hydration products on the surface of the specimen between the original and self-healing zones. Figure 2 (b) shows the X-ray map and spectra taken from rehydration products. It was found that the re-hydration products were mainly composed of high alumina silicate materials as shown in the X-ray mapping results. 3.2

(a) Cement Type I Japan Portland cement was used in all cementitious composite and concrete mixtures.

RESULTS AND DISCUSSION

Self-healing capability of self-healing concrete

In case of self-healing concrete, the crack was significantly self-healed up to 28 days re-curing. Figure 3

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Figure 3. Process of self-healing on self-healing concrete at water/binder ratio of 0.47

Figure 1. Process of Self-healing on [OPC90% + CSA5% + Geo-materials5%] pastes at normal water/binder ratio of 0.45 (the three-component system)

shows the healing process of cracked self-healing concretes made at the same mixing condition as the conventional concrete. A 0.15 millimeter crack was self-healed after 3 days re-curing, as shown Figure 12 (b). After re-curing for 7 days, the crack width decreased from 0.22 millimeters to 0.16 millimeters. Furthermore, it was almost completely self-healed at 33 days as shown in Figure 3 (d).

4

CONCLUSIONS

In this study, the self-healing properties of concrete using geo-materials were investigated. 1. Self-healing capability was significantly affected by aluminosilicate materials and various modified calcium composite materials. 2. The essential properties such as swelling and recrystallization of geo-materials with pozzolanic reaction for self-healing were analyzed and discussed.

Figure 2. Self-healing phenomenon of crack by crystallization of aluminosilicate phases on the cementitious pastes incorporating CSA and Geo-materials [Surface Analysis of specimen]

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Surface protections to prevent Alkali-Aggregate Reactions (AAR) in concrete structures L.F.M. Sanchez University of São Paulo

S.C. Kuperman University of São Paulo and DESEK

P.R.L. Helene University of São Paulo

ABSTRACT: It is very difficult and expensive to repair or rehabilitate coencrete structures suffering from AAR, and for these cases, a preventive solution is desirable, both technically and economically. However, when a reactive aggregate is used to produce a concrete and its deleterious effects are discovered after its placement, some care should be taken. Until now, there is no consensus on how to repair structures suffering from AAR; however, several types of repairs as surface protection can be used. Some researchers have been trying to study surface protections to inhibit the surface moisture and the water entrance. Their thought is that the expansion due to AAR can eventually be stopped and the concrete structure can cohexist with this problem without serious effects. Some materials have already been tested as epoxi, latex and acrilic coatings. In addition, some repellents coatings were tested too. This work studies four differents types of coatings that can be used in concrete structures with the attempt to discover if the type of coating is effective or not to prevent AAR. The types of materiais used were a silicone coating, an epoxi coating, a cimentitious coating modified with latex and a flexible poliuretane coating. To perform this study, the Accelerated Concrete Prism Test (ACPT) was carried out. This test method can classify an aggregate as innocuous or reactive in 3 months for ordinary concrete and 6 months for concrete with admixtures. Four reactive aggregates were used in this research (two granites—from São Paulo and Recife, one granite/gneiss—from Recife and one quartzite from São Paulo). The results showed that silicone coating and cimentitious coating modified with latex have a great potential to be used as surface protection against AAR. However, flexible polyurethane has not shown good behavior.

1

2

INTRODUCTION

The alkali-aggregate reaction (AAR) is a chemical reaction between some siliceous or carbonate aggregates and the alkalis that are provided from cement. This chemical reaction produces an alkaline gel that when absorbing water expands and cause deleterious effects to concrete structures. It is known that the best way to avoid AAR is the prevention. However, there are some structures that are suffering from AAR and need to maintain their intended function. For these structures, surface protections can be used with the intention of stopping or decreasing the expansion caused within the concrete. These surface protections work inhibiting the water ingress and then decreasing the rate of expansion of a concrete element. If the concrete element does not have any structural problem, a good surface protection can control the problems caused by AAR.

SURFACE PROTECTIONS

There are many types for surface protections that can be used for concrete structures, among them: silicones, siloxanes and silanes, cementitious coatings modified with latex, polyurethanes coatings and epoxi coatings. To evaluate which type of coating can mitigate AAR, this work compares the behavior of different coatings, studying the expansion vs. time and rate of expansion vs. time of concrete prisms casted with known reactive aggregates. 3

EXPERIMENTAL PROCEDURES

For this research, four Brazilian aggregates that are known to react with alkalies were used (granite from the city of São Paulo, granite from the city of Recife, granite-gneiss from the city of Recife and a quartzite

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from de city of Sao Paulo). Three concrete prisms with each aggregate were casted and a surface protection was applied in two of them (with each aggregate), as follows: • Granite from Sao Paulo—Silicone; • Granite from Recife—Epoxi coating; • Granite–gneiss from Recife—Cimentitious coating modified with látex; • Quartzite from Sao Paulo—Polyurethane coating; The tests were carried out using ACPT that consists of ASTM C 1293 procedures with the specimens subjected to other conditions (60oC and relative humidity of 100% during three months). Expansions were measured every 14 days. The results show that: Granite from Sao Paulo: The unprotected prism had an expansion of 0,055% in 90 days. This expansion demonstrates a deleterious behavior of the aggregate. However, the prisms that were protected with silicone had expansions less than the limit proposed in this test method (0,04%). So, the system (concrete + silicone) had an acceptable behavior. The results show that a surface protection with silicone reduced the expansion by 75% on average. Granite from Recife: The unprotected prism had an expansion of 0,1% at 90 days. This expansion demonstrates high deleterious behavior of the aggregate. In the same way, the prisms that were protected with epoxi coating had expansions greater than the limit proposed in this test method, however, the epoxi coating decreased the expansion in the whole test. The results show that a surface protection with epoxi reduced the expansion by 50% on average. Granite-gneiss from Recife: The unprotected prism had an expansion of 0,35% at 90 days. This expansion demonstrates high deleterious behavior of the aggregate. Prisms that were protected with cimentitious coating modified with latex had expansions a litle greater than the limit proposed in this test method (of 0,04%). Even though the results show that a surface protection with cementitious coating modified with latex reduced the expansion by 82% on average, this fact did not reduce the expansions enough and the behavior (concrete + cimentitious coating modified with latex) cannot be classified as acceptable. Quartzite from Sao Paulo: The unprotected prism had an expansion of 0,29% at 90 days. This expansion demonstrates high deleterious behavior of the aggregate. In the same way, the prisms that were protected with flexible polyurethane had expansions greater than the limit proposed in this test method;

however, this coating decreased the expansion in the whole test. A surface protection with silicone reduced the expansion by 72% on average.

4

CONCLUSIONS

Silicone has a great potential to prevent expansions due to AAR, decreasing the expansions about 75% on average. At the end of the ACPT test the system (concrete with reactive aggregates + silicone) could be classified as acceptable. Epoxi coating presented a good potential to prevent expansions due to AAR, decreasing the expansions about 50% on average. It is known that epoxi is not an elastic coating and if the expansions cause cracks in the surface protection, the coating will fail. However, in the tests that were performed this problem could not be detected. Flexible polyurethane did not demonstrate good behavior to inhibit AAR. Even though the protected prisms have demonstrated less expansion than the unprotected prism, mainly in the end of the test, formation of cracks could be observed in the coating. This formation of cracks increased the rate of expansion of the protected prisms because it allowed the water entrance (increasing the expansions in the end of the ACPT test). Cementitious coating modified with latex has demonstrated a great potential to prevent expansions due to AAR, decreasing the expansions by 82% on average. However, in the end of the ACPT test the system (concrete with reactive aggregates + cementitious coating modified with latex ) could not be classified as acceptable. It is important to notice that the four aggregates that were used do not have the same reactivity. Due to this fact, even though the modified cementitious coating presented the major decrease in expansion, the system concrete + surface protection could not inhibit the expansion due to AAR. In case of this aggregate (granite–gneiss from Recife) it seems that even a good surface protection cannot inhibit the expansion to acceptable degree. Because of the use of different aggregates, it is not possible to say in the present work that one type of coating is better than the other. But it is possible to notice that some types, like silicone and cimentitious coating modified with latex, have a great potential to prevent expansions caused by AAR (depending on the reactivity of the aggregate). On the other hand, coatings like flexible polyurethane did not demonstrate potential to work well against AAR expansions.

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Micro-scale alterations of cementitious surfaces subjected to laser cleaning process and their potential impact on long-term durability A.J. Klemm & P. Sanjeevan Glasgow Caledonian University, Scotland

P. Klemm Technical University of Lodz, Poland

ABSTRACT: The paper presents part of the larger study on laser cleaning process. A special emphasis is placed on the modification of geometrical microstructure and chemical composition of cementitious composites resulting from cleaning process. A wide range of samples characterised with different internal microstructures, surface roughness, and moisture content were subjected to laser cleaning process, and a subsequent assessment. Removal of mortar, cracks and glassy patches were the characteristic features of all laser-cleaned areas. Cracks of wider openings may also form as a consequence of the expansion of water/air inside pores, predominantly smoother/denser mortars. The experimental results prove that the laser cleaning process leads to an increase of Si element concentration and/or decrease in Ca concentration. The observed alterations should be associated with changes in the relative distribution of melted cement paste and aggregate and therefore relative elemental abundance in the near-surface layer.

1

INTRODUCTION

The relationship between laser cleaning processes and substrate parameters is a two-way relationship. The great variation in absorptivity of highly developed surfaces of cementitious materials results in substantial differences in their responses to laser irradiation. The laser cleaning process is unarguably associated with some moderations of mortar’s geometrical microstructure and chemical composition. This study is focused on the effect of laser exposure on the modification of surface geometrical microstructure and chemical composition and the potential implication on long term durability of cleaned surfaces. The cleaning of spray paint from mortar surfaces was done with application of Nd:YAG laser. Effectiveness of cleaning is assessed SEM and EDX techniques.

ness (set C) was subjected to 600 alternated freezing/ thawing cycles (−20ºC to +20ºC). Exposure resulted in increased surface roughness Ra = 15.79–18.13 µm. Surface roughness was measured by a stylus device and represented by the average surface roughness (Ra). Each sample was cleaned by Nd:YAG laser with laser fluence 3.06 J/cm2.

3

SUMMARY

The main challenge in this study was the great variation in absorptivity of highly developed surfaces of cementitious materials, resulting in substantial differences in their responses to laser irradiation. Even though Q switched Nd:YAG laser can be successfully used to remove paint from mortar surfaces, there are Table 1.

2

Mixes and their selected properties.

EXPERIMENTAL DETAILS Moisture

The subjects of this investigation were laser-cleaned mortar samples with different internal microstructures (HP—high porosity; LP—low porosity), surface roughness: A (Ra = 2.28–2.49 µm), B (Ra = 7.70–8.49 µm) and C (Ra = 15.58–17.89 µm), and moisture content (DRY, WET) (Table 1). Mortar specimens had cement/sand ratio of 1:1 by mass and w/c 0.4. HP samples contained air-entraining admixture 1.3% (1/100 kg OPC). Samples were cured in lab conditions for 6 months, painted black with spray paint and laser cleaned after 2 weeks. A separate set of samples of highest rough-

Mix

Compr. Flexural Bulk DRY WET strength strength density Porosity 2 2 g/ml εHg % % N/mm N/mm

LP-A LP-B LP-C HP-A HP-B HP-C FT-Co

79.4 79.4 79.4 57.5 57.5 57.5 76.2

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8.3 8.3 8.3 7.4 7.4 7.4 8.1

2.16 2.16 2.16 1.8 1.8 1.8 2.15

11.9 11.9 11.9 26.4 26.4 26.4 12.4

2.3 2.3 2.3 3.2 3.2 3.2 4.2

8.2 8.2 8.2 10.7 10.7 10.7 10.9

always some residual surface alterations, associated with the removal of mortar itself, formation of cracks and glazing (melted mortar). These effects are generally microscopic and only visible at high magnification. Nevertheless, their importance should not be underestimated, especially on national heritage buildings as they may expose surfaces to further accelerated environmental deterioration. Based on experimental investigations, the following conclusions can be formulated: • The changes of surface roughness, with the number of pulses applied, depend mainly on the initial surface roughness of the mortar. Increased surface roughness can be divided into two or four clearly distinguished zones as shown in Figure 1. In the case of smoother surfaces (Ra = 2.28–2.49 µm and Ra = 7.70–8.49 µm), surface roughness of the mortar increases with the number of laser pulses applied. The characteristic feature in rougher surfaces (Ra = 15.58–17.89 µm) is an increase of roughness at the diminishing rate, followed by a sudden decrease. The average change of surface roughness per pulse is high for rough, highly porous, and/or dry surfaces. The changes in surface roughness are mainly due to removal of mortar. • As initial surface roughness of mortars increases, the alterations in roughness resulting from laser cleaning become more pronounced and the tendency towards crack formation reduces. Whenever change in surface roughness is great, changes in crack density are low. Decrease in surface tensile strength results in the removal of mortar, which reduces crack formation. Laser-cleaned areas proved to be generally denser and more consolidated than the reference surface, due to vitrification of some parts of mortar following the laser cleaning process as shown in Figure 2.

Figure 1. Variations in surface roughness of mortar surfaces with respect to the number of pulses applied.

Figure 2. Laser crater; a) Optical micrographs of LP-A (DRY) after 6 pulses (80x); b) Different zones within the laser crater.

Figure 3. The value of Ca/Si ratio on mortar surfaces.

• Thermal stresses on the cementitious materials, resulting from the application of the laser may lead, in extreme cases, to severe crack formation, particularly around the pit-holes on highly porous surfaces. • Cracks of wider openings may also form as a consequence of the expansion of water/air inside pores. The second mechanism is more prominent on smoother/denser mortars. • Laser-cleaned surfaces exhibit a high concentration of Si element and/or low concentration in Ca element, compared with the original mortar surface. The value of Ca/Si on laser-cleaned areas is around 50% lower than on the reference surfaces Figure 3. This is most probably due to changes in the relative distribution of melted cement paste and aggregate and therefore relative elemental abundance in the nearsurface layer. Lower values of the Ca/Si ratio could also be attributed to the removal of CaCO3 or movement of SiO2 to mortar surfaces. Variations of other elements, such as Al, Fe, and S, seem to be insignificant. Furthermore, higher concentration of Carbon (the main element of paint) in the laser-cleaned area proved that the removal of paint is always associated with some modification of mortars.

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Evaluation of anti-fouling strategies on aerated concrete by means of an accelerated algal growth test W. De Muynck Magnel Laboratory for Concrete Research, Ghent University, Belgium Laboratory for Microbial Ecology and Technology, Ghent University, Gent, Belgium

A. Maury & N. De Belie Magnel Laboratory for Concrete Research, Ghent University, Gent, Belgium

J. De Bock Image Processing and Interpretation Research Group, Ghent University, Gent, Belgium

W. Verstraete Laboratory for Microbial Ecology and Technology, Ghent University, Gent, Belgium

ABSTRACT: This paper describes the effectiveness of two strategies in preventing algal growth on aerated autoclaved concrete. A first strategy comprised the use of water repellents in order to affect the bioreceptivity of the material. In the second strategy, products inhibitory to the growth of algae were applied. For the evaluation of these strategies an accelerated growth test was developed. The experimental setup consisted of several units, which allowed for individual testing of several products at the same time. Qualification of algal growth was obtained with image analysis and spectrometric measurements. Both water repellents and biocidal surface treatments resulted in a delayed fouling rate. Nevertheless, complete coverage of these specimens was also obtained after 1 or 2 additional weeks of accelerated tests.

1

INTRODUCTION

2

The aesthetic quality of outdoor exposed concrete is seriously impaired by the development of biological stains. Biological stains result from the growth of microorganisms among which phototrophic organisms result in the occurrence of the first visible stains. In addition to environmental conditions, the rate of stain development largely depends on the bioreceptivity of the material. For building materials such as stone and concrete, bioreceptivity relates to the surface roughness, moisture content, chemical composition and the structure-texture of the material. Building materials, such as aerated concrete, that have a high roughness and high macroporosity therefore show a high bioreceptivity. Nowadays, different treatments are available on the market for the prevention of biological stains. In this research the effectiveness of two kinds of treatments in preventing algal growth on aerated concrete was investigated. The use of water repellents aimed at reducing the bioreceptivity of the material, while the use of biocides aimed at reducing the biological activity. The qualification of the behaviour of these treatments towards biological growths was done by the use of an accelerated fouling test. The choice of the design was governed by the ease of operation, low costs and possibility to test several biocidal products at the same time.

2.1

MATERIALS AND METHODS Concrete specimens

Aerated concrete specimens were provided by a manufacturer and cut to prisms with the following dimensions: 160 mm × 80 mm × 10 mm. 2.2

Surface treatments

For this research two types of treatments were evaluated (Table 1). A first series of treatments aimed at reducing the availability of water. The second series of treatments aimed at the inhibition of the microbial growth and activity and comprised of metal based biocides. The biocides were applied in a final concentration of the metal of 100 mg/m² (M1) or 250 mg/m² (M2.5). 2.3

Accelerated fouling setup

Accelerated fouling of building materials was achieved by means of a water run-off test. A modular setup was built to allow both simultaneous as separate evaluation of different test series. The setup (Fig. 1) consisted of several stainless steel compartments supported by a wooden frame at 45° inclination. Circulation of the algal cultures was achieved by means of an aquarium pump immersed in the reservoir underneath

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Table 1. Anti-fouling treatments under investigation. Treatment Untreated Water repellents Admixture Surface treatment Water repellent 1 Water repellent 2 Metal based biocides Biocide conc. 1× Biocide conc. 2.5×

Composition

Code

Area coverage (%)

100

* 80

– –

60

Silicone Aluminum stearate Silane—Siloxane

40

– – M1

20

0 0

M2.5

*

Code is abbreviation used in graph legends; (–): code is name itself.

1

2

3

4

5

6

7

8

Untreated

M1

M2.5

Water repellent 2

Untreated 2

Admixture

9

10

11 12 Test weeks

Water repellent 1

Figure 2. Evolution of the area that is covered with algae during the accelerated fouling tests.

For the image analysis, photographs of the specimens, obtained with the use of a scanner, were processed by means of ImageJ 1.38× software. For the quantification of the area covered by algae, colour thresholding was performed on the b (blue to yellow axis) coordinate of the CIE LAB colour space. 3

Figure 1. Schematic presentation of one unit of the modular setup.

each compartment. Algae were introduced at the top of the compartment by means of a plastic tube and a sprinkling rail. The run-off period was set to start every 12 h and ran for 90 min. Furthermore, the setup was submitted to a 12 h day and night regime, which started simultaneously with the run-off periods. Each test cycle lasted one week. 2.4

Due to the high porosity and roughness of the aerated concrete, the flow of water covered the complete specimen. As a result, a homogenous fouling occurred on these specimens. The presence of a water repellent however, resulted initially in distinct paths of water flow. Along these paths, intense colonization rapidly occurred. After 3 weeks of accelerated tests however, the water repellent effect was strongly reduced. Eventually a complete fouling of these specimens was also observed. The untreated specimens and the specimens treated with the lowest concentrations of biocide showed initially the highest algal growth (Fig. 2). In the first test series (week 1–4), the least rapid colonization was observed for the specimens treated with the highest concentrations of biocides. After 4 weeks of testing however, almost 90% of the surface of these specimens was also covered with algae. As metal based biocide formulations exert their inhibitory effect upon direct contact with the microorganisms, the loss of biocidal activity could be due to the piling of dead biomass upon the biocidal compound. This is in accordance with the fact that no leaching of the compound occurred. 4

Evaluation

As the problems of aesthetics are related to the proportion of the area covered with algae and to the intensity of coloring, the degree of fouling was evaluated by means of colorimetric measurements and image analysis. Specimens that were covered with algae showed a characteristic reflectance spectrum, corresponding to the absorbance spectrum of chlorophyll a, the photosynthetic pigment of algae. From these findings, the algal density was quantified by means of the difference in reflectance between the wavelength at 550 nm and 670 nm.

RESULTS AND DISCUSSION

CONCLUSIONS

Two types of treatments were investigated towards their performance to prevent algal fouling on aerated concrete. Both water repellent and biocidal product formulations were shown to slow down the rate of fouling. However, complete coverage of the specimens with algae already occurred after 4 to 6 weeks of accelerated tests. The accelerated fouling setup developed in this research allowed for a rapid evaluation of antifouling strategies on building materials. Due to its low cost and simplicity of design, this setup can be easily built in other laboratories.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Corrosion testing of low alloy steel reinforcement M. Serdar Materials Department, Faculty of Civil Engineering, University of Zagreb, Croatia

I. Stipanovic Oslakovic´ Department for Concrete and Masonry Structures, Civil Engineering Institute of Croatia, Zagreb, Croatia

D. Bjegovic´ Materials Department, Faculty of Civil Engineering, University of Zagreb, Croatia Civil Engineering Institute of Croatia, Zagreb, Croatia

L. Valek Faculty of Chemical Engineering and Technology, University of Zagreb, Zagreb, Croatia

ABSTRACT: Recent results indicate that some low alloy steels under certain conditions behave significantly better than ordinarily used black steel reinforcement. After thorough research of the steel reinforcement available on the market, different types of low alloy corrosion resistant reinforcement have been chosen for corrosion testing. The aim of this research is to evaluate the utilization of low alloy corrosion resistant steel as reinforcement in concrete when the structure is exposed to aggressive environment such as chloride contamination and carbonation. During the research two types of low Nickel steel types will be tested and their corrosion resistances will be compared to stainless steel grades 1.4301, 1.4306 and duplex steel grades 1.4362 and 1.4462. Steel rebars are tested in chloride contaminated and carbonated pore solutions using electrochemical impedance spectroscopy (EIS) and anodic polarisation, simulating concrete environment with lower pH value and high chloride contamination. In this paper results of up-to-now performed testing are presented.

1

INTRODUCTION

The aim of the present research is to evaluate the corrosion resistance of new types of corrosion resistant steels with lower content of expensive alloys, such as nickel or molybdenum in carbonated and chloride contaminated pore solution, simulating concrete in very aggressive environment, in order to investigate the possibility of utilization of these types of steel rebars as reinforcement in concrete. In order to fully understand the behaviour of different types of corrosion resistant steel in concrete research is divided into following steps:

and black steel (B500B). The chemical compositions of chosen steels are given in Table 1. The mechanical properties of chosen steels are given in Table 2. Pore solutions were prepared with addition of Ca(OH)2 and KOH to obtain given pH values. Measurements were performed using PAR VMP2 potentiostat/galvanostat with three-electrode arrangement. Working electrodes were prepared manually, counter electrode was carbon rod and reference electrode saturated calomel electrode (SCE). All potentials reported refer to SCE. 3

− electrochemical testing of steel specimens in pore solutions with different pH values and chloride contaminations; − electrochemical testing of steel embedded into small concrete specimens; − on site exposure and testing of steel embedded into larger concrete beams. In this paper, results of up-to-now performed testing will be presented. 2 2.1

EXPERIMANTAL Materials

In this paper results of corrosion testing of four steel types are presented: TOP12, stainless steel grade 1.4301, 1.4306

3.1

RESULTS Anodic polarisation

Testing was performed after one hour submersion in solution in order to allow stabilization of corrosion potential. Potentiodynamic polarization was performed in the potential range between −0.2 V and 0,8 V with scan rate 10 mV/s with current and potential measurement every second. Potential versus current curves for comparison of corrosion behaviour of four steel types in noncarbonated pore solution with 0.3% of Cl- and in carbonated pore solution with 0.1% of Cl- are shown in Figure 1 and 2 respectively. From these results it is observed that stainless steel grade 1.4306 preserves its passivity in noncarbonated and carbonated pore solution for the longest period. Steel types TOP12

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Table 1. types.

Chemical composition of the researched steel

10x100

1x100

C

Mn S

Cr

Mo

Ni

N

0,3/1,0 7,93 9,22 0,02

0,03 0,051 0,027 0,005

0,03 1,5 ≤0,015 10,5/12,5 0,024 0,058 1,46 <0,001 18,24 0,036 0,016 1,50 0,008 18,19 0,107 0,19 0,62 0,022 0,07 <0,02

log I, mA

100x10-3

TOP12 1.4301 1.4306 B500B

10x10-3

1x10-3 black steel TOP12 1.4301 1.4306

100x10-6

Table 2.

10x10-6 -0,4

Mechanical properties of researched steel types.

Type

TOP12 1.4301 1.4306 B500B

Yield strength, Rp0,2 [MPa] Tensile strength, Rm [MPa] Stress ratio, Rm/Rp0,2 Elongation, Agt [%]

566 748 1,32 8,2

882 992 1,12 10,3

697 825 1,18 5,7

-0,2

0,0

0,2

0,4

0,6

0,8

1,0

E, V

Figure 2. Anodic polarization curves of different steel types in carbonated pore solution (pH 10.1) containing 0.1% of Cl-.

500 540 1,08 5,0

40

30

2

-Zim(kOhm×cm )

10x100

1x100

log I, mA

100x10-3

20

10

10x10-3

1.4301 TOP12 1.4306

0 1x10-3

100x10-6

1x10-6 -0,4

0

black steel TOP12 1.4301 1.4306

10x10-6

-0,2

0,0

0,2

0,4

0,6

0,8

5

10

15

20

25

30

Zreal(kOhm×cm2)

Figure 3. Nyquist plot of the impedance spectra of different steel types in pore solution with pH 12.4 and with 0.3% Cl-.

1,0

E, V

Figure 1. Anodic polarization curves of different steel types in non-carbonated pore solution (pH 12.4) containing 0.3% of Cl-.

100

-Zim(kOhm×cm )

80

3.2

Electrochemical impedance spectroscopy

Steels were studied in the frequency range from 100 kHz to 10 mHz with ac potential perturbation of 10 mV amplitude. The EIS measurements were performed after 1 h exposure of the reinforcement to the testing solution. Results of experimental obtained data and fitted curves of EIS performed in noncarbonated pore solution with 0.3% Cl- and in carbonated pore solution with 0.1% of Cl- are given in Figure 3 and 4 respectively. Results of electrochemical impedance spectroscopy clearly show that pitting resistivity of steel TOP12 is between stainless steel grade 1.4301 and 1.4306, when used in noncarbonated or carbonated and chloride contaminated pore solution. 4

2

and 1.4301 have similar behaviour in noncarbonated and in carbonated pore solution contaminated with chlorides.

CONCLUSION

Comparing steel tested steel types it can be concluded that steel TOP12 has very good corrosion behaviour when used

60

40

20

1.4301 TOP12 1.4306

0

0

20

40

60

80

100

2

Zreal(kOhm×cm )

Figure 4. Nyquist plot of the impedance spectra of different steel types in pore solution with pH 10.1 and with 0.1% Cl-.

in highly alkaline media and in media with lower alkalinity. Since the price of this type of steel is lower than the price of stainless steel 1.4301 and 1.4306 the use of corrosion resistant steel TOP12 would be interesting for application as reinforcement in concrete. In order to fully evaluate behaviour of different types of low alloy steel reinforcement further research is carried out by authors of the paper.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Influence of the geometry and reinforcement content of specimens on frost de-icing salt resistance of concrete F. Dehn University of Leipzig/MFPA Leipzig GmbH, Germany

M. Orgass MFPA Leipzig GmbH, Germany

K. Pistol University of Leipzig, Germany

ABSTRACT: Within an extensive research project experiments on conventional and high-strength bridge pavement concretes have been carried out to investigate the influence of the geometry (slabs in comparison to cylindrical cores) and reinforcement content of specimens on the frost de-icing salt resistance according to the so-called CDF-test1. With the help of experiments it was investigated whether drilled cores with and without reinforcement taken out of different outdoor-stored and air-entrained bridge pavement concretes allow a reliable statement in comparison to the scaling determined on laboratory test slabs made of the same concretes. In general, a coherent verification of the test procedures should be made under practical conditions. Four different air-entrained bridge pavement concretes mixed in four different ready-to-use concrete plants were put into trial plates where from cores have been drilled. For the other specimens (slabs and cylinders) moulds were filled with the same concretes. 1

2

NTRODUCTION

The influence of a frost exposure on the durability of concrete is a complex phenomenon. Events of damage on concrete bridges pavements show that also within these pretended frost resistant structures frost and freeze-thaw losses occur, especially when concrete technological principles for production, placing and curing are only conditionally kept. To exclude any damages experimental investigations in preposition can be executed. Mostly, verifying investigations are additionally made after erecting bridges when damages already appeared. These investigations are necessary to clarify damages, to validate the damage progress and to end up in formulating and initiating the rehabilitation and retrofitting measures. Taking test specimens out of the damaged concrete bridge structure is decisive for a subsequent damage appraisal. In most cases drilled cores with a cylindrical shape are taken as samples containing certain contents of reinforcement. But for the most performance tests slabs without reinforcement are used in laboratory in order to predict the frost de-icing salt resistance of the concrete which was used.

1 CDF means Capillary suction of De-icing solution and Freeze thaw test.

AIM OF RESEARCH

In the past decades different test methods were developed to enable a judgement for concrete prescriptions deviating from standards and technical guidelines regarding the frost de-icing salt resistance. The respective procedures were particularly improved thereby in their precision and in the correlation to practical experiences. However, there is still a gap for several test methods (slab test, CDF-test, etc.) between the results obtained in laboratory and the real concrete behaviour on site. The real exposure should be simulated in the laboratory by the test methods as realistic as possible. This includes the number of freeze-thaw-cycles, the minimum and/or maximum temperature or the concentration of the NaCl-solution, etc. To prove the exposure conformity between laboratory and practice it is common to carry out field studies. The difference on the resistance side is particularly shown by the condition of an exposed concrete surface of a laboratory test specimen and the real condition of the building surface on site. To investigate the influence of practical parameters (like compaction, curing, formwork etc.) on the results of a laboratory test method drilled cores are taken out of testing walls or existing buildings in general. Then they will be compared with the laboratory test specimens. Such a comparative investigation for which the frost de-icing salt resistance at drilled cores was

135

determined is the subject of this work. It was investigated whether a change in the specimens geometry and a reinforcement content in the drilled cores have an influence on the results of the CDF-test in order to use this test method as a tool for verifying the frost de-icing salt resistance on site (Table 1).

specimens (cylinders Ø = 150, h = 75 [mm]) with and without reinforcement were manufactured as well as the usual specimens for the respective hardened concrete examinations.

3

It could be stated that the specimen geometry and the reinforcement content in drilled cores affect the CDFtest results. The laboratory-stored cylindrical specimens of air entrained pavement concretes showed a larger scaling than the CDF slabs. For cores with reinforcement drilled from the outdoor-stored trial plates smaller scaling have been determined in relation to the drilled cores without reinforcement. The influence of concrete surface zone condition (screeded vs. formed) and kind of storage (laboratory vs. outdoor) was partially substantial. Reasons for the influence of test specimens geometry can be:

TEST PROGRAMME

For the investigations four trial plates were made. Their dimensions (l/b/h = 1.50/1.50/0.15 [m]) correspond to those of a bridge pavement section (Fig. 1). Each plate was only half laterally reinforced so that both drilled cores with and without reinforcement could be sampled. For each trial plate next to the CDF-test slabs (slabs l/b/h = 150/150/75 [mm]) also cylindrical test Table 1. Variation of parameters. Parameter

comparison between specimens

Specimens geometry Reinforcement Storage

slabs vs. cylinders with vs. without reinforcement laboratory vs. outdoor storage

4

TEST RESULTS AND CONCLUSION

• different shrinkage behaviour (before the measurement), • effect of the different masses on the thermal behaviour (e.g. heat absorption) during the measurement, • different adjustment of the aggregates in the concrete surface zone. The positive influence of reinforcement could be related with: • bond between concrete and reinforcement could decelerate expansions and compressions of test specimens caused by the frost-thaw-cycles. Therefore, a crack formation could be constricted (during the measurement), • different shrinkage behaviour (before the measurement).

Figure 1. Trial plates with plan of reinforcement.

The following conclusions can be drawn from the results of this research: A transfer of the CDF-test results onto actual building situations is only possible when practical parameters are taken into consideration. The application of the CDF-criteria does not lead to a reliable conclusion regarding the frost de-icing salt resistance. Furthermore, the comparison of drilled cores extracted on site (trial plates) with CDF-test specimens (slabs) has to be critically judged. For a comparison between laboratory, practice results and conditions—in order to calibrate the CDF-test method—the influence of specimens geometry and the reinforcement content affect as systematic errors.

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Service life modelling and prediction of durability

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Life cycle management of concrete structures Part I: Birth certificate T.F. Mayer & P. Schießl Centre for Building Materials, Technische Universität München, Munich, Germany

ABSTRACT: The need to maintain an ageing stock of public buildings with very limited budget resources has promoted the development of so-called life cycle management systems. As the increased demand for tools to assist with the management of large building stocks goes along with recent achievements in the field of deterioration modeling, new generations of life cycle management systems employing full-probabilistic deterioration models for the condition prognosis are currently under development. This paper presents the “birth-certificate” as a core element of these management systems. The different stages in the development of a birth certificate, ranging from the first reliability calculations in the preliminary design stage to the quality control, update of the prognosis and inspection planning after completion, are described and illustrated by means of a case study.

1

INTRODUCTION

Life cycle management of concrete structures which are subject to severe environmental loads has gained increasing importance over the last few decades. As the increased need for tools to assist with the management of large building stocks goes along with recent achievements in the field of deterioration modeling, new generations of life cycle management systems employing full-probabilistic deterioration models for the condition prognosis are currently under development. This paper presents the so-called “birth certificate” as a basic element of a predictive life cycle management systems. An improved knowledge of current and future condition states as well as an optimized structural resistance towards environmental stresses is achieved by employing full-probabilistic deterioration models in combination with material testing during compliance stage and non-destructive test methods during acceptance testing to optimize the structural resistance and improve the condition prognosis.

2 2.1

BIRTH CERTIFICATE General

The birth certificate forms an important first step for the durability and quality control of newly completed structures. The approach followed in this paper is schematically illustrated in Figure 1. 2.2

Design stage

In contrast to load design, durability design has traditionally been carried out according to merely empirical “deemed to satisfy” rules defining minimum

Figure 1.

Stages in the development of a birth certificate.

concrete cover, maximum water-cement-ratios or cement types to be used. Recent advances in deterioration modeling nowadays allow for the use of full probabilistic deterioration models which are capable of taking actual environmental loads and material resistances determined during compliance testing into account. Probabilistic models for the depassivation of the reinforcement due to carbonation or chloride ingress have already been applied successfully over the last decade for various structures. The first step in durability design is the assessment of environmental stresses and the quantification of the target service life, limit states to be considered and corresponding safety indices required at the end of service life. Probabilistic prognosis renders requirements for specific concrete characteristics as

139

for instance the chloride migration coefficient DRCM,0 or the carbonation resistance RACC,0. These requirements can be compared to specific resistances of concrete mixes tested earlier or stated in literature. In case deterioration modeling yields that due to very severe environmental stresses the durability requirements cannot be met with concrete alone, further protective measures (coating, cladding) have to be considered.

concretes exposed to chloride ingress, carbonation, freeze- and freeze-thaw-attack. Threshold values for the minimum allowable electrolytic resistivity have to be formulated during the compliance testing stage already. If quality control shows that fractions of the concrete do not meet these requirements, possible consequences have to be evaluated and in some cases remedial actions have to be taken. 2.5

2.3

Compliance testing

Quality control after completion

Test methods used for compliance testing are intended to give a direct and reliable quantification of the concrete characteristics. However, the slow progress of deterioration mechanisms as for instance carbonation calls for accelerated test methods. Therefore, rapid chloride migration testing and accelerated carbonation testing have been established as standard procedures for quantification of chloride migration and carbonation properties respectively. For freeze- and freeze-thaw-exposure, CDF- and CIF-test appear to be adequate test methods. Due to the lack of deterioration models for these mechanisms, durability design with respect to freeze- and freeze-thaw-impact is still limited to ‘deemed-to-satisfy’ approaches. The results of compliance testing can be used for a first update of the original deterioration prognosis. If the concrete characteristics determined in the tests are not sufficient to satisfy the requirements specified during design stage, an adjustment of the original concrete mix design has to be made or changes in the structural design itself (increase of concrete cover, application of coating) have to be considered.

Results of compliance testing and quality control during construction are important indicators for the assessment of the structure’s durability. However, they cannot take the conditions on site and the execution quality (compaction, curing, etc.) into account. Therefore, an additional quality control after completion is inevitable. Especially for properties that are not changing with time (e.g. concrete cover) ample testing at this early stage is recommended. On the other hand, quality control directly after completion yields no information about the environmental stresses that the structure is subject to or structure-environment interactions, so that only the structural resistance can be assessed. Depending on the exposure of the structural elements, non-destructive techniques to be used in addition to concrete cover mapping are for instance electrolytic resistivity measurements, permeability testing or crack mapping. Visual inspections yield no quantitative results for the update of the deterioration prognosis. Still, they form an important part of the acceptance inspection as they help to identify potential ‘hot spots’ due to cracking, risk of ponding or other irregularities.

2.4

2.6

Quality control during construction

Quality control during construction also forms an important part of the birth certificate. Compressive strength testing gives a first impression of the uniformity of the concrete produced on site, but owing to the significance of durability, more specific test methods have to be provided. Especially for large structures and correspondingly huge concrete volumes, simple and easy-to-use test methods have to be chosen. Test series conducted earlier indicated that once the characteristic properties of the concrete with respect to the prevailing exposure classes have been identified, the electrolytic resistivity can be used as a sufficiently accurate conformity indicator for

Inspection planning

Inspection planning forms the last part of the birth certificate. Theoretically, two different inspection concepts are plausible: an adaptation of inspection intervals based on the expected development of the condition state or constant inspection intervals and an adaptation of inspection methods and extent according to the expected state of the structure. Maintaining constant inspection intervals helps to establish routines and responsibilities. Besides, for public structures in the scope of the German DIN 1076:1999, inspection intervals of three and six years have to be obeyed. A first extensive inspection is usually due at the end of the warranty after normally five years.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Life cycle management of concrete structures Part II: Monitoring Ch. Dauberschmidt, G. Kapteina & Ch. Sodeikat Ingenieurbüro Schießl Gehlen Sodeikat GmbH, Munich, Germany

ABSTRACT: In Europe severe damages on reinforced structures exposed to de-icing salt cause annual costs in a range of several billion Euro. There is a considerable saving potential just by detecting the deterioration with its dimension in time and subsequently carrying out appropriate intervention before damage is visible or even occurs. The monitoring of structures enables a continuous condition assessment. By the implementation of sensors the water content within the concrete can be measured, which enables to assess the proper operation of coatings and the effectiveness of hydrophobic treatment. In this context typical application examples will be given. The implementation of corrosion sensors enables to locate the depth of the depassivation front and therewith the execution of intervention measures on time before the reinforcement becomes depassivated. In this context a typical application example such as the parking garage of the Allianz-Arena in Munich will be given. Besides it will be shown how such additional information from monitoring can be used to update a service life calculation as a valuable tool for life cycle management. 1

1.1

CONDITION ASSESSMENT BY CORROSION MONITORING Deterioration process

Chloride induced corrosion of reinforcement and prestressing tendon is causing damages on structures, which amount up to several million Euro annually in Germany. In most cases the degree of damage has reached a considerable extent by the time it is visually detectable. The corresponding intervention, which can include repair as well as strengthening measures, is therefore rather extensive and complex. In this respect a huge saving potential exists in case the point of intervention is adjusted to an early damage stage. In the following the continuous condition assessment of structures via corrosion sensors is introduced. Reinforcement is protected against corrosion by the formation of a so-called passive layer caused by the alkalinity of the concrete. This protection can be destroyed due to the penetration of chlorides (Cl–) by de-icing salt and seawater or/and due to carbondioxide (CO2) leading to depassivation of the reinforcement. Usually these depassivating substances penetrate into the concrete from the surface through the concrete cover towards the reinforcement. Once the carbonation front has reached the reinforcement or/and chlorides accumulate at the reinforcement up to a critical concentration corrosion is likely to occur provided that sufficient oxygen and humidity is present, which usually apply for outward structures.

occurs at a single sensor element an electrical signal can be recorded. As the depth of the sensor element is known the location of the depassivation front can be located continuously while the structure monitored is in service. 1.2.1 Allianz Arena The Allianz-Arena is the soccer arena in which the opening match of the world championship 2006 took place. To ensure a sufficient durability for the adjacent parking garage a life cycle management has been carried out in the design stage (preliminary design). Within the framework of quality control measures the actual dimension of concrete cover and its resistance towards chloride ingress has been determined and incorporated in the calculation by an update (update after completion). As a further part of the durability concept the park decks have been equipped with 30 Anode Ladders. The readout of monitoring data and its interpretation takes place in regular intervals. Since the prediction of the depassivation is associated with several uncertainties the monitoring system gives additional information of the actual structural behaviour and hence cuts down uncertainties linked with the assumptions made. The implementation of corrosion

single sensor elements

reinforcement

1.2

Functionality of corrosion sensor

depassivation front

vertical distance from concrete surface

In the following the functionality of the Anode Ladder is briefly introduced (Fig.1). As soon as corrosion

Figure 1.

141

Schematic illustration of an Anode Ladder.

time

4

reliability index ◊ [-] reliability index β [-]

cable

3

update after incorporation of monitoring data

isolating plastic rings

update after completion

ring electrods consisting of stainless steel

2

0 0 2

10

20

30

40

50

Figure 3. Schematic illustration of a Multiring-Electrode (MRE) and a typical resistance profile obtained from a desiccative concrete.

time time t [a] [a]

Figure 2. Calculated reliability indices (preliminary design, update after completion and incorporation of monitoring data) vs. time of exposure.

monitoring data after 2 years within the probabilistic calculation via a Bayesian update leads therefore to an increased reliability (Fig. 2). The demonstrated implementation of monitoring data into a service life calculation facilitates in excess of a continuous condition assessment the possibility of an updated prognosis over service life linked to the depassivation of the reinforcement and therewith under the prevailing circumstances the onset of corrosion. This capability makes the service life calculation in connection with data achieved form the Anode Ladders an indispensable tool for the economic planning of intervention measures over service life.

2.1

electrolytic resistance [Ω]

vertical distance from concrete surface [mm]

1

preliminary design

2

concrete surface

2.5 2.5 2.5

CONDITION ASSESSMENT BY MONITORING THE WATER CONTENT OF THE CONCRETE Determination of the water content

The determination of the water content in construction materials and in particular with regard to its space and time dependent distribution is of high interest in the area of research and material development. This is due to the fact, that material characteristics as well as its durability depends decisively on the water content of the construction material. An indirect measuring method is the mapping of the electrolytic resistance of the concrete, which correlates strongly with the water content. This can be measured with the so-called Multiring-Electrode (MRE), which can be integrated in existing and to be build structures.

The measuring principle of this sensor is based on the electrolytic resistance between two adjoining stainless steel ring electrodes located in different depths (Fig. 3). 2.2

Areas of application for road structures

The Multiring-Electrode is a very effective tool when it comes to the assessment of concrete bridge coatings. In case that a local failure of the coating occurs a reduced electrolytic resistance will be measured due to the increased moisture of the concrete. Based on the same procedure the effectiveness of hydrophobic treatment can be checked.

3

CONCLUSIONS

This paper demonstrates that the implementation of monitoring data is an economic efficient and reliable method to achieve a more precise condition assessment. The introduced methods are particularly suited for the assessment of coatings and hydrophobic treatments, respectively their effectiveness. By evaluating monitoring data a failure of coating or hydrophobic treatment can be detected in time before extensive damage is caused. Thereby the point of intervention linked with an appropriate intervention measure can be identified. Embedding corrosion sensors, such as the Anode Ladder, the condition assessment of reinforced concrete structures exposed to corrosion initiating substances can be facilitated. The obtained measurement signal provides information of the depth of the depassivation front. This information can furthermore be integrated within a full probabilistic service life calculation by an update. Thereby possible intervention measures can be optimised economically with regard to the point of time to be carried out.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

A Spreadsheet Model for Service-Life Predictions E.A.B. Koenders & M. Ottelé Microlab, Delft University of Technology, Delft, The Netherlands

B. Obladen Strukton T&E, Maarssen, The Netherlands

ABSTRACT: With the ongoing developments of tender contracts towards Design, Build, Maintenance and Finance (DBMF), the performance of existing and newly build structures have to be assessed in terms of servicelife more and more. DBMF contracts incorporate responsibilities for contractors regarding their future demands for repair and maintenance costs of concrete structures. Service-life models can help to predict the condition of a structure over the years. The dominant mechanism behind these types of models is the penetration of deicing chlorides into the concrete surface. One of the most known models is this field is the DuraCrete Model which is developed within the framework of a European project. The model presented in this paper has a similar theoretical probabilistic background but is developed within an Excel-environment. The model is very practical and can easily be used to assess the service-life of concrete structures. In this paper, an overview will be provided of the theoretical backgrounds of the model and it will be shown how the model can be implemented in an Excel spreadsheet. Furthermore, a relation will be shown between the required concrete cover versus a given service-life. 1

2 THEORETICAL MODEL

INTRODUCTION

Life-time predictions of concrete structures are often conducted with help of numerical models. These numerical tools used nowadays play an indispensable role in the design process and have a substantial added value. Modern project tender contracts prescribe life-time requisites up to 100 years of service, often in combination with maintenance responsibilities. In order to deal with these issues, the European project DuraCrete (DuraCrete 2000) has been initiated with the main objective to develop a concept for durability design. The aim was to develop a performance based durability design methodology (Schieβl, Gehlen 2006) based upon realistic and sufficiently accurate environmental and material models capable to predict the behaviour of a concrete structure. The model is based on a design framework in which the probabilistic modelling limit states and modeling of the deterioration mechanisms are formulated. Emphasis of the model is the chloride ingress and the carbonation of the cover zone. However, the chloride ingress turned out to be far more influential with respect to the performance requirements and the service-life predictions. Due to this fact, in this paper the contribution of the chloride ingress as the main mechanism that affects the performance of concrete elements will be considered. Emphasis is, therefore, on the model is on the life-time prediction of structures exposed to chlorides. This paper provides an overview of how this spreadsheet model can be developed and how results can be achieved.

The core of the model for chloride ingress predictions is based on the second Fick’s law with a prescription of the time-depended diffusion coefficient:

∂C ∂ 2C = Dcl (t ) 2 ∂t ∂x

(1)

With C the chloride concentration, t time domain, x space domain and Dcl(t) the time dependent diffusion coefficient.

⎛t ⎞ Dcl (t ) = D0 ⋅ ⎜ 0 ⎟ ⎝t ⎠

n

(1)

with D0 the reference diffusivity after 28 days of hardening, t0 equal to 28 days period and n the so-called ageing coefficient. Since many countries throughout Europe require different values for the critical chloride concentration, the spreadsheet model has been evaluated with respect the sensitivity of this magnitude of this value. In Fig. 1, model results are provided for three different values for the critical chloride concentration. The impact of this value regarding the feasibility of the models predictability can be observed from the figure quite well. The results show that depending on what value for the critical chloride has been adopted, service-life predictions may range between ca. 30 (0.4) and 95 (0.8) years, for the average value (Pf = 50%). This holds for the input values as provided in the legend.

143

1

Probability of failure Cl crit. = 0.4

Cl crit. = 0.6

0.8 Cl crit. = 0.8

Set stochastic variables

0.6

0.4 Critical Cl = 0.4, 0.6, 0.8% Aging factor n=0.45 Initial Cl = 0.1%, V = 10% Surface Cl = 3%, V = 10% Cover x = 35 mm, V = 5% Dcl = 3.6 E-12 m²/s (CEM III/B)

0.2

0 0

25

50

75

100

125

Loop over cover thickness x 150

Service-life [Years]

Figure 1. Probability of failure for three different values of the critical chloride concentration, 0.4, 0.6 and 0.8, respectively.

3

Calculate Service-life x = x + ∆x

MODEL EXTENSION

The spreadsheet model as described in this paper shows enormous potential for evaluating the relevance, predictability and feasibility of results calculated by means of the chloride diffusion theory. The model provides the opportunity to simulate the chloride ingress of concrete elements which exposed to a constant level of chlorides. The most sensitive part of the model remains its dependency on the critical chloride concentration and the aging factor that accounts for the densification of the microstructure. Besides this, obtaining accurate values for the diffusion coefficient turned out to be difficult as well. Experimental methods like the Rapid Chloride Migration (RCM) method have shown to be inappropriate and sensitive to parameter variations. Strong points of the spreadsheet model are its simplicity and its ability to conduct sensitivity evaluations quickly. In line with this, the spreadsheet model has been extended with a loop that links the cover thickness to its associating design service-life. The structure of the calculation procedure is provided in Fig. 2. By means of this design graph, the cover thickness that associates with a predefined service-life requirement can be achieved quite easily. In Fig. 3, an example of a design graph is provided. The figure shows the effect of the aging factor on the thickness of the cover needed to meet a certain predefined service-life. The factors range around a value of 0.3 associating with a Portland cement type CEM I. Higher values of the aging factor, i.e. between 0.4 and 0.5, associate with a Blast Furnace Slag type of cement indicated by CEM III/B. However, in (Dalen, 2005) aging factors of 0.59 were reported for CEM III/B. The highest values of aging factors, i.e. above 0.6, associate with cements with a very dense microstructures and low permeability. In Fig. 3 results are provided representing the relationship between the cover thickness and the associating service-life for a CEM III/B.

Get Beta

Beta = 1.5 x < xmax

Figure 2. Scheme for calculating the design graph for the relationship between the service-life and the cover thickness.

90

Cover thickness [mm] n = 0.3

80 n = 0.4

70 60

n = 0.5

50 40

n = 0.6

30 Critical Cl = 0.4 % Initial Cl = 0.1%, V = 10% Surface Cl = 3%, V = 10% Dcl = 3.6 E-12 (CEM III/B)

20 10 Beta = 1.5

0 0

20

40

60

80

100

service-life [years]

Figure 3. Design graph for the cover thickness versus the associating service-life for different values of the aging factor.

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Modeling of the influence from environmental actions on the durability of reinforced concrete structures Anders Lindvall Division of Building Technology, Chalmers University of Technology, Göteborg, Sweden

ABSTRACT: This paper presents a research project in which the influence from the exposure conditions on the durability of reinforced concrete structures is studied. The focus is on structures that are exposed to chlorides and therefore may risk corroding reinforcement. The overall objective of the project is to model and quantify the influence from the exposure conditions on the durability of reinforced concrete. The output of the research will be general models, which describe the “environmental load” acting on concrete structures, applicable in available service life models. In the paper a summary of the research project and a state-of-the-art review is made. The focus is on environments with chlorides, i.e. marine environments and along roads where de-icing salt is spread. The exposure conditions in these environments are described and the main influencing factors are identified. Future developments that are planned will also be presented. 1

INTRODUCTION

Reinforced concrete structures are usually expected to have good durability and sufficiently long service life since they often require large investments. The durability of these structures is, however, influenced from the actions of different deterioration processes and the service life will be limited. As a result many structures have to be repaired or even replaced in advance, which results in extra expenditures. Therefore to reduce these expenditures there is a need to design the structures with respect to durability, to get a sufficiently long service life. This requires an understanding of the deterioration processes combined with knowledge the governing factors. One important influencing factor is how the exposure conditions are influencing the deterioration of concrete. However, despite a lot of research little is known about this factor. Therefore the research project presented in this paper is focused on how the influence from the exposure conditions can be described and modelled. The focus will be on environments where chlorides are present, i.e. marine environments and along thaw-salted roads, with the risk of chloride induced reinforcement corrosion. 2

EXPOSURE CONDITIONS AND RESPONSE FROM CONCRETE

The exposure conditions and their variations have a significant influence on the durability of reinforced concrete structures. The most accurate way is to describe them as surface actions in terms of temperature and moisture conditions, carbon dioxide concentration and chloride conditions, Nilsson (1996). There is also an interaction between the environmental actions on a concrete surface and the conditions

inside the concrete. The conditions inside the concrete also influence each other. An obvious example is the moisture conditions, where moisture will be an obstacle to transport of gas, e.g. oxygen required for reinforcement corrosion, while it is a requirement for transport of ions, e.g. chlorides. More examples of these interactions are given in Nilsson (1996). In the literature there are some examples of investigations of environmental actions and its response. Some examples can be found in the following publications − Marine conditions. Examples of measurements can be found in Fluge (1997), Wood & Crerar (1997), Buhr et al (2000), Lindvall (2003), Tang (2003), Polder et al (2003), Meira et al (2007) and Wall (2007). − Road conditions. Examples of measurements can be found in Andersen (1997), Lindvall & Andersen (2000), Lindvall (2004), Gaal (2004), Akgül & Frangopol (2005) and Fagerlund (2005). 3 3.1

FUTURE DEVELOPMENTS Qualitative model

The most accurate way to describe the influence from the exposure conditions is to describe them as surface conditions. A qualitative model has been formulated to derive the surface conditions in a systematic way and to identify important influencing parameters. The model is shown in Figure 1. The starting point for the model in Figure 1 is information about the meso conditions, i.e. the exposure conditions undisturbed by the structure. The meso conditions are then transformed, with information about the location, the surroundings and the general shape of the structure, to achieve local conditions (disturbed by the structure). Finally the local conditions are converted into surface conditions (variations

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Undisturbed of structure

Disturbed by structure

Variations within structure

Meso conditions

Local conditions

Surface conditions

Part of country

Location of structure

Surroundings of structure. General shape of structure.

Details of structure •Geometry •Materials

Surface properties: -Roughness -Colour -……… Bulk properties: -Heat transport -Moisture transport -………

Figure 1. Qualitative model used to describe the environmental actions on concrete properties. The basis for the model is a division of the exposure conditions into different levels depending on geographical dimension.

within structure) with additional information about the structure, e.g. geometry and material properties and the response of the concrete. The conditions and properties of the concrete may influence the surface conditions, e.g. surface colour and absorptance. 3.2

Plan for the research

The emphasis in the research project will be to formulate more general models for environmental actions on concrete surfaces. The models will be formulated in such way that they can be used in available mathematical models used for durability design. The focus will be on environments where chlorides are present. The output of the models will be general understandable boundary and initial conditions (characterizing the “environmental load” acting on a concrete structure) applicable in available service life models, e.g. ClinConc and DuraCrete. The developed models aim at describing the environmental actions on a micro level and how these vary over the structure. However, this gives in many cases probably a too detailed description of the environmental actions. Instead it would be more relevant to derive representative exposure conditions for larger parts of a structure. This can be made with a statistical analysis of the micro environment and its variations over the structure. Parallel with the formulation of models for environmental actions data from the literature will be

collected. The focus will be on the exposure conditions in environments where structures are exposed to chlorides, i.e. in marine conditions and along thaw-salted roads. The collected data will be used to identify and quantify the main influencing factors in the environmental actions, e.g. chloride concentrations and temperature in seawater, amount of de-icing salt spread on roads, extension of splash and spray from seawater and de-icing salt, time of surface wetness etc. The aim is to find data where both the environmental actions and the corresponding environmental response from concrete structures are described. In this way it will possible to both quantify the parameters in the developed models and verify the predictions. However, these kinds of data are scarce in the literature. Usually either the environmental actions or the environmental response are measured—it is seldom that parallel measurements are made. Some data are already collected, e.g. within the EU-projects DuraCrete and Chlortest and from the Swedish field stations along Rv40 (road conditions) and in Träslövsläge (marine conditions), but they need to be further analysed. These analyses are intended to be done in the research described here. If data are lacking in the literature complementary experimental investigations are planned to be made within the project. The purpose of these investigations will be to further clarify the influence of and quantify important factors in the environmental actions and how they influence the response by the concrete. The focus will be on how chloride and moisture conditions in the concrete are influenced. Examples of complementary investigations that may come in question are: − Significance of the time of surface wetness. − Significance of air streams around structures. − Influence of distances (horizontal and vertical) to chloride sources. − Characterization of exposure conditions along thaw-salted roads, e.g. amount of spread de-icing salt (influence from temperature passages of 0°C) and significance of road and traffic. The results from the project will be presented in papers published in journals and conference abstracts and in department reports.

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Studies on prediction models for concrete durability R. Heiyantuduwa & M.G. Alexander University of Cape Town, South Africa

ABSTRACT: In this paper the UCT Service Life Model is discussed, which deals with the lifetime prediction of concrete structures in marine environments using the South African Durability Index approach. An extensive study is being carried out both in the laboratory and on site. In this research, the UCT Service Life Model is to be re-formulated and further calibrated primarily for providing a basis for durability analysis and design of concrete structures in the marine environment taking into account the range of different available materials, as well as the different marine environmental conditions in South Africa. This should assist in obtaining better durability predictions, and result in improved design of long-term performance of new concrete structures in the marine environment. 1

in environmental conditions, concrete materials and site conditions.

INTRODUCTION

The degradation of reinforced concrete is influenced by a number of factors; material properties, cover to reinforcement, execution during construction/site practice and severity of environmental exposure. To completely understand these interactions and the differences between the different deterioration processes it is important to have proper models to describe and quantify these actions and their influence on the durability of concrete structures, as well as the response of the structures. Service life may be determined using quantitative service life models to predict the time to cracking (or unacceptable damage) for a particular element in a given environment. A number of such models exist. Many of these models adopt the two-stage service life model first proposed by Tuutti (1982) in which t the deterioration is split into two distinct phases; initiation and propagation. A simple approach to predicting the initiation period is to assume that ionic diffusion is the sole mechanism of chloride transport and to solve Fick’s second law of diffusion using an apparent chloride diffusion coefficient to characterize the concrete under consideration. The approaches adopted by the different service life models vary considerably and consequently there can be significant variations between the solutions produced by individual models. In most applications however, chloride ions are assumed to diffuse into the concrete and are modeled to follow Fick’s second law of diffusion in one form or the other. The assumptions behind these however are restrictive and simplistic. Surface concentrations of chlorides can vary over a period of time, and the rate of this chloride build-up can be influenced by binder type or surface treatments. Chloride diffusion is also affected significantly by both temperature changes and the age or degree of hydration of the concrete. It is therefore necessary to account and allow for these variations

2

UCT CHLORIDE PREDICTION MODEL

This proposed model is based on the same background as most others of Fick’s 2nd Law and taking into account the time dependent nature of the diffusion coefficient, but only essential alterations are made to the model, i.e. the use of durability indexes and incorporation of the effects of the various South African environmental conditions and range of different available materials. In this way, the model is kept as simple as possible, while functional.

3

EXPERIMENTAL WORK

The quantification of this work is based on information about the response of the concrete to environmental actions and the variation of the diffusion coefficient of concrete with time. The most significant parameters in this prediction model are divided into the following classes: • Material influences • Environmental influences • Threshold studies. 3.1

Results obtained to date, and discussions

Concrete samples were exposed to salt solution as specified in the Bulk Diffusion Test (ASTM C 1556-03). These cylinder samples were retrieved at regular time intervals and ground at different depth intervals from the surface up to a depth of about 60 mm. these powdered samples were then analysed to determine the total chloride content. For each time interval the total chloride content obtained from

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the powdered sample was plotted against the corresponding depth to obtain a chloride profile. Both the surface chloride concentration (CS) and the apparent diffusion coefficient (DC) were calculated from each of the profiles using a non-linear curve fitting method. Apparent diffusion coefficients obtained from curve fitting correlates with research conducted by Takewaka (1988). In most cases the diffusion coefficients followed a decreasing trend with duration of exposure time; however the rate of decrease varied between the various binder types and concrete grades. Diffusion is predominant in the submerged zone, when compared with other transport mechanisms and therefore as a result of continued hydration of the cement with time the porosity of the concrete is reduced which results in a reduced diffusion coefficient. Higher grade concretes were generally found to exhibit lower diffusion parameters throughout the exposure duration. Higher grade concretes have a higher binder content resulting in lower porosity and hence a lower diffusion coefficient. Irrespective of the binder types and concrete grades, the coefficients of diffusion obtained, converged to a value less than 7 × 10−12 m2/s. Table 1. Apparent diffusion coefficients and surface chloride concentrations obtained. Diffusion coefficient ×10−12 (m2/s)

Surface concentration (%)

Mix design

35 days

6 mth

1 year

35 days

6 mth

1 year

0.4 OPC 0.6 OPC 0.4 BS 0.6 BS 0.4 CS 0.6 CS 0.4 FA 0.6 FA 0.4 SP 0.6 SP 0.4 SF 0.6 SF

2.06 21.0 5.13 6.01 5.84 3.64 8.02 0.69 9.87 5.46 14.0 11.6

2.00 17.6 0.32 5.76 0.16 6.41 0.40 5.19 0.21 0.78 11.1 5.91

0.54 5.14 0.06 0.40 0.08 3.91 0.12 0.40 0.10 0.03 7.08 2.14

2.50 2.70 4.60 8.90 4.20 11.9 4.10 6.20 3.80 6.10 2.70 4.10

2.50 3.80 5.80 4.40 7.90 4.20 5.50 4.00 4.60 7.70 3.50 5.30

1.70 3.10 9.00 6.50 9.00 6.70 6.50 7.40 7.40 8.20 4.70 6.40

After one year of exposure surface chloride concentrations obtained from the curve fitting were found to increase with exposure time in most cases. Concrete mixes containing slag and fly ash binder types were found to have higher surface chloride concentrations and also recorded high concentration of chlorides up to depths of about 60 mm. This can be accounted for due to their superior binding capacity which increases their ability to hold chloride ions. Lower grade concretes also exhibited higher concentrations of chloride when compared to the higher grade concretes.

4

CONCLUSIONS

The need to provide concrete with sufficient durability has stimulated the development of models to predict service life. These models are based on parameters defining concrete permeation or resistance to penetration. In general, chloride diffusion coefficients derived from analytical solutions to differential equations are used. This approach has however several limitations as the reproduction of real boundary and initial conditions by the differential equations is not obvious or easy. Also, the approaches adopted by the different models vary considerably and consequently there can be significant variances between the solutions produced by individual models. The proposed model is primarily being developed in order to provide a more fundamental model based on characterisation work on diffusion vs. chloride conductivity (in the laboratory) verified by field studies (surface chloride concentrations). The purpose is to construct a robust model for South African materials and environmental conditions. Such a durability analysis can partly be applied to obtain a more controlled durability and long-term performance of new concrete structures, and partly be used as an improved basis for condition assessment of existing concrete structures in marine environments. The solutions provided by such a model are intended as approximations to be used as a guideline in designing a reinforced concrete structure exposed to chlorides. Assumptions must be made to simplify the model and make it accessible to engineers who may not have specific expertise in the area of chloride transport and reinforcement corrosion.

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Critical chloride content in reinforced concrete – State of the art U. Angst & Ø. Vennesland Norwegian University of Science and Technology, Faculty of Engineering Science and Technology, Trondheim, Norway

ABSTRACT: In the present work, a large number of publications in connection with critical chloride content in reinforced concrete has been evaluated. The reported results scatter over more than two orders of magnitude when expressed as total chloride content by cement weight and over three orders of magnitude when expressed as Cl−/OH− ratio. This large scatter is partly the result of different definitions and experimental techniques, but also due to many factors affecting critical chloride content. Dominating influences have been identified as follows: 1) The steel-concrete interface, 2) the pH value of the pore solution and 3) the electrochemical potential of the steel. Experimental investigation of the issue of critical chloride content is possible in a wide variety of procedures. At present, there exists no generally accepted or standardized procedure for the determination of the critical chloride content.

1

INTRODUCTION

The concept of critical chloride content in reinforced concrete is based on the general agreement that corrosion in non-carbonated, alkaline concrete can only start once the chloride content at the steel surface has reached a certain threshold value. In literature, this value is normally referred to as critical chloride content or chloride threshold value. A lot of studies have been undertaken and the reported results scatter over several orders of magnitude. This large span of results might be due to several reasons: First, different definitions for critical chloride content have been used; second, various techniques to find critical chloride contents have been applied by different researchers and, last but not least, critical chloride content is a complex matter that depends on a variety of influencing factors.

2

EXPERIMENTAL PROCEDURES

In order to experimentally investigate critical chloride contents in reinforced concrete the experimental setup has to include the following: 1. A steel electrode of interest embedded in a cement based material or immersed in a solution that simulates the concrete. 2. Chloride ions present at the steel surface, either at a constant concentration or increasing over time. 3. Detection of the depassivation of the steel. 4. Quantifying the corresponding chloride content. To achieve a setup that fulfils these four requirements, a lot of possibilities exist. A variety of options are available only for selecting a steel electrode: Different steel types (normal carbon steel, stainless steel, etc) exist, rebars can be smooth or ribbed and might

be prepared, e.g. polished, sandblasted, etc. Also the type of chloride salt and a method by which the chloride is introduced (mixed-in, diffusion, etc) at the steel surface must be selected. For the detection of depassivation several techniques are available as well and the corresponding chloride content can be determined and expressed in different ways. Not surprisingly, the multiplicity of possible test setups leads to poor comparability of the results. This has to be kept in mind when evaluating reported critical chloride contents.

3

LITERATURE EVALUATION

In the present article, nearly 40 references reporting critical chloride contents from laboratory or field studies have been evaluated. For chloride threshold values reported as total chloride by weight of binder, a large scatter was found with results from 0.02 to 3.08% chloride by weight of cement (over two orders of magnitude). Details about the experimental procedures were also evaluated but no systematic trends with regard to characteristics such as cement type, rebar type (smooth, ribbed) or chloride introduction method were identified from the totality of the data. Obviously, the effect of a single parameter on the critical chloride content is not pronounced enough to be globally apparent. The only factor that appears to have an overall effect is the electrochemical potential of the steel: In many studies, this was higher than −200 mV vs. SCE; only in two references steel potentials below −200 mV vs. SCE were reported. The corresponding critical chloride contents are on a clearly higher level than in the majority of the other publications. Figure 1 shows a selection of published chloride threshold values expressed in the form of Cl−/OH− ratios. Also this data presents a high overall scatter,

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Figure 1. Reported chloride threshold values expressed as Cl−/OH− ratios. As the values span a large range, they are divided into two separate plots with different scaling on the ordinate.

ranging from 0.03 to 45 and thus over three orders of magnitude. It is evident from Figure 1 that remarkably higher chloride threshold values were reported when investigating mortar or concrete specimens in comparison with solution experiments. This might be explained by the inhibiting effects of the interface of steel embedded in a cement matrix due to formation of a Portlandite layer at the steel surface. This is also apparent from the data of Yonezawa et al. (1988), where the formation of this layer at the interface was hindered intentionally. In addition, the results of two studies (Breit 1998a, Li & Sagüés 2001) show that the Cl−/OH− threshold ratio increases with increasing pH of the pore solution, i.e. it is no constant value and the inhibiting effect of OH− is stronger at higher pH values. The available publications have also been evaluated with regard to experimental procedures. In only 12% of the studies, steel bars were investigated “as-received”; in the other references they were prepared (sandblasted, polished, etc) or no details were given. In only 25% of the studies the use of ribbed steel was reported. In 31% of the evaluated references, chloride was added to the mix. In practice, chloride is usually not present at the steel surface initially, but penetrates into the hardened concrete during service life. Also the use of prepared and smooth steel does not give realistic conditions with

regard to the steel-concrete interface. The results of many studies are thus not practice-related. 4

CONCLUSIONS

From this literature review on critical chloride content, the following conclusions can be drawn: − The results available in literature scatter in a wide range. The reported values span from 0.02 to 3.08% total chloride by binder weight; published Cl−/OH− threshold ratios range from 0.03 to 45. − No unique chloride threshold value exists. It depends on various factors which are interconnected and variable with time. The following parameters had an overall effect on the evaluated results: 1) the quality of the steel-concrete interface, 2) the pH of the pore solution and 3) the electrochemical potential of the steel. − Experimental investigation of critical chloride contents can be performed in a wide variety of possible test setups. This leads to poor comparability of the reported results. At present, no accepted or standardized procedure for the determination of the critical chloride content exists. − Many used test setups are not practice-related.

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Use of durability indexes in probabilistic modelling for durability design of RC members R.N. Muigai, P. Moyo & M.G. Alexander University of Cape Town, South Africa

ABSTRACT: A probabilistic model for chloride diffusion under South Africa (SA) marine conditions has been developed. The model is a further improvement of a service life prediction model developed earlier on by Mackechnie (1996), as it takes into account the variability of the basic model parameters. The model parameters are classified as environmental parameters (surface chloride concentration and critical chloride concentration); material parameter (coefficient of diffusion) and geometric parameter (cover depth). All of which are described in terms of their statistical distribution, mean and standard deviation. The chloride diffusion coefficient (Ds ) data was obtained from the chloride conductivity test a Durability Index (DI) test and the surface chloride concentration (Cs ) data was obtained from field measurements. A goodness- of-fit-test on Cs data showed that it fits well to a normal distribution. Due to shortage of data in Ds non-parametric bootstrapping was carried out to obtain its statistical distribution. The statistical distributions of cover depth and critical chloride concentration were obtained from literature. An example was also illustrated of the use of Monte Carlo Simulation (MCS) techniques in calculating the reliability against depassivation of reinforcement for a given choice of the model parameters.

1

INTRODUCTION

used to model the diffusion of chloride ions into concrete and is represented by Equation 1.

Majority of South African infrastructure is situated in coastal areas exposed to marine environments (Beushausen et al., 2003). Chloride ions from marine exposure act as a critical load on RC structures and as they ingress into the concrete, they increase the potential of the reinforcement for corrosion. Eventually, a chloride threshold level value is attained above which corrosion of reinforcement occurs (Chisolm & Lee, 2001). The period up till corrosion initiation constitutes the service life for concrete structures exposed to chloride ions in most service-life prediction models. Owing to the problems associated with chloride-induced corrosion in RC structures, a reliable prediction method of chloride ingress in concrete is one of the key elements to consider in the design process if the structure is expected to be durable and cost efficient (Luping & Gulikers, 2007). For this study a combination of accelerated testing results and mathematical degradation modelling methods were used for predicting the service life. The degradation model used is based on the concept of chloride transport into concrete by diffusion and initiation of reinforcement corrosion when critical chloride content is exceeded at the steel surface.

2

DEGRADATION MODEL AND DI TEST

A modified Fickian model based on the solution of ‘Fick’s second law of diffusion’ (Crank, 1975) was

⎛ ⎛ x C( x,t ) = CS ⎜1 − erf ⎜ ⎜ ⎜ 2 D t (1− m ) i ⎝ ⎝

⎞⎞ ⎟⎟ ⎟⎟ ⎠⎠

(1)

Where, C(x,t) = chloride concentration at depth x at time t; Di = initial diffusion coefficient; m = diffusion coefficient reduction factor; and Cs = surface chloride concentration. The Fickian model is applicable for use in saturated concrete in the South African marine environment and incorporates the reduction of diffusion coefficient with time due to the process of chloride binding and pore structural changes which effectively deplete the chloride concentration with increasing depth (Mackechnie, 1996). The diffusion coefficient (Di) in the model describes the material properties of concrete and reflects the ability of the concrete to resist chloride penetration. For this study Di is obtained theoretically from the chloride conductivity (CC) test value. The CC test, (Figure 1) is one of the durability index (DI) tests developed for South African marine conditions and measures the resistance of 25 mm thick concrete samples (Ø 68 mm) to chloride conduction under the action of an applied voltage across the specimen. The effective diffusion coefficient Di is determined theoretically by applying the simplified NernstEinstein relation (Equation 2).

151

⎛ ⎛ x Ccrit − CS ⎜1 − erf ⎜ ⎜ ⎜ 2 D t (1− m ) i ⎝ ⎝

Figure 1. Chloride conductivity test (Alexander et al., 1999).

Q=

Di σ i = Do σ o

(

RELIABILITY ANALYSIS OF A RC PIER

The durability design focus using reliability-based design approach is to present the Fickian model (Equation 1) in the form of a limit state function (LSF). For this study, the limit state is initiation limit-state (ILS) described by the depassivation of reinforcing steel in concrete. The ILS will be represented by two functions: C(x,t) function which represents the action effect and; Ccrit representing the barrier effect, and which when exceeded leads to steel depassivation. The ILS condition is represented by Equation 3.

C( x,t ) = Ccrit

)

Pf = P Ccrit ≥ C(x,t) ≤ Ptarget

for t = t i

(3)

The relationship between C(x,t) and Ccrit is modelled mathematically by a function of basic variables g(.), representing concrete dimensions, material properties and the service environment of the RC element as shown by Equation 4.

(5)

Where Ptarget = allowable (target) probability of failure at the durability limit state.

4 3

(4)

The natural variability in the material properties of concrete makes it inevitable for probabilistic means to be used in the adequate prediction of service life. Hence, all the variables in the LSF (Equation 4) need to be represented as stochastic quantities. The statistical information for each parameter is exploited to provide improved uncertainty estimates of the output, which is expressed in terms of the probability of the LSF failure (Pf). The Pf calculated is then compared to a target probability of failure to assess the risk of the RC element not achieving its expected performance. Equation 5 serves to express this process.

(2)

Where Q = diffusibility ratio; Do = diffusivity of chloride ions through the pore solution; Di = steady state (effective) diffusivity of chloride ions through concrete; σo = conductivity of pore solution; σS = conductivity of concrete.

⎞⎞ ⎟⎟ = 0 ⎟⎟ ⎠⎠

CONCLUSIONS

The South African service life prediction model consists of basic parameters representing concrete dimensions, material properties and the service environment of the RC element. The concrete material property represented by the chloride diffusivity is derived theoretically from the results of chloride conductivity (CC) test, which has been adopted for concrete quality testing in SA. In the present study, a further improvement was made on the prediction model by taking into account the variability of the model’s basic parameters. The statistical distributions of the model parameters were investigated in this study. The statistical information for each parameter was then exploited to provide improved uncertainty estimates of the output, which was expressed in terms of the probability of the LSF failure (Pf).

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A simplified and practical approach regarding design for durability of reinforced concrete structures based on probabilistic modeling of chloride ingress J.J.W. Gulikers Rijkswaterstaat Bouwdienst, Utrecht, The Netherlands

ABSTRACT: There is a clear trend towards design for durability of reinforced concrete structures using probabilistic modeling. However, such an advanced approach is considered to be useful for scientists rather than practitioners dealing with real structures. Although a probabilistic approach is essentially correct, the observed lack of a sufficient amount of reliable statistical data on input parameters as well as an objective use of this data will limit a widespread application in practice. In this paper an attempt is made to develop simple mathematical expressions based on a significant number of probabilistic calculations, taking into account the variability of all model parameters. Eventually, this exercise has resulted into practical formulas for 3 levels of probability of corrosion initiation, Pi = 5, 10 and 20%. Using these formulas, for any arbitrary choice of cover depth the corresponding apparent chloride diffusion coefficient can be calculated, the combination corresponding to the desired probability level within a narrow range of 0.5% in most situations.

1

INTRODUCTION

on Fick’s second law of diffusion. The basic form is given by:

In most codes of practice it is common to find that design for durability of reinforced concrete structures is addressed by a so-called deem-to-satisfy approach. This implies that the designer chooses the appropriate exposure class and specifies the concrete to meet the limits on maximum w/c, minimum cement content and minimum grade, and the cover thickness to exceed a minimum. On this basis the designer assumes that the structure will be durable although knowledge on the actually desired design life and the corresponding serviceability limit state is not taken into account. However, during the last decade there is a clear trend towards probabilistic modelling of service life. This change in approach can be attributed to the developments made in the European project DuraCrete. Later this approach has been adopted in the fib Model Code. Experience gained so far has demonstrated that the introduction of such a probabilistic approach for service life design encounters major problems in practice. In order to overcome some of the problems a more simplified mathematical approach has been developed for chloride-induced reinforcement corrosion, thereby still allowing significant freedom to the designer. This paper outlines the basis and the assumptions for this simplification as well as the practical implications of this approach.

⎛ ⎞ x C (x,t) = Cs − (Cs − Ci) ⋅ erf ⎜ ⎟ ⎝ 2 Da ⋅ t ⎠ where C(x,t) = chloride content at depth x and time t, [%m/m]; Cs = fixed chloride content at the exposed concrete surface, [%m/m]; Ci = initial chloride content, [%m/m]; erf = error function; x = distance from the exposed concrete surface, [m]; Da = apparent chloride diffusion coefficient, [m2/s]; t = time, duration of exposure to chloride, [s]. For the design of new structures, however, there is no data available on the actual performance of the concrete under the prevailing exposure conditions. In order to take into account concrete quality and type of cement, according to DuraCrete the reference level of Da at time to has to be derived from the results of laboratory testing, according to: Da , o = ∏ ki ⋅ Dtest , o where Da,o = apparent chloride diffusion coefficient at reference age to [m2/s]; ki = factors that correct for a certain influence, [-]; Dtest,o = chloride diffusion coefficient obtained from a laboratory test performed at concrete age to, [m2/s]. 3

2

MODELLING OF CHLORIDE INGRESS

The most commonly used mathematical expression to describe chloride ingress into concrete is based

PROBLEMS ASSOCIATED WITH A PROBABILISTIC APPROACH

Experience has shown that a number of serious problems have been encountered when using a probabi-

153

listic approach for service life design of reinforced concrete structures. The majority of the problems are related to both the interpretation of the accepted serviceability limit state as well as the validation of the input and the output. At present there is no clear view on what the practical meaning is of a certain probability of corrosion initiation in terms of either amount of depassivated steel, damaged concrete surface area or maintenance cost. Essentially, in a full probabilistic approach each input parameter has to be characterised by a mean value, a standard deviation and a type of probability distribution. However, for most of the input parameters the amount of data on individual parameters is limited thus the translation into statistical terms becomes questionable. Furthermore, the output cannot be validated without the use of appropriate software, e.g. comprising Monte Carlo simulation. In order to overcome some of the problems associated with a probabilistic approach a more pragmatic approach is advocated.

on the type of structure, its economical importance, and the implications of repair and maintenance on availability, an appropriate probability level less than 50% has to be chosen. As initiation of corrosion is considered a serviceability limit state (SLS) it is suggested that the eventual probability level adopted should reflect an economic optimum. A sensitivity analysis highlights that the age factor, a, is by far the most dominant parameter in service life calculations. Consequently, it was concluded that the simplified approach should reflect the influence of the age factor. For a given design cover depth, cdesign, the required maximum mean value for the apparent diffusion coefficient Da at the end of service life can then be expressed by:

4

For an accepted probability of corrosion initiation Pi = 5% the factor α can be approximated by: α 5 ≈ 1.42 + 0.05 ⋅ a + 1.155 ⋅ a 2 Likewise for Pi = 10% and Pi = 20%: α 10 ≈ 1.32 + a 2 and α 20 ≈ 1.15 + 0.321⋅ a + 0.143 ⋅ a 2

DEVELOPMENT OF A SIMPLIFIED MATHEMATICAL APPROACH

For new structures initiation of reinforcement corrosion is generally accepted as the criterion defining end of service life. This situation is achieved when at the depth of the reinforcing steel, x = c, the chloride level reaches the so-called critical chloride content, Ccrit. However, when in a probabilistic calculation mean values are used for all model parameters this will result in a situation characterised by a probability of corrosion initiation, Pi, of approximately 50%. Dependent

scaling factor, α [-]

2,5 Pi = 5%

2,0

Pi = 10% Pi = 20%

1,5

1,0 0,0

0,2

0,4 0,6 age factor, a [-]

0,8

1,0

Figure 1. Scaling factor α as a function of the age factor a.

Da ,max

5

⎛ ⎜ 1 ⎜ = 2⋅ α ⎜ ⎜ 2 inverf ⎝

2

⎞ ⎟ cdesign ⎟ ⋅ 1 ⎛ Cs − Ccrit ⎞ ⎟ t dsl ⎜⎝ C − C ⎟⎠ ⎟⎠ s i

CONCLUDING REMARKS

A simplified calculation procedure has been developed to calculate the required nominal design cover depth based on mean values for all relevant model parameters, taking into account an assumed statistical variation. This procedure has been exemplified in a case study. Thus, the required concrete quality in terms of a diffusion coefficient according to a laboratory test can easily be derived for a wide range of service lives, critical chloride content, and age factors (equivalent to type of cement). The calculation procedure does not necessitate any probabilistic calculations. This procedure thus also allows results of probabilistic calculations to be checked. Three different levels of probability of corrosion initiation can be used depending on the inspectability, maintainability, and economical importance of a structure. The procedure can be chosen to derive performance criteria depending on the nature of the exposure conditions and the required design service life.

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Assessment of the behaviour of concrete in the initiation period of chloride induced corrosion of rebars M. Castellote & C. Andrade Institute Eduardo Torroja of Construction Science (CSIC), Madrid, Spain

ABSTRACT: Reinforcing steel bars embedded in concrete depassivate when a certain amount of chloides built up in their surrounding. In order to reach the rebar, chloride ions have to diffuse through the concrete and then accumulate in a specific amount able to depassivate the steel. In order to predict the time to corrsion onset, both steps, that in a whole are called initiation period for corrosion, have to be taken into account. In this paper, two methods that accelerate the phenomena by the application of electrical fields, are presented: The first one is the so called Multi-Regime Method (MR), that allows the determination of both diffusion coefficients, Ds and Dns, and therefore the binding ability of the matrix. The second one is the Accelerated Integral Corrosion Method (AIC) to determine simultaneously the non steady diffusion coefficient and the threshold value of corrosion of steel. Both methods are reliable, quick, very appropriate concerning cost in hours-man and resources and have been validated in comparison with traditional non accelerated methods.

1

MULTI-REGIME METHOD (MRM)

The test consists of applying a voltage drop to a test specimen between two compartments containing, upstream, a chloride solution (1 M NaCl) and, downstream, a solution without chlorides. The negative electrode is placed within the upstream compartment and therefore the external electrical potential forces the chloride ions to migrate through the specimen towards the downstream compartment. The test is based in measuring the amount of chlorides arriving in the downstream cell (anolyte) by means of measuring the conductivity of that solution instead of analysing chlorides in it. The concentration of chlorides is calculated through an empirical correlation between chloride concentration and conductivity for the specific conditions of the test (eq1).

c25 = cT + 0.02 (25 − T ) cT

S: surface area of test specimen exposed to the chloride solution (cm²). Cl = Cl− concentration in the catholyte (mol/cm3). γ = Activity coefficient of the catholyte solution = 0.657. ∆Φss = Averaged effective voltage (V) through the specimen during the steady-state period. l = thickness of the sample (cm) =1. R: perfect gas constant (1.9872 cal/mol K). T: Average temperature during the test (K). z: ion valence, for chloride, z = 1. F: Faradays constant = 23060 (cal/V eq). The calculation of the non steady-state diffusion coefficient is made from the time taken by the chloride ions to establish a constant flux, that is to say, from the so-called time-lag by application of the simplified equations (3) and (4):

(1)

Ds =

where c25 = conductivity (mS/cm) at 25ºC. Va = volume of the anolyte (l). The steady-state coefficient is calculated then from the flux of chlorides through the specimen calculated from the measurement of the conductivity of the anolyte in the anodic compartment using a modified Nernst-Planck equation (eq.2). Cl−( mmol) = (−1.71 + 11.45 * c25)Va.

(2)

where: Ds = steady-state diffusion coefficient (cm2/s). A = slope of the linear regression during the steadystate period (mol/s).

A⋅ R ⋅ T ⋅ l z ⋅ S .F ⋅ Cl ⋅ γ ⋅ ∆Φ ss

(3)

Where:

Dns =

2 l 2 ( v − 2) τ v2

(4)

Dns = non steady-state diffusion coefficient (cm2/s). τ = time-lag in the migration test (s); It is obtained from the linear correlation obtained in step 4, calculating the time for chloride in anolyte equal to zero. l = thickness of the specimen (cm). (∆φ)t = Averaged effective voltage (V) through the specimen from the beginning of the test until the

155

time lag. If the time-lag do not match with any of the available experimental points, a linear interpolation between the two adjacent points is recommended to be done. T: Average temperature during the test (K). 2

ACCELERATED INTEGRAL CORROSION METHOD (AIC)

In this method, chloride ions are driven to the rebar by migration due to an external voltage drop and the depassivation is produced at free potential of the steel. The non steady-state coefficient is calculated from the time taken by the chloride ions to depassivate the rebar. The specimens can be of any size with a rebar embedded, with a know value of area of exposition, and a container has to be glued in the face parallel to the embedded rebar. Prior to the test, the specimens have to be saturated in order to avoid the transport of chlorides by capillary absorption. To do so, the specimens have to be stored in a humid chamber (HR 100%) until constant weight. The container has to be filled with a 1M NaCl solution, where a stainless steel mesh, connected to the negative pole of a power supply, is immersed. As positive electrode, a steel plate is electrically connected to the bottom surface of the specimen by means of a wet sponge. A voltage drop of 12 V is applied between the electrodes, and during the tests, the solution is neutralized with a dropping of diluted HCl (0.05 M), due to the increase of OH- ions in the catholyte because of reduction of water, in the way that pH in the catholyte is maintained at 7±1. During the test, the potential of the rebar is monitored by means of a reference electrode placed in the surface of the specimen as closest as possible to the rebar, and the current intensity passing through the specimen is registered. The electrical field has to be switched off when changes in polarization of the steel are detected, which indicates that depassivation is going to take place, and wait until depassivation in natural conditions. Wait for a minimum period of

24 hours, measure the rate of corrosion of the rebar and in the case it had been repassivated, connect the voltage again and repeat the process. After depassivation of the steel, the specimen has to be broken and the concentration of chlorides in the mortar just close to the steel in the path of the chloride ions is analysed in order to determine the threshold of depassivation. In addition, the total chloride concentration in the surface of the specimen, Cs (%), that was in contact with the chloride solution has to be measured. The results of the chloride threshold value are directly given by the results of the analysis of the sample around the steel, and can be expressed in terms of total chloride, free chloride or chloride/hydroxyl ratio. The calculation of the non steady state diffusion coefficient is made through the total amount of current that has passed during the test, Q, and the value of the surface concentration, Cs, after the test [3]. The non steady-state diffusion coefficient (cm2/s) can be calculated according to equation 5.

ν=

z ⋅ e ⋅ ( ∆Φ )t ⎛ ( ∆Φ )t ⎞ = 1.16 E 4 ⋅ ⎜ ⎝ T ⎟⎠ K ⋅T

(5)

where Dns = non steady-state diffusion coefficient (cm2/s). Q = total charge passed (Coulombs). l = distance between the electrodes (total tickness of the specimen) (cm). Cs = surface concentration (%). ρ = density of the specimen (gr/cm3). (if not known, it can be estimated from the weight and volume of the specimen provided the weight of the steel is known). ∆Φ = drop of voltage applied (V). t = time of the test (seconds) until depassivation (with the external electrical field switched on). S: surface area of test specimen exposed to the chloride solution (cm2).

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Corrosion propagation in cracked and uncracked concrete M.B. Otieno, M.G. Alexander & H.D. Beushausen University of Cape Town, South Africa

ABSTRACT: The penetration of species that cause corrosion in Reinforced Concrete (RC) such as chloride ions, oxygen, carbon dioxide and moisture into concrete is governed by both the quality and condition of the cover concrete. The presence of cracks in the concrete cover may therefore aggravate the problem of corrosion in RC structures. This study focuses on both cracked and uncracked reinforced concrete beam specimens made using Ordinary Portland Cement (OPC) and 50/50 OPC/Corex slag blend. Crack widths of 0.4 mm, 0.7 mm and incipient cracks were investigated using 100 × 100 × 500 mm long beam specimens made with water/binder (w/b) ratios of 0.40 and 0.55. A constant concrete cover to steel of 40 mm was used for all the specimens. The tests focused on chloride-induced corrosion, using 5% NaCl solution, in order to investigate the influence of cracks on corrosion propagation. Preliminary results indicate that binder type, w/b ratio and the presence of cracks on the concrete surface influence the corrosion rate.

1

3

INTRODUCTION

One important aspect to concrete durability and performance is to attain low permeability. Long term durability of concrete is controlled by, among other factors, its ability to impede the ingress of deleterious substances into the concrete pore structure. However, it is recognized that access for deleterious agents such as chloride ions cannot be restricted indefinitely and that there is a finite service life for any reinforced concrete (RC) structure. One of the major causes of deterioration of reinforced concrete structures in the marine environment is chloride-induced corrosion of the reinforcing steel. Cracks can influence the durability of concrete structures in the marine environment by accelerating the ingress of chlorides into the concrete. The end result is to accelerate corrosion rate. The rate of corrosion directly affects the extent of the remaining service life of a corroding RC structure. The service life of RC structures may therefore be drastically reduced by the influence of cracks in combination with subsequent acceleration of degradation owing to, for example, chloride ingress. Accuracy of service life prediction can be improved if the factors that contribute to concrete degradation such as chloride-induced corrosion are well understood and taken into account.

EXPERIMENTAL

The influence of different crack widths on corrosion propagation was investigated using four concrete mixes prepared with two w/b ratios (0.40 and 0.55) and two binder types (100% CEM I 42.5 N (OPC) and 50/50 OPC/Corex slag blend). For each binder type, crack width and w/b ratio, three beam specimens (100 × 100 × 500 mm long) were cast, resulting in a total number of 48 specimens. Crushed greywacke rock (max. size of 19 mm) and Philippi dune sand were used as aggregate. A clear concrete cover to reinforcing steel of 40 mm was used for all the specimens. After 28 days of curing (under water at 23 ± 2oC) and 10 days air drying, the beams were pre-cracked using three point machine loading. The incipient cracks were produced by application of three point machine loading and then unloading. The 0.4 and 0.7 mm crack widths were thereafter maintained using a loading rig (Fig. 1). Occasional unwanted cracks were sealed using a low viscosity epoxy. 500 300 450

100

100 PVC pond

5% NaCl

40 mm cover

100

10 mm diam. bar

2

AIM OF RESEARCH

To develop accurate service life prediction models for the corrosion propagation period, an in-depth understanding on factors that affect the corrosion process is mandatory. Cracking is one such factor. This study takes a step forward to improve the understanding on the relationship between crack width and corrosion rate. This can then be quantified and incorporated in corrosion propagation service life models to predict service life more accurately.

M16 Nuts

TEST SPECIMEN

Stainless steel counter electrode

30 mm diam. steel rod

3 mm thick plate

60 M16 Bolts

Universal beam (I-Beam)

225

225 Beam cross section

40 mm cover 60

100

Figure 1. Experimental beam set-up (Loading rig and beam cross section)—Dimensions in millimeters.

157

All the beam specimens were subjected to a weekly cycle of 3 days ponding with 5% NaCl followed by 4 days drying. To ensure that the crack widths remained constant, Demec studs were placed on the concrete surface across the crack. The distance between the studs was monitored weekly using a demountable mechanical gauge at a gauge length of 100 mm. The loading rig (Fig. 1) was then used to adjust the crack widths as required. The samples were monitored weekly for half-cell potentials (Ag/AgCl), corrosion rate (coulostatic technique) and concrete resistivity (Wenner probe method).

4 4.1

RESULTS AND DISCUSSION Influence of cracks and crack width on corrosion rate

For a given binder type, cracked specimens had higher corrosion rates than uncracked ones. Irrespective of binder type and w/b ratio, corrosion rates for the 0.4 and 0.7 mm cracked specimens exceeded the 0.1 µA/cm2 mark, which is conventionally taken to mark the transition from passive to active corrosion, indicating that the presence of cracks affects both the corrosion initiation and propagation. The presence of cracks increases concrete permeability and hence eases the ingress of corrosion agents (oxygen, moisture and chlorides). Generally, corrosion rate increased with increasing surface crack width, except for the 50/50 Corex slag specimens. This may be attributed to continued hydration of the binder and chloride binding process. 4.2

Influence of binder type and resistivity on corrosion rate

Both cracked and uncracked Corex slag specimens (with higher resistivities) showed lower corrosion rates compared to the corresponding OPC specimens (with lower concrete resistivities). For a given crack width, corrosion rate was found to be binder-sensitive. Two scenarios were noted: i. For a given crack width, corrosion rate is sensitive to binder type/concrete quality; ii. For a given binder type/concrete quality, corrosion rate is sensitive to crack width.

4.3

Influence of reloading on corrosion rate

Reloading of the cracked specimens was done to reactivate the cracks, especially self-healed cracks, if any. However, after reloading, there was no noticeable change in the corrosion rates. The reloading may have been done somewhat too early for the effects of crack self-healing, if any, to be counteracted.

4.4

Half-cell potential (HCP) measurements

The Corex slag specimens had relatively high initial potential readings (> −300 mV). Such potential readings are associated to active corrosion states (> 0.1 µA/cm2) but this was not the case. The initial high potential readings for the slag specimens may be due to the limited availability of oxygen at the steel level due to its consumption in the oxidation of sulphides and thiosulphates in the slag. The interpretation of HCP readings should therefore be coupled with other corrosion test procedures such as linear polarisation resistance as was done in this study. In general, for a given binder, the uncracked specimens show lower corrosion potentials (Ecorr) than the cracked specimens. However, HCP measurement gives a better picture compared to resistivity measurement because it gives the corrosion potential state of the embedded reinforcing steel.

5

CONCLUSIONS

The following conclusions can be drawn from this study: • Cracks accelerate chloride-induced corrosion by increasing concrete penetrability. In general, corrosion rate increases with increasing crack width but is sensitive to concrete quality. For a given binder type and w/b ratio, corrosion rate increases with increasing crack width while for a given crack width, corrosion rate increases with decreasing concrete quality. • This study has shown that for concrete specimens exposed to chlorides, cracks have an effect on the corrosion rate and should therefore be taken into account in service life prediction models. • The use of half-cell potential measurements to monitor corrosion in cracked concrete was found to be more reliable than concrete resistivity measurements.

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Effect of temperature on transport of chloride ions in concrete Qiang Yuan School of Civil Engineering and Architecture, Central South University, Hunan, China Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Belgium

Caijun Shi College of Civil Engineering, Hunan University, Hunan, China

Geert De Schutter & Katrien Audenaert Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Belgium

ABSTRACT: Chloride-induced corrosion is the major durability issue of reinforced concrete structures along seacoast and in cold areas where de-icing salts are used. Various service life prediction models based on chloride induced corrosion have been developed. Temperature plays an important role in modeling chloride transport in cement-based materials. However, it is often overlooked. In this paper, the effect of temperature on non-steady-state migration and diffusion coefficients of chloride ion in concrete with water-to-cement ratios of 0.35, 0.48 and 0.6 were investigated. Non-steady-state migration coefficient was measured at 20°C and 5°C following NT build 492. Non-steady-state diffusion coefficient was measured at 5°C, 20°C and 40°C according to NT build 443. The effect of temperature on migration/diffusion coefficient is examined by using Arrhenius Equation. The results show that higher temperatures result in higher diffusion/migration coefficients. Temperatures alter the chloride penetration depth, but not the trend of chloride profile. The activation energy obtained from non-steady-state migration coefficient is quite comparable to Samson and Marchand’s results (Cement and Concrete Research, V37, 2007, 455–468), which is around 20 kJ/mol, and independent of water-to-cement ratio. However, the activation energy obtained from non-steady-state diffusion tests ranges from 17.9 to 39.9 kJ/mol, which seems dependent on water-to-cement ratio. The surface chloride concentration is also affected by water-to-cement ratio and temperature.

1

INTRODUCTION

In the physical/chemical-based model for the reinforced concrete structure subjected to chloride environment, the effect of temperature on the chloride diffusivity was taken into account, which was often expressed by Arrhenius law [Anna 1993, Tang 1996, Xi 1999]: D2 = D1e

⎛ Ea(T2 − T1 ) ⎞ ⎜⎝ RT T ⎠⎟ 2 1

(1)

Where D1 and D2 are the diffusion coefficients at T1 and T2; Ea is the activation energy, R is the gas constant. A few of the data on activation energy are available. Published data show that there is a variation in the activation energy, and most activation energies are obtained from cement paste, not concrete. Different methods using to measure diffusion coefficient might be partly responsible for the variation. There is no report was found on studying the difference between the effect of temperature on the migration and diffusion coefficients. The surface chloride concentration is an important parameter for the service life prediction of the reinforced concrete structure subjected to chloride environment, which is not an experimental

data of the chloride concentration at the surface, but a value obtained from non-linear regression analysis. The surface chloride concentration is time dependent; several empirical equations were proposed to describe the time dependence of surface chloride concentration [Amey 1998, Kassir 2002]. However, few studies were found on the effect of temperature on the surface chloride concentration. The objective of this paper is to investigate the difference between effect of temperature on migration coefficient and diffusion coefficient, and the effect of temperature and water-to-cement ratio on the surface chloride concentration.

2

EXPERIMENTAL

An ordinary Portland cement (CEM I 52.5 N), complying with EN 197-1 (2000), was used in this study. The details of concrete are shown in Table 1. The migration and diffusion tests were carried out on concrete at the age of 56 days. Migration test was performed at 5°C and 20°C according to NT build 492,

159

Details of concrete mixes.

0.8 Chloride content (concrete%)

Table 1.

3

Mix proportions (kg/m ) Mix Water Cement Gravel Sand Water reducer Slump (mm) Density (kg/m3) Air content Strength at 56 d (MPa) Porosity accessible to water (% by volume)

B6 B48 218 182 363 380 1162 1217 559 627 – – Properties of concrete 250 172 2478 2487 1.2% 1.1% 37.8 46.7

B35 140 400 1281 660 0.94%

15%

10.2%

14.1%

Mix

Temperature

0.4 0.3 0.2

0

2

4

6

8

10

12

14

16

18

Depth (mm)

Figure 1. Chloride profile of concrete at different temperatures.

dependent on water-to-cement ratio, and are lower than the published data (Goto 1981, Page 1981, Atkinson 1983, Collepardi 1972, McGrath 1996).

Activation energy (kJ/mol)

3.2

8.5* 14 15.5 20 18.15 B48 8 5.98 26.7 19.5 9.55 B35 7.5 3.46 24.9 20.6 5.65 *due to the joule effect, the temperature increased slightly.

Diffusion test was conducted at 5°C 20°C and 40°C by following NT build 443.

3.1

0.5

0.0

B6

3

5ºC 40ºC 20ºC

0.6

0.1

200 2481 0.6% 81.7

Table 2. Chloride ion migration coefficient and activation energy. Migration coefficient (×10−12 m2/s)

0.7

Surface chloride concentration

The surface concentrations obtained from diffusion tests were obtained from non-linear regression analysis. For a given temperature, the chloride surface concentration seems increases with increased waterto-cement ratio. This is partly because that concrete with higher water-to-cement ratio has more pore voids than concrete with low water-to-cement ratio. Free chloride in the pore voids is counted into total chloride. Another possible reason for this is that higher waterto-cement ratio results in higher chloride binding. The surface chloride concentration also shows temperature dependency. Increased temperature seems result in higher surface chloride concentration.

RESULTS AND DISCUSSION 4

Migration/diffusion coefficients

The results of the migration tests are given in Table 2. The results show that a higher temperature results in a higher migration coefficient. The migration coefficients measured at 5°C and 20°C were used to calculate the activation energy of chloride ion transport in concrete according to Equation 1, as shown in Table 2, which is very comparable to the data obtained from migration test ranging from 17.9 to 21.2 kJ/mol by Samson and Marchand (2007). A comparison of the chloride profiles at different temperatures is shown in Figure 1. It can be clearly seen that the trend of the chloride profiles don’t change with temperature. Only the penetration depth decreased with decreased temperature. The activation energy obtained from non-steady-state diffusion tests are 23, 17.9 and 39.9 kJ/mol for concretes with water-to-cement ratios of 0.35, 0.48 and 0.6, respectively, which seem

CONCLUSIONS

Based on above analysis and experimental results, the following conclusions can be drawn: 1. Higher temperature results in higher diffusion/migration coefficient. Temperatures alter the chloride penetration depth, but not the trend of chloride profile. 2. The activation energies obtained from migration tests and diffusion tests are different. For the concrete without supplementary cementing materials, the activation energy around 20 kJ/mol was obtained from migration tests, which is independent of water-to-cement ratio. However, the activation energy obtained from diffusion tests seems water-to-cement dependent. 3. Chloride surface concentration obtained from nonlinear regression analysis after immersion testing increases with water-to-cement ratio. Increased temperature seems result in higher chloride concentration.

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Repair of architectural concrete and related modelling of carbonation-induced corrosion H.S. Müller, E. Bohner & M. Vogel Institute of Concrete Structures and Building Materials, University of Karlsruhe (TH), Germany

ABSTRACT: The repair of architectural concrete should aim for the preservation of the original appearance of the concrete surface. The related method of concrete restoration became known under the term of “gentle concrete repair” and can be characterised by conservation and preservation of concrete structures based on local repair rather than treatment of the entire concrete surface. Such an approach requires detailed investigations of the structure as well as an estimate of the remaining life time of still undamaged sections of the structure. The reliable assessment and the prediction of the damage development are based on a probabilistic approach since it allows for the definition of boundary conditions being subject to appropriate safety levels. The procedure of prediction of the probability of carbonation-induced corrosion is demonstrated by calculating the estimated residual durability of a building facade consisting of architectural concrete.

1

INTRODUCTION

The original appearance of concrete facades and, connected to that, the impression of entire buildings is often endangered if concrete repair methods strictly follow the conventional basic principles mentioned in well established guidelines and regulations. Therefore, a so-called gentle method for repair of architectural concrete has to be applied. This particular repair method is normally applied locally and is in line with the state-of-the-art. Its use has to be especially proved. In many cases it is a more economic and sustainable alternative compared to conventional methods that are characterised by extensive and irreversible procedures such as the use of surface coatings, which underlie ageing and need to be renewed after approx. 10 to 20 years.

2

Figure 1. Detail of a locally and gently repaired concrete surface at the Norishalle in Nürnberg, Germany (built 1966–69). Colour and texture of the repair mortar are adjusted to the surrounding original concrete surface. The repaired area is marked with a dashed rectangle.

GENTLE CONCRETE REPAIR

The developed gentle concrete repair method, being required for architectural concrete, e. g. at historical concrete buildings or monuments, reconciles technical needs with requirements of preservation. Hence, the repair of damage and the rehabilitation of durability have to comply with the following conditions: − minimisation of intervention, − preservation of the original construction, − preservation of the architectural and optical appearance of the structure with its original surfaces. It is obvious that methods are unsuitable where surface coatings or overlays made of polymer or mineral materials are applied to the entire surface of the structure.

The performance of a gentle concrete repair includes some specialties. Among others, the development of suitable repair mortars that are adjusted to the mechanical and visual properties of the old concrete is mandatory. Figure 1 shows an example of a locally and gently repaired concrete surface. Colour and texture of the repair mortar are adjusted to the surrounding original concrete surface. 3

PREDICTION OF DURABILITY AND SERVICE LIFE

In cases where the deterioration of the concrete surface is caused by weathering only, an estimation of the dete-

161

relative humidity [%]

100

potential corrosion of reinforcement

corrosion probability of the reinforcement 90 …95% RH 85 …90% RH maximum relative humidity minimum relative humidity

carbonation 40

degree of optimal moisture conditions for corrosion in section c

frequency of the relative humidity in section c

concrete cover degree of depassivation concrete

depth below the concrete surface (exposed to atmospheric conditions)

reinforcement

Figure 2. Scheme for correlation of the parameters carbonation depth, concrete cover, relative humidity and corrosion probability, which all are required for the prediction of corrosion of the reinforcement in concrete.

rioration progress can be done easily by using power law models. An accurate prediction of the corrosion progress of the reinforcement is more difficult. Such a prediction has to be done both for areas that should be locally repaired and for areas that remain original. The frequent presence of moisture and oxygen is one prerequisite for corrosion of the reinforcement. Another prerequisite is the deterioration of the passivation layer of the reinforcement. At historical buildings a loss of the passivation is caused only rarely by a local ingress of chlorides, but usually by the carbonation of the concrete. Thus, for the prognosis of the future corrosion progress it is necessary to predict the carbonation development and the hygro-thermal behaviour of the concrete. The progression of the carbonation depth within the concrete cover can be estimated and extrapolated according to well-known carbonation models. If data about the concrete cover are considered additionally, the degree of depassivation of the reinforcement can be estimated by the correlation of the sample related distribution curves. The hygro-thermal behaviour of the concrete at the building facade is preferably evaluated by a numerical analysis. For the operation of the numerical analysis, the cross-section of the facade has to be modelled and realistic climatic conditions of an outdoor weathering have to be implemented. Furthermore, the material properties need to be determined in detail, usually by means of laboratory investigations. As a result of the simulations, the distribution of the moisture content is given in the course of an entire year in relation to the depth below the weathered concrete

surface (see curved lines at left centre of Fig. 2). Usually, significant changes and high moisture contents only appear close to the concrete surface. Significant corrosion rates are observed only, when the ambient relative humidity is at least 85%, but still below 100%. Therefore, noteworthy corrosion will not take place if the moisture content stays below a critical value, which can already be the case in a relatively low depth below the surface. For predicting the durability of the concrete structure, the above mentioned processes need to be combined (see Fig. 2). This procedure, i. e. the entire analysis of a service life prediction and its practical application for the facade of a historical building is shown in detail in the full-length paper.

4

CONCLUSIONS

The gentle repair method described in the full-length paper can be used very often for the restoration of historical buildings and monuments. The gained experiences show as well, that a gentle repair does not only result in an enhancement of the building’s durability and appearance but also represents an economical way of restoration. Normally, the gentle repair method is more economical and more sustainable than a conventional repair, since repair takes place only locally and extensive surface coatings, which require regular maintenance or renewal, are avoided. These circumstances do not limit the use of the gentle repair method to historical buildings but make it a suitable repair approach for any damaged architectural concrete surface.

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Determination of critical moisture content for carbonation of concrete M.O. Mmusi, M.G. Alexander & H.D. Beushausen Department of Civil Engineering, University of Cape Town, South Africa

ABSTRACT: Carbonation of concrete is one of the causes of deterioration in reinforced concrete structures. Moisture is of primary importance in the carbonation reaction, facilitating the dissolution of calcium hydroxide in the cement paste. Experimental work was carried out in order to study drying characteristics of concrete in combination with carbonation. Concrete samples of varying water/binder ratio and binder type were dried at a relative humidity of approximately 60%. The samples were simultaneously exposed to carbon dioxide at 5% concentration in air. Drying profiles of the samples were measured using the degree of capillary saturation test (DCS). The effect of drying on carbonation of the samples is observed and discussed. This paper explores the issue of critical moisture content at which carbonation ceases to occur. It concludes that moisture content becomes a governing parameter that controls the rate of carbonation on concrete exposed to drying. Testing and evaluation of results are ongoing and the results discussed here represent a preliminary analysis of the work carried out so far.

1

INTRODUCTION

1.1

Carbonation of concrete is one of the major causes of concrete deterioration in drier regions of the world (Richardson, 2002). Concrete provides a protective layer to the embedded rebar against aggressive agents. Due to its inherent nature concrete has pores through which moisture and gases such as carbon dioxide penetrate and react with calcium hydroxide present in the hardened cement paste. The carbonation reaction can be written as shown in equation 1. Ca(OH)2 + CO2 → CaCO3 + H2O

(1)

A protective film of gamma ferric oxide forms around the rebar due to the high alkaline environment at a pH of about 12.5. The reaction of calcium hydroxide and carbon dioxide yields calcium carbonate. The presence of calcium carbonate lowers the pH of concrete to approximately 9. The carbonation reaction can be divided into two primary mechanisms i) diffusion of carbon dioxide into concrete, ii) dissolution of calcium hydroxide in moisture. The carbonation reaction is known to progress favourably, between a relative humidity of 50%–75%. The moisture content in concrete has a significant role on the rate at which carbonation progresses. High amount of moisture in the pores hinders the diffusion of carbon dioxide and if the pores are too dry, insufficient dissolved calcium hydroxide will be available for carbonation.

Predicting the carbonation depth

Current models developed for predicting the rate of carbonation-induced corrosion are primarily based on the effective buffering capacity of the cementitious binder expressed as an equivalent Portland cement content as the principal parameter affecting the rate of carbonation (Bamforth and Summers, 1994). In these models the following parameters are considered as inputs i) the equivalent buffering capacity of the cementitious components ii) the C3A content of the Portland cement iii) equivalent period of wet curing iv), mean relative humidity of the environment. The models are based on many input parameters as such they have become increasingly complex. The concept of effective buffering capacity is fundamentally not new and has been used before by Hobbs, (1988) and (Parrott, 1994), but not as the unique parameter for predicting carbonation rates as shown by long-term carbonation data by Wierig, (1984), Hobbs, (1988 and 1994) and Skjolsvold, (1986). Therefore the w/b or some other determinant of the physical properties of the concrete would also be influential. This study investigates the influence of moisture on the rate of carbonation as a governing parameter. This is achieved by exploring the possibility of the presence of critical moisture content at which carbonation progresses. DCS is used to characterize the moisture content. Studies on moisture in concrete and mortars have been carried out by (Nilsson, 1980), (Sellevold, 1997) and (Relling, 1999). In their work the degree of capillary saturation (DCS) is used, which is simply the

163

ratio of the actual pore water content to the maximum pore water content, and can be expressed as (2).

60

DCS (%)

(2)

0. CE 75 M 1/ FA 0. 45 CE M 1/ FA 0. CE 6 M 1/ FA CE 0. M 75 1/ SL A G CE 0. 45 M 1/ SL A CE G M 0. 6 1/ SL A G 0. 75

Carbonation

Carbonation depth measurements were carried on the 100 mm by 100 mm concrete cubes. The influence of w/b and binder type on Carbonation depth was investigated. The carbonation depth increased with an increase in w/b as shown in figure 1. CEM1/FA binder type has a consistently higher carbonation depth for the different w/b. At 4 weeks of exposure, there was no carbonation found on the CEM1 and CEM1/SLAG at a w/b of 0.45 as shown in figure 2. For all three binder types the carbonation depth is higher at a w/b of 0.75, and this is consistent for the 3 different durations of exposure to CO2.

2.2

60 50 40 CEM1 CEM1/FA

30

CEM1/SLAG 20 10 0 w/b,0.45

w/b,0.60

w/b,0.75

Figure 1. Carbonation depth and w/b relationship after 12 weeks of exposure to CO2.

Critical moisture contents

Moisture profiles with depth were plotted for the different exposure durations, with the moisture content expressed as the degree of capillary saturation. The critical DCS values were then established for each binder type by obtaining the corresponding DCS value at the carbonation depth. After 8 weeks of exposure, the highest critical DCS of 64% is obtained with a CEM1/SLAG at 0.75, while the lowest critical DCS of 18% is obtained with a CEM1/FA at a w/b of 0.75. 2.3

d c [mm]

CE M 1

0. 45 CE M 1

Figure 2. Critical DCS values obtained after 8 weeks of exposure to CO2.

RESULTS AND DISCUSSIONS

2.1

0. 6

10 0

W1 = initial mass, W2 = mass after 10 days in water (suction saturated mass), W3 = mass after 1 week drying at 105ºC. 2

50 40 30 20

CE M 1

DCS = (W1 – W3)/(W2 – W3)

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Critical moisture content for carbonation

The concrete prisms were allowed to dry under the different relative humidity. The mass loss was measured daily until constant mass was reached. Carbonation of the concrete prisms was measured after 3 weeks of exposure to CO2. Carbonation had occurred only in the 80% RH environment while there was no carbonation observed in the other RH environments. Carbonation occurred rapidly at the lower relative humidity of 80%, due to the availability of CO2 to the dissolved calcium hydroxide. From the above findings it can be suggested that 80% RH is the critical moisture content for carbonation.

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Effect of environmental parameters on concrete carbonation. DURACON collaboration (Mexican results) E.I. Moreno & C. Vinajera-Reyna Universidad Autónoma de Yucatán, Yucatán, México

A. Torres-Acosta Instituto Mexicano del Transporte, Sanfandila, Querétaro, México, DURACON Mexican Coordinator

J. Pérez-Quiroz & M. Martínez-Madrid Instituto Mexicano del Transporte, Sanfandila, Querétaro, México

F. Almeraya-Calderón & C. Gaona-Tiburcio Centro de Investigaciones en Materiales Avanzados, Chihuahua, México

P. Castro-Borges & M. Balancan-Zapata Centro de Investigación y de Estudios Avanzados-IPN, Yucatán, México

T. Pérez-López & M. Sosa-Baz Universidad Autónoma de Campeche, Campeche, México

E. López-Vázquez Instituto Tecnológico de Oaxaca, Oaxaca, México

E. Alonso-Guzmán, W. Martínez-Molina & J.C. Rubio-Avalos Universidad Michoacana de San Miguel de Hidalgo, Morelia, Michoacán, México

L. Ariza-Aguilar Universidad Veracruzana, Boca del Río, México

B. Valdez-Salas Universidad Autónoma de Baja California, Baja California, México

D. Nieves-Mendoza Universidad Autónoma de Tamaulipas, Tampico, México

M. Baltazar Universidad Veracruzana, Jalapa, México

O. Troconis-Rincón Universidad del Zulia, Maracaibo, Venezuela, DURACON International Coordinator

ABSTRACT: This work is part of the DURACON Ibero-American project, aimed to characterize concrete durability under environmental conditions by natural exposure in two different atmospheres (marine and urban) for each of the 13 countries involved in the project. Specimens were exposed in 15 Mexican sites (9 urban and 6 marine atmospheres). Concrete specimens were 15 × 15 × 30 cm with two concrete mixes (water/cement ratios of 0.45 and 0.65). Six reinforced and six plain concrete specimens were placed on each exposure site. Environmental data was collected, including rainfall, relative humidity, time of wetness, temperature, wind velocity, and carbon dioxide/chloride concentrations in the atmosphere. Concrete resistivity, carbonation depth, and chloride concentrations in concrete were measured. Electrochemical characterization of the rebar was also performed by means of corrosion rates and half cell potentials. This investigation is focused on the effect of the environmental parameters inducing corrosion due to concrete carbonation during the first four years of exposure. A discussion regarding corrosion initiation period, with clear influence of the specific atmospheres tested, is also included.

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1

Table 1. Carbonation coefficients from concrete specimens (w/c of 0.65) after three years of exposure.

INTRODUCTION

Corrosion of reinforcing steel is one of the main causes of deterioration in concrete structures. Every year, a large amount of money is spent on maintenance and repair due to unexpected corrosion damage. Investigations related to corrosion of reinforcing steel have been particularly focused on aspects such as causes and mechanisms of deterioration, development of electrochemical techniques, and methods of protection. The need of quantification of the structure’s remaining service life affected by corrosion has been recognized recently. Thus, the interest has been focused on the development of models which allowed us to predict the service life of reinforced concrete structures in order to effectively program maintenance actions. Still, the localized environmental loading may create microclimates where the performance of a concrete element may be different from one region to another. Without taking into account this effect, service life prediction would be inadequate. In order to determine the impact of the localized environmental loading, a Project called DURACON was developed. Eleven countries are participating including Mexico, which has 15 stations in 13 cities, 8 urban and 7 marine. This work presents the results relative to concrete carbonation after three years of exposure. The progress of the carbonation front has been modelled as: x22 − x12 = k2 ⋅ (t2 − t1)

(1)

2.1

METHODOLOGY Materials for concrete specimens

Two concrete mixes were made with w/c of 0.65 and 0.45. The amount of water used for mixing was 185 lts/m3. 2.2

Physical characterization of reinforced concrete specimens

In each station, 6 reinforced and 6 plain concrete specimens were placed. The troweled surface was placed downside to avoid preferential ingress of the aggressive agents. One lateral face was directly exposed to the predominant winds. 2.3

Station

Exposed face

Opposite face

Mérida Chihuahua† Oaxaca Morelia† Mexico City Querétaro‡ Toluca Mexicali Progreso Veracruz‡ Campeche-1 Campeche-2 Campeche-3

3.0 6.1 5.5 6.7 5.9 1.7 5.0 2.7 4.0 3.6 3.2 2.5 3.2

2.8 5.6 5.8 6.9 3.8 2.5 5.5 1.0 5.6 2.3 3.7 2.8 3.5

†

Four years of exposure, ‡ Two years of exposure.

the methodology given in ISO 9233 for the aggressiveness classification of the atmosphere. The most important parameters to classify the environment were the relative humidity, the time of wetness, the carbon dioxide concentration, wind velocity and direction, rainfall, and temperature. 3 3.1

where x1 is the carbonation depth at time t1, and x2 is the carbonation depth at time t2. Concrete structures with k lower than 3 mm/year½ are considered of high quality, meanwhile concrete structures with k higher than 6 mm/ year½ are considered of low quality. Based on the impact of the environmental loading during the first year of exposure an analysis of carbonationinduced corrosion potentiality is presented, followed by an evaluation of the corrosion initiation time using the carbonation coefficients. 2

Carbonation coefficient (mm/year½)

Characterization of environmental parameters

The evaluation of the environmental parameters in each station during the evaluation period was performed following

RESULTS AND DISCUSSION Environmental and meteorochemical parameters

Based on the environmental parameters from the first year, carbonation-induced corrosion is likely to happen first in Mexicali, followed by Chihuahua, Toluca, Mexico City, Queretaro, Oaxaca, Morelia, and Merida. 3.2

Evaluation of carbonation and corrosion parameters

Taking into account only the exposed face, it is observed that the maximum carbonation depth occurred for the specimens exposed in Morelia, followed closely for those exposed in Chihuahua and Mexicali, followed by those exposed in Mexico City, Toluca, Oaxaca, Queretaro and Merida. Considering the carbonation coefficients, it can be observed that the highest carbonation rates are from Morelia and Chihuahua, with k values above 6 mm/ year½, followed by those exposed in Mexico City, Oaxaca, Toluca, and Merida, with values between 3 and 6 mm/year½. Mexicali and Queretaro are presenting values lower than 3 mm/year½. 4

CONCLUSIONS

• For the same type of concrete with a w/c ratio of 0.65 and 15 mm of concrete cover, the estimated initiation time for active corrosion of the reinforcement varied between 5 and 41 years, depending on the exposure site. • This study emphasized the need to characterize the environmental loading in order to predict the service life of concrete structures.

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Improvements in durability design of concrete bridges in Croatia P. Sesar, S. Pastorcic & Z. Banic Civil Engineering Institute of Croatia, Zagreb, Croatia

ABSTRACT: Seventies, of the twentieth century, in Croatia, characterize the significant increase in design and in velocity of construction processes of bridge structures built of precast elements. During this period concrete has been considered almost eternal and indestructible construction material, and a very little attention was paid to durability of bridges and details design. During eightieth years, awareness of disadvantages of concrete structures change as well as the approach to durability design. Ninetieth years bring accelerated highway construction as well as work on the large number of structures. During this period, main criterion in bridge structure design is rapidity and simplicity of construction in spite of obvious disadvantages. Nowadays, new European codes have been accepted and expected service life of bridge structures is between seventy and hundred years. The pros and cons of different structures considered are discussed and improvements applied in design, based on rebound information of structural behavior and durability will be presented.

1

INTRODUCTION

The period between the WWII and the end of the 60’s was marked by reconstruction and betterment of roads and the aim to establish normal traffic between Croatian towns. At that time, the Adriatic motorway, which connects the most important coastal towns Rijeka, Zadar and Split with the capital Zagreb, is being built. These construction undertakings represent the beginning of a period of intensified construction, introduction of new technologies and viewing of concrete as indestructible material. At the end of the 1980’s, durability design and regular maintenance of structures started to receive more attention in Croatia. This upswing in the development of bridge management systems was at its very beginning arrested by the war for independence (1990–1995). During the war, the traffic infrastructure maintenance funds were almost non-existent. During past five years new bridge management system, developed by Croatian experts, is introduced in practice.

2

DESIGN AND CONSTRUCTION DURING 1970’S

At the beginning of the 1970’s, a boom in road and bridge construction occurred in Croatia. During this boom, a great number of prestressed concrete bridges were constructed. Apart from building on scaffolding, which was a predominant method until then, combined construction method with precast elements was also used. Since this period was characterized by the construction of concrete bridges, and there were few specialized companies, new companies dedicated exclusively

Figure 1.

Structure without closed drainage system.

to bridge construction are starting to emerge. These circumstances will leave its trace on the structures, so later inspections often reveal lack of experience in construction, which consequently led to reduced structural durability. During this period, concrete is considered as almost indestructible material. The only measures for ensuring structural durability are the determination of concrete cover, type of cement, water—cement ratio and crack width limitation. This type of designing proved unsatisfactory in terms of bridge service life, especially on structures that were built badly or where details were designed inappropriately (Fig. 1.).

3

DESIGN AND CONSTRUCTION DURING 1980’S

Regarding a growing number of structures that had to be built in a short period of time, preference was given to the use of as many precast bridge elements as possible, which are manufactured in a factory, delivered to the site and easily placed.

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Figure 2.

Inappropriately designed and placed equipment.

During the 1980’s, there was a marked tendency to reduce element dimensions to their statically necessary minimum. Longitudinal and transverse layouts of the elements are often adjusted to available girders, and not to the characteristics of the obstacle to be bridged. Details of joints, supports and bridge equipment are sometimes designed inappropriately (Fig. 2.). On some structures, the wish to build quickly and cheaply and use their own technology on the part of some constructors prevailed at the expense of quality. The problems of durability due to non-uniform quality of the construction of individual bridges were observed. These trends are not specific only for Croatia, they have been noticed worldwide.

4

DESIGN AND CONSTRUCTION FROM THE 1990’S UNTIL TODAY

life with satisfying performance level, bridges should be designed as continuous over intermediate supports, except in special cases. In order to minimize damages from run-off water, closed drainage systems have been used in new projects. Great attention should be placed on maintenance and in-time replacement of expandable (rubber and steel) parts in expansion joints. Bearings have to be properly designed with sufficient bearing capacity.

6

Taught by the events from the 1980’s, and faced with the challenge to construct new modern roads, or precisely 15 new sections of the motorway in the length of 350 km, the investors decided to embark upon a project of typification of structures. It was determined that in these conditions, when such a great number of structures was being designing, it would not suffice to give only the basic guidelines available before, but that it is necessary to work out standard designs, up to the level of the main design.

5

Figure 3. Characteristic bridge built as typificated structure.

IMPROVEMENT IN DESIGN

Experience gained in past few decades, refers on shortened durability of bridges made of series of precast simply supported girders. Nowadays, tendencies are to design and built construction with lesser number of expansion joints and to integral construction. Disadvantages on some constructions are shown very early and they were removed afterward by improved design solutions. Continuous systems have proved more durable then simply supported structures. In order to achieve service

CONCLUSION

Today, special attention is paid to achieving durability. It affects the reliability of existing structures, and the amount of maintenance costs, which grow exponentially in low quality structures. If we consider the number of bridges, their age and expected service life, the funds needed for such repairs in Croatia are considerable. Typification of culverts, bridges, viaducts and overpasses was carried out in order to standardize the quality of designing, construction, quality control and maintenance. Standard designs were made so they would serve as a baseline to determine the standard level of design elaboration, while standard solutions allow better preparation and use of equipment available to the constructor and better trained personnel. Furnishing details have a significant impact on deterioration of the whole structure. Traditional design methods regarding durability are based on allowable limit principle such as minimum concrete cover, maximum water/binder ratio, minimum cement content and limitation of cracks. New strategies are based on minimising deterioration threatening, for a specified period of use, by selecting optimal material composition and structural detailing.

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Fire engineering in Croatia D. Bjegovic´ & M. Jelcˇic´* Faculty of Civil Engineering, University of Zagreb*/Civil Engineering Institute of Croatia, Zagreb, Croatia

M. Carevic´ Faculty for Safety Measures at Work, Zagreb, Croatia

M. Drakulic´ Brodarski Institute, Croatian institute for advanced technologies, Zagreb, Croatia

B. Peroš Faculty of Civil Engineering and Architecture, University of Split, Split, Croatia

ABSTRACT: Fires obviously cannot be prevented, but structures can be designed, built and maintained in a better way. In order to achieve this, an adequate system should be established which takes decades and includes many segments. This primarily refers to the development of legislative and technical regulations, development of an education system (education of future design engineers and persons responsible for system implementation and development of scientific research), creating of a system for monitoring of the designed measures, etc. For the above reasons, the Civil Engineering Faculty of the University of Zagreb and the institutions it cooperates with organized one-year (two-semester) specialist post-graduate study of Fire Engineering which will be described in this paper. This multidisciplinary study will provide students a complete overview and necessary knowledge of Fire Engineering which includes structural, material, sociological, physical, economical and ecological aspects as well as risk assessment methodologies.

Unwanted fire is destructive force that causes many thousands of deaths and billions of property loss each year. According to data of World Fire statistics series, published by Geneva Association, fires, including its prevention, protection and repression costs, charge economies about 1% GDP, but these costs vary widely between countries (International Association for the Study of Insurance Economics 2007). The similar situation is in Croatia where 30–40 people are injured in fires and material losses are assumed to be 40–50 mills.€ every year (Croatian Ministry of the Interior 2006). That is why the effect that fire has on building structures has been given the same status in newer European legislative and technical regulations as other standard structure loads. Obviously, fires cannot be prevented, but structures can be designed, built and maintained in a more proper way which would increase their safety. This makes safety in case in fire more and more important issue related to construction works and products. Disastrous consequences of fires caused early development of legal and technical regulations concerning fire safety in European and other world countries. Each country defined its regulations generally based on its own experience of the fire safety problem (historical consideration, previous experience, expert judgment etc.). Due to the limitations of such prescriptive design methods, a lot of European research

projects concerning fire safety were performed in last ten years. As results of those researches, knowledge about fire safety became very complex and grew up into a new engineering discipline called Fire Safety Engineering. Fire safety engineering is an application of scientific and engineering principles to the effects of fire in order to reduce the loss of life and damage to property by quantifying the risks and hazards involved and provide an optimal solution to the application of preventive or protective measures (Purkinss 2007). The main objective of specialist study of Fire Engineering is to develop appropriate scientific methods for evaluation of fire effect on human beings and structures in particular situation, and in the same way to increase fire safety of human beings and structures. It is of great importance to define the border for so called acceptable and real risk on one and economy of applied measures on the other hand. It is especially visible when compare the fire safety measures prescribed by current regulation with those assessed by Fire Engineering methods. As a general rule, there is not evident selectivity in regulation where certain measures are prescribed by legislator ˝sense˝ which often leads to exaggeration (especially in fire resistance of structure elements) or to mutual discrepancy within the same regulation and often within fire safety regulations as whole. In that sense, Fire Engineering methods are consistent, because

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they are based on scientific principles and offer the best safety solutions that are economically acceptable for investors. From the academic year 2006/2007, postgraduate specialist study Fire Engineering exists also in Croatia. It is founded at Zagreb University, Faculty of Civil Engineering Faculty in cooperation with other faculties and institutions from Croatia. The obligatory courses are: Thermodynamics of fires, Fire performance of construction materials and elements, Fire safety of structures, Architectural and civil engineering measures of fire protection and Active fire protection systems (fire detection systems, electro-installations, ventilation and water suppression systems) The elective courses are: Fire modeling, Human behavior in fires, Building code requirements for fire protection, Fire risk management and Research methodology. Structural Fire Engineering is one important aspects of a comprehensive fire protection design. New approach in fire protection area encompasses determination of structure fire resistance where new scientific methods for real fire calculation are applied. These issues are described in European standards for load bearing structures, while principles and recommendations are covered by Eurocodes (part 1–9). The postgraduate specialist study Fire Engineering gives educational support in area of fire resistance of structures through the courses Fire safety of structures and Fire modelling. The tendency is to encourage the engineers to apply advanced calculation models of structure behaviour in case of fire occurrence in compare to present approach based on standard fire testing. The core of new principles is that the influence of real fire (high temperatures) must be considered as a possible action in process of analysis of bearing structures safety. For Fire Engineering education, application of calculation methods for structure behavior in fire is demonstrated on various buildings, such as various types of garages, shopping centers, sports hall, etc., (Figure 1). All activities mentioned above are of great importance for everyday ongoing projects as well as for large future infrastructural projects in Croatia and Region. For example, through the realization of the very ambitious plans from document titled “Strategy of development for road network in Croatia” (1996), many kilometers of modern highways is already built

Figure 1. Example—Calculation of heat release rate (HRR or RHR) in one garage.

Figure 2.

Fire tests in Mala Kapela tunnel (5.760 m).

and will be built in Croatia, enable the best possible connections with all European road directions, putting Croatia in the first plan as important point of Pan-European road network (Figure 2). The mentioned projects, which only represent the part of the actual investments in the civil engineering and industry area in Croatia and Region, give the good and prospective base for our fire safety specialists to apply their new knowledge and skills. Fire Engineering principles are based on prevention and management. But if the prevention cannot be achieved, the exposures to fire must be addressed and fires must be managed. In worst case, if structures are damaged by fires, great knowledge of repair of firedamaged constructions is required. That’s why fire safety science is rapidly expanding multidisciplinary field of study that requires the integration of many different fields of science and engineering.

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Modelling of reinforcement corrosion

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

DFG Research Group 537: Modelling of reinforcement corrosion – An overview of the project P. Schießl & K. Osterminski Centre for Building Materials of TU München, Germany

ABSTRACT: A strong interest for the durability of reinforced concrete structures currently exists in industry and research. Against the background of immense costs for maintaining reinforced concrete structures and repairing damages caused by corroding reinforcement steel, this interest lead to a German joint research project. The aim of this network-based (www.dfg-for537.de.vu) research group is, to develop reliable models for the corrosion propagation and to make a probabilistic service life design tool available for engineers so that a complete design for durability, concerning reinforcement corrosion, will be possible.

1

MOTIVATION

The durability of reinforced concrete structures is a key task for engineers throughout planning and execution. To ensure their durability reinforced concrete structures are currently designed by means of minimum requirements for concrete quality and concrete cover. For critical environmental actions, like chloride attack on horizontal concrete surfaces, only a requirement for special measures like surface coatings exists. A performance based design for durability, analogous to the design for loads, is currently missing. During the last few years, new probabilistic service life design models have been developed, allowing performance based design related to depassivation caused by carbonation and chlorides in uncracked concrete. These design models have already been successfully used in design tasks (Schießl & Gehlen 1998, Schießl & Gehlen 2000). For the period after the depassivation—the corrosion process of steel in concrete—no practical design models can actually be found. On the one hand, electrochemical models with a high grade of complexity (Sagües et al. 1993, Noeggerath 1990, Bažant 1979) exist. These models are useless for the planning engineer because of the lack of information concerning model parameters for real construction situations. On the other hand, rules of thumb exist which do not portray the real corrosion process properly, e.g. a design model only based on the influence of the resistivity of concrete (Alonso et al. 1988). Using these models will lead to a false interpretation of the corrosion state in many practice situations. On the basis of this situation, durability often is ensured by exaggerated measures in case of new construction or repair. In many cases, chloride containing concrete has been excessively removed or the structure has been claimed to be demolished partly or completely. This shows that the knowledge concerning the damage propagation caused by corrosion of

steel in concrete is insufficient. There are also other cases in which the applied measures and construction details do not suffice to ensure the intended service life. Critical situations are underestimated which may lead to severe problems. International research in the field of corrosion of steel in concrete of the last few years has shown a number of investigations concerning the basic knowledge of the mechanisms. As already mentioned, practice oriented design models based on these results are still missing. In practice different corrosion elements with different geometries occur (micro- and macro elements with extremely varying anode/cathode ratios). To achieve a realistic picture of the state of reinforcement corrosion, it is necessary to model the same or similar geometrical parameters. Apart from some exceptions, this has not been done so far. Against this background, the German Research Foundation (DFG) funds a research group in order to quantify all influences on reinforcement corrosion and to develop a service life design model. Environmental influences as well as the construction’s resistances and the resulting corrosion propagation are accompanied by large scatter. That is why a design model can only be successful and practicable if it is on a probabilistic basis. Consequently, a probabilistic design model can assure a common and compatible safety level like the modelling of reinforced concrete structures for loads, e.g. EUROCODE 0 (Basis of Design). Herein the constructions resistances as well as the stresses have to be quantified statistically. The task of the research group is to quantify probabilistically all influences of the reinforcement corrosion, so that integration into this safety concept is possible. By the end of this joint research project and in combination with the existing probabilistic design models for the depassivation phase of uncracked concrete, a complete picture of the corrosion propagation can be given and an entire service life design of reinforced

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concrete structures—including all limit states—will for the first time be available.

2

STRUCTURE OF THE PROJECT

The participating research institutes have particular knowledge and experiences in special fields of the corrosion of steel in concrete. The institutes cover the whole spectrum of competence that is necessary for the work in this research group. A structure of the whole research project is given in Figure 1. In project area A, a quantification of all influence parameters on the corrosion propagation will be done for the first time. Here, particular attention will be paid to the time-dependence of the system parameters of the electrical circuit model presented in Equation 3 of the full length paper. The resulting corrosion rate and the dynamically developing parameters of the corrosion systems will be inserted into the model developing process. The volume of the necessary investigations in project area A leads to a division into four subprojects. Those results will later on be combined in a model of corrosion propagation. Subproject A1 is processed at the Federal Institute for Materials Research and Testing (BAM), subproject A2 at the Centre for Buildings Materials (ibac) at RWTH Aachen, subproject A3 at the Centre of Building

Materials (cbm) at TU München and subproject A4 at the Materials Testing Institute (MPA) of University of Stuttgart. The consequences which result from the ongoing corrosion process will be examined in Project area B which is divided into two subprojects. Subproject B1 is processed at Institute of Concrete Structures and Building Materials (IfMB) at University of Karlsruhe, whereas subproject B2 is handled at Institute for Construction Materials (IWB) at University of Stuttgart. All knowledge gained in both project areas A and B will be merged into a durability design model for the damage propagation of reinforced concrete structures in project D which is mainly processed at Centre for Building Materials (cbm) of TU München. Detailed descriptions of the projects listed in Figure 3 will be given in separate papers of the proceedings to this symposium. ACKNOWLEDGEMENT Further information about the DFG research group 537 modeling reinforcement corrosion can be found on the internet at http://www.dfg-for537.de.vu. The authors wish to thank the DFG for funding the project and supporting the research issue. REFERENCES

Figure 1.

Structure of the DFG research group 537.

Alonso et al. 1988. Relation between resistivity and corrosion rate of reinforcement in carbonated mortar made with several cement types. Cement and concrete research, 18(1988)5, pp. 687–698. 1988. Bažant 1979. Physical model for steel corrosion in concrete sea structures—Theory and application. Journal of the structural division 105. ST6(1979)2, pp. 1137–1153, pp. 1155–1166. 1979. Noeggerath 1990. Zur Makroelementkorrosion von Stahl in Beton: Potential- und Stromverteilung in Abhängigkeit verschiedener Einflussgrößen. Dissertation, ETH Zürich. 1990. Sagües et al. 1993. Computer modelling of corrosion and corrosion protection of steel in concrete. Concrete 2000, pp. 1275–1284. 1993. Schießl & Gehlen 1998. DuraCrete: Modelling of degradation. Document BE95–1347/R4–5. Schießl & Gehlen 2000. Betondauerhaftigkeit—Um-setzung des Wissensstandes in Bemessungskonzepte. 14th ibausil, 20.–23. September 2000, pp. 1–001—1–021. Weimar. Tuutti 1982. Corrosion of steel in concrete. CBI research (1982). No. Fo 4:82, Stockholm. 1982.

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DFG Research Group 537: Modelling reinforcement corrosion – Investigations on the mechanism of cracking and spalling E. Bohner, N. Soddemann & H.S. Müller Institute of Concrete Structures and Building Materials, University of Karlsruhe (TH), Germany

ABSTRACT: Cracking and spalling of the concrete cover due to the corrosion of the reinforcement in concrete define important limit states of durability and stability of concrete structures. The mechanism of fracture and the stresses, which cause cracking of the concrete are still unknown and can only be simulated incompletely so far. Thus, a realistic and quantitative description as well as a reliable prediction of the damage process due to corrosion of the reinforcement is not possible. Therefore, the aim of the presented work is the development of an analytic prediction model, which describes the time dependent damage process of cracking and spalling due to corrosion under realistic conditions. Since the accuracy of such an analytic model depends primarily on the quality of the implemented constitutive models, a main target of the work is to determine the mechanical properties of the corrosion products by applying novel experiments and inverse analyses. By means of numerical investigations, a model of the damage process was developed, which allows for the detailed analysis of the stresses and strains, as well as for the formation of cracks within the concrete cover. Based on parameter studies of several specimens, which are subjected to different corrosive conditions, the analytic prediction model for the time dependent damage process can be derived. This model will enable the prediction of damage development under conditions of practical relevance and serves as a part of a complete probabilistic design approach for durability of reinforced concrete structures.

1

INTRODUCTION

Although research on reinforcement corrosion in concrete has been extensive in the last decades, research on the corrosion effect on structural serviceability remains unsatisfactory. So far, most of the analytical investigations in this field were carried out by the use of finite element analyses based on models, which describe the process of concrete cover cracking only in a qualitative way. Aspects of fracture mechanics and time dependent effects as shrinkage and creep of the concrete were mostly neglected. A major deficiency in previous investigations is the ignorance of a constitutive model for rust. Thus, it is impossible to evaluate realistic time intervals for the appearance of corrosion damage. The experimental investigations, which imply novel experiments, aim to determine the mechanical properties of the corrosion products that develop under different conditions at the reinforcement. Afterwards, a numerical model is presented, which allows for a realistic evaluation of the stresses and strains inside test specimens that have been produced within the experimental program. 2

EXPERIMENTAL INVESTIGATIONS

By means of specific experiments, the time dependent crack development due to reinforcement corrosion is currently monitored and quantified at 275 concrete cylinders (denoted as corrosion cylinders in the following) with centrically embedded reinforcement bars. Next to different corrosive agents (chloride and carbonation-induced corrosion), particularly

the influence of material parameters (cement type, w/c-ratio) and geometrical characteristics (diameter of the reinforcement bar, concrete cover thickness) on corrosion are in the focus of the investigation. A homogeneously distributed corrosion of the reinforcement bar was realized by means of adding chloride to the fresh concrete or by an accelerated carbonation of concrete. Finally, all specimens were directly exposed to cycles of drying and wetting. Upon selected specimens the tangential deformations on the surface of the concrete as well as the chronology and intensity of the internal corrosion pressure are measured. The measurements help to identify and quantify the effects of reinforcement corrosion and allow for the determination of the mechanical properties of the corrosion products. The measurements are performed on two different kinds of specimens. At first, tangential deformations due to corrosion of the centrically embedded reinforcement bars are measured with strain gauges that are glued to the concrete surface along the perimeter of the corrosion cylinders. In addition, further measurements take place at modified corrosion cylinders (denoted as hollow cylinders), where the reinforcement bar is replaced by a thin-walled copper tube (wall thickness 0.5 mm) that is filled with hydraulic fluid (see Fig. 1). As with the regular corrosion cylinders, the tube causes a resistance against shrinkage of the concrete resulting in a residual stress condition inside the specimen. While at the regular corrosion cylinders the internal pressure is caused by the corrosion progress, it is applied

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ycorr effective expansion of the corroding reinforcement bar

linear loading function bilinear loading function multilinear loading function

filling of cracks filling of voids

time t shrinkage and creep loading from corrosion

Figure 2. Linear, bilinear and multilinear loading function.

Figure 1. Test facility for application and control of a hydraulic pressure inside a thin-walled copper tube being centrically embedded inside a concrete cylinder. A surrounding steel frame prevents longitudinal deformations.

hydraulically inside the hollow cylinders. The level of the hydraulic pressure has to be chosen in such a way that similar deformations of the specimens occur as measured and observed at the associated corrosion cylinders simultaneously. By means of the assumption, that the hydraulic pressure and the internal pressure due to corrosion cause the same load on the concrete, the time-dependent effects from shrinkage and creep can be eliminated. Thus, the chronology as well as the level of the stresses caused by corrosion can be determined. 3

NUMERICAL INVESTIGATIONS

The numerical investigations are on the one hand carried out to determine the stresses and strains that can be expected during the experiments. On the other hand, they allow for a detailed analysis of the governing processes during cracking and spalling for different test parameters, and an evaluation of any possible reinforcement configuration of practical relevance. The corrosion process and hence the loading that results from the corrosion is a complex procedure,

which is governed by the corrosion rate and the volume of the rust relative to the volume of the steel (volume ratio). This difference in volume of the corrosion products compared to the original steel leads to an impeded increase of the reinforcement bar diameter and therefore to stresses in the surrounding concrete cover. It is however possible, that part of the corrosion attack does not lead to stresses in the surrounding concrete as corrosion products migrate into voids around the steel surface (transition zone) or into emerging crack spaces in the concrete. Based on the knowledge of the corrosion rate, the volume ratio and the volume of voids in the transition zone a bilinear loading function can be determined. By analysing the stable crack propagation during the numerical calculations a multilinear loading function that additionally considers the migration of corrosion products into significant cracks can be developed (see Fig. 2). 4

CONCLUSIONS AND OUTLOOK

The presented theoretical and experimental approach allows for estimating the level of stresses and the chronology of appearance of cracks inside the concrete cover, which are caused by the corrosion of the reinforcement. The approach is based on novel experimental investigations being supplemented with numerical analyses. By means of parameter studies with numerous specimens a prediction model for the time dependent damage process can be derived. It serves as a part of a complete probabilistic design approach for durability of reinforced concrete structures, which is a future target of the DFG research unit No. 537 of the German Research Foundation (DFG).

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DFG Research Group 537: Modelling of reinforcement corrosion – Macrocells and time dependence M. Raupach, J. Warkus & J. Harnisch Institute of Building Materials Research (ibac), RWTH Aachen University, Aachen, Germany

ABSTRACT: Subproject A2 of this joint research project investigates macrocell formation with the main emphasis on geometrical influences. Another important aspect that will be investigated within the next years is the time dependence of the anodic activity and the macrocell current. Hence, this subproject provides key data for the final design model. The investigations on macrocells were carried out in laboratory experiments and additionally in numerical analyses to increase the available database especially with respect to geometrical influences. Thereby all specimens were designed to represent practical conditions, i.e. slabs or beams with a localised depassivation. Based on these results, it is possible to model the macrocell current under steady-state conditions. Chloride induced corrosion at real structures, however, is influenced by many time-dependent factors, i.e. temperature and polarisation behaviour of depassivated reinforcement parts. In order to implement the time dependence into the probabilistic design model extensive laboratory tests on about 110 macrocell corrosion specimens are actually carried out. During a time span of two years, electrochemical measurements and visual inspections are conducted on the test specimens to investigate especially the time-dependent behaviour of the anodically acting parts. The determination of the characteristic corrosion patterns and metal removal of the anodic areas is achieved by a unique three-dimensional optical technique using high-resolution digital cameras. 1

INTRODUCTION

The corrosion process—especially in case of macrocell formation—depends on a lot of different boundary conditions which have to be considered to develop a design model for the remaining lifetime after depassivation. A major issue is given by different geometrical arrangements of diverse structural members like beams or slabs. As it is almost impossible to cover this wide range of geometrical varieties by a laboratory test programme additional investigations by numerical methods have been carried out. First the numerical models were successfully calibrated by laboratory measurements. Afterwards parameter studies were carried out to investigate the influence of selected parameters. A numerical simulation allows to study galvanic systems under constant steady-state conditions, but for a damage model one has to consider the time dependence of the corrosion process. Especially the anodic polarisation behaviour is time-dependent because corrosion pits at the surface of depassivated steel bars can change their quantity and size. In terms of a time-dependent damage model these issues will be investigated in a comprehensive testing series. 2

2.1

MACROCELL CORROSION AND GEOMETRICAL EFFECTS

constant for all specimen. The concrete composition is conform to the standard composition of the research group and the exposure conditions were chosen at 20°C and 95% RH to prevent the specimens from desiccation. 2.1.2 Specimens To cover a wide range of different structural members, specimens with different geometries were fabricated. The test series varies from simple beam-like specimens with two parallel reinforcement bars to more complex slab-like specimens or girders with three layers of reinforcement and additional stirrups. In this paper one example of a slab-like specimen with a length and width of 140 cm, a thickness of 20 cm and two layers of reinforcement is presented. In a single corner of the upper reinforcement layer chlorides were added to the mixing water. 2.1.3 Simulations All simulations were carried out by use of a boundary element programme called BEASY CP. As one result the potential distribution can be visualised (figure 1). For an estimation of corrosion rates and the remaining lifetime the macrocell current is of main concern. A comparison shows that the difference between calculated and measured macrocell current is in most cases less than +/–10%, indicating that the galvanic systems were simulated correctly.

Laboratory specimen

2.1.1 General As the focus is directed on geometrical effects the concrete composition and exposure conditions are kept

2.2

Parameter study

Based on these results it becomes possible to study macrocells under different boundary conditions to

177

Figure 1. Calculated potential distributions of the slab-like specimen.

Table 1.

Figure 2. Cell factor vs. resistivity (slab-like specimen, U = 400 mV).

Parameter study.

Geometry

Slab-like

Resistivity [Ωm] Driving voltage [mV] Cathode/anode surface area ratio –

100, 500, 1000, 10000 300, 400, 500 53:1, 96:1, 192:1

derive needed parameters for the engineering model of the research group. Subsequently, a brief example for the slab-like geometry (fig. 1) is presented. For this purpose the boundary conditions have to be varied within a practical range. Therefore a parameter study with the values given in Table 1 was carried out. The working hypothesis of the research group is subsequently briefly repeated: U I macro = ra rc + +k e ⋅ ρe Aa Ac

Figure 3. 3D optical scan of a steel surface damaged due to pitting corrosion (left), software based analysis of pit distribution and depths (right).

geometries, further research work will focus on simulations of other practical geometries. (1) 3

where Imacro = macrocell current; U = driving voltage; rx = specific polarisation resistance; Ax = electrode surface area; ke = cell factor: ρe = concrete resistivity. The resistances and the cell factor ke can easily be derived from the simulation output files. Figure 2 shows the cell factor for different concrete resistivities and cathode to anode surface area ratios. The dependence of the cell factors on the resistivity and the cathode to anode surface area ratio is comparably weak. Thus the macrocell current beneath a concrete resistivity of 1000 Ωm (high corrosion rates) can be estimated on the safe side using equation 1 with a cell factor ke of 3,0 m−1. This is an example, how numerical analyses can help to derive necessary input parameters for the working hypothesis of the joint research group. As the presented results are only valid for one specific

TIME-DEPENDENT EFFECTS

In terms of a lifetime assessment the time-dependent system parameters like driving voltage or polarisation resistances will be investigated. In order to achieve this an extensive laboratory test series has been initiated. Based on a defined reference case selected boundary conditions such as temperature, cement type, water application and chloride content will be varied. The influences on the above mentioned time-dependent model parameters and on the actual damage will then be studied by continuous and periodical electrochemical and destructive investigations. The detailed surface analysis of the dismounted anode bars will be achieved by means of a high resolution 3-D camera technology. The created data set of the scanned anode surface will then be analyzed using suitable software packages with respect to total mass loss, pit distributions and pit depths, see figure 3.

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DFG Research Group 537: Modelling of reinforcement corrosion – Validation of corrosion model S. von Greve-Dierfeld, C. Gehlen & K. Menzel MPA Materialprüfungsanstalt Universität Stuttgart (Otto-Graf-Institut(FMPA)), Stuttgart, Germany

ABSTRACT: Corrosion of steel reinforcement in concrete is substantially governed by macro cell activity whereas the effective range of cathodic areas is strongly dependent on the electrolytic resistivity. As the corrosion rate depends on the effective cathode to anode area ratio, electrolytic resistivity influences the corrosion rate of anodic spots and must be one of the main parameters in modelling the corrosion process. This paper focuses on concrete resistivity only and is based on a series of measurements performed during the latest four years from exposure tests. Herein the factorial approach for calculating electrolytic resistivity will be introduced followed by the validation of a factorial approach for the probabilistic calculation of electrolytic resistivity. The factorial approach is one part of the corrosion model which predicts values in regard to corrosion current.

1

FACTORIAL APPROACH

exposure tests in Stuttgart, are summarised in the following table.

To predict resistivity values of concrete that will develop under certain environmental conditions, a factorial approach, given in Equation 1, was quantified by DuraCrete (2000).

ρ = ρ0 (t ) ⋅ kT ⋅ k(ToW , RH ) ⋅ kCl ⋅ kt

(1)

with ⎛ t hydra ⎞ ρ0 (t ) = ρ0 ⎜ ⎝ t0 ⎟⎠

n

(2)

and kT = ρ0

R(T ) 1 = R(T = 20) 1+ K ⋅ (T − 20)

(3)

= resistivity of water saturated concrete at a concrete age of 28 days measured with the two electrode measurement method (TEM) thydra = concrete age at time t t0 = concrete age of ρ0 (28 days) n = age-factor kT = temperature factor K = factor within the temperature factor kToW = humidity factor for unsheltered conditions depending on time of wetness (ToW) kRH = humidity factor for sheltered conditions depending on relative humidity (RH) kCl = chloride factor kt = test factor for specific test method. All given k factors represent single influences upon the resistivity. Most of these factors have been quantified by DuraCrete (2000). The factors, which are needed to validate the aforesaid factorial approach with data measured within

2

QUANTIFICATION OF MEASURED DATA FROM EXPOSURE TEST

To validate the above shown factorial approach, electrolytic resistivity was measured with samples from exposure tests in Stuttgart. Measurements were performed with an internal AC four-point measurementmethod using embedded rods as electrodes (“Geohm”). One part of the samples was sprayed daily with chloride solution (kCl, k RH100% ). The other part was wetted to simulate rainy conditions (k ToW). Data, recorded in 2007 and 2008, were analyzed statistically. An example is shown in Figure 1. Herein the measurement results of the electrolytic resistivity from exposure tests for CEM III/A with all influences of time of wetness and temperature are given as cumulative frequency. The cumulative frequency fits well with lognormal distribution. The estimation of parameters was done by Maximum-Likelihood method. The results of all statistical quantifications for each condition (type of cement, with or without chloride, humidity) are listed in Table 2. 3

VALIDATION FACTORIAL APPROACH

The goal of the factorial approach is to predict the electrolytic resistivity on site for special concrete mixtures in dependence on age and environmental conditions. Therefore probabilistic calculations for each of the conditions of Table 2 were accomplished with the program package Strurel (A structural reliability analysis program-system) from RCP GmbH to obtain the distribution of the calculated resistivity.

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Table 1.

Quantified factors of the factorial approach. Distribution

Symbol

Table 2. Statistical quantification of the measured variables to validate the factorial approach.

*1)

Variables

Factors depending on the concrete nCEMI ND (0.23/0.04) ρ0CEMI ND (77/12) nCEMIII/A *2) ND (0.31/0.08) ρ0CEMIII/A*2) ND (177/15)

ρ CEM I ρGEOHM, CEMI, Cl Chloride content > 2%, kRH = 1 ρ CEM I ρGEOHM, CEMI, ToW No chloride, kToW ρ CEM III/A ρGEOHM,CEMIII, cl Chloride content > 2%, kRH = 1 ρ CEM III/A ρGEOHM,CEMIII, ToW No chloride, kToW Near surface climate Temperature T

Factors depending on environmental conditions ND (0.025/0.001) kT (K) kToW sLN (0.62/0.33/0.79) kRH100% D (1) kCl ND (0.72/0.11) Factors depending on the measurement method ρGEOHM = A*ρWER + εv*2) A ND (0.96/0.02) εv ND (0/44) ρWER = ρ0/A/1.2 ND (0.78/0.18) A

0.9

Ωm

LN (82/29)

Ωm

LN (181/77)

Ωm

LN (293/112)

Ωm

LN (623/285)

°C

ND (13/7)

In parentheses parameter of the distributions LN = Lognormal Distribution (µ/σ), ND Normal Distribution (µ/σ).

Probability Log Normal-Distribution

0.8 0.7

Table 3. Quantified data from exposure test versus calculated data from factorial approach. Measured data Conditions [Ωm]

Calculated data [Ωm]*1)

CEM I Chloride ToW

Weibull(min) (129/70/0.02) Weibull(min) (253/138/0.05)

LN (82/29) LN (181/77)

CEMIII/A Chloride LN (293/112) ToW LN (623/285)

0.6 0.5

Unit Distribution*1)

*1)

*1) In parentheses ND = normal distribution (µ/s), s LN = shifted lognormal distribution (ξ/δ/τ), D = deterministic. *2) Factors, quantified within this task.

1.0

Symbol

Cumulative frequency

0.4

Weibull(min) (357/139/90) Weibull(min) (700/271/183)

0.3 *1)

0.1

In parentheses mean, standard deviation and parameter (µ/σ/τ).

0.0 200 400 600 800 1000 1200 1400 1600 Valid Observations of ρGEOHM [Ωm] for CEM III/A, ToW

Table 4.

0.2

Figure 1. Quantification of the electrolytic resistivity obtained from exposure test measurements.

The results are summarized in Table 3. The model factor, which compensates all uncertainties between measured and calculated resistivity, was calculated from all validation runs as multiplicative factor for CEM I and CEM III/A separately. The results are shown in the following table. The above shown model factor for each concrete enters the factorial approach as further factor as shown in Equation 4.

ρ = ρ0 (t ) ⋅ kT ⋅ k(ToW , RH ) ⋅ kCl ⋅ kt ⋅ M

(4)

M = Model factor for CEM I and CEM III/A separately. 4

CONCLUSION AND PROSPECTS

Validation with quantified data from measurement on site shows the practicability of the factorial approach by using a model factor.

Model factor.

Conditions

M [–]

CEM I CEM III/A

LN (0.69/0.31) LN (0.87/0.35)

By using the factorial approach the calculated resistivity will be described statistically. Also resistivity on site is a random variable within all influences of age, temperature, humidity and corrosive load and is described statistically, too. Different distributions with different standard deviations lead to the possibility of both overestimation and underestimation of resistivity on site. But overestimation of resistivity may presumably lead to underestimation of corrosion current within the corrosion model. However overestimation of high resistivity values has probably less influence on the underestimation of corrosion current then in the case of overestimation of low resistivity. These influences have to be evaluated within validation of corrosion model. Therefore the next step will be to validate the corrosion model which predicts values in regard to corrosion current in dependence on the used concrete, the environmental conditions and the used test set-up.

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DFG Research Group 537: Modelling reinforcement corrosion – Corrosion measurements on cracked reinforced concrete beams K. Osterminski, A. Volkwein, W. Tian & P. Schießl Centre for Building Materials of TU München, Germany

ABSTRACT: The alkalinity of concrete forms a protection for the embedded steel. When cracks occur this protection is disturbed and the migration of corrosion-promoting substances into the crack is possible. In this case macro element corrosion with high rates usually takes place. This process is basically influenced by the conditions in the crack (moisture content, oxygen, chloride and pH), the electrolytic resistivity of the concrete and the ratio of the anode to cathode surface. The aim of subproject A3 of the DFG research group 537 is to analyse these influences so that quantification is possible. Therefore the basic parameters for the rate of corrosion, like the distribution of potential and resistivity will be investigated depending on mix design and environmental conditions. The paper presents first results of concretes made with Portland cement and blast furnace slag cement after about 55 days of cyclic water and chloride solution exposure. Active corrosion and differences in corrosion behaviour due to concrete composition were found. 1

INTRODUCTION

Nowadays, the time until the initiation of reinforcement corrosion can be described satisfactorily. As a next step for modelling service life the period of active corrosion should be taken into account. An electrical circuit model can be used as physical model, Figure 1 and Equation 1. Thereby, the corrosion rate can be defined as the quotient of the driving potential and the sum of all system resistances. I corr =

U0 + I corr , 0 ra rc + + ρe ⋅ k e Aa Ac

(1)

with Icorr as total corrosion current in [A], Icorr,0 as micro corrosion current in [A], Uo as driving voltage in [V], ra and rc as area specific polarisation resistances of anode and cathode in [Ωm²], Aa and Ac as areas of anode and cathode in [m²], ρe as resistivity of concrete in [Ωm] and ke as cell constant in [m/m²].

The increase of corrosion depth ∆xcorr(t) can be obtained by using the corrosion rate and Faradays law as shown in Equation 2. ∆x corr ( t ) =

∫ (i ∆t

corr

( t ) ⋅ ∆t ⋅ Y ) ⋅ dt

(2)

Herein, ∆xcorr(t) is the increase of corrosion depth at a given time in [m], icorr(t) is defined as the quotient of corrosion current Icorr and anodic area Aa in [A/m²], ∆t as time in [s] and Υ constant which is 36.866·10–12 [m³/C]. A first quantification of the main influences on the corrosion current Icorr has been done during the first period of funding of the research group DFG-FOR 537 for a homogeneous that means uncracked concrete (Schießl & Osterminski 2006, Beck et al. 2006, Warkus & Raupach 2006, Osterminski et al. 2006, Büteführ et al. 2006 and Bohner & Müller 2006). In the present paper a description of the specimen design, measuring technique and research programme in order to investigate the influence of cracks on the corrosion behaviour of the reinforcement are presented. A presentation of first results is also included.

moist concrete Re

2 Rc

Ra Rs U anode

Figure 1.

Electrical circuit model.

cathode

CONCLUSIONS IN THE PAPER

In order to analyse the influence of the used cement type and the cover depth of reinforcement on the corrosion behaviour in cracked concrete, specimens with a well defined crack of w = 0, 30 mm have been produced. Amongst all these specimens only five anodes showed first activity concerning corrosion of reinforcement in the first 55 days of cyclic chloride exposure. The results of these specimens lead to the following conclusions.

181

1. Chloride induced corrosion is found to be more critical for concretes with Portland cement than for those produced with blast furnace slag cement. 2. The cathodes that are the closest to the anode are participating the most in the corrosion process. 3. The cover depth and the moisture state of the concrete are heavily influencing the participation of the cathodes.

ACKNOWLEDGEMENTS Subproject A3 is part of the DFG research group 537 “Modelling Reinforcement Corrosion”. The authors wish to thank the German Research Foundation (DFG) for supporting this project.

All in all a further quantification and monitoring of the specimens is needed.

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DFG Research Group 537: Modelling reinforcement corrosion – Numerical modelling of bond strength of corroded reinforcement C. Fischer, J. Ožbolt & C. Gehlen University of Stuttgart, Institute of construction materials, Stuttgart, Germany

ABSTRACT: In the present paper the first results of numerical studies on bond strength in dependency on different corrosion levels on a beam end specimen are shown and discussed. The calculations are performed by employing a 3D nonlinear finite element programme with the use of a discrete bond model. The obtained bond stress-slip curves and maximum bond strengths as a function of crack width are presented and discussed.

1

INTRODUCTION

The corrosion of steel in concrete is a serious cause of deterioration of concrete structures. The rebar corrosion affects (i) the steel by reducing the bar diameter, (ii) the concrete by cracking due to the volumetric expansion of the corrosion products and (iii) the interaction between steel and concrete due to the loss of bond. Within the research group this project is consecrated to study the effect of rebar corrosion on bond strength and behaviour. The present paper discusses the results of numerical studies of corrosion effects on bond of ribbed bars on a beam end specimen.

2

NUMERICAL MODEL

The calculations presented in the paper are performed by the use of the 3D finite element (FE) code MASA3. MASA3 was chosen because of the possibility to use the discrete bond model, which is implemented in the code. The discrete bond model allows bond characteristics to be defined by bond stress-slip-curves. For this purpose 1D bar elements (reinforcement) will be connected to the surrounding 3D solid elements (concrete). Only the degree of freedom in the bar direction (slip) is considered. The connection with the concrete perpendicular to the bar direction is assumed to be perfect. Different failure modes (pullout or splitting) are accounted automatically by the model. Corrosion is modeled by a radial expansion of the concrete elements, which are connected to the bar elements. These elements form a cylinder around the bar elements with 2.4 times the actual bar diameter.

3

SPECIMEN AND DISCRETISATION

The beam end specimen with four bars placed in the corners has a dimension of 200 × 200 × 300 mm3

Figure 1.

Cross section of the specimen.

(see Figure 1). The horizontal support at the pull-out face of the specimen has a height of 100 mm whereas the vertical support is 90 mm wide and placed on the rear top. At this position the bar is covered by a plastic sleeve to avoid an enhancement of the bond strength due to the vertical support forces. Thus bond is provided over a length of 180 mm.

4

RESULTS

The calculations were performed with four different specimen types (see Table 1). Each type was calculated without expansion (reference calculation) as well as with five different expansion values. The expansion was simulated before the application of the pull-out load. By comparing the specimens excluding stirrups with the ones including stirrups a clear trend to higher residual bond strength can be seen, as found by different researchers as well. The absolute increase of bond strength between type 1 and type 3 specimens is negligible. The experiments of different researchers show principally the same tendency. The bond strength-crack width curves indicate that prior to crack widths of 0.1 mm the bond strength of specimen without stirrups remains at the same value as the reference calculations (see Figure 3) or even

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Table 1.

Specimen types.

8.00

d mm

c mm

c/d

Stirrups mm

1 2 3 4

12 12 16 16

20 20 35 35

1.67 1.67 2.19 2.19

– 6/90 – 6/90

2

8.00 d=12; c=20; w/o stirrups d=12; c=20; with stirrups

2

average bond strenth [N/mm ]

7.00

d=16; c=35; w/o stirrups d=16; c=35; with stirrups

7.00 average bond strenth [N/mm ]

Type No.

6.00 5.00 4.00 3.00 2.00 1.00 0.00

6.00

0

0.1

0.2

0.3

0.4

0.5

maximum crack width [mm]

5.00

Figure 3. Numerical results of average bond strength over maximum crack width for type 3 and type 4 specimen.

4.00 3.00 2.00

5

CONCLUSION AND PERSPECTIVE

1.00 0.00 0

0.1

0.2

0.3

0.4

0.5

maximum crack width [mm]

Figure 2. Numerical results of average bond strength over maximum crack width for type 1 and type 2 specimen.

above (see Figure 2). If crack widths become wider than 0.1 mm the bond stresses starts to decrease almost linearly in the absence of confining reinforcement. The influence of confining reinforcement can clearly be seen. In the post peak regime the bond strength reaches values of around 4 N/mm2, i.e. bond strength does not reduce to zero. Furthermore, the confining reinforcement appears to have influence on the maximum bond strength for reference calculations, which differs from experimental results found in literature.

The performed numerical studies at a beam end specimen on bond strength and behavior show realistic results. The characteristic bond stress-slip behavior was obtained for the reference state (without corrosion expansion) and for different corrosion states. Furthermore, the influence of confining reinforcement was shown in dependency of the corrosion state. Even though the numerical model has a geometrical exception, due to modeling the corrosion expansion with a concrete cylinder of 2.4 times the actual bar diameter, the presented results show good agreement with results known from the literature. Within the project, experiments will be conducted on the specimen studied here. By having these results, the model will be calibrated and improved. Finally, using the improved model, extensive parametric study will be carried out in order to formulate an engineers model for bond, which will account for the influence of corrosion.

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DFG Research Group 537: Modelling of reinforcement corrosion – Modelling loss of steel cross sectional area and design for durability K. Osterminski & P. Schießl Centre for Building Materials of TU München, Germany

ABSTRACT: Actual models for reinforcement corrosion can not reliably predict the durability of reinforced concrete structures. Some of the existing models ease the complexity of the corrosion process by neglecting the influence of important parameters, whereas some other models reach a complexity, which can solely be solved by numerical programmes. Against this background, the aim of project D is the transformation of physical damage models developed in the subprojects of the research group into user friendly probabilistic design models. Herein the limit states of loss of steel cross section, cracking, spalling and loss of bond strength caused by corrosion will be taken into account. To comply with actual design standards (e.g. EC0), the proposed models will be probabilistic, so that for the first time a complete design for durability of reinforced concrete structures will be possible.

1

INTRODUCTION

The main deterioration aspect for reinforced concrete structures is the corrosion of steel in concrete. Nowadays, a lot of existing infrastructure buildings like bridges or park decks show damages like cracks and spalling of concrete cover due to an ongoing corrosion. A strong interest in predicting damages and the corrosion propagation therefore exists. In general five preconditions must be given, to make reinforcement corrosion possible: 1. Missing passivation (by low pH or chloride ingress) so that anodic reaction can take place, 2. sufficient oxygen in the surrounding concrete of the reinforcement for a cathodic reaction, 3. electric connection between anode and cathode for an electron transport, 4. potential difference (driving potential) between anode and cathode and 5. electrolytic connection for a transport of ions from cathode to anode. In order to start the corrosion process a depassivation of the steel embedded in concrete has to take place first (compare condition 1). The time period until a depassivation takes place is known as the initiation phase and can be modeled probabilistically (DURACRETE 2000, Gehlen 2001). Herein two causes for depassivation can be identified. On the one hand the carbonation front reaches the surface of the reinforcement and reduces the pH-value significantly and on the other hand the chloride concentration exceeds the critical corrosion inducing threshold. Both mechanisms destroy the thin passive layer on the steel surface leaving an unprotected reinforcement.

For the period of damage due to reinforcement corrosion, no practicable models exist. Current models are not capable of doing the splits between the complexity of the corrosion process and their usability. A numerical approach to a solution like proposed in e.g. (Maekawa et al. 2003) or (Song et al. 2005) may be a good approximation of the corrosion propagation but reaches an impracticable complexity. However, simple models strongly ease the complexity of the corrosion process e.g. models solely based on the resistivity of the concrete. In this course a factor of proportionality accompanied by a huge scatter is introduced to fit the modeled curve with the measured results, Equation 1. i corr =

A , ρe

(1)

with icorr as density of the corrosion current in [A/m²], A as factor of proportionality [V/m] and ρe as resistivity of concrete in [Ωm]. Further on, an add-on of Equation 1 has been proposed (DURACRETE 2000). Here probabilistic factors were introduced in order to model the corrosion propagation as a function of resistivity, Equation 2, where resistivity itself is a parameter modeled by a factorial approach, Equation 3. A complete quantification of the single factors (influencing parameters) considered in these formulae has not been processed yet.

185

i corr =

A ⋅ FCl ⋅ FOxid ⋅ FGalv ⋅ FO2 , ρe

(2)

with Fi as influence parameters for chloride, oxide at the anode, relationship of the areas (anode/cathode) and oxygen supply at the cathode. All factors Fi are dimension free [-].

moist concrete Re

Rs

(3) U anode

with ρ0 as resistivity of concrete in [Ωm] at an age of t0 [d] (usually 28d), tHydr as age of concrete in the moment of determination of resistivity in [d] (approximately 1 year proposed by DURACRETE 2000), n as ageing factor of the resistivity in [-] and ki as influencing factors of resistivity describing different measuring techniques, curing techniques and environmental influences (temperature, relative humidity and chloride). As well as the factors Fi before, all factors ki are dimension free [-]. 2

BASIS OF THE DESIGN MODEL

An electrical circuit diagram is used to model all subprocesses that take place in corrosion, Figure 1. With respect to Ohm’s law and the negligence of steel’s resistance the electrical circuit diagram can mathematically be formulated, Equation 4.

∑R

i

=

U U ⇔ I corr , macro = I corr , macro Ra + Rk + Re

Rk

Ra

n

⎛ t Hydr ⎞ ρe = ρ0 ⋅ ⎜ ⋅ k t ⋅ k c ⋅ k R , T ⋅ k R , RH ⋅ k R , Cl ⎝ t 0 ⎟⎠

(4)

with Icorr,macro as corrosion current of an ideal macro cell in [A]. The paper presents the methods of project D so that the complex physical model describing the corrosion process can be transformed into user-friendly design

cathode

Figure 1. Electrical circuit diagram of reinforcement corrosion (Re: Resistance of concrete, Ra, Rk: Polarization resistances of anode and cathode, Rs: Resistance of steel → negligibly small and U: Driving potential).

models. The following working steps have been worked out: 1. A quantification of all influencing parameters of the physical model. 2. In order to identify type of statistical distribution and its parameters, all results have to be processed statistically. 3. An analysis of correlation between parameters has to be done. 4. A study of dominance and sensitivity in the physical model should identify the relevant parameters for a design model applicable in practice and sufficiently accurate. Furthermore, factorial approaches are used to quantify the influences of single parameters. The approach will be compatible with the design for loads so that an integration of the service life design procedure in current design standards is possible. In combination with existing models for the prediction of depassivation a complete design for durability of reinforced concrete structures will be possible.

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DFG Research Group 537: Modelling reinforcement corrosion – Observation and monitoring of self-corrosion processes in chloride contaminated mortar by X-ray tomography M. Beck, J. Goebbels, D. Meinel & A. Burkert Federal Institute for Materials Research and Testing (BAM), Berlin, Germany

1

INTRODUCTION

Corrosion of steel reinforcement in concrete exposed to chloride containing environments is a serious problem in civil engineering practice. Electrochemical methods, e.g. potential mapping, (SIA 2006, DGZfP 1990) provide information whether the reinforced steel is still passive or depassivation has been initiated. By applying such techniques no information on the type of corrosion, its extent and distribution of corrosion products is available. Until now it is impossible to collect such information without destroying specimens after electrochemical testing has taken place. Within the scope of a german research project, investigations were performed, to determine the part of self-corrosion of the total corrosion. In the selfcorrosion measurements a number of fluctuations of the electrochemical values of the same series were recorded (Beck et al. 2006a, Beck et al. 2006b). Aim of this examination is to correlate the electrochemical values to the real steel surface. For this purpose the dismantled steel surface of one specimen (after a certain period of time), is correlated to the electrochemical values of the whole series. From the observations obtained so far it can be concluded that due to the different conditions of each specimen this procedure has to be used with caution. To overcome this problem it was tried to study the steel surface inside the mortar specimens by X-ray tomography (CT).

2 2.1

ded bars were predamaged periodically by static galvanostatic polarisation (10 µA/cm²). After 5 cycles of pre-damaging the specimens were investigated by X-ray tomography. Subsequently, the mortar cylinders were destroyed and the bar was dismantled, cleaned and documented by photography. 2.2

CT-Examination

The measuring system for the 3D-computer tomography consists of a X-ray tube, a manipulator and a detector. The experimental arrangement is shown in Figure 1. This is realised by rotation of the test object in the X-ray cone beam. A radiography of the test object for many different viewing angles is taken. The radiography is an image of attenuation of the primary X-ray beam by differences in material density and material thickness. The specimen volume is reconstructed from all radiographic images. 3

RESULTS

By using the data from CT examination, it is possible to give a realistic impression from inside the specimen. Therefore the cover mortar was removed and the dismantled bar was cleaned from mortar and corrosion products. In Figure 2 the lower area of the bar is shown in two views. On the left side the result of the 3D X-

EXPERIMENTAL Electrochemical pre-damaging

For the examination by X-ray tomography mortar cylinders with a height of 90 mm and a diameter of 40 mm were used. The mortar was produced and treated according DIN-EN 196–1 (1995) with 6.0% chloride by weight of cement. Steel bars (S 235 JR G2), with a length of 100 mm, a diameter of 8 mm, were embedded in the center of the samples. All embed-

Figure 1. Principal configuration of computer tomography.

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ray tomography of the embedded bar and the right side a photo of the dismantled bar in the same position are shown. It is obvious that a realistic picture of the embedded bar is given by the X-ray tomography. Some significant points on the steel surface (1–10) were marked. In order to get information on the extent and size of some representative pits their dimension were measured by microscopy. The results are presented in Table 1. So it is obvious that it is possible to give a realistic picture from inside the specimen by this technique, without destroying the specimen. Moreover it was possible to highlight different levels of the specimen by X-ray tomography and to survey the local pits.

4

DISCUSSION

Within the scope of these investigations it could be shown for the first time, that X-ray tomography is suitable to visualise corrosion pits on rebars, embedded in mortar with a cover thickness of about 20 mm. Even pits as small as 50 µm could be detected by X-ray tomography. The results of the X-ray tomography investigations were verified by inspection of the real surface of the dismantled bar and by different metallographic sections. Furthermore, for the first time this method enables to observe corrosion pits and its growth in-situ without destroying the specimens. In addition it can be made a realistic plane geometry of those pits in different levels by sections from CT-data. 5

CONCLUSIONS

Within the scope of examining steel specimens embedded in chloride contaminated mortar X-ray tomography was used to analyse the areas, damaged by chloride induced corrosion. Damaged areas in a scale of 100 µm (diameter) and a depth of about 45 µm could be detected and surveyed. The results of the X-ray tomography were verified by inspection of the surfaces of the bars and metallographic sections after removing the cover mortar. REFERENCES

Figure 2. Comparison between embedded steel bar (CT, left) and dismantled steel bar (photography, right).

Table 1.

Diameter and depth of selected pits.

Location

diameter µm

depth µm

4 5 6 8 9

610 950 310 202 680

158 120 150 50 271

Beck, M., Burkert, A. & Isecke, B. 2006. Modellierung von Bewehrungskorrosion Betontechnologische Einflüsse auf elektrochemische Parameter. Proceedings ibausil Weimar 2006: 1199. Beck, M., Eichler, T. & Isecke, B. 2006. Modelling of reinforcement corrosion- influence of concrete technology on electrochemical parameters. Proceedings eurocorr 2006 Maastricht. Beck M., Goebbels J. & Burkert, A. 2007. Application of X-ray tomography for the verification of corrosion processes in chloride contaminated mortar, Materials and Corrosion, 58: 207. Berger, W., Kalbe, U. & Goebbels, J. 2002. Fabric studies on containment mineral layers in composite liners, Applied Clay Science 2002. 21: 89–98. DGZfP Berlin 1990. DGZfP-Merkblatt über elektrochemische Potentialmessungen zur Ermittlung von Bewehrungsstahlkorrosion in Stahlbetonbauwerken. SIA 2006: Durchführung und Interpretation der Potentialmessung an Stahlbetonbauten, Interessengemeinschaft Potentialmessung Stahlbeton (IG Pot), Schweizerischer Ingenieur- und Architekten-Verein.

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Theme 2: Condition assessment of concrete structures Degradation assessment and service life aspects

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

An Austrian experience with identification and assessment of alkali-aggregate reaction in motorways E.K. Fischboeck & H. Harmuth University of Leoben, Leoben, Austria

ABSTRACT: Drilled cores from two different Austrian concrete motorways suspected to be damaged by Alkali Aggregate Reaction (AAR) were investigated. The aim of the investigation was to reveal the cause of cracking and thereby apply published methods for investigating AAR and adapt them to Austrian needs. More specific, visual examinations of the cores after wetting-drying cycles as well as microscopic investigations of thin section and polished sections and SEM examinations were performed. For mechanical testing, ultrasonic pulse velocity and resonant frequency were measured and dynamic Young’s modulus determined. Additionally static Young’s modulus and compressive strength measurements were conducted and an extensive statistical analysis was carried out with the data from mechanical testing. Damage Rating Indices developed by Dunbar and Grattan-Bellew were determined and adapted to improve the correlation with the results from mechanical testing. The overall chemical analysis of the cores and the water-soluble alkali, silicon and chlorine content were surveyed for hints for an ongoing AAR. It could be shown that for the cores from one motorway AAR was clearly the cause of the damage, whereby the damage was most severe in the area of the first lane. The cores from the other sampling site showed few to no signs for AAR damage so far. But, the samples showed a potential for AAR in the future since the preconditions for AAR could be observed.

1

INTRODUCTION

Two motorways (A2, A9) situated in the region in the south of the Austrian province Styria are subject of the present study. Visible damages—mostly longitudinal cracks and map cracks—were noticed in case of A2 three years after reconstruction (at which the old concrete was recycled and used for the subconcrete) and 13 years after construction in case of A9. Alkali-aggregate reaction was not considered as major reason for the damage at first. Nevertheless, after first investigations of the motorways in 2003 it became more obvious that AAR could be the reason for the damage. Thus, investigations were carried out in order to reveal the cause of the deterioration with special regard to AAR (more precisely alkali-silica reaction ASR). The results will be presented in this paper. For the study three cores 100 mm in diameter were drilled from the first lane (C1, C2 and C3) of the A2. From the A9, with a more advanced damage, four cores were drilled from the emergency lane (edge: C6, C7; centre: C8, C9) and two from the first lane (C4, C5).

2

EXPERIMENTAL

of resonant frequency, static Young’s modulus, determination of damage rating indices (DRI), X-ray diffraction, light microscopy and scanning electron microscopy (SEM) of polished, thin sections and fracture surfaces as well as chemical analysis. 3

From visual examination a ranking of the cores according to the damage could be achieved. C4 and C5 showed the highest damage—followed by C7, C8 and C6, C7 from the A9. Features typical for an ASR could be observed (e.g. cracks running through cement paste and grains being partly filled with gel, reaction rims around grains, translucent drops of gel on grains at the superficies surface and air voids filled with gel). The samples from the motorway A2 showed only little signs for ASR. There were fine map cracks on the surface, minor deposits of gel and very few grains with reaction rims. Samples of reaction products taken from the samples could be identified as alkali-silica gel with electron probe microanalysis (EPMA). 3.1

In the laboratory the following investigations were performed: visual examination of the cores after wetting and drying cycles, measurement of ultrasonic pulse velocity (USPV) in radial and longitudinal direction,

RESULTS AND DISCUSSION

Mechanical testing

A comparison of the results showed that the methods for determining static and dynamic Young’s moduli yield in different results—the higher the damage the higher the diversity. Nevertheless, statistical tests

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Figure 1. Crack and air void lined with alkali-silica gel in sample C4 (transmitted light microscopical investigation of a thin section).

Figure 2. Crack within a quartzite grain partly filled with alkali-silica gel in sample C8 (backscattered electron image of a polished section).

showed that all methods were applicable to detect damage and yield a ranking according to damage equal to visual examination. The determination of dynamic Young’s modulus from USPV was of special interest as length profiles of the dynamic Young’s modulus could be obtained which enabled the location areas of higher or lower damage.

by sample preparation could be observed which led to the additional SEM investigation of fracture and sawn surfaces. Ettringite could be found in all samples from the A9 but its contribution to the damaged was considered to be not relevant. 4

3.2

Petrographic examination

The results of the determination of the DRI, showed a significant ASR damage for C4 and C7 (with DRIs of 136 and 126 respectively). C8 and C2 were affected but not yet damaged by ASR (DRI 39 and 35 respectively). Due to inconsistencies of the results from visual examination and the DRI, modification of the DRI was performed which also resulted in an improvement of the correlation of the DRI with the mechanical properties of the concrete. With microscopic/SEM investigations the type of aggregates could be determined and reactive aggregates identified. For both motorways the deleterious components were quartzite and gneiss. Generally, the impression of the damage observed reflected the results from the other conducted investigations. Nevertheless, a significant loss of gel caused

CONCLUSION

With the presented methods it was possible to diagnose an ASR damage of varying extent for the samples from the A9 and an incipient ASR which has not yet yield damage respectively for the A2. ASR could thereby be observed in a direct and an indirect way. Directly, visual examination, determination of DRI and petrographic methods allowed identifying and quantifying the damage. Indirectly the effects of ASR on the concrete could be measured by mechanical testing (USPV, resonant frequency, Young’s modulus). ACKNOWLEDGEMENT We would like to thank the Austrian Federal Ministry of Transport, Innovation and Technology for funding the present project (no. 3.312).

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Chloride ion propagation in onshore zone of Recife-PE R.B. Pontes Association Institute of Technology of Pernambuco, Recife, Pernambuco, Brazil

E.C.B. Monteiro, R.A. de Oliveira & S.C. de Paiva Catholic University of Pernambuco, Recife, Pernambuco, Brazil

ABSTRACT: On exposing concrete structures to saline mist, they can suffer chloride attack, which could cause reinforcement corrosion. The structures directly facing dominant winds are most susceptible, especially those with greater surface areas in relation to volume, as pillars and beams. The chloride ions reach surface of structures at different levels of concentration, depending on the distance from their origin, in this case, the sea. The purpose of this research is to evaluate the level of marine aggressiveness, in the onshore zone of Recife-PE, in function of distance from the sea, measure the value of chloride deposition using the wet candle method (NBR 6211, 2001), using environmental management as support. The results indicate that chloride deposition decreases, in exponential relation, when there is an increase in distance from the sea, and that aggressiveness is significant up to 400 m from the marine edge.

1

2

EXPERIMENTAL PROCEDURE

This work has the objective of making a survey of environmental aggressiveness, through chloride depositions, in the coastal zone of the Recife Metropolitan Region, specifically in the Boa Viagem district, employing the wet candle method, in accordance with NBR 6211 (2001) criteria, having the climate characterization of the environment as back up.

2.1

Exposure conditions

The monitoring was conducted at five stations, situated at 7, 100, 169, 230 and 320 m in relation to the sea.

Table 1.

Environmental conditions

In Table 1, the environmental conditions are presented for the twelve months of study (Aug./2005 to July/2006).

2.2 1.1

RESULTS

Behavior of marine aerosol

2.2.1 Effect of distance in relation to the sea The reduction in chloride deposition, at each monitoring station, while there exists a distance from the sea, behaves in an obvious way, in agreement with Figure 1.

Summary of climatic data in the period of study.

Predominant Wind velocity Month/Year direction of wind (m/s)

Precipitacion (mm)

Relative humidity (%)

Temperature (ºC)

Radiation (h)

Aug./2005 Sept./2005 Oct./2005 Nov./2005 Dec./2005 Jan./2006 Feb./2006 Mar./2006 Apr./2006 May/2006 Jun./2006 Jul./2006

290.8 45.3 59.7 8.2 174.2 12.3 32.4 156.8 322.8 338.1 431.4 222.5

84 77 74 71 73 71 71 75 81 86 84 82

24.2 25.2 26.0 26.9 26.7 27.1 27.7 27.6 26.6 25.8 25.0 24.5

181.1 234.3 277.4 270.7 241.1 259.5 221.9 258.3 194.1 192.3 137.3 201.5

SE SE SE E E E E NE E SE SE SE

2.0 2.7 2.7 2.8 2.7 2.8 2.3 2.1 1.7 1.8 2.2 2.2

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The model that best adjusted to the results obtained was the exponent of type y = aebx, where y is the marine aerosol chloride depositions, and x is the dis-monitoring station, the concentration was higher tance in relation to the sea (Figure 1). In Figure 2, the linearized model is shown, and in Table 2, a model prediction is made.

Chloride deposition (Aug./2005 to Jul./2006)

Chloride concentracion 2 (mg/m .day)

700

-0.01x

y = 624.86e 2 R = 0.9204

600 500 400 300 200 100 0

Figure 3. Wind channeling diagram at fifth monitoring station.

0

100

200

300

400

Distance in relation to the sea (m)

Figure 1. Relation between average chloride deposition and distance in relation to sea, in the period of Aug./2005 to Jul./2006.

7 ln (Dep.) = 6.4375 - 0.01x 2 R = 0.9204

6,5

Figure 4. Wind channeling effect.

ln (Dep.)

6 5,5 5 4,5 4 3,5 3 0

50

100

150

200

250

300

350

Distance in relation to the sea (m)

Figure 2.

Table 2.

Linearized model.

From Figure 1, observe that, the farther in distance from the sea, there is a decrease in chloride concentrations in the atmosphere, however, at the 5th monitoring station, the concentration was higher than the 4th station, in nine of the twelve months of study, this behavior can be explained by the fact that, there is a channeling of wind direction, this Chloride deposition (Aug./2005 to Jul./2006) condition does not exist at the other monitoring stations. In Figures 3 and 4 the channeling is shown.

Model prediction.

Distance in relation to sea (m)

Chloride deposition (mg/m2.day)

7 100 160 230 320 400 500 600 700 800 900 1000 1100 1200 1300 1400 1500

584.06 230.44 126.47 62.8 25.53 11.47 4.22 1.55 0.57 0.21 0.08 0.03 0.01 0.004 0.001 0.0005 0.0002

3

FINAL CONSIDERATIONS

The following considerations were raised in relation to saline mist behavior (Pontes, 2006): a. the chloride depositions decrease while there is a distance in relation to the sea, in an exponential relationship, of type y = ae−bx, where y is the chloride deposition, expressed in mg/m2.day, and x is the distance in relation to the sea; b. in a zone of marine aggressiveness there are higher, distinct levels of aggressiveness with approximation to the sea; c. the chloride depositions are in agreement with model prediction (Table 2), given that, in the first 400 meters they reach significantly high values and after 700 meters from the sea they reach lower values.

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Investigation of the deteriorations of educational institutes Milad M. Alshebani & Sanousi Azhari Department of Civil Engineering, Alfateh University, Tripoli, Libya

ABSTRACT: Two public schools showed significant signs of distress, and one of it had to be closed until the building’s safety was assessed. A thorough examination of the schools building was carried out which included field and laboratory investigations. The pattern of deteriorations were typical: concrete cover peeling off, steel reinforcement corrosions, shrinkage cracks, and weak concrete strength. Geographical, the two schools are located in two different regions each with its own characteristic environment. Thus, the surrounding environments had a direct link to the level of deteriorations of each building. Specifically, the aggressive environment surrounding the school located near the sea shore played an important role in the wide spread defects of the concrete members. Furthermore, it was found that lack of professional engineering supervision during construction contributed largely to the current situation of both school buildings. It was also speculated that sea water may have been used in mixing concrete ingredients during the construction process of the near sea school. The paper presents the findings of field inspections, field examinations and laboratory tests that had been carried out on the main building elements of the two schools.

1

building, differences were noticed in columns footings dimensions and in footings steel distribution.

INTRODUCTION

Concrete structures under service loads exhibit different signs of defects with time. Although causes for such defects differ even within the same structure, they can be related broadly to three main reasons: (i) inappropriate concrete mix, (ii) absence of routine inspection and maintenance, (iii) aggressive surrounding environment and (iv) lack of professional supervision during construction. This paper presents a description of the overall deteriorations of two educational buildings, each located geographically in two different environmental region.

2

FIELD INSPECTIONS

The two buildings are typical concrete beam-column structures with isolated column footings. It was obvious from field inspections that the type of concrete used was in poor quality and that raised questions on the concrete mix design, placing of fresh concrete and concrete treatment. Reinforced bars were sometimes displaced and changed in numbers and in total area. It has been observed that fluctuations in concrete cover thickness and cover peeling off are common deterioration in the near sea buildings. In some location in the coastal building, the concrete cover was less than 20 mm thick. Insufficient cover thickness was main contributor to the deterioration of the building concrete elements. Reinforcement corrosions were dominant in the coastal building while in the desert building the cracks were the dominant scene. Ties spaces were not kept to design standards in both columns and beams. In fact, different tie sizes were found in the same concrete member with no obvious reason. In the older

3

FIELD AND LABORATORY TESTS

Typical field and laboratory tests were conducted on both buildings including destructive tests and chemical analysis. 1. Concrete cores extracted from beams elements. cores showed the quantity of sand was more than required and aggregate was not cleaned from impurities before mixing. Voids were found in some of the cores of both buildings, indicating unsuitable aggregate size distribution and lack of appropriated compaction. 2. Compressive strength showed great fluctuations in the concrete quality, and in many instances the strength does not meet the minimum requirements. 3. Chemical tests revealed deep penetration for both chlorides and sulphates compounds in concrete

Figure 1. Variation of chloride percentage through the concrete thickness.

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elements of the coastal building, Fig (1). A lower PH value suggests a lower cement content in the concrete mix.

4

CONCLUSION

1. In the two building examined, concrete quality fluctuates substantially within the same building’s elements. This would suggest: improper concrete mix design, violation of quality control by the contractor and/or aggressiveness of surrounding environment.

2. The humid surrounding environment with saturated salty air at the coastal building was responsible for the deep penetration of chlorides into the concrete body. Thus causing a wide spread of reinforcement corrosions and concrete cover failures. While in the desert building, hotdry atmosphere is likely the cause for the appearance of cracks. 3. It is believed that the contractors of both building used unskilled labor combined with little or no supervision. This may explain the tremendous variations in concrete cover thickness, structural elements dimensions, and improper steel placement.

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Durability evaluations on the bridge over the Ruwais lagoon at Jeddah M. Arici & M.F. Granata Dipartimento di Ingegneria Strutturale e Geotecnica,Università di Palermo, Italy

The durability of concrete bridges in service life depends on a large number of parameters. First of all in the design stage a particular attention must be paid to the stress and strain patterns in service life in order to avoid concrete cracking or excessive deformations due to time-dependent phenomena (creep, shrinkage, steel relaxation). The construction methods involved, the chosen materials, the environmental conditions, the bridge management and maintenance are all elements which contribute to the definition of the technical service life and influence the durability of the structure. An example of a multi-span concrete slab bridge, 29 meters wide and resting on central piers, placed in a very aggressive area is here presented to make possible evaluations on the durability of this kind of structures and to have knowledge of the durability problems encountered in similar conditions for prestressed concrete bridges. The bridge over the Ruwais Lagoon at Jeddah is a multi-span concrete slab bridge, horizontally curved and doubly variable in depth with a longitudinal and transverse catenary symmetric profile (figure 1). The deck is made of prestressed concrete slabs solidly joined to the piers and longitudinally connected by means of Gerber saddles. The environmental conditions over the lagoon (a hot, very humid and salty atmosphere) played a decisive role in construction stages and in service life determining the durability of the structure. The design choices in order to limit the effects of the aggressive agents were the following: the slab cross section has the minimum perimeter with the maximum concrete area to limit the exposed surface; prestressing in two directions was used into the deck to control the concrete cracking under loads, avoiding completely tensile stresses for dead loads and strongly limiting the width of cracks for live loads. This condition has been reached through the complete balancing of bending deformations due to dead loads and prestressing, with a null value of the resultant deflections due to bending moments, section by section. This result can be reached by applying the concepts of the “load balancing method” by balancing dead loads with prestressing equivalent loads, in order to obtain only uniform compressive stresses for dead load condition in every slab section. So, it is possible to obtain a funicular behavior of the bridge for dead loads, by a careful evaluation of prestressing

Figure 1. Bridge construction. a) Gerber saddle and transverse prestressing tendons anchorages. The longitudinal and transverse shapes are evident. b) Completed spans and continuous scaffolding for the formworks.

force values, tendons layout and deck variable profile. In the paper the conceptual design of the bridge is examined by considering the problems related to concrete cracking and fatigue under live loads. Because of the construction methods involved, all concrete casting was made in situ. Moreover it was not used high strength concrete and coated or galvanized reinforcements. The safety against steel corrosion has been faced through tensile stresses limitation for live loads and careful attention during casting and construction works. A sensitive point for the durability of the structure is the presence of the Gerber saddles, because them could be preferential ways for concrete degradation due to the waterproofing loss of the expansion joints. In this case the prestressing degree has been chosen in agreement with the requested safety level against cracking. In fact, the concept of full prestressing, in which no tensile stresses were allowed into the

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prestressed section for every load combination, has been considered less advantageous, leading to high quotes of compressive stresses. The reduction of cracks width due to tension can be achieved by increasing the ordinary reinforcements and by decreasing the quote of prestressing. This leads to the concept of partial prestressing, in which a low level of tensile stress was allowed for the maximum value of loads acting on the structure. When partial prestressing is used, compressive stress due to service loads decreases in every section and so creep effects also decrease, limiting the time-dependent strains. By using the concepts of partial prestressing and by avoiding bending deflections through the balancing load method, the presented study shows the design criteria chosen to improve the safety level against concrete degradation and steel corrosion of the slab bridge. The chosen shape for the variable concrete slab, the degree of prestressing and the considerations about the cycle loads leading to fatigue in steel and concrete are the elements examined and discussed. The design criteria adopted and exposed here have been followed in order to strongly reduce the effects of time-dependent strains due to creep. By balancing the dead load at the time t = t∞ through the right quote of prestressing force, it is possible to avoid the total strains due to bending (both elastic and delayed ones). This aspect is very important for the subsequent durability conditions, especially in the case of very large transverse cantilevers due to the deck width, which are subjected to excessive deflections in time due to creep-induced strains and in the case of Gerber saddles which are the sites of concentrated relative rotations. These deformations could lead to joints misalignments with waterproofing loss and consequently dangerous local degradation. The paper refers to the general structural conditions registered after 25 years from the bridge construction and

Figure 2. The bridge today. a) A general vision under the deck. b) The saddle.

to the durability of the structure, exploiting the bridge over the Ruwais lagoon as a case-study (figure 2). Durability evaluations on the Jeddah slab bridges are exposed and the results of an accurate visual inspection of the bridge are presented. The results show that the bridge is perfectly aligned and prestressing appears effective in each point; no joints misalignments, no relative rotations into the saddles and concrete or steel degradation are evident, confirming the consistence and effectiveness of original design choices.

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Assessment of post-fire reinforced concrete structures. Determination of depth of temperature penetration and associated damage C. Alonso Institute of Construction Science Eduardo Torroja, Madrid, Spain

The assessment of concrete structures affected by the fire should allow understanding what happened during fire and developing rational criteria for the evaluation of the security of the structure. The result of the assessment should be used to take the decision on further repair, partial or total demolition of the structure. Two main goals to be answered during the assessment analysis are: 1) The determination of the depth of concrete exposed to temperatures that produce irreversible alteration of concrete components, and 2) the mechanical strength decay of concrete that can affect the load bearing capacity of the element or the structure itself. In-situ non-destructive techniques are widely employed for distinguishing between damaged and apparently non affected areas of concrete elements, such as resilience, hammer-tapping, ultrasonic pulse etc. However, the determination of the penetration of the damage and the differentiating between the type of damage (chemical, physical or mechanical) are important to ascertain the consequences of the thermal gradients, which require the further use of concrete cores for laboratory testing, usually for mechanical strength, petrography analysis and thermal alteration of cement paste. The isolated use of any method does not give reliable results because of the gradient and nonhomogeneity of the damage that cause differences between the external and internal strength. In addition the assessment of post-fire reinforced concrete structures is complex due to the overlapping of chemical and physical phenomena inducing different type of damages. The assessment of the damage of concrete in postfire structures that have suffered different scenarios of fire are performed in one underground structure 48h in fire and a tall building, Windsor Tower, 18h in fire. In-situ inspection and laboratory tests have been performed. The heterogeneity in the distribution of the damage is identified, and the depth of concrete affected by the fire is determined, that has allowed identifying the gradient of temperatures reached in the concrete. The depth of concrete for the critical temperature of 500ºC has been estimated. Finally, a protocol for assessment of concrete structures affected by fire is given including non destructive and destructive methods and micro-macrostructure analyses of the damage.

As commented above, although there are several techniques to determine the depth of concrete altered by the temperature, their use alone do not allow accurately determining the depth of concrete affected by the fire, and some of them are not able to discriminate between the type of damage, physical or chemical. None of the methods, neither NDT nor DT is able to determine the loss of mechanical strength even with the direct measurement of the strength, because of the presence of a gradient of damage in the concrete elements. The ultrasound velocity has the advantage of a NDT technique able to differentiate among damaged and non-damaged zones, however the determination of the depth of concrete cover affected is difficult from this method, as the ultrasound velocity variation in a fired concrete structure changes not only due to the dehydration of cement paste, but also because of the crack formation, Besides, the presence of cracks does not necessarily mean a fire alteration of the concrete components; in fact, most of cracks are due to the thermal stresses during fire. In the case of the 48h structure in fire the ultrasound rate varied with: a) the height of the column, indicative of the heterogeneous distribution of the damage, b) the presence of cracks and c) the distance to the crack. These measurements put in evidence that increase in cracking correspond well with the lower ultrasound rate measurements; however the penetration of the damage and the depth of chemical alteration of components due to fire could not be deduced directly from these measurements. Ultrasound velocity measurements taken in Windsor Tower showed that the direct ultrasound velocity measurement data did not allow discriminating with respect to the depth of damage, and laboratory tests on concrete cores were needed to calibrate the measurements on-site to give an overestimation (up to a depth of 10 cm) of the depth at which the compressive strength of the concrete was considered irreversibly affected. Thermo-gravimetric tests are able to identify the local chemical changes in cement paste and the level of dehydration of CSH gel of cement paste, which is the component responsible for concrete strength development. The chemical transformations in CSH of cement paste in concrete cores allowed identifying the depth of concrete exposed to temperatures up to 350ºC. Variations in CSH transformation and

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microstructure changes in concrete elements from different floors of the Windsor Tower indicated that the alteration of concrete due to fire penetrated up to a maximum of 3 cm, which did not agree with the values predicted with ultrasound measurements. In the case of the building 48h under fire, the alteration of CSH reached up to 7 cm. Microscopic techniques in general and SEM in particular are used to confirm the type of damage. The SEM allow differentiating between the microstructure changes in the paste, aggregates, the interface bonding loss and the cracking, but experience is needed to identify the origin of the damage. In fact, in the case of the concrete from the prestressed beams of the underground building 48h in fire, intense micro cracking in the mass of the concrete was observed; both in aggregates and cement paste; but the cement paste did not show chemical alteration. The damages were attributed to the explosion due thermal and pore pressures accumulated inside the concrete. Relationships have been found between several indicators of concrete damage, including microhardness, porosity and CSH transformation. A transformation of at least 55% CSH is needed to induce relevant changes in the other properties.

The damage induced by the fire in the concrete elements is very heterogeneous; the reason is attributed to the fact that in a real scenario the fire is not homogenous within the structure, and even within the same floor or column. The surface temperature varies between one place and another and also the duration of the intensity of the fire, so that the effect of the concrete damage also will be different. In the case of the Windsor Tower the extension of the damage varied from one floor to another, i.e. in floors 12 to 19 the differences were very relevant and the results showed that in the higher floors the depth of concrete affected by the fire did not penetrate more than 1.5 cm, while in the lower floors the penetration of damaged reached a depth of 3 cm. The use of a set of tests allows differentiating between several levels of damage and to identify indicators of damage in post-fire structures. Integrating the results of the different indicators used in the assessment it was possible to build up the isotherms curves of the temperature depths inside the concrete columns applied to the case of the fire in the Windsor Tower and the isotherm for the depth of 500ºC was determined, which can be used for further calculation of residual load bearing capacity of the structural elements.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Computer integrated knowledge systems for the assessment and diagnosing distress in concrete structures in Persian Gulf F. Moodi Concrete Technology and Durability Research Centre, Amirkabir University of Technology, Tehran, Iran

ABSTRACT: The maintenance and repair of existing concrete structures in the field of civil engineering is very important and it needs engineers with broad experiential knowledge. The difficulties of structural damage assessment, diagnosing distresses and obtaining recommendations on materials and procedures for repair are sometimes more complicated. Seeking to apply a single form of knowledge representation is insufficient to handle all cases. Thus, it makes an excellent environment to develop a framework that provides a means for an engineer to use expert systems with having facility for database management system in diagnosing of structural damage problems and to be able to recommend maintenance strategies. This paper outlines proposed computer integrated knowledge-based expert systems, called SEMARECEXPERT, Bridge Slab-Expert and 4.T.K, to address the needs of diagnostic-related issues, identification and system knowledge users involved in repair and maintenance activities, such as condition assessment, material failure analysis, material selection and rehabilitation recommendations in concrete structures in Persian Gulf region. These programs are created to disseminate knowledge of the concrete distresses in Persian Gulf and recent advancements in concrete repair problems. Also, they have facility for sophisticated data management so it is coupled to an independent database management system individually for obtaining recommendations on materials and procedures for repair and rehabilitation methods. Keywords:

Expert System, Concrete Distresses, Database Management System, Knowledge Management.

The Expert systems in this paper are designed to assist engineers in diagnosing distresses and in the assessment of current conditions in concrete structures in Persian Gulf. Each of them is coupled to an independent database management system (i.e. REPCON) for obtaining recommendations on materials and procedures for repair and rehabilitation methods. Recognizing the importance of transferring information to practicing concrete engineers, every effort is made to incorporate new concrete technology in these expert systems. Examples of distress that occur in concrete structures in Persian Gulf region are those that are induced as a result of exposure to adverse environmental conditions e.g. sulphate attack, chloride induced corrosion of the embedded steel and aggregate-alkali reaction. For the durability recommendations, knowledge regarding the selection of concrete constituents for different environments is included. Material selection and repair methods include knowledge that relates to the selection of materials and procedures for various repair and rehabilitation methods. An important factor in developing these expert systems was the need to limit its scope so that design criteria can be applied effectively. The system initially accomplishes well-defined goals and then allows for the addition of new knowledge as it becomes available or as the system matures. Such additions would

make the system more comprehensive and useful for concrete decision making for both the present and the future. These expert systems are designed to assist various users in the field. For instance, it is expected that field inspectors or engineers would use its concrete diagnostics parts of DEMAREC-EXPERT (CRACON, SURCON or MISCON) to identify distresses and to determine their causes. The concrete repair and rehabilitation methods would be useful to concrete designers who need recommendations regarding materials and methods for concrete needing repair or rehabilitation. The durability recommendations are useful for anyone involved in concrete repair problems. Bridge Slab-Expert is designed to identify causes and distresses in bridge concrete slabs in Persian Gulf region. It can predict realistically the condition and status of concrete bridge decks, including the determination of remaining service life. 4.T.K. expert system is developed for assessment of deterioration of concrete cap pile in a commercial port in south of Iran. Therefore, these computer programs are considered as a decision-making tool and to be comprehensive computerized expert systems that give recommendations on concrete structures in Persian Gulf region. The M.4 development tool used to develop the DEMAREC-EXPERT is an example of commercial software available Expert System Development Shell

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integrated into the Visual Basic Graphic User environment through the Visual Basic Controls. The Bridge Slab-Expert and 4.T.k are developed using Visual Basic Programming and Visual C# programming respectively and based on user-friendly GUI environment. These knowledge-based expert systems contain all parts of a typical expert system program. The structure of the DEMAREC-EXPERT system comprises M.4 as an expert system development shell, knowledge bases, the Visual Basic user-friendly interface, and report generation modules. The DEMAREC-EXPERT is independently coupled to the REPCON database with a subsystem connection for taking recommendations in durability, repair materials and methods. 4.T.K is simply coupled to a database which is designed in SQL software with a subsystem connection for storing and retrieving the concrete cap pile distresses. The initial step of the development strategy for the knowledge base is to organise the knowledge and then modified it by the author’s experience using the approach an expert would take to address the problem or activity. Secondly, a narrative description of the knowledge is developed in a question-and-answer format and conclusions and recommendations are also included at this stage. Then the diagnostic (hierarchical) trees are generated to provide a logical sequence to the knowledge and to represent how the knowledge is linked together. The integrity of the knowledge base lies in the development of the diagnostic trees. The pieces of knowledge in the diagnostic trees are arranged in a manner allowing for an uninterrupted flow of knowledge upon completion of the tress. In developing DEMAREC-EXPERT, the diagnostic trees, served as the vehicle for communication between the experts who interpreted and organised the knowledge in a hierarchical structure and the knowledge engineer who initially record the knowledge into a question-and-answer sequence form along with a network diagram. 4.T.K expert system is developed for assessment of deterioration of concrete cap piles. The assessment of cap piles is based on the cracking in concrete, surface distresses and structural distresses. 4.T.K expert system prepares a user friendly interface via Persian question-and-answer session hence facilitates access and assessment of defects of concrete cap piles. The rules and the parameters in the DEMARECEXPERT knowledge bases contain confidence level (CL) values to indicate the degree of reliability about the acquired knowledge. The confidence levels (CL)

included in the DEMAREC-EXPERT are words and descriptive functions with pre-determined CL value which are used to describe the knowledge. Words such as “Certain”, “Very Sure”, “Sure”, “Quite Sure”, “Possible Case” and “Innovative Idea” conceal specific uncertainty within the meaning. The confidence levels (CL) included in the Bridge Slab-Expert and 4.T.K are words such as “Definite”, “Sure”, and “Possible”. The systems usability and results are validated by three case studies taken from actual cases of concrete diagnosis and repair. In these case studies, it is shown that the results of this research could have been used to enhance the process of determining the cause of the failure and in selecting the repair material and method. This system is a valuable tool in automating the acquisition of field information, in presenting a hypothesis on how known distresses relate to site specific problems, in preserving the knowledge of experts and in providing a record of the condition of structures at different ages. It also provides much needed guidance for practitioners while serving as a decision support system for other experts and specialists in the field. The system usability and the DEMAREC-EXPERT results are validated by three case studies taken from actual cases of concrete diagnosis and repair and from the literature. The Bridge Slab-Expert system usability and results is evaluated and validated by three bridge structures taken from actual cases of concrete diagnosis in southern of Iran near Persian Gulf. The results allow the correct diagnosis of concrete decks, realistic prediction of service life and the description of condition and the recommended action for repair. Also a case study was performed on cap piles of Aluminum Jetty of Bandar Emam to confirm validity of the system. The aim of the expert system is to provide a rapid and consistent method to make decisions from the assessment of concrete structure impairment. These expert systems are designed to provide the user with recommendations related to the best course required for maintenance, the realistic prediction of service life and in the selection of materials and methods for the repair of concrete structures. For this purpose, these computrised systems are coupled to an independent database management system with the knowledge that offer information relating to selection of materials and methods for repair and rehabilitation and give the recommendations to enhance durability. Therefore, it is a valuable tool for engineers and other professionals dealing with the maintenance and repair of concrete structures.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Mechanical behaviour of corroded rebars and influence on the structural response of R/C elements S. Imperatore & Z. Rinaldi Department of Civil Engineering, University of Rome “Tor Vergata”, Rome, Italy

Corrosion of reinforcement is a common form of degradation of reinforced concrete structures. The chemical attack varies the mechanical properties of both the steel rebars and concrete, and the bond characteristics. The study presented is mainly devoted to the analysis of the corrosion effects on the steel rebars and at the steelconcrete interface, and their influence on the local and global behavior of simple r/c elements. At this aim, experimental analyses are developed, for the calibration of the constitutive relationships of both the steel rebars and the bond-slip response, in presence of corrosion. Finally the obtained laws are implemented in an analytical model, developed by the authors, in order to evaluate the influence of the corrosion on the global behaviour of simple r/c element beam elements. 1

collapse governed by the section with the small area, while a very sharp reduction of ductility is observed.

CONSTITUTIVE RELATIONSHIP OF CORRODED REBARS

The influence of the corrosion on the constitutive law of steel rebars is studied on the basis of an experimental campaign carried out in the Laboratory of the University of Rome “Tor Vergata”. In order to study the actual behavior of corroded rebars, it is simulated the degrade by carbonation and pitting. Preliminary tests have shown that bare bars, after the artificial treatment, exhibit a corrosion comparable with bars naturally corroded by carbonation. On the contrary, steel rebars embedded in a concrete prism, are affected by localized corrosion, with marked pits. The morphology of the bare steel rebars after the corrosion process and load-displacement relationships are shown in Figure 1. When a bar embedded in a concrete specimen is corroded, the pressure exercised by the oxides cracks the cover concrete, the corrosion localizes, and the reinforcement presents marked pits in this zone. As clearly shown in Figure 2, the embedded bars have a peak load greater than the bare one, with the

2

BOND

The influence of the corrosion on the bond characteristics is studied on the basis of an experimental campaign carried out at the “Tor Vergata” University of Rome. In order to analyze the phenomena connected to the bond, two types of specimen are tested, defined average or locals in function of the bonded length (70 mm and 30 mm respectively). The experimental results are expressed in terms of load—slip displacement and depicted in Figure 3. Similar patterns are found for the average and local bond. 3

CASE STUDY

The influence of the corrosion on the global behavior of a simple beam is finally analyzed. The behavior of the beam is simulated with an analytical,

40 35

0.00%

Load [kN]

30 25

6.96% 12.51%

20 15

15.83%

10

31.60%

5 0

0

2

4

6

8

Displacement [mm]

Figure 1.

Effect of corrosion on the geometry and load-displacement relationship of bare bars.

203

10

12

14

35 uncorroded bare (9.52%)

30

Load [kN]

25

embedded (9.92%)

20 15 10 5 0 0

2

4

6

8

10

12

14

16

18

Displacement [mm]

Figure 2.

Effect of corrosion on the embedded bars.

20

a)

45 40

16 14 0.10%

30

Load [kN]

Load [kN]

35 25 20 3.17%

10

0.00%

Figure 3.

4

6

0.96%

8

3.56%

1.50%

2

3.88%

2

10

4

1.01%

5

0

12

6

0.80%

15

0

b)

1.76%

18

8

0.00%

0

10 12 Slip [mm]

14

16

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20

2

4

6

8

10 12 Slip [mm]

14

16

18

20

Load-Slip curves; (a) average bond; (b) local bond.

F [kN] F/2

F/2

B-5% sound

175

B-10%

150

A-5%

2φ 12

125 3φ 16

A-10%

100

F/2

75

F/2

50 25 0

Figure 4.

5

10

15

20

25

30

35

40

δ[mm]

Geometry of structural scheme, strut and tie model and analytical load-displacement curves.

non-linear model. Two corrosion levels are considered related to a 5% and 10% of diameter loss. The force displacement diagrams are obtained by considering the diameter reductions only, with (A-5%, A-10%) and without any variation of the mechanical properties of the steel (B-5%, B-10%). In a first case no reduction of bond is considered. It appears not negligible the contribution of the corrosion influence on the constitutive behavior of the steel.

Finally in order to account for the influence of corrosion on the bond properties, according to the experimental tests, the limit case of the complete loss of bond for corrosion level of 10% is simulated. In this case a strut and tie model is studied. The ultimate force is related to the achievement of the concrete crush in the compressed strut. This scheme provides a bearing capacity higher than the related flexural scheme in presence of bond.

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NDE/NDT and measurement techniques

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Nondestructive testing methods for hardly accessible parts of structures H. Anton Civil Engineering Institute of Croatia, Osijek, Croatia

I. Netinger Faculty of Civil Engineering, Osijek, Croatia

E. Evic´ Civil Engineering Institute of Croatia, Osijek, Croatia

ABSTRACT: The paper presents some less known NDT methods applicable for condition assessment of some hardly accessible parts of structures which are commonly used in the Republic of Croatia. The dynamic test methods for integrity assessment of pilots/deep foundations are described. The pilots and deep foundations have also been tested for bearing capacity acting at horizontal force. Also the working principles for the each method have been mentioned here and the required instrumentation described. Finally, the paper describes the advantages and disadvantages of the methods’ application, as well as the responses of a structure to the each described method.

1

INTRODUCTION

In case of very unapproachable parts of the structure, like deep foundations or pilots, the non-destructive test methods are almost not offered as an alternative, but the same appear to be the must. Owing to the fact that it is a matter on elements of the foundation structure which might not subsequently be visually controlled as to determine eventual deficiencies/defects which took place during their work-out, there had been developed dynamic test methods which grant the opportunity of subsequent control of the pilot’s integrity and/or their bearing capacities. 2

DYNAMIC TEST METHODS

Based on energy level, the dynamic test methods might be divided into two groups, i.e.: test methods of a strong impact and those having a weak one. Test methods of a strong impact, for such impact intensity, usually implement the equipment for the pile driving. Owing to the fact that in recent times, within the area of deep foundation there prevail the drilled piles made in situ, such equipment is not directly available on the construction site upon work-out. Because of the mentioned reason, for the condition estimate, there are being considered the methods of a weak impact, which give a more comprehensive view of integrity, rather than the bearing capacity of the fitting. 2.1

Sonic—echo method

Method of the Sonic—echo presents a rather efficient mode of the foundation/pilot control, and which

implicates detection as well as qualitative-quantitative estimation with appearance of phenomenon like: inclusions, disruptions, thickenings and narrowing, and soil stratification, too. Applying the, electronic trigger equipped hammer blow upon the upper part of the structure, with this method there is being measured the time necessary for the wave to pass its path from the top all down to the bottom of the foundation/pilot, and back all up to the receiver. The time of the blow and vertical displacement of the element upon the blow, are recorded by oscilloscope or digit device which records the data on the time. In case the data on the velocity being available, there might be calculated the length of the element. Except with the estimation of the element depth, the wave propagation velocity might also be used for calculating the concrete density. The compressive strength might be estimated roughly, also. 2.2

Mobility method

With this method, owing to the hammer blow upon the head of the pilot/upper layer of the foundation equipped with a measuring cell, there is produced a wave of the frequency range depending on the hammer type. The cell located on the head of the pilot, measures the inlet strength, while the vertical impact of the pilot’s head might be viewed on the geophone. Entered force and signal velocity are recorded by the digital device, and are computer processed with a fast Fourier (FFT) of the algorithm that converts the data into frequency. Further, the velocity is divided by the force in order to achieve the response, or the function of transmission, which is graphically presented as mobile property of the pilot comparing to the frequency.

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2.3

Induction resistance

This method is the combination of the Sonic—echo method and the Mobility method, where the increased time reaction of the Sonic—echo is combined with a characteristic resistance on the bottom of the foundation/pilot, measured by the Mobility method. In spite the force applied to the pilot’s head with methods of reflection is temporary with methods of reflection, it appears not to be as such with the wave created from the flow. This wave contains data related to the changes of the pilots’ resistance as it moves downwards, and such a data reflects to the pilot’s head. Reflectogramme obtained in such a way with testing by the Sonic— echo, might not be expressed quantitatively. But, with a support of modern equipment for data recording, sampling of both wave reflections and reflection characteristics in test pilots might be feasible. Measurements of the force and velocity are stored as time data, by quick sampling in rather extensive transit filter, where there are taken into consideration resolutions of weak and strong reactions. Analysis of frequency (resistance), obtained by testing the reaction to impulse, confirm the pilot’s length and give the dynamical rigidity of the pilot, as well as characteristic resistances. 2.4

3

Crosshole sonic logging

This method requires a series of parallel plastic tubes which should be placed inside the structure prior to concreting, or requires the holes to be drilled upon concreting. Measuring probe, which is placed on the bottom of one of the tubes, emits ultrasonic impulse which might be detected applying the recipient probe located on the bottom of another tube. Recording device measures the time of the ultrasonic impulse through the concrete, between the tubes. The probes are sealed devices, and the tubes are filled with water in order to enable the tubes to be linked with concrete. Cables of the probes should be driven across the instrumented wheal which measures the cable length and the probe depth, or these cables might be marked entirely lengthwise. Continuous measuring the impulse should be performed during driving. This procedure gives a series of measurements, which data might be printed in order to obtain a vertical profile of the material laying between the tubes. Presence of defects is evidenced by absence of receiving signal. 2.5

pilot’s head/upper part of foundation is not anymore feasible without destructions. A hole of a small cross section should be drilled in the soil parallel and adjacent to the pilot that is supposed to be tested. Drilled hole should be made underneath the known or estimated pilots’ depth, and is usually marked by a plastic hose in order to retain water acting as water-acoustic element. Acoustic probe should be placed inside the tube starting from the top, and the structure should be stroked by a hammer as close to the pilot’s head as possible. Signals from the hammer and receiver are recorded in the device as time of the wave transmission through pilot and adjacent soil all up to the receiving probe. After that, the probe should be driven down in identical height elevations and the process should be repeatedly performed, and each time, the actions should take place on the same point. Recorded data should be put into vertical profile in each time of the wave transmission from the point of action to each and every position, all up to the approaching tube. Each significant disruption or inclusion into the pilot shall cause signalization around the same, increasing the path length and the transmission time.

Parallel seismic-method

Parallel-seismic method has been developed specifically for situations where direct approach to the

PILOT TESTING PER TO THE ASTM D 3966-90 NORM

This test method is used for testing the side bearing capacity of the system pilot-soil. Test area should be dig out prior to the pilots’ testing. Before starting the bearing capacity testing, the area around the upper part of the pilot should be filled with sand or any other appropriate material—the same material and the same accompanying methods should be applied for all of the pilots. Upon finalization of all of the preparation activities, there should be deposited the force—incrementally—up to the entire presumed value, and values of associated displacements should be recorded. One of such tests is performed on the bridge over the Drava river, and the results are given in this paper. 4

CONCLUSION

In spite of being lawfully liable, in case of a pronounced suspect in quality, lacking their own ordinances on testing the deep foundations/pilots, the engineers within the Republic of Croatia perform tests in compliance with adoptable worldwide recognized methods.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Vibration based assessment of shear connectors in concrete composite bridges B. Sibanda, P. Moyo & H.D. Beushausen Department of Civil Engineering, University of Cape Town, South Africa

ABSTRACT: Concrete composites consisting of pre-cast prestressed standardised beams and a cast in-situ deck slab have been used for the construction of short to medium span bridges for the past four decades worldwide. The pre-cast beams and cast in-situ slab are commonly connected using shear connectors. The failure of these connectors would compromise the composite action of the structure, thus reducing the load carrying capacity and hence its efficiency. The research described in this paper seeks to assess the integrity of such shear connectors using dynamic-based testing. A scaled bridge model was constructed and 10 mm bolts were used to link the beams and slab in order to simulate the shear connectors in the prototype bridge. Different damage scenarios were introduced by loosening some of the connectors. Vibration-based techniques are applied to detect the artificial damage. The main objective of the work is to develop a practical vibration testing scheme to assess the integrity of shear connectors in concrete composite bridges. The preliminary results show that the natural frequencies of the system detect the damage of shear connectors in a global sense. The Modal Assurance Criteria (MAC) values show that the failure of shear connectors affects the torsional stiffness of the system more than its bending stiffness. However, there were no visual differences between the mode shapes extracted for the undamaged and damaged structure. Very little information was deduced from the damping ratios as there was no clear variation. Practical vibration-based techniques might be of help to most bridge management systems that currently depend on visual inspection techniques for the condition assessment of these structures.

1

INTRODUCTION

Highway bridges constitute significant and critical components of transportation systems and are among the most expensive investment assets of any country’s infrastructure. However, there is a growing number of damaged bridge structures owing to the increase in allowable axle loads and to ageing of these structures (Parkash et al, 2006 and Seracino et al, 2004). In South Africa it was estimated, in 2000, that the annual cost of overload road damage was R650 million and the accrued backlog for the upgrading, maintenance and repair of over 13000 bridges was R37 billion (Sowman and Poree, 2000). For these reasons, damage assessment of these structures is required especially for inaccessible parts such as horizontal shear connectors. Despite the technological advancements, most bridge management systems (BMS) still rely on visual inspections for condition assessment. This means that damage in inaccessible parts of the structure may go undetected until it is expensive to repair or catastrophic failure occurs.

earlier, these techniques are not applicable for assessing inner damaged parts of the structure such as shear connectors. However, there are localised non-destructive techniques (NDT) that can be applicable for detecting damaged shear connectors in bridges. These include ultrasonic techniques, radar method, impact testing, magnetic based methods and proof load tests amongst others. Nevertheless, these techniques are limited to small areas and are time consuming. Currently, there is little work in the literature related to the use of these techniques in detecting damaged shear connectors in concrete beam and slab bridges. During the past few decades, vibration-based damage detection techniques (VBT) have emerged as promising tools in assessing and detecting damage in structures (Yong et al, 2007 and Wahab and De Roeck, 1999). The basis of VBT is that a change in the structural stiffness due to damage results in the change of modal parameters i.e. natural frequencies, damping ratios, mode shapes and curvatures. These techniques can therefore be used to detect internal structural damage such as damaged shear connectors.

3 2

BRIEF REVIEW OF DAMAGE DETECTION TECHNIQUES FOR SHEAR CONNECTORS

Visual inspection techniques are widely used for condition assessment of bridge structures. As mentioned

EXPERIMENTAL WORK

A 1/6 scale model of a typical 25 m span concreteconcrete composite bridge consisting of I-beams and slab bridge was constructed for this research. The shear connectors were simulated using 10 mm bolts.

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Three damage scenarios were inflicted on the system by loosening the shear connectors and these were investigated using vibration techniques. These are the intact state (X0), Light damage scenario (X1) and a severe damage case (X2).

The damping ratios had no clear trend. This finding correlates with the work by other researchers (Wahab and De Roeck, 1999 and Yong et al, 2007) who noted that damping ratios are difficult to measure and are therefore not suitable for most damage detection works.

4

4.2

RESULTS AND DISCUSSIONS

Four techniques were used to assess the inflicted damage on the structure. These are the modal frequencies and the damping ratios, the mode shapes and the MAC values. 4.1

Modal frequencies and damping ratios

The first four natural frequencies and damping ratios are shown in Table 1 for both experimental and theoretical work. The frequencies decrease from the intact state to the severe case for all the experimental modes. On the other hand, significant changes in frequencies were noticed on the severe damage on the shear connectors for the theoretical case. These changes indicate that failure in shear connectors reduce the global stiffness of the system.

Table 1. Comparisons of experimental and theoretical natural frequencies. Experimental work

Numerical

Mode

Case

Hz

%

Damping

Hz

%

1st bending

X0 X1 X2 X0 X1 X2 X0 X1 X2 X0 X1 X2

20.0 19.7 19.6 30.2 28.6 27.9 65.2 64.5 61.7 75.5 64.0 73.4

– 1.0 2.0 – 2.6 1.3 – 1.4 1.2 – 1.1 5.8

0.08 0.04 0.05 0.50 0.43 0.74 0.10 0.12 0.03 2.30 0.00 0.01

19.8 19.8 19.5 30.2 30.1 29.8 65.1 65.1 63.6 80.9 80.8 80.6

– 0 1.5 – 0.3 1.3 – 0 2.3 – 0.1 0.4

Torsional

2nd bending

Combined

Modal assurance criterion (MAC)

MAC values were calculated to quantify the correlation between mode shapes for different damage tests. Good correlation exists between the first and second bending modes for both the undamaged-light damage and undamaged-severe damage cases. On the other hand, the torsional and combined bending and torsion modes give lower MAC values. This might be attributed to the loss of connection between the girders and the slab. 4.3

Mode shapes

Very little information was deduced from the mode shapes shown without further analysis, however. The torsional and combined bending and torsional modes were expected to indicate the positions of the removed connectors as explained above.

5

CONCLUDING REMARKS

The frequency change method was found to be a good tool for detecting the change of stiffness in a structure resulting from the partial loss of shear connector integrity. The damping ratios, on the other hand, were not consistent and therefore did not indicate whether the change is due to removal of the connectors or other parameters. This agrees with work by Yong, et al 2007. The MAC values indicate that the failure of shear connectors affects the torsional modes of the structure. These values, in the same way as natural frequencies, show a global change of the structural stiffness and cannot localise the damage. The use of vibration techniques might be a useful tool in assessing and locating damage in civil engineering structures. However, local vibration techniques have to be employed if damage of the shear connectors in concrete beam and slab bridges is to be found.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

New approach to in-situ evaluation of watertightness of reinforced concrete underground structures Andrey Zakorshmenny ZAO “Triada-Holding”, Moscow, Russia

ABSTRACT: All existing techniques of watertightness testing of reinforced concrete structures are based on modeling the process of water absorption by a structure. This process changes physical and chemical properties of structural material and causes its gradual deterioration. All testing techniques have certain limitations: sensitivity to humidity and temperature deviations; change of air permeability in relation to the applied pressure; dependence of the measured results on the quality of surface preparation, etc. When we use the technique implying drilling of testing boreholes, the drilling process can cause some discontinuities in concrete around the area to be tested. Unfortunately, this process cannot be controlled. Besides, cracks can appear in concrete and they will facilitate water/air migration within a structure. All techniques for underground structures are based on testing only the inner surface of concrete elements, whereas their outer surface and quality of concrete are not checked. Distinction of our new method (developed by us and patented) is that it determines watertightness throughout the whole structure. Our new testing instrument VBK-1 drills a testing borehole with a diamond drill bit; it is equipped with a special water cooling system to reduce potential cracking, increase drilling rates and minimize bit wear. In order to eliminate the interference of concrete wetting with the accuracy of testing results as well as to be able to make corrections, it is necessary to measure electrical resistance of concrete.

1 1.1

INTRODUCTION

1.3

Significance of the subject

Nowadays it is absolutely impossible to be successful in construction of modern underground complexes if you do not possess a quality control system which has been carefully thought over and is able to guarantee a trouble free operation. Normally concretes with high watertightness grades (W8-W14) are to be used to build underground structures. However, if we try to evaluate what we actually get with regard to all manufacturing defects the grade is not higher than W2-W4. Besides there is always a residual water ingress which often exceeds what has been anticipated. All these result in an accelerated wear of structures.

Methods of control based on laboratory testing of samples as well as their results have but very little in common with the actual state of structures. Subject to the influence of various factors both during construction and subsequent operation, structural state deteriorates and meets design requirements no longer. It brings us to the conclusion that periodic watertightness control of underground reinforced concrete structures is a must during the whole period of operation.

2 2.1

1.2

Problems of watertightness check of underground reinforced concrete structures

The standards consider watertightness of the concrete cover only (up to 50 mm) which provides primary protection of reinforcement. After a certain period of operation, though, being subject to the action of chlorides and CO2 the concrete tends to demand repair. Choice of testing procedure and subsequently the measured results are influenced by the quality of the surface to be tested, especially if it has some structural defects. Many regulations do not take into account temperature and humidity fluctuations as well as the influence of inner reinforcement on structural watertightness.

Necessity of periodic checking

BASIC PRINCIPLES Analysis of the existing methods

Our analysis of the existing testing methods of concrete watertightness shows that methods based on air permeability measuring principle are quick and easy unlike testing methods involving the use of water, though there is rather a big spread in the measured values and one has to perform a lot more tests as compared with the direct testing methods. When we speak about methods involving the use of water, it must be mentioned that it is very important to provide an absolutely tight contact between the testing apparatus and the concrete surface. On the other hand, one must consider a wide range of various parameters, such as signal frequency; geometry of samples; choice of testing areas with

211

regard to reinforcement location; ambient temperature during testing; concrete composition, etc. Stationary laboratory testing equipment cannot be used for in-situ concrete watertightness testing of underground structures. 2.2

New technique to check watertightness of underground reinforced concrete structures

After we analyzed all the methods available, we developed a fundamentally new technique to evaluate in-situ concrete watertightness (patented in Russian Federation). If we take, for example, tunnel linings all the existing techniques allow to evaluate the quality of the surface (∏) from inside the tunnel (a) with the concrete cover up to 50 mm and it differs from the watertightness level of the whole structure (Fig. 1). Watertightness of the outer surface (∏1) of the tunnel lining from the outside (∏1) as well as of the adjacent areas is never checked. The outer surface of the lining is constantly subject to moistening whereas inside the tunnel high concentrations of oxygen and CO2 are observed. If it were possible to check watertightness of the outer surface (∏1), it could help to detect steel reinforcement corrosion at early stages. Moreover, it would also help to save substantial resources which are needed if structures in preemergency or emergency states have to be repaired. 2.3

2.4

Analytical background and equipment needed to check structural watertightness

With regard to the principal procedure of watertightness testing we developed a special device to determine structural water absorption (VBK-1). Obtaining of the results is based on the general principles of the filtering theory. Analytical model takes into account Darcy linear law and equation of flow continuity. When the device is being operated, the pressure in a borehole is kept on the same level and the medium is assumed being isotropic. 2.5

Parametric considerations

Thus, measuring the amount of water absorbed by a structure we can calculate filtering coefficient of concrete. Our method does not take into account carbonation and presence of chlorides penetrating into the structure from the surface as surface layer does not belong to the working range of our device BÞK-1. Factors that can influence the results of watertightness testing are compensated by additional evaluating of porosity, relative water content, reinforcement ratio of structures, irrespective of the state of concrete surface layer as well as viscosity of water at various ambient temperatures. This allows to check quality of concrete in the presence of aggressive media, maintaining accurate results.

Preparations to watertightness checking

The peculiarities of the method developed by us can be described as follows. A testing hole is drilled with the help of a diamond drill bit with a water-cooling system. Such technique allows to cut concrete without inducing dynamic loads. Water-cooling makes the process of drilling 3–4 times more profitable as compared to air-cooling as well as helps to increase drilling speed, lower drilling bits wear, preserve the texture of concrete (unlike percussive-rotary drilling).

3

CONCLUSIONS

The most trustworthy criterion of structural reliability is constant control of structural state during the whole period of its operation. It brings down the risk of emergency situations which may be provoked by loss of bearing capacity due to reinforcement corrosion, uncontrolled water ingress into a structure or some other reason.

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Experimental evaluation of indirect sonic wave transmission technique in the diagnosis and monitoring of concrete slabs G. Concu, B. De Nicolo, F. Mistretta & L. Pani Department of Structural Engineering, University of Cagliari, Cagliari, Italy

ABSTRACT: An experimental program has been started with the purpose of evaluating the reliability of the sonic Indirect Transmission Technique (ITT) in the damage diagnosis of reinforced concrete elements. Four concrete slabs, two lightened by polystyrene and two entirely made of reinforced concrete, have been tested carrying out on them the ITT before and after a punching-load test. The ITT has been performed settling the transducers on concentric circumferences instead of on nodes of a squared mesh as usually done in standard operating procedure. This innovative procedure permits to intercept anomalies oriented in different ways, thus leading to a richer information concerning the tested areas. Results show that the ITT is a suitable technique for crack-damage detecting.

1

INTRODUCTION

Non-destructive techniques based on sonic wave propagation are often used in structural diagnosis. They are applied directly in the field for investigations of a wide range of structures and infrastructures, and in laboratory for the characterisation of materials. Non-destructive sonic methods (SMs) are based on measurements of the velocity v of acoustic waves propagating through the material. They are preferentially carried out applying the Direct Transmission Technique (DTT), in which the wave is transmitted by a transducer (Emitter) through the test object and received by a second transducer (Receiver) on the opposite side. This allows to measure the time t that the wave needs to travel through the object’s thickness, from the emitter to the receiver, along a path of length l; the average velocity of the wave is then obtained from the ratio l/t. The wave velocity is directly related to structure’s elastic parameters, thus its analysis provides information crucial for inspections of structures’ stability and lastingness. The DTT is very effective, since the broad direction of wave propagation is perpendicular to the source surface and the signal travels through the entire thickness of the item. Standards concerning the determination of waves velocity in structures—e.g. European EN 12504–4—suggest, therefore, the application of this kind of signals transmission. Nevertheless, there are many kind of structures, e.g. slabs, retaining walls, piers, in which the DTT cannot be performed, because only one side of the item is accessible. In these cases the Indirect Transmission Technique (ITT), in which both the emitter and the receiver transducers are placed on the same side of the investigated object, might be used. ITT is less effective than the DTT because the amplitude of the received signal is lower, and the pulse propagates in a concrete layer just

beneath the surface. These remarks have since now not allowed the ITT systematic development, and the scientific literature concerning ITT use is still quite poor. Despite that, ITT skills of ease to be performed, high potential to evaluate the quality and the characteristics of concrete covering on site, immediacy and low cost, claim to thorough examine its suitability in concrete diagnosis on site, and then to develop studies concerning the standardisation of its application. In the light of this, an experimental program has been started with the purpose of evaluating the reliability of ITT in the diagnosis and monitoring of concrete elements. 2

EXPERIMENTAL

The research has been carried out analysing the propagation of sonic waves along the surface of four concrete slabs. The sonic test has been performed on the slabs before and after that a punching test has been carried out on them. In this way, the evaluation of ITT effectiveness in detecting the type and level of damage has been evaluated. Two sets of reinforced concrete slabs, having the same dimension, mix design, reinforcement type and boundary conditions, have been tested. The A set consists of two slabs lightened by the interposition of a polystyrene layer 13 mm thick, whilst the B set consists of two slabs entirely made of reinforced concrete. The square slabs have side of 1200 mm and thickness of 50 mm, and they are reinforced by two welded grids (side of 100 mm, ∅ of 5 mm, step of 8 mm) settled close to the top and the bottom surfaces respectively. The punching shear resistance of the slabs has been evaluated. The sonic test has been performed on the four slabs using the ITT. The transducers have been settled on concentric circumferences instead of on nodes of a

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squared mesh as usually done in standard operating procedure. This novel approach has the advantage of considering other routes of signals propagation beside the traditional one obtained intersecting vertical and horizontal paths. This allows to intercept anomalies oriented in different ways, thus leading to a richer information concerning the tested areas. In detail the adopted measurement schema is the following: 48 receiving points settled on 6 concentric circumferences having radius respectively of 100, 200, 300, 400, 500, 600 mm, and one emitting point settled on the circumferences centre. Data have been collected recording each measurement carried out hitting the circumferences centre with an instrumented hammer and receiving the signal at one of the receiving points. Data processing has been performed considering two sets of measurements: set C refers to the six circumferences, whilst set L refers to the eight straight lines that start from the slab centre.

Figure 1. Set C: average velocity (m/s) before the load test.

3

RESULTS

Achieved results point out that:

Figure 2. Set L: average velocity (m/s) before the load test.

− the ITT is able to qualitatively identify the lightened slabs, which indeed are characterized by lower values of the sonic velocity; − after the load test, the sonic velocity always decreases, and the decrement is much more significant in the slabs entirely made of reinforced concrete rather than in the lightened slabs; − the use of measurements paths arranged in concentric circumferences and in straight lines that start from the slab centre allows one to detect the areas where the damage due to the load test is most significant. Figure 3. Set C: average velocity after the load test Va (m/s) compared with average velocity before the load test Vb (m/s).

4

Figure 4. Set L: average velocity after the load test Va (m/s) compared with average velocity before the load test Vb (m/s).

An experimental program has been started with the purpose of evaluating the reliability of the sonic indirect transmission technique (ITT) in the damage diagnosis of reinforced concrete slabs punching-loaded. Four slabs, two lightened by polystyrene and two entirely made of reinforced concrete, have been sonic-tested before and after the punching-load test. The experimental data collected so far show that the ITT is suitable for crack-damage detecting. Future research will investigate the method effectiveness applying it to different concrete structures, and integrating the sonic velocity with other waves features, e.g. pick to pick amplitude, elastic energy, signals spectrum.

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CONCLUSIONS

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Ultrasonic imaging of post-tensioned concrete elements: New techniques for reliable localization of grouting defects M. Krause, B. Gräfe, F. Mielentz, B. Milmann, M. Friese & H. Wiggenhauser BAM—Federal Institute for Materials Research and Testing (BAM), Berlin, Germany

K. Mayer University of Kassel, Germany

ABSTRACT: Quality assurance of grouting and reliable localisation of defects is an important testing problem. A new possibility for this task is to consider the phase information in ultrasonic imaging. Applying this method, the reconstruction calculation allows the distinction, if a reflector is metallic, or if the reflection is caused by an air inclusion. Exampless for this new imaging technique are described. For making ultrasonic echo measurements faster new techniques are presented: Linear array measurement and air coupled ultrasonic transducers. The organisation of the short version of the paper is: 1. Ultrasonic imaging including phase evaluation; 2. Imaging and modelling of grouting faults; 3. Linear Array technique; 4. Air coupled ultrasound.

1

ULTRASONIC IMAGING INCLUDING PHASE EVALUATION

Many testing tasks for concrete elements to be solved with elastic waves are related to multilayer systems. Especially this is important for the investigation of tendon ducts. Much effort was invested and success was achieved in the last decade by applying ultrasonic imaging techniques in this field. When the ultrasonic waves transmit a prestressed concrete element, several interfaces appear: Concrete/steel sheet of tendon duct; steel sheet/grouting mortar; grouting mortar/steel wires (strands) and so forth. The scatter and reflection depend of the thickness and form of the steel sheets and the form of the steel wires. Additionally to the grouting faults of interest sometimes air pores strongly appear in the surrounding of the interfaces and influence the wave propagation. A new feature of the evaluation was recently installed in the acoustic imaging procedure, resulting from the cooperation between BAM and University of Kassel. Considering the phase jump of the elastic shear waves being reflected at interfaces of a material having higher acoustic impedance, air interfaces can be indicated by their opposite sign of the pulse shape. In figure A this principle is applied to the reconstruction calculation of the data set using the usual SAFT-evaluation. The negative sign of the pulse maximum now indicates the air inclusions, whereas the metallic reflection pulse has a positive sign. In this way an unintended delamination appearing in the specimen is identified, which was not clearly possible regarding at the reflection intensity.

Colour table in a. u.

Figure A. Indication of the sign of the pulse shape. Light gray (negative sign) corresponds to air reflection, dark gray (positive sign) corresponds to steel reflection. Thus a unintended delamination is indicated down right.

2

IMAGING OF GROUTING FAULTS WITH PHASE EVALUATION

In 2002 a Large Concrete Slab (LCS) was designed and constructed at BAM in order to realize typical testing tasks for the comparison of different NDT methods and their validation. One part of this concrete slab contains 11 tendon ducts in the diameter range from 40 mm to 120 mm having concrete cover between 80 mm and 200 mm with artificial grouting faults. The grouting was performed similar to practical methods. For the present paper the results are presented for one typical duct (named D3), which exemplarily

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demonstrates the potential of phase evaluation for localizing grouting faults in tendon ducts. It is a steel bar in a duct having a diameter of 35 mm. The experiments were carried out with a 55 kHz point contact transducer working with automated scanner like for specimen containing metal plates described in paragraph 1. Figure B depicts the results of ultrasonic imaging experiments of the tendon D3 in comparison with the plan of artificial grouting faults (shaded in figure Ba). There location was verified applying γ-radiography. (Figure Bc) depicts the magnitude of ultrasonic reflection intensity as longitudinal section along the tendon duct (ultrasonic B-scan). There is no significant change in reflection intensity at ungrouted areas. Otherwise the phase evaluation shown in (Figure B) shows a very clear difference between air filled and well grouted areas. The false colour scheme indicates a shift of the phase value of 180° between grouted and ungrouted areas. This means that the phase evaluation clearly indicated the difference between air filled and grouted areas, and allows to distinguish between reflection at air and reflection at steel. This is not possible from the intensity representation in this case. The void at the right end of tendon D3 is only partly indicated: The reason is that this void was produced only at one half inside the tendon duct.

4 AIR-COUPLED ULTRASOUND FOR FAST SCANNING

3 TENDON DUCT EVALUATION WITH LINEAR ARRAY TECHNIQUE

Figure C. Linear array with dry contact transducers (top), principle of data acquisition (bottom).

In order to accelerate ultrasonic echo measurements and to allow an even faster performance as it is possible with scanner application using dry contact transducers, air-coupled ultrasonic (ACU) transducers can be used. It has been shown that ultrasonic echo measurements of 200 mm thick concrete slabs can be carried out with ACU transducers as well as the imaging of voids.

Air Void

Air Void

partial grouted

Figure B. Results for localizing grouting faults for tendon duct D3 in a large test specimen: a) Sketch of the specimen with desired grouting faults (air filled voids) c) Result of scanning with 55 kHz shear waves: magnitude representation from 3D-SAFT reconstruction e) Phase values of reflecting pulses allows the clear distinction between air inclusion (ϕ = 190°) and steel bar (ϕ = 10°).

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Practical experience of geo-radar technique application in the course of an integrated inspection of historical buildings Alexey Kirilenko & Arseni Baukov ZAO “Triada-Holding”, Moscow, Russia

ABSTRACT: The report describes the advantages of geo-radar testing combined with other inspection techniques to examine various buildings and structures. The report contains a case study (slabs and floors in the Pushkin State Museum of Fine Arts built almost 100 years ago in the centre of Moscow). This museum is one of the most prominent museums not only in Russia but in Europe. The report gives the results of the geo-radar technique application to solve such original tasks, as defining the inner structure of building elements as well as groups of elements when it was impossible to use traditional destructive methods for the purpose; discovering of rooms, which were not mentioned in the construction documentation or shown in the drawings; locating of structural defects due to vibrations from trains running in the tunnels of a near-by metro line, etc. The report analyses advantages and limitations of geo-radar applicability for in-situ quality control of structures. We managed to establish optimum inspection parameters as well as inspection modes and procedures of obtained data analysis. It should be pointed out that the combined application of various inspection techniques, on one hand, provides an overall implementation of each of them whereas on the other hand, allows us to obtain as much detailed data as possible on the inspected project as a whole.

1

2

INTRODUCTION

Nowadays a lot of advanced techniques are used for inspection of various engineering structures. One of them which is being developed very actively is geo-radar sub-surface probing. The essence of this technique can be described as follows: emission and propagation of electromagnetic metric waves and microwaves through a dielectric medium to be tested and subsequent detection of signals reflected at the border between layers with different electro-physical properties. This technique was first tested in the seventies of the past century, though since then it has not generated a stable interest. Application scope was restricted to geophysics, i.e. investigation of various strata, location of ground water, etc. In last decades thanks to development of advanced instrumentation and procedures for comprehensive data analysis geo-radar techniques started being applied for structural inspection as well. Though until quite recently the range of tasks was limited to reinforcement location (mainly upper level), tracing of electric cables and water pipelines, measuring structural thicknesses, search of leakages. Such limited application can be explained by the fact that geo-radar technique was not used in conjunction with other investigation techniques. We tried to combine various techniques, including geo-radar, thus expanding the scope of its applications.

2.1

PRACTICAL APPLICATION OF GEO-RADAR TECHNIQUE IN THE COURSE OF COMPLEX PROGRAMME OF TECHNICAL STATE ASSESSMENT OF (PUSHKIN STATE MUSEUMOF FINE ARTS) Project description

One of the examples of successful applications is the Pushkin State Museum of Fine Arts, one of the most prominent museums not only in Russia but in Europe (Fig. 1). This museum was built almost 100 years ago in the centre of Moscow, not far from the Kremlin. Complex investigation program included inspection of slabs to define their composition, floor foundation, load-bearing structures both of slabs and floors. 2.2

Aim of inspection

The reason was that there were lots of cracks on the surfaces of floors and ceilings. Crack pattern was so quaint that nobody could give any definite explanation why the cracks had appeared. Not a single out of a great many of technical reports on structural state assessment contained any information on the composition or quality of slabs and floors. Another task was to locate unused areas under the rooms of the ground floor. There was no valid data on composition/state of soil under the building foundation.

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It must be said that we were not allowed to disrupt or break any structures because the museum is a historic building. Ceilings are decorated with paintings and stucco. Self-leveling floors preserved their authentic marble aggregate decoration of the beginning of the last century. All the above mentioned dictated the use of nondestructive methods of structural investigation. Georadar sub-surface probing was specified as the basic technique for inspection as it could provide continuous survey of structures, the majority of which had a one-sided access for testing. 2.3

Instrumentation and software. Working procedure

We used geo-radar “OKO” of Russian manufacture with a set of screened antennas (emitting frequencies of 400, 700 and 1700 MHz) (Fig. 2). We chose those frequencies based on electrophysical parameters and geometry of objects to be tested. To make the data obtained more comprehensive we used different antennas to survey same locations. The data was processed with the help of RadExpro-2 software. It allowed to set adjustment for discrepancies, correction filtering, migratory data conversion, pulse deconvolution and conversion. Depending on the task an individual graph was made after each survey based on the obtained and processed data. 2.4

To check the results obtained in the course of geo-radar probing we drilled several micro boreholes. Not to spoil decorative elements, drilling was performed either through cracks wider than 10 mm or in the spots where the tiles were spalled. Several holes were drilled in places “hidden” from the visitors by the exhibited art works. The results of the drilling fully confirmed the data of our geo-radar survey. To survey slab structure between the basement and the ground floor we used screened antennas of 700 and 1700 MHz. The results showed that they consisted mainly of concrete arched slabs of various cross-sections (cylindrical, cross-vaulting). Arched slabs were supported by metal beams and columns. The results were confirmed by the use of Profometer-5 (rebar locator manufactured by Proceq SA, Switzerland). During our investigation we managed to discover unused rooms in the basement, partially filled with earth. These rooms can be used in future for museum needs. Geo-radar probing in the basement allowed to reveal the structure of soil under the foundation of the building. The basement floor (concrete screed) was15–20 cm thick. It rested on a soil cushion of 0,5 m under which there was soil mixed with construction wastes and material debris of earlier buildings (up to 2 m deep). Several man-caused leakages were also discovered.

Inspection results

The results showed that load-bearing structures of the 1st floor were represented by metal I-beams supported by brick walls. Secondary metal beams were placed on top of the load-bearing beams as a supporting structure for self-leveling floors. Ceilings in the rooms of the ground floor were false. Total thickness of the slab structure (excluding decorative elements of the ground floor ceiling) was about 1 m. Reflected wave speed allowed to state that the space between the ground floor ceiling and the floor of the 1st floor was infilled with air. Conducted geo-radar probing did not show any major structural defects. Typical defects were cracks in self-leveling floors and ceramic floor tiles. When we compared the results of visual inspection and georadar probing, it became clear that location of cracks was associated with the location of metal beams.

3

CONCLUSIONS

Now that we finished the project we can definitely state that only complex investigation we carried out could give such comprehensive results. Among these are: slab structure between the floors and in the basement; location of structural defects; associating of cracks to the position of bearing structural elements; revealing cavities and unused rooms as well as infrastructural mains under the basement floor, etc. To conclude our report, we would like to underline once again that only a combination of various inspection methods could provide such completeness of obtained results, which would hardly be possible with the use of a single technique, no matter how smart it was.

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Wireless monitoring of structures including acoustic emission techniques C.U. Grosse, M. Krüger & S. Bachmaier Materialprüfungsanstalt Universität Stuttgart (Materials Testing Institute of the University of Stuttgart), Germany

ABSTRACT: The inspection of building structures is currently made by visual inspection or by wired sensor techniques, which are relatively expensive, vulnerable to damage, and time consuming to install. In contrast, wireless sensor networks are easy to deploy and flexible in application so that the network can adjust to the individual structure. Different sensing techniques can be used with such a network, but acoustic emission techniques have been rarely utilized. With the use of Acoustic Emission (AE) techniques it is possible to detect internal structural damage from cracks propagating during the routine use of a structure or the break of wires of pre-stressed elements for example. Most of the existing AE data analysis techniques are not appropriate for the requirements of a wireless network, especially power consumption. Sensors with low price are required for AE systems to be accepted. To fully utilize the power of the acoustic emission technique on large, extended structures, recording and analysis techniques need more powerful algorithms to handle and reduce the immense amount of data generated. In particular, this paper deals with the optimization of the network to record different type of data including AE data. The basic principles of a wireless monitoring system equipped with MEMS sensors is presented along with a first prototype able to record temperature, moisture, strain and other data continuously.

1

WIRELESS MONITORING

Existing monitoring systems use traditional wired sensor technologies and several other devices that are time consuming to install and relatively expensive (compared to the value of the structure). Typically they are using a large number of sensors (i.e. more than ten) which are connected through long cables and will therefore be installed only on a few structures. A wireless monitoring system with MEMS sensors could reduce these costs significantly. MEMS are small integrated devices or systems combining electrical and mechanical components that could be produced for about 50 € each or less. Wireless sensor networks consist of many nodes (motes) having one or several different sensors on board. After the recording and a preliminary analysis of the data in the mote, the data has to be transmitted using, for example, a radio transmission system to a base station or supervisor for further data processing or proper generation of alarm messages. For the transmission of data using sensor nodes in a network of motes several topologies exist including the star and the multi-hop topology. The main advantage of multi-hop techniques are the transmission power efficiency, because only a fraction of energy is necessary to transmit data compared to other techniques; the data are transmitted just to the neighbor nodes and not necessarily to the sink. This reduces also the danger of interference since a node communicates only with a few others. However, this requires sophisticated network protocols including ad hoc configuration capabilities as well as self-

configuration, calibration and encryption. A next step is a clustered multi-hop technology. Motes in a cluster share the data of all sensors attached to these motes. A pre-processing of the data is done in the cluster prior to transmission via the other clusters in the multi-hop network. Intelligent data processing in the motes or clusters enables pattern recognition algorithms, which can additionally reduce the power consumption. Only meaningful data are transmitted to the sink. The data sink is further extracting information out of the data using knowledge-based algorithms sending afterwards the information to the responsible person (construction engineer, owner) using automated email messages or short message systems to mobile phones. A sensing system based on wireless motes has several more advantages. Such a system is easy and cost efficient to be applied to structures. It can be used on one structure for a while and when or if the stakeholder decides to have enough data collected at this particular structural part the system can easily be deployed somewhere else. Additionally, a variety of sensors can be used to get information about the status of the structure. Further on, comparison of time series acquired by recording different physical quantities results in a drastic improvement of reliability and lowers the detection threshold of deterioration. Establishment of a correlation between data and structural performance is difficult and should be based on the data interpretation expertise of the user, implying a natural application of Bayesian statistics. Embedding some local processing capabilities within the sensor networks is desirable. For example, the temperature

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In-Mote Signal Detection and Analysis Procedure Analog Signal Conditioning & Detection (Hardware: e.g. Gain, Filtering, Analog Threshold)

A/D Conversion

Cluster Analysis

In-Mote Data Analysis (e.g. Denoising, Filtering, Waveform Analysis, Signal Discrimination)

Clustering Procedure

Cluster Analysis (e.g. Event Localization, Pattern Recognition, Statistics)

Threshold Level Gain

Reference Data Basis (Stored in Mote)

Reference Data Basis (Stored in Mote)

Sensor

Resolution Sample Rate

Structural Analysis Postprocessing (if required)

User Notification

Data updating

Data updating

Sensor

Structural Analysis

Model Updating

Rules

e.g. Filter Frequencies Correlation Coefficients Reference Signals Signal Parameter

Model Updating

Structural Data Basis (Stored in central data base)

Update of Reference data (Regressive Models)

Model Updating

e.g. Reference Cluster, Reference Pattern

Remote Control User Interaction & User Intervention

Figure 1. Principle sketch of signal detection and analysis procedure optimized for low power operation [Krüger et al 2007].

data gathered from numerous sensors could be fed into one or more other sensors on the network for processing. A weighted average could then be calculated and transmitted to the user, significantly reducing the amount of data flying around the network. Finally, two other advantages of wireless sensor networks have to be stressed. Scalability can be an issue if the stakeholder wants to extend the monitoring area or need more data. Existing WSN techniques enables for self-organization of such networks so that sensor nodes can be added or removed at any time without time consuming user guided reorganization of the WSN. Additionally, the implemented pre-processing algorithms might need an update from time to time to adjust to the user requirements or for a more efficient data reduction. Most of the developed sensor nodes have the capability to be reprogrammable, i.e. that the user can change the algorithms implemented in each sensor with pressing a button. Fig. 1 shows the hardware setup as well as the stepwise analysis procedure for interpreting the acquired data considering low power consumption. The procedure starts with the in-mote detection and signal analysis followed by a cluster analysis and at least a structural analysis. The analysis techniques could include an updating of the implemented models as well as the data used for comparative analysis. Therefore, the sensor network could become a neural network.

2

CONCLUSIONS

Structural health monitoring deals with the more or less continuous recording of data obtained from the structure. Based It is obviously very helpful not to base an analysis on one physical quantity alone or on one sensor. The reliability of the monitoring system is fairly enhanced combining the information obtained at different sensor nodes. Further on, comparison of time series obtained by recording different physical quantities results in a drastic improvement of reliability and lowers the detection threshold of deterioration. Establishment of a correlation between data and structural performance is difficult and should be based on the data interpretation expertise of the user. This combination can be done even in terms of an intelligent pre-processing of data in the mote or in a cluster of motes which can additionally reduce the power consumption. A wireless sensor network system based on MEMS and hybrid sensors is developed by a team of scientists from different institutions (MPA, UC Berkeley, Smartmote). The network is equipped with motes and will be available for a very low budget. Since prototypes are already available, the system is now undergoing an optimization process regarding power consumption, data acquisition and data aggregation, signal analysis and data reduction.

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OSSCAR – Development of an On-Site SCAnneR for automated non-destructive bridge testing A. Taffe, T. Kind, M. Stoppel & H. Wiggenhauser Federal Institute for Materials Research and Testing (BAM), Berlin, Germany

ABSTRACT: An intelligent combination of methods along with advanced data processing makes best use of today’s capabilities of non-destructive testing in civil engineering (NDT-CE). 3D images resulting from such testing show the engineer the interrelation between testing results and the construction plan. Automated data collection is necessary to carry out such investigations economically. Scanning systems developed at BAM combine automated data collection, method combination and intelligent data interpretation to present 3D reconstructed data of an investigated concrete member. In the OSSCAR-project (On-Site SCAnneR) funded by the Federal Ministry of Economics and Technology in the InnoNet research programme a device for automated bridge testing is designed for typical testing tasks in bridge assessment. This new development will be able to carry out all testing tasks with one device and one computer program to run the scanner, to collect data, to carry out data interpretation and to image the results.

1

BRIDGE TESTING ACCORDING TO GERMAN STANDARD DIN 1076

Engineering structures in connection with federal highways are tested according to German Standard DIN 1076 in regular intervals. In case of complex damage patterns additional information is necessary by detailed analysis on the object level (OSA). Non-destructiv testing methods are part of this further analysis. Many bridges date from the 1960 and 1970ies with typical damages of the prestressed steel due to incomplete grouting of the tendon ducts, low degree of reinforcement and cracks in the coupling joints. On-site scanning systems for automated data collection and imaging of results are a powerful tool for the engineer to make decisions for a durable and safe concrete repair. 2

AUTOMATED NDT-DEVICES

The imaging of results of the inner structure below a tested area requires a dense measuring grid between 2 and 5 cm. The spacing between each point has to be very precise to be processed by reconstruction calculation. This leads to an amount of several thousands measuring points. Such data collection can only be carried out economically by automated testing. BAM has developed since 2002 several applications ready for horizontal areas and overhead testing as well as applications ready for vertical areas even in narrow areas like under a rim beam. These so called “On-Site Scanners” can be used for radar, ultrasonic and impactecho testing. Ultrasonic and impact-echo can be carried out at the same time. With radar and contact-free positioning of the antenna 15 m²/h are possible. A 2 cm grid with both acoustic methods ultrasonic and impactecho at the same time allows only 0.4 m²/h.

Figure 1.

Scanner with vacuum plate for overhead use.

For non-destructive fixing of the scanner vacuum plates have been developed (Fig. 1). They can be used on vertical areas as well as overhead. 3 3.1

NEW DEVELOPMENTS IN AUTOMATION The OSSCAR research project

In 2008 the research project OSSCAR (On-Site SCAnneR) funded by the German Ministry of Economics and Technology was started. Like shown in Fig. 2 three research institutions (BAM, Fraunhofer IZfP, BASt) develop together with partners from industry (Bilfinger Berger Vorspanntechnik) and eight small and medium-sized enterprises a scanning system ready for on-site use with controller for data collection, data assessment and imaging of results in one platform. Testing tasks to be carried out with the scanner are: • Imaging of the geometry. • Location, depth, diameter of multi-layer reinforcement. • Location and depth of multi-layer tendon ducts.

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• Quality assurance for complete grouting of tendon ducts. • Location of grouting defects in existing constructions. 3.2

Requirements

Within the OSSCAR project a lightweight, flexible and easy to handle scanner should be developed. The new frame provides testing area of 50 × 100 cm² with a quick installation and easy changing the measuring position. Due to the small dimensions it fits in narrow corners and can be used even for columns. It should be easy to transport, quick and save to mount on the structure. Therefore a vacuum system is under development. 3.3

First Results and further development

A carriageway of a post tensioned concrete bridge has been investigated with ultrasonic-echo from the bottom side. The reinforcement in the areas near the surface consists of three layers with a diameter from 16 to 25 mm with 15 cm spacing. After applying

reconstruction calculation (SAFT) the track of the tendon duct is visible up to a depth of 1 m (Fig. 3). Even the top side of the tendon duct can be distinguished from the bottom site. In this case no evidence for an incomplete grouting has been detected. The projection of several cross-sections a long a line of 1 m shown in Fig. 4 reveals that the back wall can be detected at depth of 1.75 m. This is 50% more than the limit of detection that was published at ICCRRR in 2005 for thickness measurement at foundation slabs with a comparable device. For a reliable detection of grouting defects in the OSSCAR-project a very promising technique of data assessment is used: the so-called phase sensitive characterisation of reflected signals. It allows distinguishing whether a reflection comes from a steel/concrete interface (sound area) or a steel/air interface (grouting defect). This technique will be applied to data from artificial grouting defects measured at specially constructed specimens or real constructions regarding the fact that real grouting defects typically are partly grouted.

Figure 4. Projection of several cross-sections with a back wall reflection of depth of 1.75 m. Figure 2. The OSSCAR-consortium.

Figure 3. Carriageway slab of a post tensioned concrete bridge. Top: Longitudinal section of the investigated area. Bottom: Results from ultrasonic-echo with SAFT-reconstruction. Top and bottom side of the tendon duct can be distinguished up to a depth of 1 m.

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Development of a portable LIBS-device for quality assurance in concrete repair A. Taffe, D. Schaurich & G. Wilsch Federal Institute for Materials Research and Testing (BAM), Berlin, Germany

F. Weritz Federal Ministry of Economics and Technology, Berlin, Germany

ABSTRACT: The Laser-Induced Breakdown Spectroscopy, LIBS, has been applied to determine the chlorine and sulphur content of concrete structures like bridge decks, parking garages and sewage treatment plants. LIBS as a laboratory method has already been presented at ICCRRR 2005. For quality assurance in the concrete repair process it has proved to be necessary to get LIBS results on-site. The on-site use of LIBS helps the engineer to come to the decision of sufficient concrete removal on-site under ongoing work and by testing a dense grid of measurement points, e.g. one per square metre. This procedure ensures that chloride or sulphate contaminated concrete is fully removed—as much as necessary and not more. BAM has developed a portable LIBS-device using the expertise of LIBS-testing in the laboratory. This device is designed to test horizontal and vertical concrete surfaces that can be reached by the probe. The device is ready to measure chloride and sulphate contents on concrete surfaces. Results will be presented at the conference.

1

NECESSARY QUALITY ASSURANCE DURING CONCRETE REPAIR WORKS

For a durable concrete repair in case of external chloride or sulphate ingress it is necessary to remove the contaminated concrete. A depth profile gained with LIBS laboratory equipment in a mm scale helps the engineer in the planning process to decide the depth of the concrete removal. A portable LIBS equipment for on-site use will help to make a quick decision whether concrete removal is sufficient or not. Only access to the prepared surface that should be tested is necessary.

2

MEASURING PRINCIPLE, EXPERIMENTAL SETUP AND CALIBRATION

The measuring principle of the laboratory or the onsite application is the same: An intense laser pulse is focused on the surface of the specimen generating a plasma plume. During the relaxation process element specific radiation is emitted. The radiation is guided through an optical fiber to the detection unit. The light intensity is measured as a function of the wavelength, i.e. the spectrum. Only the content of elements can be measured. The content of the damaging species can be calculated straightforward from elemental content. Typical LIBS equipment consists of a laser, an optical system focusing the laser beam onto the sample surface, a sample stage, an optical fiber guiding the light to the detection unit, and the detection unit consisting of a spectrometer and a CCD-camera.

To deduce chloride contents from the measured chlorine intensity a calibration has to be carried out. Also the content of sulphur components like SO3 can be calculated from the measured sulphur intensity. There routines for the automated data assessment of the measured data have been developed.

3 3.1

PORTABLE LIBS Portable device

The portable LIBS setup, see Fig. 2, is based on the BAM laboratory set-up like shown in Fig. 1. A rugged Nd-YAG laser is used for plasma generation. The laser pulse is focused on the surface under investigation by

Figure 1.

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Experimental LIBS set-up.

Figure 4. Element distribution of sulphur at a concrete core of a sewage plant (light colour: high sulphur content; dark colour: aggregates).

Figure 2.

Principle of the portable LIBS application.

Figure 5. Depth profile of sulphur measured at concrete four different cores from a sewage plant.

Figure 3. Photos of the portable LIBS prototype. Left: horizontal use; right: vertical use.

a lens that is located in the optical head in front of the laser. To take the heterogeneity of the material into account the laser is mounted on a pan-tilt unit that allows scanning of regions of size 10 cm by 10 cm during a measurement. The pan-tilt unit is mounted on a tripod that is designed to permit measurement of horizontal and vertical areas. Measurements directly on the surface of a column or on the floor of parking garage or bridge are possible (Fig. 3). Additional measurements on cores to get depth information are feasible. The plasma radiation is collected in the direction of the laser beam. A mirror is used to focus the plasma radiation in an optical fibre. For detection different types of spectrometer/detector combinations can be easily connected to the end of the fibre. For an increase in the sensitivity for chlorine detection it is possible to purge the region around the plasma with helium.

The complete beam path from the laser to the surface is for laser safety reasons covered with special bellows. Electrical contacts are used to stop laser pulses if the bellows are moved from the surface. The labview-based software is used for control of the set-up and also for data evaluation. Results are both maps of element distributions like shown for sulphur in Fig. 4 and depth profiles of sulphur ingress measured on cores like shown in Fig. 5. The same results are also possible for chlorine measurements. 3.2

Quality assurance with portable LIBS

In case of chloride contamination the complete removal of contaminated concrete is essential for a durable concrete repair. With a portable LIBS-unit the chloride content of the concrete surface can be deduced. So it is possible to decide on-site whether further concrete removal is necessary or not. The use of portable LIBS in quality assurance: • ensures that contaminated concrete is fully removed and only as much as possible, • helps to make an on-site decision whether further removal is necessary and • helps to save money for costly devices for concrete removal.

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High speed chemical analysis of concrete surfaces using the LIBS method within the ILCOM project M. Bruns & M. Raupach Institute of Building Materials Research (ibac), RWTH Aachen University, Germany

C.D. Gehlen & R. Noll Fraunhofer Institute for Laser Technology (ILT), Aachen, Germany

G. Wilsch & A. Taffe Federal Institute for Materials Research and Testing (BAM), Berlin, Deutschland

ABSTRACT: For the assessment of the status of concrete structures the chemical analysis of the concrete surface (composition, chloride and sulphate content) is of major importance. Laser Induced Breakdown Spectroscopy (LIBS) offers a powerful method to analyse concrete surfaces with a high local resolution and high speed. Concrete surfaces can be scanned line by line and relevant chemical elements can be semi-quantitatively analysed. This allows a quick determination of two-dimensional plots of these elements with much higher local resolutions than by the traditional methods like powder sampling and subsequent chemical analysis. A research and development project called ILCOM has been initiated with several partners from research and industry to develop a laboratory instrument with improved resolution and a mobile instrument to be used on site allowing a quick and adequate analysis of relevant elements like chlorides and sulphates of concrete surfaces. Within a few weeks the laboratory instrument will be finalised and extensive calibration works using prefabricated concrete specimens with defined compositions will be carried out. In parallel the on-site instrument is under development. It is expected that both instruments will offer considerable new possibilities for an effective diagnosis of concrete structures as well as for quality control of concrete works. The project with its partners and aims as well as the actual status of the measurements are presented and discussed.

1

INTRODUCTION

A decisive parameter at the building condition assessment and the repair planning of structures exposed is the depth-staggered determination of the chloride content of the existing concrete. At overcritical chloride contents, it is specified on the basis of the determined chloride profiles, for instance, in which areas and up to which depth the concrete must be removed and replaced. The more information there are concerning the chloride distribution, the better the areas to be removed can be specified. An alternative to the at present customary time and cost intensive wet chemical determination of the chloride content on drill dust specimens taken in staggered depths could be a new method for a fast and highly spatially re-solved analysis of the chemical element distribution in the concrete. This method is currently further developed in collaboration with the Federal Institute for Materials Research and Testing (BAM), the Fraunhofer Institute for Laser Technology (ILT), the Institute of Building Materials Research (ibac) of RWTH Aachen University as well as different small and medium-sized enterprises and industrial partners. The main target of the development is first the fast, spatially resolved, two-dimensional analysis of

the chloride and sulphur distribution in the concrete. Here, as method of analysis, the Laser-Induced Breakdown Spectroscopy, LIBS, is applied. 2

LASER-INDUCED BREAKDOWN SPECTROSCOPY (LIBS)

The functional principle of the imaging LIBS method is schematically displayed in Figure 1. At the LIBS method, a small amount of material is “vaporised” on the specimen surface by means of a laser pulse and a plasma is generated. With the spectral analysis of the plasma emission single chemical elements existing in the vaporised volume can be identified by their element-specific spectral lines. The evaluation of the measured intensity of each spectral line enables the quantitative analysis provided that there is a calibration relation between element content and intensity of the spectral lines. By scanning for instance the surfaces of drilling cores, “element maps” can be prepared within a short period of time. These maps display the local distribution and the varied concentrations of chemical elements across the surface. Within the framework of this research project, two LIBS demonstrators are developed. Demonstrator I (Fig. 2) enables the laser

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analysis of concrete specimens under laboratory conditions at flexible choice of the scan sizes (up to about 15 ⋅ 15 cm2) and of the set of chemical elements to be analysed. It shall be used to develop the method as well as for the validation. Demonstrator II (Fig. 3) is a portable system designed for in situ application, which shall in particular allow for a fast chloride analysis on site, e.g. at repair works.

3

EXAMPLES OF LIBS-ANALYSIS

Figure 4 shows the results of the LIBS measurements on a concrete drilling core (28 mm ⋅ 41 mm) from practice (maritime building) compared to the photography. Starting at the surface, the drilling core was

Figure 4. Areas containing chloride from a drilling core of a maritime building detected by LIBS.

laser

Ar spectrometer

computer

signal electronics

Figure 1. Functional principle of the imaging LIBS method.

Figure 5. Depth profile of the Cl-penetration measured at a drilling core taken from a structure. For comparison, the values determined with standard procedures are displayed as grey bars.

Figure 2.

ILCOM Demonstrator I.

Figure 3.

Laser module of Demonstrator II.

linearly scanned. One LIBS measurement was made per millimetre. The distance between two lines was 2 mm. In a first step, the single measurements were assigned to the aggregate (black points in Fig. 10, right) or the hardened cement paste matrix (grey points in the same figure). Subsequently, the measuring points lying on the hardened cement paste matrix were examined regarding the existence of a chlorine signal. When chlorine was measured, the points were coloured in light grey. By the determination of a weighted mean value of the chlorine content per measurement line, a depth profile of the chloride content arises with a depth resolution of 2 mm (Fig. 5). For comparison, the values determined by standard procedures are depicted. Still in 2008, both demonstrators shall prove their effectiveness in practice. If this proves successful, there will shortly be a new, fast and highly spatially resolved method of analysis for practice and science to determine the chemical element distribution in concrete.

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On the possibility of rapid nondestructive determination of compressive strength of cement materials Nevenka Kamenic´ Civil Engineering Institute of Croatia, Zagreb, Croatia

Tomislav Matusinovic´ & Juraj Šipušic´ Faculty of Chemical Engineering and Technology, Zagreb, Croatia

ABSTRACT: In this work we were able to correlate the compressive strength of calcium aluminate cement mortar specimens to the features of the through-transmitted longitudinal ultrasonic pulse. The procedure is based on the description of the first part of through-transmitted ultrasonic pulse as a single Gaussian wave packet. A group of three parameters of Gaussian wave packet (Aω/tC), namely envelope amplitude, A [V], angular frequency, ω [rad s−1], and wave packet center transit time, tC [µs] are proportional to the compressive strengths measured. Based on the simple model of one-dimensional longitudinal ultrasonic wave propagation, we were able to show the physical meaning of the Aω/tC group of parameters as a maximum compressive stress of the material accompanying longitudinal ultrasonic wave propagation. We propose that these findings could be valid for other cement materials and also contribute to the development of the methods for rapid characterization of building materials. Keywords:

∆l

INTRODUCTION

Hydration of cement materials, because of its practical and economic importance, as well as its complexity, triggers a lots of research. The trend is toward better understanding of both basic hydration mechanisms and structure evolution—property relations. The nondestructive testing methods, and especially ultrasonic NDT methods serve in this respect. The advance of computers and instrumentation in past years enabled in-depth investigation of cement materials by various ultrasonic techniques. Consequently, the focus of investigation, at least in laboratory conditions, moved toward better understanding of the whole ultrasonic signal received, or extraction of some useful parameters from it. Those parameters, e.g. amplitude, duration, frequency shift, wave energy etc. aid in better detection of early damage of the material, determination of porosity, water to cement ratio or improved strength correlations. This work will outline a simple procedure for extracting several meaningful parameters from the transmitted ultrasonic signal. The combination of parameters is proportional to the maximum stress accompanying longitudinal ultrasonic wave propagation through the material and to the measured compressive strengths. 2

THEORETICAL

The simple model of one dimensional longitudinal wave propagation (Alfirevic´ 2003) is shown in Fig. 1.

l = v∆t

FORCE

1

cement, compressive strength, ultrasound, Gaussian wave packet

x

Figure 1. Simple model of one dimensional longitudinal wave propagation in homogenous material. Force acting on the rod displaces its left end for an increment ∆l, and causes the longitudinal wave to propagate along the x-axis.

We generate the disturbance on the left side of the rod, and study it after a small time increment, ∆t. It can be observed that when the left side of the rod is pushed for an increment ∆l, the generated disturbance propagates a certain distance during the time ∆t, which is equal to: l = v∆t. Thus, after time ∆t, only the left portion of the rod of length l is subjected to compressive stress. Applying Hookes law and recognizing the relative strain as ε = ∆l/l, where ∆l << l, one obtains: ⎛ ∆l ⎞ σ = εE = ⎜ E ⎝ v∆t ⎟⎠

(1)

Taking into account that elasticity module is proportional to ρv2, and that the quotient (∆l/∆t) i.e. the velocity of the movement of the left side of the rod is proportional to velocity of the movement of the

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Aω tC

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σ∝

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RESULTS AND DISCUSSION

Calcium aluminate cements belong to the class of special cements and are known for their corrosion resistance and rapid hardening characteristics. With the addition of lithium-carbonate it is possible to prepare also rapid setting and hydrated cement materials. The addition of setting accelerator into the sample SL significantly shortens the induction period duration when compared to the reference sample S. Rapid hydration of CAC samples S and SL had been studied also by ultrasonic NDT measurements. The final result of this work is comparison of the evolution of mechanical properties of the samples S and SL obtained by standard compressive strength measurement and by nondestructive in-situ ultrasonic measurement (Figs. 2 and 3). From the experimental results presented we suggest that it could be possible to use ultrasonic NDT method presented as an additional aid for rapid characterization of cement material. The combination of

50

Aω/tC

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05:00

06:00

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compressive strength [MPa]

4

0 04:00

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Figure 2. Comparison of time evolution of compressive strength (•) (sample S) and ultrasonic parameters Aω/tC ( ).

0,0 02:00

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compressive strength [MPa]

particles of the rod, i.e. Aω, and that the distance between transducers is also constant the simplified expression is given by Eq. (2).

time [ h ]

Figure 3. Comparison of time evolution of compressive strength (•) (sample SL) and ultrasonic parameters Aω/tC ( ).

parameters extracted from the ultrasonic signal has some physical background, but it is the experimentally determined compressive strength that suggest possible correlation to the ultrasonic parameters measured.

4

CONCLUSION

Rapid strength gain of calcium aluminate cement materials prepared has been studied by compressive strength and nondestructive ultrasonic throughtransmission measurement. The focus was on the extraction of some useful parameters from the ultrasonic signals that could help in the estimation of compressive strength of the material. The ultrasonic signal received had been successfully fitted by two Gaussian wave packets, but a far simpler approach can be adopted when fitting only the first portion of the signal received by a single Gaussian wave packet. This part of the signal is more likely to be free of side echoes and thus has a greater potential for practical use. The combination of parameters extracted from the ultrasonic signal, Aω/tC, is proportional to the maximum stress in the material during longitudinal wave propagation, and it is well correlated to the measured compressive strength of cement materials. We suggest that these findings could help, along with other methods, in rapid evaluation of cement materials in practice.

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Locating reinforcing bars in concrete with Electrical Resistance Tomography K. Karhunen, A. Seppänen, A. Lehikoinen & J.P. Kaipio Department of Physics, University of Kuopio, Kuopio, Finland

P.J.M. Monteiro Department of Civil and Environment Engineering, University of California Berkeley, Berkeley, California, USA

ABSTRACT: In this paper Electrical Resistance Tomography (ERT) is applied to locating reinforcing bars (rebars) in concrete. Location of rebars is an important issue in validation of concrete structures, especially in estimating the corrosion risk. ERT is a nondestructive imaging modality in which the internal conductivity distribution is reconstructed using surface voltage measurements. Due to ill-posed nature of the ERT reconstruction problem and the high conductivity anisotropies over the rebar-concrete interface, sophisticated mathematical methods are needed. In this paper we introduce a new computational method for locating the reinforcement bars using ERT. The approach is tested both numerically and with real measurements. The results show that ERT can yield valuable information about the location of rebars.

attention has been directed towards electrical tomography methods. Electrical Resistance Tomography (ERT) is a nondestructive imaging modality in which array of electrodes is attached on the surface of the target, and alternative currents are injected through the electrodes. The same electrodes are used to measure potential differences on the surface during the current injections. Based on these voltage measurements

6

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4 y−coordinate (cm)

y−coordinate (cm)

Non-destructive testing (NDT) of concrete structures is becoming increasingly important due to aging and deterioration of infrastructures.Wide variety of NDT techniques have been developed for assessing the condition of reinforced structures. In locating the reinforcing bars, methods such as electromagnetic covermeters, magnetometer-based systems, radar imaging, radiography imaging and computerized tomography are currently in use. However, not much

2 0 −2 −4 −6

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5

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Figure 1. Estimated locations of reinforcing bars from simulated data: A rebar 16 mm in diameter near the boundary (left), 16 mm rebar closer to the center (right). The true boundaries of the rebars are marked with thick lines, and the final estimates with thin lines. The estimates were computed using two different initial locations; the evolutions of the estimated rebar center locations corresponding to different initial guesses are drawn with (solid and dashed) lines. The dots represent the estimated rebar center locations corresponding to iteration steps.

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z (cm)

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Figure 2. Left) Finite element mesh corresponding to the initial guess for the rebar location. Right) The mesh after the iterative method is converged.

Figure 3.

Measurement configuration.

various electrical parameters of the target can be estimated. The most typical parameter is the conductivity distribution. In this paper we treat the rebars as (passive) internal electrodes. This approach allows us to model accurately the high conductivity of steel and the contact impedance between the reinforcing steels and the concrete. The locations of the internal electrodes (i.e. rebars) are considered as unknown parameters, and they are estimated on the basis of ERT measurements. The feasibility of the proposed method for estimating rebar location was tested with numerical simulations and one real experiment. In the reconstruction the diameter and orientation of the rebar were assumed to be known. Therefore, the rebar location can be parameterized using only the (x, y)-coordinates of the rebar long axis. These parameters were estimated simultaneously with the homogenous conductivity distribution and electrode contact impedances. The solutions for the simulational study are presented in Figure 1. The estimates were computed starting from two different initial (x, y)-locations to test the convergence to the right minimum. In Figure 2 the

Figure 4. Estimated locations using two initial starting points superimposed on a photograph of the measurement situation.

computational domain and the varying rebar location is illustrated. The thickness of the modeled concrete cylinder was 3 cm and the diameter 15 cm. The measurement configuration of the experiment is depicted in Figure 3. A steel bar 45 mm in diameter was embedded in the concrete cylinder with same dimensions as in the simulations. The estimated locations are presented in Figure 4. The results indicate that locating rebars with ERT is possible, at least to some extent. The accuracy of the estimates depends on the diameter and the location of the rebars: the bigger and closer to injecting electrodes the rebar is the better the estimates. In all example cases the location of a single rebar with known diameter and orientation was estimated. However, the proposed approach is directly applicable to cases of various rebars with unknown diameters and orientations. Furthermore, the method is not restricted to cases on cylindrical geometries only.

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Electrical resistance tomography imaging of concrete A. Seppänen, K. Karhunen, A. Lehikoinen & J.P. Kaipio Department of Physics, University of Kuopio, Kuopio, Finland

P.J.M. Monteiro Department of Civil and Environmental Engineering, University of California Berkeley, CA, USA

ABSTRACT: In this paper we apply Electrical Resistance Tomography (ERT) to imaging of concrete. ERT is a nondestructive imaging modality based on electrical measurements acquired using an array of electrodes attached to the boundary of the target. In ERT alternating current is injected to the target using the electrodes, and the resulting voltages are measured using the same electrodes. These electrical boundary measurements are used for reconstructing the internal (3D) conductivity distribution of the target. In concrete e.g. metals (reinforcement bars and fibers), cracks, air voids, moisture and chlorides carry contrast with respect to conductivity, and hence ERT might serve as a tool to get information of those properties/structures. Advantages of ERT are that it is a relatively inexpensive, quick and safe modality. We applied ERT to imaging a set of concrete samples. The results suggest that ERT can be used as tool for nondestructive testing of concrete.

1

EXPERIMENTS AND RESULTS

In order to test the feasibility of ERT to imaging of concrete we cast flat cylindrical specimens (diameter 15 cm, height 3 cm), and measured the boundary voltage data by using Stanford Research Systems Lock-in amplifier. Electrical contact to concrete was obtained by using wet Cu-CuSO-electrodes. Table 1 lists the objects cast inside the specimens. The transparencies (specimens 4 and 5) were used for modeling resistive cracks in concrete. The ERT reconstructions are shown in Figures 1–5 together with photographs of the specimens. The resistive bar (specimen 1) and the steel bars (specimens 2 and 3) are relatively accurately tracked, although the real conductivity of steel is much higher than the estimated conductivity in Figures 2 and 3. For a new approach to locating reinforcing bars in concrete, see (Karhunen et al. 2008). The cracks are also accurately located (Figures 4 and 5), and the results show a clear difference between cracks with different depths. Table 1.

Description of the specimens.

Specimen

Encased object

1 2 3

Plastic bar (vertical), diameter 4 cm Steel bar (vertical), diameter 5 cm Steel bar (horizontal), diameter 1 cm, length 7 cm Transparency, width 3 cm Transparency, width 6 cm

4 5

Figure 1. Photograph of the specimen 1 (top) and the reconstruction (bottom).

The experiments were carried out in a rather simple geometry, but the method is straightforwardly applicable to more complex geometries. The potential applications of ERT in nondestructive testing of concrete include locating rebars, detection

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Figure 2. Photograph of the specimen 2 (top) and the reconstruction (bottom).

Figure 3.

Figure 4. Photograph of the specimen 4 (top) and the reconstruction (bottom).

Reconstruction of the specimen 3.

of cracks and air voids, and estimating inhomogeneously distributed physical/chemical concrete properties that carry contrast with respect to conductivity (moisture, carbonation, chlorides, etc.).

REFERENCE Karhunen, K., Seppänen, A., Lehikoinen, A., Kaipio, J.P. & Monteiro, P.J.M. 2008. Locating reinforcement bars in concrete with Electrical Resistivity Tomography, 2nd International Conference on Concrete Repair, Rehabilitation and Retrofitting, Cape Town, South Africa, November 24–26.

Figure 5. Photograph of the specimen 5 (top) and the reconstruction (bottom).

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Materials and structural assessments

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Quality assessment of a 40-years old tunnel E.A.B. Koenders, M. Ottelé, O. Çopurog˘lu & K. van Breugel Microlab, Delft University of Technology, Delft, The Netherlands

ABSTRACT: A submerged traffic tunnel, named “Coentunnel”, as part of the A10 beltway that encircles the city of Amsterdam, has been assessed in terms of its service life, after 40 years of service. The tunnel has been built in 1967 and exists of two tubes that underpass the North Sea Channel that runs from North Sea coast line at IJmuiden towards the Amsterdam harbour, in The Netherlands. The tunnel is made of insitu cast concrete containing Blast furnace slag cement. Due to the ongoing intensifying car traffic around Amsterdam, it was decided to extent the current capacity of the tunnel by building a second tunnel next to the existing one. In terms of service life it has been required that both, the existing and the newly build tunnel, should last for at least a 100 years. This requirement has lead to an extensive durability assessment of the current 40 old tunnel with emphasis on the current performance of the microstructure and on the chloride ingress into the tunnel walls. Cores have been drilled from the tunnel walls and were analyzed in the Microlab of Delft University of Technology. In the paper, results of the quality assessment of the 40 years old tunnel will be provided. The effect of the ongoing hydration of the slag particles on the development of the material properties will be shown as well as the chloride ingress due to de-icing salts into the concrete walls. The paper ends with the overview of the ongoing strength gain that which is the result of the ongoing hydration for 40 years.

1

2

INTRODUCTION

Assessment of the actual durability state of existing infrastructure remains a very difficult and complicated issue. The definition of the durability of, in this case, concrete structures is covering a large number of material related topics that all have to be examined with respect to their actual behaviour. This assessment of the actual state of the material is therefore a function of the performance of the individual aspects of the material. With respect to the 40 years old Coentunnel project in The Netherlands, the question was about assessing the current state of durability of the existing tunnel and to conduct a prediction about the potential behaviour of the tunnel in the coming 60 years (Koenders et al.). These questions were on what the research program could provide a most objective view on the durability performance of the existing tunnel. With respect to this, the material-related aspects of the tunnel walls were selected to be evaluated whereas the structural behaviour was excluded from the scope of this research project. The material-oriented research program has been developed consisting of: ESEM investigations, ASR verifications, carbonation tests, chloride penetration tests, permeability tests and finally, the compressive strength and tensile splitting strength were tested within the scope of this research project. In this paper, a glance of the results of this durability assessment project will be provided.

PROJECT DESCRIPTION

The Coentunnel is a prefabricated submerged tunnel, which has been built as part of the national high way network enrolment, in 1967. The tunnel exists of two tubes that underpass the North Sea Channel that runs from North Sea coast line at IJmuiden to the Amsterdam harbour, in The Netherlands. The tunnel, which is part of the belt way A10 that surrounds the city of Amsterdam (Figure 1) is build up from sections which are prefabricated in a dock near the site. The concrete used for the sections contained Blast furnace slag cement comparable with a today’s CEM III/B (300 kg/m3) with a water cement ratio of 0.5. After finishing the tunnel sections, the dock was filled with water and the tunnel sections were floated into position and submerged towards their final destination. At the north and south side of the tunnel, the sections were connected to the entrance heads of the tunnel which were cast in situ. In terms of service life assessment, it has been required that the existing tunnel with a current age of 40 years should extend its service-life with another 60 years in order to last for a design life of 100 years. 2.1

Chloride penetration

The ingress of chlorides have been measured for the cores that were drilled from the tunnel wall locations. Chloride (bindings), inside the concrete cover zone were measurement in accordance with the method

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Figure 2. From the figure it can be observed that the data points range from almost 4% close to the surface of the wall to almost zero at a cover depth of about 60 mm. The fitted curve of the actual data turns out to cross the critical chloride concentration (0.4%) at a cover thickness of about 40 mm. While using the aging opportunity of the diffusion coefficient a prediction could be made for the profile of the chloride concentration after 100 years of service. For this, a diffusion coefficient of 2.55E-12 m2/s and an aging factor of 0.5 was used. 2.2

Figure 1.

The results achieved from the 40-year old cores drilled from the tunnel walls have been compared with the 28-day compressive and tensile strength achieved from cores drilled from the reference specimen. From these results it can be observed that the average compressive and tensile splitting strength has increased substantially during the 40 years of service. The tensile splitting strength has increased by 74% whereas the compressive strength gained 87% more strength. The increase of strength is most likely be the result of the ongoing hydration induced by the secondary slag reaction.

Detailed view of the Coentunnel location.

Chloride [% cement by weight] 5 40 year (actual) 100 year (prediction)

4

Tensile splitting- and compressive strength

40 year Experiments 3 critical chloride concentration (0.4%)

3

CONCLUSIONS

2

A glance of the results achieved from a 40-year old tunnel are presented in this paper. From the experimental tests performed to evaluate the actual state of durability it can be seen that good insight in the condition of the microstructure could be achieved. In combination with the mechanical tests a reliable prediction of the remaining service life could be achieved.

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Figure 2. Chloride concentration measurements, prediction curves and critical chloride concentration for a tunnel wall.

ACKNOWLEDGEMENTS Volhardt (NEN-EN 5984), which is a titration-based chemical method. Samples are taken from the core sections and dried up to a temperature of 105°C. After that, the remaining of the drying are crushed and treated with nitric acid. The residuals are filtered and the chloride concentrations are determined using titration according to the Volhardt principle. The chloride concentrations are determined for different sectioned parts of the core. Sections with a thickness of 20 mm were used to measure the course of the chloride concentration into the concrete, representing the quality of the concrete cover zone with respect to chloride ingress. The elapse of the chloride concentration into the cover of a concrete wall for a certain location in the Coentunnel is provided in

The authors are very grateful to the Dutch contractors Strukton and Ballast Nedam for providing the data from the Coentunnel project necessary to write this paper.

REFERENCES Koenders, E.A.B. and Çopurog˘lu, O. (2007), Duurzaamheidsonderzoek Coentunnel, Stevin report nr. 25.5-06-14, Delft University of Technology, Delft The Netherlands. Ottele, M. (2007), Betrouwbaarheid van duurzaamheidsvoorspellingen van bestaande betonconstructies (in Dutch), MSc-report, Delft University of Technogloy, Delft, The Netherlands.

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Deficiencies in concrete structures “The Sherlock Holmes” factor Albert N. van Grieken Facades and Remedial Technology, Connell Wagner, Melbourne, Australia

ABSTRACT: Reinforced concrete structures can deteriorate due to exposure conditions, develop defects due to overload or become damaged due to unusual loads or forces. Incidental or programmed observations and inspections usually reveal cause and extent. Yet in some cases investigators remain baffled. This paper describes a number of case studies where the cause of the deficiency was an unexpected exposure or load condition, the “Sherlock Holmes Factor”. Investigation guidelines are offered.

1 THE SHERLOCK HOLMES FACTOR 1.1

Case studies

Investigations fall into two categories namely: In response to incidental observation of a deficiency or part of a planned activity including sale or purchase, refurbishment or ongoing maintenance and compulsory local regulations requirements. Whilst most investigations produce sufficient and logical results to allow correct conclusions and recommendations to be made, some investigation findings and results will baffle the investigator. Following are some case studies.

1.2

Unexpected concrete slab deterioration

A two storey building of historic significance, had been unoccupied for some time and was to be converted to a restaurant. Removal of the suspended ceiling revealed to the surprise of all, extensive areas of sheet type delamination to the soffit of the suspended concrete slab and extensive areas of corroding reinforcement. Considering that the slab was rather thicker than expected for these spans and application and considering the dry interior exposure conditions, the sheet delamination due to corroding reinforcement was unexpected and could not be explained. Accordingly archives were researched and persons who had some knowledge of the building were consulted. This revealed the startling fact that the building had originally been used as horse stables involving water and urine. As the soffit of the slab was heavily carbonated, as expected for internal conditions, the extensive corrosion in the reinforcement was due to this unusual set of circumstances. The slab was repaired (Figure 1) and the restaurant is now fully operational and successful and the previous guests, the horses, a distant memory. 1.3

and was subjected to a range of inspections and diagnostic tests. However the existing structure included insitu reinforced concrete struts which exhibited regular perimeter cracking, which is typically due to tensile forces. Furthermore tensile flexural cracking was observed on the compression faces of the fenders. It was concluded the waves created by the a pleasure craft passed through the gaps in the fenders and underneath the fenders, then bounced off the solid rear wall underneath the broadwalk, and hit the rear face of the fender. Thus the fenders where subjected to the impact of the rear face waves which were more severe than front face waves. This reversed the flexural stresses in the fenders, reversed the stresses in the compression struts to tensile stresses and reversed tension in the ties to compression. 1.4

High rise public housing building in Melbourne

Over time, the facades to these buildings and their external walkways developed extensive concrete deterioration. These are subjected to an extensive range of tests and it is typically concluded that the defects are largely due to corroding reinforcement and corroding connections. However at one building the vertical edges of panels had developed an unusual number and size of defects (Figure 2).

Reinforced concrete broadwalk

A broadwalk structure had developed extensive cracking due to chlorides related corrosion of the reinforcement

Figure 1.

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Level 1 slab soffit repairs in progress.

This could not be readily explained based on the general observation and called for a further study of the test results which are summarised below. This table shows the unusual or higher than typical concentration of chlorides at the panels. Furthermore the 20 mm+ concentrations were higher. After reviewing construction records and studying the sequence of the building construction it became clear that the internal vertical walls were critical path elements. These would have been constructed using concrete set accelerators which contained higher levels of chlorides to meet the construction program. 1.5

Jakarta building column defects

Figure 4.

During the construction of a insitu reinforced concrete frame for a two storey building in the far outskirts of Jakarta, a high incidence of face cracking was Range of chloride ion test results as cl: mg/kg Location

0–20 mm

20 mm+

Panels to window bays Access balcony slabs Exposed ends of load bearing panels

110 to 185 95 to 370 710 to 1065

55 to 90 <20 to 95 1390 to 1680

Column after removal of cracked ‘concrete’.

noted. Due to the inconclusive results local exploratory investigations were undertaken (Figure 3) involving opening up of small pockets at the face of the columns. To the surprise of all involved the ‘concrete’ readily broke away and large particles of clay type material were taken out of the face of the column. Figure 4 shows the result after removal of all ‘weak concrete’ and cleaning with water. The key to this baffling finding was provided by locals who revealed that the traditional means of repairing walls was by the use means of mud. Thus in this case the columns were clearly not well compacted leaving large voids. When the form work was removed these voids were filled with mud and finished off with a cementitious slurry. 2

LESSONS LEARNT

There are no doubt many more stories and cases like these. The following guidelines are offered for those who are investigating deficiencies in concrete structures or involved in deficiencies investigations in general: − Expect the unexpected and remain fully vigilant. − Be prepared to change the scope of the investigation. − ‘Ask does this make sense’ if not sure or if it does not make sense ask locals and research archives. Figure 2. Extent of removed defective concrete at ends of load bearing panels.

Figure 3.

Extraction of cores.

3

CONCLUSIONS

The investigation into deficiencies of concrete structures requires experienced investigators. The scope of the investigation in terms of percentage of areas examined and number type and location of tests is of key importance. In the majority of the cases credible conclusions are reached and recommendations can be made with options for consideration by clients. However in a small number of cases additional data is required. This may come from of additional tests, exploratory investigations or from research data and in some cases simply through chance. Such information enables one to conclude the investigation with the satisfying thought ‘elementary my dear Watson’.

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Contribution to diagnosis of alkali-silica reaction in a bridge structure S.O. Ekolu School of Civil & Environmental Engineering, University of the Witwatersrand, Johannesburg, South Africa

ABSTRACT: This paper describes an examination of concrete cores drilled from an old motorway bridge that had manifested deterioration and was in need of repair. Site observations and the nature of cracking indicated that alkali-silica reaction was suspect. Cores taken from the damaged structure were analyzed by using optical petrography for potential presence of reactive phases in aggregates, and chemical analyses to estimate the residual alkali content in the concrete. Results from the test methods were consistent with existence of alkalisilica reaction damage mechanism in the concrete. The level of severity of alkali-silica reactivity appeared to be moderate. The diagnostic features determined from tests indicate with some certainty, that alkali-silica reaction might have partly or solely contributed to distress in the bridge structure.

1

INTRODUCTION

Several techniques and methods are now available for diagnosing the presence of alkali-aggregate reaction in concrete, developed through concerted research over the years since Stanton, 1940 first identified the mechanism 70 years ago. Interest was also driven by increasing field experiences and reports in several countries around the world. In certain regions, the mechanism cannot be avoided as the only locally available aggregates are reactive. As concrete assumed dominance in construction and infrastructure systems, the industry by default undertakes national economic and social contributions that rely on performance and durability of concrete as a material. Alkali-aggregate reaction has gained prominence against this backdrop. In South Africa, the mechanism was first reported in the 1970’s in the Cape Penisula (now Western Cape). Presently, the damage mechanism is known to exist in areas of South Western Cape, Eastern Cape, and Gauteng. The work presented in this paper was conducted to determine suspected presence of ASR in a motorway bridge in Gauteng located along the Johannesburg to Pretoria N1 national highway and lies in a belt where quartzites are the predominant aggregate type, perhaps of the Pretoria Group or Witwatersrand Supergroup.

2

SAMPLES AND TECHNIQUES

Cores of 90 mm and 200 to 300 mm length extracted from the bridge were used for examination of concrete using laboratory techniques. Petrographic and chemical methods were employed in the analysis to detect the presence of ASR. The rationale was to utilize petrography to detect the presence of ASR in two ways:- (1) examining the aggregates used in concrete

for potential reactivity, (2) possible presence of ASR gel product and reaction rims. The chemical methods on the other hand would provide complementary information from concrete mix design to be used to evaluate availability of sufficient alkali contents in the mix necessary to produce disruptive ASR in the concrete. Ultrasonic pulse velocity (UPV) test was also done to assess cracking and voidage, its severity and general soundness of the concrete examined.

3 3.1

OBSERVATIONS AND ANALYSIS Visual inspection, UPV and stereomicroscopic examination

Examination of the core surfaces with low magnification stereomicroscope revealed presence of a transluscent gel at the interface of some coarse aggregates and the surrounding mortar. When observed by unaided eye, the feature exhibited sight of permanent dampness along the stone-mortar interface. Close examination at 10X magnification clearly showed it to be an extrusion of a colourless gel product lining the interface. 3.2

Optical microscopy

Thin sections prepared from cores were examined by optical microscopy under plane-polarized light (PPL) and cross-polarized light (XPL) to confirm the potential presence of alkali-silica reaction. Coarse aggregates were mostly quartz and the fine aggregates predominantly consisted of minerals of quartz, plagioclase feldspar and microline. Optical examination indicated a presence of reactive and potentially reactive aggregates consisting of optically strained quartz with undulatory extinction, microfractured and microcrystalline quartz grains. These mineral characteristics are indicative of the presence of alkali-silica reaction

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Table 1. Alkali content of hardened cement paste.

under appropriate conditions. Photomicrographs of these optical features are shown in figures 1 to 2.

Core 1 Core 2

3.3

UPV (m/s)

Na2O (%)

K2O (%)

“Residual” Na2Oe

4125 4275

0.14 0.13

0.37 0.40

0.38 0.39

Alkali content

In this investigation, the ‘residual’ amount of alkalis were determined as shown in Table 1. Results indicate that the current level of ‘residual” alkalis available in the cementitious system, of about 0.4% are in the range of low-alkali content.

4 Figure 1. Strained quartz grains in potentially reactive aggregates, XPL, 100X.

Figure 2. Highly strained quartz grain exhibiting dark and light bands within the same grain, XPL, 50X.

CONCLUSIONS

The foregone laboratory-based analysis for detection of ASR in concrete cores drilled from a deteriorated bridge structure demonstrates the use of two complementary techniques of optical petrography and chemical determinations to sufficiently provide analysis for diagnosis and prognosis that may be of interest in design of repairs. The reactive grains were found to be optically strained, microfractured and microcrystalline, features that are essentially characteristic of ASR prone quartz aggregates. The presence of ASR gel was first observed, albeit inconclusively, by stereomicroscopy on cored concrete surfaces, and also seen in thin sections infilling fractures within internally map-cracked reactive aggregates. Available alkalis in the cementitious system were found to have been depleted to a present amount of about 0.40%, a low-level not normally likely to promote further extensive ASR damage except at high cement contents.

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Condition assessment and repair alternatives for Lucko bridge overpasses of Zagreb K. Mavar, S. Skaric Palic & A. Balagija Civil Engineering Institute of Croatia, Zagreb, Croatia

ABSTRACT: Research teams from the Civil Engineering Institute of Croatia performed investigation works on eleven overpasses (length of the biggest being 552.0 m) on junction ‘Lucko’, which constitute core of the Zagreb bypass. All of these overpasses, that have been under traffic use since 1979, are important for the capital of Croatia because they serve both to re-direct city traffic to suburban areas as well as entrance to and exit from the city. All works had to be planned carefully, on a daily basis due to heavy traffic on the eleven overpasses. This paper describes tests performed on the bridges along with results of tests. The results of tests and visual inspection all of the structures were the basis of assessment. This paper also presents an evaluation of repair alternatives for protection and repair works, and life cycle costs for these alternatives.

1

2.1

INTRODUCTION

Zagreb, capital of the Republic of Croatia is the main highway intersection in the State. The following highway routes interweave within Zagreb territory: A1 Zagreb to Dubrovnik (650 km), A6 Zagreb to Rijeka (150 km), A3 Bregana to Zagreb to Lipovac (280 km), A2 Zagreb to Macelj (60 km), A4 Zagreb to Gorican (100 km) and A11 Zagreb to Sisak (45 km) currently under construction. Since Lucko junction is a major structure and most important connection of traffic routes all directions and also for the city of Zagreb with over 60000 vehicles per day in summer tourist season, all activities on each facility require detailed consideration. Inspection and repair of each junction structure first of all demand that issues of coordination of activities and traffic circulation during these construction works be resolved. The Bypass was built in 1979.

2

Visual inspection

Based on visual inspection, preliminary categorization of defects was performed. Also, characteristic locations were selected for verification of visually assessed impairments and defects with detailed testing of mechanical, physical and chemical properties of materials. 2.2

In situ testing and sampling

The following in situ testing was preformed:- Drilling of cores for laboratory testing of 100 mm and 50 mm diameter according to HRN EN 12504-1; Determination of carbonation depth by in-situ method, HRN CR 12793; Taking of dust for determination of chloride concentration and depth of chloride penetration

INVESTIGATION WORKS

Investigation works were conducted on all 11 overpasses so that the condition of materials could be determined and design project of rehabilitation works (DPRW) made accordingly. Investigation works consisted of visual inspection of all structures with recording of defects and registering of cracks, establishing locations for taking of specimens and in-situ testing (Fig. 1), taking of concrete specimens and opening of asphalt surfacing for the purpose of condition assessment of materials and laboratory testing on samples made out of cores drilled from the structures.

Figure 1. Head cantilever, leaking from the pavement structure and vehicle with a basket which was used for inspection.

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Estimation of LCC for one Overpass for three repair alternatives (RA) EUR 1.000.000 900.000 800.000 700.000

RA 1/10 y RA 2/25 y RA 3/50 y

600.000 500.000 400.000 300.000 200.000

Figure 2. Equipment for testing of concrete strength and gas permeability.

100.000 0 0

into the concrete (for making of chloride profile); Determination of degree of reinforcement corrosion by measuring of potential and electrical resistivity of concrete, ASTM C 876. Cores were taken to the laboratory, examined in detail and described. Samples were made from these cores and tested. 2.3

Laboratory testing

A detailed scheme for taking of samples from drilled cores (Fig. 8) was made for:- Determination of compressive strength and density of concrete (Fig. 2), HRN EN 12504-1; Determination of chloride content through the depth of concrete, prEN 14629; Determination of specific capillary action coefficient, HRN U.M8.300; Determination of gas permeability coefficient (Fig. 2), EN 993-4. 2.4

Preliminary and final categorization

Final categorization of impairments and defects was made based on approximate visual categorization, detailed in situ and laboratory testing, and for the purpose of making DPRW.

3

REPAIR ALTERNATIVES AND EXECUTION REHABILITATION DESIGNES

All principles and methods for reinforced concrete repair and protection were conducted according to standard EN 1504-Part 9; chapter 6 Basis for the choice of products and systems. The next step after investigation and condition assessments are finished is the assessment of repair alternatives for repair and protection works for each overpass. Various repair alternatives are based on service life estimation after the repair works are performed. Usually aimed life cycles are 5 to 10 years, 25 years and 50 years service life.

Figure 3.

5

10

15

20

25 years

30

35

40

45

50

Durability of RA and flow of cost (LCC).

Using this information, the authors indexed the cost of works for certain groups of structural elements, with increasing factors for quantity of works, due to progression of degradation processes. In this way we established possible flows of expenses for different scenarios of repair protection works, depending on extent of the first rehabilitation process (Fig. 3).

4

CONCLUSION

After 30 years of use and multiple increase in traffic loading of Junction ‘Lucko’, the need for rehabilitation of all structures (overpasses) arose. For the purpose of evaluation and assessment of all structures investigation works were performed on certain structure elements, materials and equipment. Total number of taken cores is 171, out of which 216 samples for compressive strength and density of concrete, 331 chloride tests, 18 samples for water absorption and 17 for gas permeability. Based on performed investigation works, repair alternatives were made with principles and strategies for repair protection which assumed undisturbed life cycle for regarded structures in time periods of 5 to 10 years, 25 years and 50 years. Estimated costs for one of the 11 overpasses were: for 5–10 years 160.000 Euro, for 25 years 490.000 Euro and for 50 years 720.000 Euro. Although the repair alternative for 50 years is most expensive, after estimation of life cycle costing it is shown that it is most costeffective in long term. Based on the owner’s decision about the strategy for rehabilitation of the structures and considering availability of financial resources, the DPRW will be further developed along with a traffic study for the period when rehabilitation works will be performed.

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Service life extension of existing precast concrete girders D.I. Banic, D. Tkalcic & Z. Banic Civil Engineering Institute of Croatia, Zagreb, Croatia

ABSTRACT: Segmental bridges have been the most commonly built type of road structure in the last 30 years, as a part of modern roads and motorways in Croatia. Fast construction of road network was possible only with segmental concrete bridges. In most cases bridges were built as simple supported systems. During regular and preventive bridge inspections that were carried out on a great part of motorways, dangerous defects on overpasses were recorded. These defects were mostly found on one type of bridge structure. The paper shows a detailed description of this bridge structure, problems in inspection and maintenance, the results of damage inspections, characteristic defects as well as repair methods and proposals for future maintenance. These bridges were tested, using both static and dynamic full scale load tests, before and after rehabilitation. The paper also presents test results and conclusions. As a result, some improvements that would upgrade structural behaviour and load bearing capacity were suggested.

1

INTRODUCTION

In the period between 2000 and 2007, the Civil Engineering Institute of Croatia carried out detailed visual inspections, field investigation works and laboratory testing of specimens and based on that made a condition evaluation of over 200 bridges on Croatian motorways. Conclusions reached in this study will help state authorities to make a modern system of bridge management on Croatian roads. The analysis of the results of investigation works showed that bridges with the same or similar structures suffered the same type of damage, but in various degree of progress. Since some structures no longer satisfied the required level of load bearing capacity and reliability, additional investigation works were performed on them, as well as control static and dynamic calculations. This referred especially to the overpasses across the motorway, which sustained damage due to the traffic of heavy vehicles under and over the structure. The majority of these overpasses was built with precast hollow core prestressed girders of SAN 75 type. 2

OVERPASSES WITH SAN TYPE GIRDERS

Overpasses built with SAN girders have proved as very suitable for accelerated construction. The girders are manufactured in a factory, as precast elements and delivered to the site, where they are placed simply and quickly, by means of truck cranes, due to their small weight of approx 40 tons. Extensive investigation works that were performed on all structures included a visual inspection, field and laboratory testing. Field methods included pull off, mapping of electrochemical potential, registering the position and width of cracks,

reinforcement layout, and determination of concrete strength with rebound hammer. Laboratory testing was carried out on drilled concrete specimens. 3

STRUCTURAL CONDITION

The results of investigation works showed that the repair of individual parts of the structure is possible, including the removal of damaged concrete and local repair of concrete surfaces. The first repair method would consist of hydrodemolition of damaged concrete surfaces, placement of additional reinforcing bars where necessary and the repair of damaged concrete surfaces. Since girders were prestressed with adhesion, they should be repaired segment by segment, using high quality mortars with high early strength and compensated shrinkage. It would be necessary to wait 5 to 7 days in order to move to the next phase, in order to prevent loss of prestress. At this rate, the duration of works on one span would be 45 days, which means that the repair of one structure would last 180 days in ideal conditions. Extension of works would reduce the traffic flow and increase the possibility of accidents. The cost of this rehabilitation would amount to 1.7 million euros. The second method would be a complete replacement of the structure, which means that it would be necessary to stop the traffic on the motorway completely, during disassembly of certain parts, especially during the construction of the new structure. The cost would be 800,000 euros. The third method would be the replacement of the entire overpass superstructure and the repair of head beams, piers and abutments. The evaluated cost of this repair method is around 300,000 euros. After consulting the owner about the repair costs, temporary traffic regime during the works, the duration of

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works and the designed service life of the structure, the third repair method was chosen. 4 4.1

STRUCTURAL REPAIR Substructure

Parts of the substructure were made with reinforced concrete, so it was necessary to perform standard repairs of individual elements, in order to remove damaged or carbonated concrete, repair or replace reinforcement bars, depending on the local damage. These works involved the removal of 1.5 to 2.0 cm thick layer of moist and chloride saturated surface concrete, in order to allow good adherence between the old and new concrete. 4.2

Superstructure

The works on the superstructure included the replacement of main girders, removal and replacement of sidewalks, expansion joints, guard rail, kerbs and the deck structure. In order to prevent greater differential displacements, main girders were connected with a deck slab. It had changeable thickness from 10 to 20 cm, so as to attain sufficient fall for rainwater drainage. Total duration of all repair works was 60 days, as it was anticipated. Overall repair costs exceeded the estimated costs in the study by somewhat more than 10%. 5

STRUCTURAL LOAD TESTING

In order to confirm the efficacy of scheduled and performed works on rehabilitation of structures, load testing of rehabilitated overpass was performed. Test results were compared to the behaviour of the structure before rehabilitation and with results of testing carried out on similar structures. Calculated values coincide better with the measured values, so it can be concluded that connecting main girders by placing a reinforced-concrete slab improves not only the performance of the structure under asymmetrical load but overall stiffness as well. This follows from the fact that under the maximum load, the maximum deflection of the structure before rehabilitation amounted to the average 7.5 mm, and after rehabilitation to the maximum of 4.3. mm for the same test load. It can be concluded that overall stiffness for vertical loading increased by more than 50%.

5.1

Dynamic testing

Based on detailed analysis, it was concluded that dynamic properties significantly improved in rehabilitated structures. The frequency of free oscillations was increased by 50% i.e. the period of natural oscillation of the first tone was reduced in comparison with the values before repair works. The maximum acceleration also decreased from 0.2 g to 0.11 g, and linear to that vertical displacements due to dynamic loading were reduced as well. Critical damping coefficient was increased by 60%, i.e. before the repair it amounted to 2.3, and later to 3.3.

6

CONCLUSION

Due to errors repeated in construction, characteristic damage patterns with expected consequences occur on Croatian motorways. Under specific conditions, seeping of salty rainwater leads to severe and serious damage, which in turn significantly affects the safety and load bearing capacity of the structure. Design, repair and construction require the understanding of the problem, of the physical and chemical processes in concrete, steel reinforcement and repair materials. xtensive investigation works are necessary, especially if they are not performed frequently, in order to precisely determine the actual condition of the structure in terms of constituent materials, concrete, steel etc. Test loading of the structure is a very useful method for assessing the actual behaviour of the structure. It is also useful to perform it on a new structure before opening to traffic, in order to establish the reference condition of the new structure for condition assessment in the future. The owner can use data obtained by this testing in determining appropriate measures to ensure the extended service life of the structure.

REFERENCES Mavar, K., Barisic E. Petrak I International Conference on Bridges, May 21–24, Dubrovnik, Croatia. 2006. Examples of use principles and methods according to HRN ENV 1504-9 in concrete bridge repairs. CEB-FIP Bulletin, 2003. Monitoring and Safety Evaluation of Existing Concrete Structures—State of the Art Report, Task Group. Tkalcic, D., 2004. “Concrete Bridge Condition Assessment”, Master’s thesis.

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Approaches for the assessment of the residual strength of concrete exposed to fire E. Annerel & L. Taerwe Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Ghent, Belgium

ABSTRACT: Generally, concrete structures have a high fire resistance. After fire, it is of economical interest to reuse the structure after appropriate repair based on a reliable assessment of the residual strength. This paper deals with some fundamental aspects of a scientific and systematic methodology to assess the damage and to estimate the residual concrete strength on the basis of the change in colour and the crack development. This method seems to be promising, but the number of cracks and the change in colour are influenced by the test set up. Furthermore, these relationships change when the concrete ages after heating. Other methods such as water immersion, the Rebound Index and microscopy also provide an adequate basis for the assessment of the temperature in heated concrete.

1

INTRODUCTION

The effect of fire on the strength of traditional concrete has been studied for a long time and typical models are described in codes. However, the behaviour of self-compacting concrete under fire load has not yet been widely investigated. One of the main topics in fire research today is the assessment of the residual strength after fire exposure. Concrete heated to high temperatures develops cracks and changes in colour from red (300–600°C) to whitish grey (600–900°C) and buff (900–1000°C). Both alterations can be linked to the residual strength, since they are all temperature related. Assessing the temperature based on the alterations should therefore provide enough information for estimating the residual strength of the heated structure. However, the residual strength is influenced by the test conditions: the heating rate of the furnace, the time spent at target temperature and the cooling method seem to play an important role. Furthermore, a certain degree of strength recovery of the heated concrete depending on the type of storage after fire is noticeable. All these parameters are important for a practical analysis, because they determine the value of the actual residual strength, which is necessary for structural calculations. This paper also discusses 3 other methods to assess the temperature history of the concrete: the surface hardness determined with a Schmidt Rebound Hammer, weight increase due to water immersion and microscopic research. 2

RESIDUAL STRENGTH

Cubes (150 mm) from 3 types of concrete, namely a self-compacting concrete (SCC), a traditional

vibrated concrete with siliceous aggregates (TC) and with calcareous aggregates (TCk), were tested for compressive strength immediately after heating to different target temperatures. The heating occurred at a rate of 3.5°C/min, the target temperature was held for 750 minutes and the cubes were cooled in ambient air. To avoid explosive spalling, the selfcompacting concrete was dried at 105°C. The curves of the strength decrease are situated around the Eurocode curve for normal siliceous concrete (EN 1992-1-2). To determine the influence of the test conditions on the residual strength, TC and SCC 150 mm cubes were heated up to 350°C and 550°C. The standard test conditions as mentioned above are chosen as reference. One of these conditions is altered, while the others remain the same. The heating rate is changed from 3.5 to 10°C/min for TC and 20°C/min for SCC, whereas the duration at the target temperature is increased from 750 to 3600 minutes and the cooling regime is modified from a slow cooling into a rapid cooling by immersion under water. The cubes are not previously dried. The results show that the cooling method is the most important factor, since water cooling introduces an extra drop of the residual strength of 30–35%. The effect of storage conditions after heating is tested with SCC and TC 150 mm cubes, which were heated to 350°C and 550°C, and stored after heating in air and under water for 7, 28 and 56 days. The strength seems to drop till a minimum around 7–14 days with an additional strength loss of 20–30%. From there the strength slowly recovers, which is faster for cubes stored under water. However, the strength after 56 days of storage is still lower than the strength immediately after cooling down to ambient temperatures.

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3

3.1

ASSESSMENT OF THE TEMPERATURE PROFILE Colorimetry

The temperature can be assessed based on the colour change induced by heating. Cores were drilled out of 150 mm SCC and TC cubes, which were sawn and polished. These discs were heated to different target temperatures (up to 1160°C), after which the colour was measured with a X-rite SP60 spectrophotometer according to the CIE Lab-colour space. The coarse aggregates were masked with black ink to minimize the effect of the colourful aggregates. During heating the colour describes an elliptical path in the a*b*colour space. Differences can be seen between the elliptical path of the SCC and TC samples, which can be attributed to differences in the number of masked aggregates and cement stone. The surface colour of 150 mm TC and TCk cubes, heated to different target temperatures, was measured during a test period of 90 days. Again, the elliptical path is clearly visible for the TC concrete surface, while it alters slightly for TCk. Notice that these concrete surfaces don’t have colourful aggregates and that they are not polished. When concrete ages a shift of the colours can be seen towards the inner part of the elliptical path, which can be linked to the moisture absorption. 3.2

Crack development

For the macroscopic determination of the porosity, the surfaces were completely blackened and the pores and cracks were filled with white BaSO4 powder. The sample was scanned by a flatbed scanner, from which the total porosity and the size distribution was analysed with the standard image processing programme ImageTool. A curve fitted to the measured points shows a constant value until a transition temperature, from where it increases almost linear. TC and SCC have a different transition temperature, respectively 200°C and 400°C, which is probably due to the fewer coarse aggregates in SCC. 3.3

Water immersion

The increase of porosity of the concrete can also be determined with water immersion. Tests show a

linear relationship with the porosity determined on the polished samples. This method is very promising, because the reference weight is the investigated (heated) sample it self, preferably after drying. The search for a non heated sample with the same initial properties as the concrete in the heated zone is thus not necessary. This storage under water of samples from a drilled core could be a quick and cheap alternative to the method of polished samples. The weight increase due to storage in air can also be used, but has lower differences in value. This method is investigated on TC discs (Ø 80 mm, 15 mm thick) sawn from cores and on 150 mm cubes. The discs were heated at target temperature for 1 h and 4 h, while they were cooled in the furnace, in air or quenched in water.

4

CONCLUSIONS

– The residual strength should be measured at least 7 days after the end of the fire or the heating period. Water cooling may induce an additional reduction of the compressive strength of 30 to 35%. – When concrete has been heated, the colour change of the matrix follows an elliptical path in the a*b*-colour space and shifts to the inner part as a function of moisture absorption. – A distinction should be made between matrix cracks and interfacial cracks; the total porosity seems to be a good indicator to estimate the maximum temperature which was reached in the heated concrete. – Water immersion shows promising results to reflect the temperature history of the concrete, since the weight of the heated sample itself can be used as reference. – Other methods such as the Rebound Index, PFM and ESEM yield also useful results.

ACKNOWLEDGEMENT The authors would like to thank the Fund for Scientific Research in Flanders (FWO) for the financial support; and are thankful for the help of TNO Delft and TU Delft (Netherlands) for their contribution to the microscopic research.

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Residual strength of R.C. buildings after a fire: A case study I. Venanzi & A.L. Materazzi University of Perugia, Perugia, Italy

M. Zappia Fire Department, Florence, Italy

ABSTRACT: In the present paper a procedure for the assessment of the residual strength of R.C. structural members after a fire is proposed. The procedure has been applied to the evaluation of the load bearing capacity of a storage and apartment building which experienced a real fire. The damage provoked by the fire was evaluated through a survey and an experimental campaign carried out using non destructive techniques. Then, the analytical assessment of the residual strength using a simplified fire model based on the zone method was performed. On the basis of the analysis results, a retrofitting programme was defined, with the aim at restoring completely the serviceability of the building.

1

2

INTRODUCTION

The assessment of the residual load bearing capacity of reinforced concrete structures after their exposition to fire requires the estimate of the actual strength of both concrete and steel and the subsequent computation of the sections’ capacity. In most cases the knowledge of materials strength, even after accurate and extensive testing, is not sufficient for the purpose. In fact in-situ testing cannot be extended over all the sections depth, while every single spot of the sections contributes to the overall bearing capacity with a strength that depends on the local maximum temperature reached during the fire. The present paper is aimed at giving a contribution to solve this problem, proposing an assessment methodology based on a suitable analytical procedure. The proposed procedure was applied to a casestudy consisting in a seven-storey building which was badly damaged by a severe fire that occurred in its basement and ground floor during the night between the 23rd and the 24th of March 1991. The building is a reinforced concrete structure, made of seven floors and the basement. It has an L shape in plan, with a symmetry axis at an angle of 45° from the direction of the external walls. The part of the structure damaged by the fire was used for storage and selling of clothing. The high temperature reached in the structural elements in the compartments where the fire developed provoked transversal cracks, separation and crumbling away. As a consequence of concrete spalling, in some cases the steel reinforcement was exposed directly to the fire and so plastic strains, loss of bond and localized yielding occurred. The most damaged structural elements were those at the basement. Minor damages were observed at the ground floor.

ANALYSIS OF THE RESIDUAL STRENGTH AFTER FIRE

The load bearing capacity of the structural members after the fire was assessed by evaluating the interaction diagrams between the ultimate compressive force Nult and the bending moment Mult of the relevant cross-sections. The Mult-Nult diagrams were computed considering the residual strength of the materials after their cooling down after the fire. The residual strength of steel was assumed not to be affected by the fire. On the contrary the residual compressive strength of concrete depends on the maximum temperature reached at each point within the cross-section during the fire. As the maximum local temperature is hardly available, it was estimated through the analysis of the thermal field, induced by a proper time history of the air temperature, representing the actual fire. The variation of the air temperature with time during a real fire was analytically evaluated using a “zone” model, which operates dividing the compartment in two zones, an upper and a lower one with respect to the surface of separation between the hot and cold gas, that interact between each other through mass and energy exchange. The zone model operates solving, for each region in which the compartment is divided, a system of differential equations that express the conservation of mass and energy, assuming valid the ideal gas law and defining the density and the internal energy. The solution of the differential equations governing the problem was carried out using the CFAST software, contained in the package HAZARD I distributed by the NIST. The compartment was preliminarily divided into seven sub-compartments. After having identified the sub-compartment where the fire developed (no. 2), it was hypothesized a way of propagation towards the

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other sub-compartments, dependent on the time and the ignition temperature of the materials. Then, the thermal field of the cross-section during the fire was evaluated by solving the Fourier equation through a finite element approach. For this purpose the cross-section of each member exposed to fire was discretized in rectangular or triangular finite elements. The temperature was reasonably assumed constant in each element, hypothesizing no variation over the length of the member. The variation with the temperature of the thermal properties of the materials, as the thermal conductivity, the specific heat and the density of the concrete were taken from EN 1992-1-2. After having solved the thermal problem in correspondence of several fire exposition times, the maximum temperature reached by each concrete element during the fire was evaluated. Finally, the Mult-Nult diagrams were obtained assigning over the considered cross-sections the full set of the strain diagrams corresponding to all possible failure conditions and computing the corresponding ultimate internal forces. To evaluate Mult,i and Nult,i it was necessary to integrate the residual strength of the concrete fibers corresponding to the assigned strain distributions. For the variation of the residual compressive strength of concrete as a function of the maximum temperature reached during the exposition to fire by each element in which the member’s cross-section is divided, the graph suggested by the ACI 216.1-97 for unstressed carbonate aggregate concrete was adopted. As the residual compressive strength of concrete depends on the maximum temperature reached at each point within the cross-section during the fire, the temperature computed during the analysis of the thermal field was used. For every strain value, the corresponding stress was evaluated by means of the constitutive law of concrete having the appropriate temperature-dependent strength. For the constitutive laws of concrete and reinforcing steel, the stress-strain relationships proposed by the EN 1992-1-1 were adopted. It was assumed that the maximum strength of steel does not depend on the temperature reached during the fire. The safety of the structural members was ensured if the points representative of the design loads (Md, Nd) are internal to the edge of the interaction diagram. Naturally the loads are factored applying the standard load factors used for the design in cold condition. The outline of the proposed procedure is shown in Figure 1.

3

Figure 1. Schematic representation of the procedure for the assessment of the residual load bearing capacity.

column 43 and beam 34–42 at the basement, that were directly exposed to fire. Both the analyzed members were located in the compartment where the fire developed. From the interaction diagrams it was observed that for the analyzed members there was a significant reduction of the load bearing capacity after fire, due to the decreased residual compressive strength of concrete. The application of the procedure to a real case study has shown that, in some cases, the residual load bearing capacity of the structural members is not sufficient to sustain the design loads and a retrofitting programme has to be defined on the basis of the evaluated bearing capacity after the fire.

RESULTS AND CONCLUSIONS

The proposed procedure was applied for the analysis of the residual strength of two structural members:

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Assessment of fire-damaged concrete structures and the corresponding measures Yngve Anderberg Flight & Safety Design International AB, Lund, Sweden

ABSTRACT: The possibility to reuse a concrete structure after fire is mainly dependant on the residual load-bearing capacity and on acceptable residual deflections and/or deformations. The degree of damage due to fire, possible repair measures and the corresponding repair costs are decisive parameters in the decision to reuse or demolish the structural member. An approach is described in order to evaluate the residual load-bearing capacity due to fire by using a combination of visual observations, thorough inspection, interviews and a fire simulation and analytical predictions. Sometimes there may also be a need to make supplementary residual strength tests of concrete and steel as well as loading tests of the structure. This paper presents a practical and efficient way of analyzing the damage and which repair measures that can be applied to reasonable costs. A striking example of an early collapse, although designed in accordance with the prescriptive fire code, is presented as a learning lesson for future design against fire.

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• Changes in the visual appearance as sooty and colored surfaces etc

INTRODUCTION

The residual load-bearing capacity of concrete structures after fire is both from an owner’s and from an insurance technical point of view of great importance. The owner has an interest in restarting the business or activity as soon as possible and not to suffer too much even if the insurance company must compensate for the immediate loss but not for unpredictable future loss. Also the insurance company wants to keep down the interruption and repair costs and reuse the building as soon as possible. You are able to assess the residual load-bearing capacity after fire and evaluate and make a decision of possible repair measures for reuse of the structure and a corresponding cost analysis.

2 2.1

REUSE OF A STRUCTURE AFTER FIRE Overall procedure

A very important step in the assessment of the damage is to get a clear picture of the intended function of the structural member. The damage of the structure and its consequences for load-bearing capacity must be judged correctly. The typical damages of a concrete structure due to fire are: • Loss of prestress • Residual deformations, cracking, loss of material by spalling etc • Detailing and connection failure especially between steel and reinforced concrete members but also at supports

In the assessment of the extent of damage the following input are important: • Detailed information from the owner or user about the activity in the fire compartment, fire load density, storage of fire hazardous material etc • Visual observation and information from witnesses and fire brigade about the fire • Possible need to make strength tests of reinforcement as prestressing wires as well as for the concrete • Loading test of the structure is performed but it is often expensive and difficult to carry out. • Fire simulation and analytical prediction of temperature distribution and residual load-bearing capacity When all possible information about the damage and the remaining load-bearing capacity is on the table the repair measures are evaluated. Possible measures can be a new concrete cover for the reinforcement or an increased concrete section with additional reinforcement or an extra insulation layer as an improved fire protection or replacing some of the concrete members in the structure etc.

3 3.1

PROCEDURE TO ESTIMATE THE DAMAGE Search for information about the fire

The most important basis for estimating the damage is to investigate the fire process in the building.

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Important visual observation by witnesses and fire inspector can be • sooty concrete surfaces which means that the temperature has been below 500°C • pieces of charred wood where the charred depth may give you information about the duration of the fire (charring depth of about 20 mm is equivalent to 30 min ISO 834 fire) • the degree of combustion of the plastic cover of electrical wires may give you hints about the temperature level • other pieces of combustible materials in the fire compartment of known burning and/or melting point temperature • color changes on the concrete surface may indicate very high temperatures 800–1000°C

3.2

Inspection of damage of the structure due to fire

Detailed drawings of the geometry and location of reinforcement of structural members exposed to fire as well as all detailing of connections and supports and strength properties are needed. Hot-rolled steel has an insignificant loss of residual strength but prestressing wires has lost up to 40% of its original strength if exposed to a temperature of 600°C. Other important inspections are unacceptable deflections or deformations, possible damages on anchorage of prestressing wires and on detailing. Sometimes the thermal elongation of a slab may cause deflections at the top of columns or walls. This creates additional bending moments and shear forces which must be considered.

4

FIRE SIMULATION AND ANALYTICAL PREDICTION

After gathering all information and the result of inspection of damages the fire scenario is evaluated. The temperature distribution of the structural member is preferably predicted by dedicated and userfriendly software for fire design (TCD –Temperature Calculation and Design). Thermal properties of concrete as function of temperature are normally taken from EN 1992-1-2 and serve as necessary input for the calculation. The temperature distribution of the cross-section serves together with the mechanical

properties of steel and concrete as function of temperature as input in the structural analysis. 5

INTEGRATED ASSESSMENT OF THE DAMAGE

All information collected from witnesses, fire brigade, visual observations and inspection by specialists is finally compared with the analytical predictions. By comparing the two sources of results an integrated assessment starts in order to identify the possible lack of load bearing capacity, improvement of connectors, supports, loss of concrete layer due to spalling and material falling off during cooling. If still uncertainties exist tensile strength tests of prestressing wires and compressive tests of drilled concrete specimens in the compressive zone can be a supplement. Loading tests are not very common but can also contribute to a more reliable assessment. Sometimes the judgment of no possible reuse due to too heavy fire exposure, too much damage or too great deformations or deflections is easy to make. 6

REPAIR MEASURES

If the temperature exposure has been above 800°C for more than 60 min the surface part of the concrete 20–30 mm i.e. the concrete cover for reinforcement is damaged and will be knocked down and replaced by new concrete. This is a very common measure for beams and columns. It is not possible to increase the load bearing capacity of a prestressed beam but very often the load utilization degree is much less than 100% and this means a reserve load bearing capacity which is positive for reuse. The load bearing capacity for columns after fire can sometimes be improved by increasing the cross-section with a new layer of concrete and additional reinforcement. 7

CONCLUSIONS

Concrete structures have normally a very good fire resistance. Fire-damaged concrete members have a good potential for reuse by adequate and not to costly repair measures. This paper has highlighted the procedure to assess the damage and the residual load bearing capacity in a substantial way by combining site inspections and investigations combined with analytical predictions.

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Different methods to investigate and to repair the damages of a post-tensioned deck of a high-way bridge after a fire incident due to a traffic accident A.-W. Gutsch & F. Herschelmann Braunschweig Civil Engineering Materials Testing Institute (MPA BS) and Institute for Building Materials, Structural Concrete and Fire Protection, Technical University of Braunschweig, Braunschweig, Germany

ABSTRACT: As result of a traffic accident on a high-way a truck started to burn just under a post-tensioned reinforced concrete bridge. The rapid development of the fire caused significant damages like concrete spallings of the concrete cover at the underside of the bridge deck. The question was whether the degree of damages was so critical that a repair of the bridge deck was possible or not. In this report the systematic way to investigate the bridge and to classify the damages after the fire incident will be presented. New methods were used to estimate the temperature exposure of the pre-stressing steel during the fire. Only with these methods the decision to repair the bridge was possible. Finally, the applied repair concept with different steps, methods and materials will be shown.

1 1.1

1.2

INTRODUCTION Situation

A truck crashed a heavy noise barrier which was installed along a high-way and also under a bridge crossing the high-way. As result of the accident the truck started to burn just under the crossing bridge. Due to the fuel in the tank of the truck and its load the fire lighted very fast and high temperature occurred under the bridge deck before the fire brigade started to extinguish the fire. After the fire incident it could be reconstructed that the fire brigade started to extinguish the fire about 10 Minutes later than they got the alarm call and only 14 Minutes later than the accident happened and the fire lightened. The fire was dammed about 40 Minutes later than it started. Just after the fire was extinguished the bridge deck was inspected visually on damages due to the fire incident. It was the first estimation that the grade of damages of the bridge deck due to the fire was very high up to the danger of collapse of the bridge deck. Due to this first estimation it was not possible to allow any traffic on the high way and on the crossing bridge itself before detailed investigations on the grade of damages of the bridge deck were executed. But these investigations were critical in time as the high-way is the most frequent high-way in West-East-direction in Germany with more than 120,000 vehicles per day. A blockade of the high-way would yield to traffic jams in a greater area around the bridge. Thus, detailed investigations were necessary to estimate the stability of the bridge.

Classification of damages

When the burnt out truck was removed from the fire place under the bridge the deck was inspected visually in detail. A wide range of damages became obvious. The damages of the bridge deck after the fire exposure were classified into different classes to get a certain pattern of the damages: • Class 1: Concrete not damaged, only black soot on the concrete surface, black soot is a indication for lower temperature due to fire (T < 400°C). • Class 2: Spalled off concrete cover, reinforcement still concrete covered. • Class 3: Spalled off concrete cover and uncovered reinforcement. • Class 4: Spalled off concrete cover up to greater depth in the cross-section; no soot on the surface is a indication for very high temperature due to fire (T > 500°C). • Class 5: Cracks in the concrete due to the fire incident. The cracks occurred near the post-tensioning elements. The main questions to estimate the stability of the bridge deck and to decide whether the deck could be structural repaired or had to be reconstructed were: • Which is the residual compressive strength of the concrete? • Which is the residual strength of the reinforcement (ordinary and post-tensioning steel)? If possible, the bridge deck should be repaired and not reconstructed as the limitation of the traffic seemed to be less and the costs were estimated to be lower.

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2.1

INVESTIGATION OF DEGREE OF DAMAGES Concrete, Reinforcement and Pre-Stressing Steel

Concrete specimen were taken as drilling cores (d = 50 mm) from locations belonging to different damage classes. The flamed surface of the drilling cores were taken into account for the compressive strength. The mean value of the compressive strength was 71.5 MPa. Concrete seemed not to be affected by the fire incident. Reinforcement samples were also taken from different damage classes. The yield strength and the ultimate tensile strength was not affected by the fire either. The steel quality was BSt 420/500 which is not critical in case of hot fire exposure and re-cooling. 2.2

3.2

Investigation of temperature exposure of the pre-stressing steel

Strand were used for the post-tensioning elements. In general drawn wires and strands have a critical mechanical behavior in case of hot fire exposure. For strands a temperature T > 300°C is critical for their mechanical properties. As it was not possible to take samples from the strands to determine the residual tensile strength after the fire attack other ways to estimate the temperature of the post-tensioning elements during the fire had to be found. The temperature of the pre-stressing steel was estimated by a temperature field calculation on the one hand and by the mineralogical investigation of the concrete next to the strand on the other hand. Both methods yield independent from each other to a maximum temperature of the pre-stressing steel of T < 180 °C during the fire which was not critical for the mechanical properties of the strands. These results were the key-information for the decision to be able to repair the bridge. 3

3.1

The loose concrete layers could not be removed by sand blasting alone. Other methods like chiselling or high-pressure water jet method were necessary. The high-pressure water jet (HPWJ) seemed to be preferable as concrete microstructure, reinforcement and prestressing steel could not be damaged by this method. Thus, the loose concrete layers were removed with high-pressure water jet (2,000 bar) over about 400 m2 concrete surface area. The concrete surface treated by HPWJ was rough enough for a direct application of mineral re-profiling materials. For safety and quality reasons the bond strength was tested before applying re-profiling material. The bond strength of areas treated with HPWJ were in the range of 1.29–2.05 MPa, without HPWJ between 0.14–1.82 MPa. This proves the benefit of HPWJ to prepare the concrete surface.

REPAIR CONCEPT FOR THE STRUCTURAL REPAIR

Shotcrete and Concrete Repair System

Dependent on the degree of damage and the necessary thickness of the concrete re-profiling two different methods for the repair and two different materials were chosen: • Damage class 1 and 2: Only thin layers for reprofiling (thickness t < 3 cm) necessary: Concrete repair system from polymer modified cement mortar. • Damage classes 3 to 5: Thick layer (thickness t > 3 cm) for re-profiling necessary: Shotcrete. Dry shotcrete was chosen for the repair work as relative small shotcrete charges were necessary. The concrete repair system was applied according to the references of the producer and according to the national guidelines. The shotcrete was applied on the rough and prepared concrete surface in two layers minimum. On plain concrete surfaces a concrete thickness of 4 cm minimum was chosen to receive a proper bond strength to the bulk concrete surface. The fresh shotcrete was not troweled but still left rough as generated by the spraying. For curing the shotcrete was covered with a foil.

Preparing the concrete surface

At first the soot was removed from the concrete surface. Then the damaged concrete surfaces were prepared by sand blasting to make them ready for the re-profiling. But at this point further areas with loose concrete layers without any bond to the bulk concrete structure became obvious. These bondless concrete layers resulted from the thermal stresses due to different temperature across the cross-section during the rapid fire attack.

3.3

Quality Control

For the quality control separate shotcrete samples were sprayed. Specimen were cut out of sprayed specimen to determine the compressive strength. Additionally the bond strength between the shotcrete and the bulk concrete of the bridge deck was determined at several locations. The bond strength was in all cases greater than 1.3 MPa.

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Structural behaviour of beams under simultaneous load and steel corrosion G. Malumbela, P. Moyo & M. Alexander University of Cape Town, South Africa

ABSTRACT: This paper presents the results of an experimental study conducted to characterize the structural behaviour of beams corroded whilst subjected to a constant sustained load. Corrosion on tensile steel bars was induced by an accelerated corrosion process using a 5% solution of NaCl and a constant impressed current. Four RC beams were tested, each with a width of 153 mm, a depth of 254 mm and a length of 3000 mm. Beams were tested whilst under self-weight, under 10% of the ultimate load and under 33% of the ultimate load. Longitudinal tensile strains and longitudinal compressive strains were monitored during the corrosion process. Measured strains were used to determine the depth of the neutral axis, the curvature and the moment of inertia of beams. The results indicate that the longitudinal strains, depth of the neutral axis and curvature depend on both the level of corrosion and the applied load whilst the moment of inertia only depends on the level of corrosion.

1

INTRODUCTION

2

Corrosion of steel bars embedded in concrete is a worldwide problem that affects numerous reinforced concrete structures Roberge (1999). A great deal of research has been done on the effects of corrosion of tensile steel bars on the structural performance of RC members. Majority of the work done has focussed on corrosion damage under no sustained load, Ballim et al (2003). In real structures however corrosion normally takes place whilst the structure is under a sustained load. The limited work that was done on structural behaviour of corroded beams under a constant sustained load mainly investigated the interaction between central deflections of beams, degree of corrosion and the sustained load, Ballim et al (2003), El Maaddawy et al (2005) and Yoon et al (2000). In spite of the limited work, the results by the researchers clearly indicate that deflections of beams increase with an increase in degree of corrosion and the magnitude of the applied load. Whilst deflections of beams can be used as an indicator of structural performance, models of structural behaviour of beams (including deflections) require variations of strains, depth of neutral axis, curvatures and stiffness as input parameters. Hence research is needed to clarify the variation between these input parameters with degree corrosion in the presence of a sustained load. This paper presents an experimental programme and a discussion of results on the interactions between longitudinal strains, depth of neutral axis, curvatures and stiffness of beams with magnitude of sustained load and corrosion of tensile steel bars.

2.1

EXPERIMENTAL PROGRAMME Test programme

Four beams were used in the test programme. Beam 1 was tested under self-weight; Beam 2 was tested under a constant sustained load that was equivalent to 10% of the ultimate load capacity of the beam (uncracked condition); and beams 3 and 4 were tested under a constant sustained load that was equivalent to 33% of the ultimate load capacity of the beam (cracked condition). Beams 1 to 3 were corroded under their respective loading systems whilst beam 4 was not corroded. A summary of the test programme is shown in Table 1.

2.2

Sustained loading

Figure 2 shows a schematic of the loading frame used in the research programme to test beams 2, 3 and 4. Support columns of the frame were bolted to a strong floor to provide adequate reaction force. Weights were hung on a loading beam and transferred to the load distribution beam using a frictionless bearing support and pinned struts. From the load distribution beam, the load was transferred to the test specimen to produce four point bending with a constant moment in the middle third of the beam, using rollers. The loading beam had a lever arm of 1 to 14 to magnify the hung weights. Ball joints were used at the supports of the test specimens to allow for free rotations.

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3 3.1

curvature can therefore be attributed to transverse cracks due to applied load and longitudinal cracks due to corrosion of steel bars.

RESULTS AND DISCUSSION Longitudinal strains

Not surprisingly Figure 3 shows that longitudinal tensile strains on a beam that had transverse cracks from applied load (beam 3) are higher than strains on corroded beams without flexural cracks (beams 1 and 2). After 55 days of corrosion, tensile strains on beam 3 were about 1.5 times higher than strains on beams 1 and 2. Unexpectedly, tensile strains on beams 1 and 2 were almost the same but significantly higher than strains on beam 4. Whilst longitudinal compressive strains on beams 1 and 2 like longitudinal tensile strains increased monotonically with time, compressive strains in beam 3 increased for the first 20 days after which slightly decreased and became constant. After 25 days of testing, compressive strains in beams 1 and 2 were higher than compressive strains in beam 3. 3.2

Depth of the neutral axis

Figure 4 shows that for beams 2 and 4, there is a marginal change in the depth of the neutral axis whilst for beam 3 there is about 50% reduction in the depth of the neutral axis. This is consistent with the general behaviour of cracked and uncracked beams corroded under a constant sustained load. After 40 days of testing, the depth of neutral axis of beam 3 is almost equal to the cover depth of compression bars and becomes constant till the end of the test. This indicates that after 40 days, steel bars in beam 3 that were initially subjected to compressive strains become subjected to tensile strains. 3.3

Curvatures

As expected from the variation of strains, curvatures in beam 3 increase for the first 25 days after which become constant whilst curvatures in beam 2 increase monotonically at a decreasing rate till the end of the test. The curvature in beam 3 is significantly higher than the curvature in beams 2 and 4. The figure shows that there is a sudden increase in the curvature of beams 2 and 3 after 14 days and 5 days respectively. These times coincide with the times of appearance of visible corrosion cracks on the tensile face of the beams. The increase in

3.4

Moment of inertia

Figure 6 shows the moment of inertia of beams 2 to 4 derived from the curvatures in Figure 5, the applied moment and measured elastic moduli of beams shown in Table 1. The figure shows that whilst the moment of inertia of beam 4 fluctuates around the same value, the moment of inertia of beams 2 and 3 reduces with an increase in degree of corrosion. At the start of the test, the moment of inertia of beam 2 is as expected much higher than the moment of inertia of beams 3 and 4 because beams 2 is free from flexural cracks due to applied. Unexpectedly, after 20 days of corrosion, the moment of inertia of beams 2 and 3 are almost equal. This indicates that the stiffness of beams tested under simultaneous load and corrosion is the same for beams with transverse cracks and for beams without transverse cracks due to applied load.

4

CONCLUSIONS

1. Longitudinal tensile strains are influenced by the applied load but mostly by the level of corrosion of steel bars. 2. Longitudinal tensile strains of corroded beams increase monotonically with time at a decreasing rate. 3. Depth of the neutral axis is independent of the level of corrosion for beams free from flexural cracks and beams free from corrosion but significantly reduces with an increase in degree of corrosion for corroded beams with flexural cracks. 4. Curvatures of corroded beams increase monotonically with degree of corrosion but at a decreasing rate. 5. The effective moment of inertia of corroded beams decreases monotonically with time of electrolysis but at a decreasing rate. 6. For corroded beams, the effective moment of inertia of beams with flexural cracks is almost the same as the effective moment of inertia of beams free from flexural cracks.

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BETOSCAN – An instrumented mobile robot system for the diagnosis of reinforced concrete floors Michael Raupach & Kenji Reichling ibac—Institute of Building Materials Research, Aachen University, Germany

Herbert Wiggenhauser & Markus Stoppel BAM, Federal Institute for Materials Research and Testing, Germany

Gerd Dobmann & Jochen Kurz IZFP, Fraunhofer Institute for Non-Destructive Testing, Germany

ABSTRACT: A huge number of reinforced concrete floors, e.g. of parking garages or bridge decks are exposed to de-icing salts and thereby subject to a high risk of chloride induced reinforcement corrosion. To be able to properly maintain such concrete floors the whole concrete surface needs to be assessed regarding the following key parameters: concrete cover, carbonation depth, chloride profile, cracks, spalls, hollow areas, condition of the reinforcement, etc. As optimal basis for such an assessment several measurements have to be carried out over the whole concrete surface including mapping of electrochemical potentials and the parameters as mentioned above. For large structures with areas of often some thousand m2 the costs for a full evaluation are usually not accepted by the clients resulting in reduced diagnosis works and design decisions for protection and repairs far on the safe side due to limited data. To overcome this problem a research and development project has been initiated with several partners from research and industry to develop a robotic system which is able to drive over large concrete surfaces and take all relevant data automatically in one step. Furthermore the data collected from the measuring systems can be evaluated together resulting in different synergetic effects: Evaluations on the optimal repair method can be carried out automatically for single points and certain areas enabling the engineers to separate the whole concrete surface into areas subject to different measures from doing nothing over e.g. applying protective coatings up to traditional repair by removing the concrete cover. Furthermore detailed maps are available for design and execution of repair works including cracks and locally damaged concrete. Actually a prototype of the system is under development. This paper focuses on the possible fields of applications for the BETOSCAN system. Keywords:

1

Corrosion, diagnosis, steel, concrete, reinforcement.

INTRODUCTION

A large number of floors from parking structures or bridges are suffering from severe corrosion problems world-wide. Mainly due to the action of de-icing salts or insufficient concrete quality of elder structures the steel reinforcement starts to corrode causing cracking, spalling and losses in cross section leading finally to static problems of the whole structure. To evaluate the condition of such structures adequately usually extensive investigations are necessary. However, to save the costs for the required works often only simple investigation programs are carried out, which are not suitable as a basis for a professional design of measures of maintenance, repair or protection of the structures. Subsequently often improper measures are carried out leading finally to higher total costs than targeted actions based on a reliable database of the structure.

This situation has been the basis for a research and development project to develop a robotic system which is able to drive autonomously over large floors and measure the relevant parameters of the concrete surface simultaneously. The collected data are stored for each investigated point of the structure allowing complex evaluations of the data regarding assessment of the condition, prognosis of the future state, design of measures for protection and repair as well as quality control. A prototype of the BETOSCAN-system is actually under development. 2

CONDITION SURVEYS OF FLOORS: ACTUAL STATE

Especially for concrete structures exposed to de-icing salts the first step of a condition survey should be

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available sensors have been chosen to be integrated in the robotic system. Each parameter can be investigated with a corresponding sensor before it is saved with the right coordinates in a database where all relevant informations are collected. The operator can track the gained data at the same time via WLAN and can generate first maps on site. With these informations further investigations can be planed, prepared and performed without time delay. After the measurements the operator can analyse the data by means of a software tool which will be developed within the scope of the BETOSCAN project.

4

CAPABILITIES OF USING THE BETOSCAN-SYSTEM

The BETOSCAN-system opens several new possibilities for the management of concrete structures: – Diagnosis: Complete assessment of the condition of a structure. – Prognosis: Basis for a detailed evaluation of the future state of the structure. – Service life management: Possibility of repeated or cyclic diagnosis to update and sharpen the prognosis. – Planning of measures: Selection of suitable measures for protection and repair. – Quality control: Determination of key parameters of protection and repair measures after application.

Figure 1. Front and rear view of the BetoScan robotic system (3D Model).

5 a potential mapping of the whole concrete surface. The potential maps show only critical areas, but do not allow to decide upon if and which measures are required. For this purpose further parameters like the distribution of concrete cover, cracks, hollow areas etc. need to be determined. The idea of the BETOSCANsystem (see Fig. 1) is to measure as much additional relevant data as possible in parallel.

3

FEATURES OF THE BETOSCAN-SYSTEM

To investigate the above-named parameters of a reinforced concrete structure, different commercially

SUMMARY AND OUTLOOK

The investigations of large reinforced concrete floors, e.g. of parking garages or bridge decks, are time-consuming and expensive works. Various nondestructive testing instruments are available in which every single investigation result forms a jigsaw piece of the actual state. By combining the different methods in one procedure the analysis will be accomplished faster and more accurate as up to the present. The introduced system can be applied in addition to the analysis to work out various damage zones which can be differentially repaired. In order to manage the service life of a component or to control the quality of a repair achievement, the operator can gain accordant data in an economic manner.

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Accelerated high water corrosion K.P. Mackie Keith Mackie, Consulting Coastal & Harbour Engineer, Cape Town, South Africa

ABSTRACT: Recently the phenomenon of Accelerated Low Water Corrosion (ALWC) has been identified in structural steel members that pass from air, i.e. above high tide level into sea water to below low tide level leading to a pronounced peaking of corrosion rates in the region just below spring low water. This pattern, however, does not manifest in the corrosion behaviour of the embedded steel in reinforced concrete components that similarly pass from air into sea water in tidal waters. These structures commonly exhibit a peaking of corrosion rates just above the high tide level in the region commonly referred to as the “splash zone”. The implication is that this zone has a characteristic structure that could be termed “Accelerated High Water Corrosion” (AHWC) that is just as significant in limiting the lifespan of reinforced concrete marine structures as ALWC is for steel structures.

1

BACKGROUND

This matter is peculiarly associated with marine structures in harbours—in particular structural members that pass from air, i.e. above high tide level into sea water to below low tide level. There is a long tradition in corrosion science and in coastal and harbour engineering that the “splash zone” is a particularly corrosive environment. More recently there has been the recognition that, in the case of steel structures passing from air into the sea, there is a particularly severe peaking of corrosion intensity just below the low water line. This does not occur with the reinforcement to reinforced concrete structures and, in fact, these areas appear particularly free of corrosion. Contrariwise, in reinforced concrete structures, there appears to be a potential for a particularly severe peaking of the intensity of the corrosion of the reinforcement just above the high water mark. In other words, the “splash zone” is not just an amorphous region of elevated corrosion but is in itself formally structured—in particular with respect to reinforced concrete structures. Fig 1 published in the CIRIA report on ALWC shows a classic expression of the corrosion of reinforcement above high water. The term “Splash Zone” implies, in simplistic terms, that the concrete is being splashed with sea water—alternately wetted and dried. This wettingdrying process is the presumptive primary mechanism driving AHWC in that it acts as a pump to drive ions into the concrete. Presumably lack of oxygen below low tide explains the cessation of corrosion in areas permanently submerged.

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Figure 1.

2

AHWC in concrete reinforcement (CIRIA).

GEOMETRIC AND GEOGRAPHICAL PROVENANCE

Fundamentally, AHWC is a positional proposal—it falls within the general concept that corrosion is

a function of position with respect to the environment. Since we are dealing here with marine corrosion, the critical positional datum is the sea shore and we need to understand the location of the sea shore. Superficially, it would appear to be a simple matter of geometry and it is common practice in marine corrosion studies or codes of practice with respect to marine corrosion to delineate offset zones landward of the sea shore. In fact the situation is, geometrically, vastly complex. The main factors are the tides, storms and the fractal nature of coastlines. On the one hand these topics can become very complex and irrelevant to corrosion studies, on the other a naïve understanding can be quite inadequate to understand those aspects that do influence the patterns of marine corrosion. A simplified summary of these issues is needed and is presented here. The effects of local topography, weather and climate—wind, rain, sunshine and ambient temperature and humidity and quality of construction also play a major role in the corrosion processes. The net effect is to make the use of simple geometric zones totally irrelevant. Strictly speaking the problem of corrosion

is a functional one, not a geometric or geographical matter. When the geometric constraints are coupled to the environmental and the tidal, the whole concept of neatly mapped zones collapses. The conclusion, ultimately, is that an alternative, functional, non-geometric approach is needed. A very similar function-geometry conflict arises in the law where the sea shore is used as a geographical property boundary. The approach adopted by legal systems based on Roman law and dating back to the Institutes of Justinian in 533 AD is to eschew the geometric in favour of the functional: “the sea shore is as far as the winter tide reaches its fullest extent”. An appropriate position can always be identified by inspection on site. It cannot be computed from scientific data. This functional based approach is the only way to discuss the spatial variation of corrosion intensity. 3

CONCLUSION

Examination of a variety of examples suggests that AHWC is a real phenomenon but its mechanisms and characteristics need to be investigated and mapped.

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Condition assessment of prestressed concrete structures with static and dynamic damage indexes X.Q. Zhu & H. Hao CRC for Integrated Engineering Asset Management School of Civil and Resource Engineering, University of Western Australia, Crawley, Australia

ABSTRACT: This paper studies the feasibility of using changes in the static and dynamic properties of the prestressed concrete structures under static loads to detect damage. An experimental study is carried out on a prestressed concrete beam of Tee cross section subjected to the static loads. The crack damage is created by applying the four point loads on the beam. A static damage index, based on the flexural rigidity of the beam, is proposed for condition assessment of the prestressed concrete beam structures under static loads. A dynamic damage index based on the change of the natural frequency is also proposed. Experimental results show that the two indexes are effective to detect the damage in prestressed concrete structures. Unlike results obtained for structures of isotropic and homogeneous materials, the identified damage level of the current prestressed concrete beam is found dependent on the static load it carries. This implies that the prestress reduces the crack damage in a prestressed concrete beam after the applied load is removed.

1

2

INTRODUCTION

Quantitative condition assessment of structures has been traditionally performed using the proof load test that leads to an indication of the load-carrying capacity. Alternative approaches such as the ultrasonic wave propagation method, guided wave propagation method, and the vibration testing methods are usually performed on the unloaded state of the structure. At different loading levels, the anomalies in a structure may not be fully mobilized in the load resisting path and thus the measured responses on structures at unloaded condition may not be able to reflect the true damaged status of the structures. This paper investigates the possibility of using changes in the static and dynamic properties of the prestressed concrete structures under static loads to detect damage. An experimental study is carried out on a Tee-section prestressed concrete beam subject to the action of static loads. The crack damage is created using a four point load system. A static damage index, based on the flexural rigidity of the beam, is proposed for condition assessment of prestressed concrete beam structures under static loads. A dynamic damage index based on the change of the natural frequency is also proposed. Experimental results show that the two indexes are effective to indicate the damage in prestressed concrete structures. Unlike results obtained for structures of isotropic and homogeneous materials, the identified damage level of the prestressed concrete beam studied is found dependent on the static load it carries, indicating the prestress reduces the damage after the static load is removed.

EXPERIMENTAL SETUP

The experiment specimen is a Tee-section reinforced concrete beam of 6 m long, with 5.8 m simply supported span. Laboratory experiments were conducted on this simulated bridge deck model with or without damage. The beam section is 500 mm high with 700 mm wide flange and 300 mm web thickness. The beam was resting on the top of two 412 mm high solid steel brackets connected to the concrete floor, as shown in Figure 1. The four-point loading system is used to create the crack damage in the prestressed concrete beam. The loading positions are 1/3 L and 2/3 L from the left end. L is the length of the concrete beam. The detail loading stages are listed in Table 1. Five displacement transducers and four load cells are used to monitor the loading procedure. Two of the four load cells monitor post-tensioned force and the rest monitor the applied loads. The data is recorded and saved to the computer by a data acquisition system. The sampling frequency is 1 Hz for each test. At each loading stage, the impact test is carried out on the beam after the static load is removed from the beam. Nine accelerometers are located on the beam evenly to measure the impact responses. The impact excitation is applied at 1/3 L from the left support using a 12 lb model 5803 A (sledge hammer) of Dytran Instruments, Inc. The data is obtained and saved to the computer by a dynamic data acquisition system, as shown Figure 1. The sampling frequency is 500 Hz for each test.

259

Table 2. αst from the deflection of the beam under different static masses. αst (%)

Figure 1. Table 1.

Experimental setup.

Load (kN)

0

2

3

4

5

6.97

0.00

1.64

2.20

20.88

40.11

10.58

0.00

8.79

14.29

20.33

44.51

Average

0.00

5.22

8.24

20.60

42.31

Loading stages.

Loading Stage

0

1

2

3

4

5

6

Load at 1/3 L (ton)

0

2.3

6

5

6

7

8

Load at 2/3 L (ton)

0

2.3

4.11

5

6

7

8

Table 3.

Load (kN)

3 3.1

RESULTS AND DISCUSSION Static damage index

When there is a static load P on the uniform beam, the deflection of the beam can be obtained. The equivalent flexural rigidity of the beam EIeq can be calculated from the static deflections for the loading configuration. A static damage index can be defined in terms of this flexural rigidity as

α st = 1 −

EI eq

(1)

EI int act

where EIintact = flexural rigidity of the intact beam. Table 2 shows the static damage index from the corresponding equivalent flexural rigidity. When a 6.97kN static load is placed on the beam, the static damage index increases from 1.64% for the loading stage 2 to 40.11% for the loading stage 5. For the case with a 10.58kN static load, it increases from 8.79% for the loading stage 2 to 44.51% for the loading stage 5. In the loading stages 2 and 3, the damage index corresponding to the case with a large mass is larger than that with a small mass, and they are almost the same for loading stages 4 and 5. The results also show that the crack is closed by the post-tension force in the loading stages 2 and 3 when the static load is removed and the small mass is not enough to open the crack. For the loading stages 4 and 5, the crack remains full open even when the static load is removed. 3.2

Dynamic damage index

The equivalent flexural rigidity of the beam can also be computed from the fundamental frequency of the beam, and the damage index from the dynamic approach is defined as

α dyn = 1 −

⎛ f ⎞ = 1− ⎜ EI ⎝ f int act ⎟⎠

EI eq

2

(2)

αdyn of the Beam for different loading stages. αdyn (%) No damage 4.11T 5T

6T

7T

8T

No Load 0.0

1.87

4.17 6.78 9.69

6.97

0.0

3.12

4.18 7.75 11.86 16.48

13.48

10.58

0.0

2.48

2.24 5.85 11.64 15.17

where fintact=fundamental frequency of the intact beam. Table 3 shows the dynamic index for different loading stages. αdyn also increases with the loading stage no. This increasing trend is due to the reduction of the bending stiffness of the specimen as a result of cracking. The index increases from 1.87% for the loading stage 2 to 13.48% for the loading stage 6 when there is no static mass on the beam. For loading stages 3 and 4, the dynamic damage index for the case of no static mass or a small mass on the beam is larger than that with a large mass on the beam. However, the index is almost the same for loading stages 5 and 6. This is again due to the crack close by the post-tension force and the post-tension force is not large enough to close cracks at the loading stages 5 and 6.

4

CONCLUSIONS

An experimental investigation on the static and dynamic properties of the prestressed concrete beam under static loads is reported. Damage in the form of groups of cracks or crack zone is created from incremental static damage loading. Two damage indexes, based on the flexural rigidity and fundamental frequency of the beam respectively, are employed as indicators to describe the presence of crack-type damage in a beamlike concrete structure. Both indexes increase with the crack damage and the static damage index is considered to be more sensitive for large damage cases. The static damage index is much more sensitive to the crack damage than the dynamic damage index when the static load on the beam is large enough.

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Debond Detection in RC Structures using piezoelectric materials X.Q. Zhu* & H. Hao* School of Civil and Resource Engineering, The University of Western Australia, Australia

K.Q. Fan* School of Civil and Resource Engineering, The University of Western Australia, Australia School of Information Technology, Wuyi University, Guangdong, China

Y. Wang* School of Civil and Resource Engineering, The University of Western Australia, Australia School of Civil Engineering, Harbin Institute of Technology, China

J.P. Ou* School of Civil Engineering, Harbin Institute of Technology, China

ABSTRACT: This paper presents a new technique to detect the delamination between the steel bars and concrete in the reinforced concrete structures. The piezoelectric components are mounted on reinforcing bars that are embedded in RC structures as sensors and actuators to generate and record the signal, which is sensitive to the delamination between the steel bars and concrete. The experimental study is carried out on a concrete slab with different debonds between the rebars and concrete. The test results show that the delamination between the rebars and concrete can be detected with the embedded piezoelectric sensors and actuators. 1

INTRODUCTION

Reinforced Concrete (RC) structures are widely used in civil infrastructure systems because of low construction cost and long service life under various conditions. The behaviour of reinforced concrete structures under both static and dynamic loads is highly dependent on the interface between the concrete and the reinforcing bars. When the interface is seriously damaged, such that a macro-crack is formed, debonding takes place and large slip occurs, and the load-transferring capacity of the interface will drop dramatically. This paper presents a novel technique to detect the delamination between the steel bars and concrete in the reinforced concrete structures. The piezoelectric components are mounted on reinforcing bars that were embedded in RC structures as sensors and actuators to generate the signal, which is sensitive to the delamination between the steel bars and concrete. The experimental study is carried out on a concrete slab with different debonds between the rebars and concrete. The test results show that the delamination between the rebars and concrete can be detected with the embedded piezoelectric sensors and actuators. 2 THEORETICAL FRAMEWORK Two parameters are changed when the wave propagates along the steel bar in or outside of concrete: the wave

speed and amplitude. The speed of the wave propagation along the steel bar in concrete are related to the property of the interface between steel bars and concrete. Measurements of the wave speeds using the embedded piezoelectric sensors provide a technique to assess the delamination in the interface. Supposing that the distance between the tip of actuator and the tip of the receiver is L and the time for the wave to travel this distance is t, the average speed of the wave is v = L/t . If the wave speeds along a bare steel bar or a bar inside concrete without delamination are vs and vc, respectively, the scalar parameter can be defined as follows

α speed = 1 −

v − vc v s − vc

(1)

where αspeed is a scalar parameter. αspeed = 0 corresponds to the wave propagation along the bare steel bar and αspeed = 1 is for the steel bar in concrete without delamination. The wave amplitude is another parameter to be affected by the interface between steel bars and concrete. When the wave travels through a medium, its intensity diminishes with the distance. Attenuation that includes the combined effect of wave scattering and energy dissipation is the decay rate of wave as it propagates in a solid. For a single frequency wave, the amplitude change of a decaying plane wave can be expressed as

* CRC for Integrated Engineering Asset Management.

261

A = A0 e− α x

Amplitude

where A0 is the amplitude of the wave at the actuating point and A is the reduced amplitude after the wave travelled a distance x from the initial location. α is the attenuation coefficient of the wave travelling in the x-direction. Similar to the scalar parameter αspeed, there is another parameter corresponding to the changes of the wave amplitude

α amplitude = 1 −

A − Ac As − Ac

200

200

-1 0

0.5

1 Time (ms)

No debonding

Amplitude

0.4

1.5

2

37 mm

58 mm

0.2 0

-0.2 -0.4

0

0.5

1 Time (ms)

1.5

2

casting. One actuator and four piezo film elements with different distances (400 mm, 600 mm, 800 mm and 1000 mm) are mounted on the surface of the reinforcement bar without debond, and other rebars are with one actuator and two piezo film elements at 400 mm and 1000 mm, respectively as shown in Figure 1. EXPERIMENTAL RESULTS

Figure 2 shows wave propagation along the rebars with different debonding lengths. From these results, the following observations can be obtained. 1. From figure 2, the time delay of the response signal reduces with the debonding length. Corresponding to the no debonding case, the time delays are 0.401 ms for 37 mm debonding, 0.363 ms for 58 mm debonding. The distance between the actuator and sensor is 1000 mm. From Equation (1), α speed are 0.97, 0.87 for 37 mm and 58 mm debonding, respectively. α speed reduces with the debonding length and it could be a good indicator of the debonding damage. 2. In Figure 2, the amplitude of the response signal increases with the debonding length. The amplitudes for no debonding, 37 mm and 58 mm debonding are 0.063, 0.146 and 0.215, respectively. By Equation (3), αamplitude are 0.84, 0.71 for 37 mm and 58 mm debonding, respectively. αamplitude also reduces with the debonding length, and it could be another indicator of the debonding damage.

Piezo Film Elements

5

200

58 mm

(b) Response signal

4

The experimental system includes two parts: a) the actuating part is to provide the excitation or input of the system. It includes the actuator of piezoelectric strips and the power amplifier that provides the power supply to the actuator. b) The piezo sensing part is to measure the response. This part includes the piezo film element and its charge amplifier. A reinforced concrete slab (1500 mm × 500 mm × 100 mm), shown in Figure 1, are constructed for debond tests in the laboratory. The slab is supported at two ends. The slab includes 5 reinforcement bars (round bar with diameter 16 mm) with 50 mm cover of concrete to reduce the effect of the concrete thickness. The distance between two rebars is about 100 mm. In the slab, there are 5 reinforcement bars with different debond sizes between the reinforcement bar and concrete, b = 0 mm, 21 mm, 37 mm, 58 mm and 99 mm. Debonding is simulated by a plastic tube sealed at two ends so that concrete can not enter the tube during the

250

37 mm

0

Figure 2. Wave propagation along the rebars with different debond lengths.

EXPERIMENTAL SETUP

PZT Debonding Concrete Actuator b

No debonding

1

-2

(3)

where As, Ac are the amplitudes of wave propagating along the bare steel bars or in concrete without delamination, respectively. αamplitude = 0 or 1 indicates that there is no concrete around the steel bar, or the steel bar is in the concrete without delamination in the interface, respectively. 3

(a) Input signal

2

(2)

200

200

250

1500

100 50100 100 100 10050

Figure 1. Two concrete slabs for debond tests.

RC Bar

CONCLUSIONS

A new method has been presented to detect the delamination between the steel bars and concrete in the reinforced concrete structures. The experimental study is carried out on a concrete slab with different debonds between the rebars and concrete. Two scalar parameters are defined according to the changes in the received signal to detect debonding damage. The test results show that they could be good indicators of the debonding damage. Further study is needed to quantify the debonding damage.

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Development of a new test method for mineral based composites – Wedge splitting test K. Orosz & B. Täljsten Technical University of Denmark, Kgs., Lyngby, Denmark

K. Orosz Norut Technology Ltd., Norway

ABSTRACT: The well-known wedge splitting test, often used for characterizing brittle materials has been modified and adapted to testing MBC-reinforced concrete under splitting load. MBC (Mineral Based Composites) is a newly developed strengthening system for existing concrete structures where FRPs, mainly CFRP grids are externally bonded to the concrete surface by means of cementitious bonding agents. Crack development, crack patterns, Crack Opening Displacement (COD) versus splitting load and fracture energy are investigated and evaluated. Development of a suitable test specimen and test setup has been accomplished. Bond provided by both mortars was excellent leading to CFRP rupture. By applying PVA-reinforced ductile ECC as bonding agent, improved performance, significantly higher fracture energy, multiple cracking and enhanced ductility were observed, caused by improved bond between grid and mortar due to the refined grain structure, the bridging effect of the embedded fibres working against crack opening and via direct mechanical interlock with the grid.

1

STRENGTHENING WITH MINERAL BASED COMPOSITES

The use of concrete has been widely spread over the world and is the most used construction materials for structures. Nowadays, there is an increasing need to restore or strengthen civil engineering structures worldwide, in particular related to transportation because of ageing, deterioration, and misuse of facilities, lack of regular maintenance and repair, and the use of inappropriate materials, construction techniques or both. Increased or changing loads may also lead to a need of upgrading. Externally bonded FRP (Fibre Reinforced Polymer) systems have been proven to be an effective strengthening method in repairing or strengthening structures. In traditional FRP strengthening, the bonding agent is normally an epoxy resin. FRP most frequently is either glass or carbon fibre, from unidirectional reinforcing bars, strips to 2D textiles, sheets, plates or grids made from separate fibre tows. A new FRP-based strengthening system by applying the so-called MBC materials (Mineral Based Composites) implements the advantages of traditional FRP strengthening (light weight, corrosion resistance, high tensile strength, high modulus etc.) but eliminates major disadvantages and some limitations of those. Problems with epoxy-based systems are diffusion closeness, poor thermal compatibility with base concrete, sensitivity to moisture at the time of application, hazardous working environment for the manual worker and the problem of minimum temperature of assemble in cold climates.

In this recently developed innovative strengthening system, the epoxy is being replaced by cementitious matrices to bond the FRP material to the concrete surface. MBC therefore is a composite material made by replacing part of the cement hydrate binder of conventional mortar with polymers which, with the addition of conventional FRP becomes a high-performance external strengthening system for existing concrete structures. The FRP in MBC applications normally is a two-dimensional grid. As matrix, premixed, commercially available, polymer-modified quasibrittle mortars are typically used. Several tests have been carried out on MBC-systems, focusing on flexural strengthening by Wiberg (2003), Becker (2003) and more recently, shear strengthening by Christiansen and Jürgensen (2006), or Blanksvärd (2007). 2

RESEARCH SIGNIFICANCE

Before applying the MBC as for strengthening material in conditions where the structural member is subjected to tensional or splitting forces (or bending combined with axial forces), it is necessary to properly characterize the tensile properties of it. To the authors´ knowledge, at the present time there is no literature published on tensional behaviour of MBC except for uniaxial tests carried out on MBC dogbones by (Andersen and Jespersen, 2006). Other suitable test methods are being searched for, which quantify the tensile properties of MBC directly (dogbone tests) or indirectly (WST, giving the splitting tensile strength). Knowledge of related parameters such

263

as the softening curve makes possible estimations for instance about brittleness in compression and tension or shear capacity. Brittleness of concrete-like materials is usually evaluated by means of the post failure behaviour in tension governed by the stress versus crack width relation (σ-w), the so-called softening behaviour which is a basic property of a concrete described by the tensile strength, the maximum crack width and the fracture energy, which corresponds to the area under the stress versus crack width curve. WST (Wedge Splitting Test), originally introduced by Tschegg and Linsbauer in 1986 and further improved by Brühwiler and Wittmann in 1990 is a suitable method, traditionally used for brittle materials, for obtaining both splitting tensile strength and an estimation on the fracture energy. WST is a test method which ensures post-cracking stability and captures the softening part of the load-displacement curve after peak load. Since both components of the Mineral Based Strengthening System behave in a brittle manner, WST has been adapted for investigating the tensile contribution of the strengthening system. Tests presented here also involve a more ductile, long PVAreinforced “mortar” exhibiting strain-hardening under tension, ECC (Engineered Cementitious Composites). Different combinations of brittle grid reinforcement and cementitious bonding agents of different ductility are investigated and evaluated through recording and comparing crack widths, crack patterns, splitting load versus crack opening displacement curves and fracture energy for the MBC-strengthened specimens. 3

TENSIONAL BEHAVIOUR OF CONSTITUENTS

Brittle matrices, such as plain mortar and concrete, lose their tensile load-carrying capacity almost immediately after formation of the first crack. Polymer-modified quasibrittle mortars used in this test behave in a similar way. In contrary, ECC or in general, HPFRCC (HighPerformance Fibre-Reinforced Cementitious Composites) are defined by an ultimate strength higher than their first cracking strength and the formation of multiple cracking during the inelastic deformation process. In a structural member subjected to tensile or splitting forces, after the first macro-cracks have appeared, the tensile load still continues to increase. The contribution of a cementitious matrix to the load-deformation response of reinforced concrete or ECC is the so-called “strain hardening” or “plastic hardening” (Naaman and Reinhardt, 1995). It guarantees that a structural element made of such a

high-performance material will maintain its stability also after cracking. Behaviour of CFRP reinforcement is linear elastic up to failure with no plastic reserve. In tension, failure is characterized either by slippage in the fibre tows instead of yielding of those, or by linear elastic deformations until FRP ruptures in a brittle manner when reaching the tensile failure strain of the reinforcement (Brameshuber, 2006). 4

TESTING

Traditional WST uses a steel wedge which is being pressed downwards at a constant rate. The vertical load is being converted into horizontal splitting forces by means of roller or needle bearings in order until the specimen has split into two halves. The “scaled up” test set-up involved re-designed loading devices with needle bearings and larger specimens to ensure enough anchorage length for the CFRP in the mortar. Dimensions of the base concrete were set to 400 × 400 × 100 mm with a starter notch length of 50 mm at the top of the specimen. One medium-sized (grid spacing: 42 × 43 mm) epoxy-coated CFRP grid in two possible positions (perpendicularly to the expected vertical governing crack, and rotated in 45°) and two different mortars have been used to strengthen the base concrete. The grid is cast between two layers of mortar. Improved bond between base concrete and mortar is achieved by a sandblasted concrete surface and treatment with a primer first. The premixed mortar is a high strength, quasi-brittle mortar while the more ductile ECC mortar contains long embedded PVA-fibres to improve ductility and a large portion of fly ash. All possible combinations were tested giving a full test matrix. All specimens had at least 28 days of curing before tested. The MBC reinforcement was applied on both sides. Specimens were simply supported at the bottom and have a starter notch at the top where the splitting load was applied to initiate cracking from the weakened cross-section. Tests were run deformation-controlled by keeping the crack opening displacement constant by means of a clip gauge mounted at the tip of the starter notch at the top of the specimen. Vertical load and COD are recorded. ACKNOWLEDGEMENTS The research work presented in this paper was performed at the Technical University of Denmark and financed by the Norwegian Research Council through the strategic institute program RECON at Norut Technology Ltd.

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Evaluation of fibre distribution in concrete using AC impedance technique B. Srinath & Manu Santhanam Indian Institute of Technology Madras, Chennai, India

ABSTRACT: Distribution and orientation of fibres in concrete has a profound effect on the hardened mechanical properties. AC Impedance Spectroscopy (AC-IS) is a useful tool to characterize bulk changes in the electrical properties of materials; thus, its application to fibre reinforced concrete is only natural. In this study, the impedance characteristics of fibre reinforced self compacting concrete and normal concrete were measured using an impedance analyzer. Ultrasonic signals were also collected for the same specimens. Beam specimens were used to create a condition of one-directional flow of concrete during casting. The impedance response was measured in the direction of pouring concrete, in the direction of flow and in the orthogonal direction. The results from the response were presented in terms of the normalized matrix conductivity, the fractional function, the equivalent single fibre representation, and the dispersion factor. All results indicate a preferential alignment of the fibres in the direction of the flow, and in the plane normal to the direction of pouring. As the fibre volume fraction decreased, the degree of orientation in the flow direction reduced, while the dispersion of fibres increased. The reduction of specimen width in the orthogonal direction led to the increase in alignment in the pouring direction. Similar trends were observed for self compacting and normal fibre reinforced concrete. The ultrasonic method was not found to be sensitive enough at low fibre volume fractions (below 1%) to detect significant changes in the amplitude of signals in the direction of flow. The results of impedance spectroscopy were corroborated by the evidence from image analysis and manual counting, which indicated a higher orientation number in the direction of flow. Furthermore, the load carrying capacity and deformability of the composite was also found to be higher in the direction of flow.

1

INTRODUCTION

2

Self compacting concrete (SCC) can compact under its own self weight, without any external vibration. This is made possible by the choice of suitable ingredients to enable the concrete flow and also resist segregation. Research has shown lately that because of the highly flowable nature of SCC, more fibres can be incorporated into it as compared to conventional concrete, without affecting the workability much. Thus, there is a good potential for the use of fibre reinforced self compacting concretes (FRSCC). Subsequently, there is a need for studying the orientation of fibres in FRSCC. The distribution of fibres in concrete is influenced primarily by the casting and placement technique, specimen size, fibre size, geometry, volume fraction, and maximum aggregate size, in addition to other factors. Recently, AC impedance spectroscopy has been used for studying fibre orientation and distribution in concrete composites. AC-impedance spectroscopy (AC-IS) is uniquely effective as a characterization tool for a general class of short or discontinuous fibre-reinforced composites (FRCs).

METHODOLOGY

After hardening, the beam specimens were demoulded and cut along the direction of flow (long direction) into five equal cubes of 150 mm length, using a diamond tipped concrete saw; these sections were labelled as 1, 2, 3, 4, and 5 (where section 1 is the end where concrete is poured and section 5 is the other end). Sections 1 and 5 were discarded to avoid end effects, while Sections 2, 3, and 4 were selected for the AC-IS and ultrasonic pulse velocity studies. The cubes were tested in all the three directions. The specimen details are shown in Figure 1. 3

RESULTS AND DISCUSSIONS

There is almost negligible difference in the normalized matrix conductivity values at different sections of the beam with respect to a particular direction i.e. X, Y and Z directions. This shows that fibre orientation in beam remains more or less constant throughout the beam with respect to a particular direction. However, when the normalized matrix conductivity values are compared with respect to X, Y and Z directions,

265

Figure 1. Specimens for fibre orientation studies, cut from the 750 mm long beam.

it is observed that in the X direction, normalized conductivity value is maximum, while the minimum normalized conductivity value is observed in the Z direction. Hence, it can be concluded that the fibres are primarily oriented in X direction, and in the X-Y plane. Similar observations were made by Ozyurt et al. (2006). It can also be observed that there is not much difference in the values of normalized matrix conductivity for high slump concrete and self compacting concrete. Hence, it can be concluded that orientation behavior of fibres is almost the same in vibrated and self compacting concrete; this shows that the effect of vibration technique on fibre orientation is almost negligible in this case. In the beam with the decreased width (the width is reduced in the Y direction) there is a general trend of increase in normalized conductivity values in the Z direction compared to the Y direction; hence, it can be concluded that there is an increase in the fibre orientation in the Z direction compared to the Y direction. But for the beam specimen with unaltered width, no such trend can be observed. This observation clearly indicates that the width of the specimen has an effect on the fibre orientation. The above observation compares well with the investigations made by Austin et al. (2003). Fibres are better dispersed at lower fibre volume fractions when compared to higher volume fractions. This is expected since at higher volume fractions

there are greater chances of clumping of fibres. There is also not much difference in the fibre dispersion for the self compacting and vibrated concretes with 0.25% fibre volume fraction. The dispersion factor is, however, affected by the width of specimen, as it can be observed that the average dispersion factor for a specimen with decreased width is less than for the normal specimen. Orientation numbers for self compacting concrete with a fibre volume fraction of 1% in the X direction (or YZ plane) are higher compared to the Z direction (or XY plane), which indicates a preferential orientation in the X direction. It can be observed that even for smaller fibre volume fractions, fibres are primarily oriented in the X direction compared to the Z direction. There are no significant effects of either changing the mode of vibration or reducing the specimen width on the general pattern of the results. Hence, it can be concluded that the fibres are preferentially oriented in the X direction compared to Z, and these results compare well with the AC-IS method.

4

CONCLUSIONS

• Studies on fibre distribution using AC-IS revealed that the composite resistance significantly varied in the X (direction of flow), Y (direction perpendicular to the flow) and Z (pouring direction) directions. The orientation of fibres was observed to be preferentially in the X direction and in the X-Y plane. • When the specimen dimension was reduced in the Y direction, the orientation of fibres in the Z direction increased compared to the Y direction. • With a decrease in the volume fraction of fibres, the dispersion factor increased. There was nearly a 15% increase in dispersion factor for the concrete with 0.25% fibre volume fraction when compared to a 1% fibre volume fraction. • Fibre dispersion was not significantly different for high slump vibrated concrete and SCC (the difference in dispersion factor was less than 5%). • Image analysis results showed that the orientation number was highest in the X direction, thus corroborating the results of the AC-IS studies regarding the preferential orientation of the fibres in the X direction.

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Simple hydration equation as a method for estimating water-cement ratio in old concrete S.O. Ekolu School of Civil & Environmental Engineering, University of the Witwatersrand, Johannesburg, South Africa

ABSTRACT: Hydration studies done several years ago by R.H. Mills, built on the work of T.C. Powers and other researchers to derive a relationship between the ultimate degree of hydration and water-cement ratio (w/c). This paper reviews the αult-w/c hydration equation for possible application to w/c estimation in hardened concrete. Unlike current methods, the empirical method based on hydration equation has the merit of simplicity but its accuracy has not been tested and proved. A χ2-test statistic showed a good agreement between values of original w/c and estimated w/c for pastes hydrated for 448 days. Accuracy seemed particularly good for mixtures of portland cement only. The method is sensitive to the C3 A content while use of extenders in mixtures appears to reduce accuracy. These and other shortcomings are discussed including the effects of advances in cementitious material systems. The need for modifications and development towards an improved method is highlighted.

1

INTRODUCTION

It is usually essential to determine the actual w/c in already hardened concrete regardless of prior knowledge of its original mix design. W/C is a key parameter of influence in mix design of concrete and subsequently, its long-term durability. Determination of w/c for new concretes may be needed to verify adherence to specifications during concreting or to examine effects of adverse environmental conditions encountered during casting such as rainy conditions. In certain situations, knowledge of w/c is required to settle disputes. For old concretes, determination of w/c can be important in preparation for repairs. W/C is a parameter possessing plenty of information associated with durability. Accurate measurement of w/c in hardened concrete is however elusive and the current methods available have several limitations.

2

STANDARD METHODS FOR ESTIMATION OF W/C IN HARDENED CONCRETES

Current methods for w/c determination in hardened concrete are based on chemical analyses or microscopy techniques. Problems common to the methods are poor accuracy, expense, and practical limitations. Chemical methods require analysis for cement content and water content followed by calculation of w/c ratio. It is worth noting that the standardized method BS 1881: Part 124 is not considered suitable for analysis of old concretes of age more than 5 years but even more importantly, the method does not apply to poor concretes, distressed concretes or concretes

containing extenders. Yet some of these issues may be the very reason(s) for requiring determination of w/c value. Accuracy of the method is suspected to be as poor as ± 0.1.

3 A DIFFERENT APPROACH TO W/C ESTIMATION 3.1

Early hydration studies

Studies on hydration conducted by Powers and other researchers around 1930’s to 1960’s divulged incredible understanding of micro-mechanics and characteristics emanating from cement hydration. It can be appreciated that w/c plays a significant role in the important properties of concrete. If for arguments sake, all water mixed with cement paste reacted to full hydration (which is known to be unlikely), the quantity of products of hydration would provide a direct account of the cement and water contents used. Complexity is introduced when in reality, water takes different forms during hydration and a fair amount of it is not at all involved in chemical reaction. Another difficulty encountered is the lack of accurate means of quantifying hydration products. The hydration product calcium silicate hydrate (CSH or tobermorite) gel is amorphous while Ca(OH)2 may be semi-crystalline. The use of x-ray diffraction is unable to provide adequate quantification of both products present in the system. To get around the fore-mentioned difficulties, it suffices to assume that the extent of hydration attainable in a closed hydrating systems of cement paste (containing sufficient water for full hydration), can be

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monitored through the degree of hydration ultimately achieved or achievable. The ultimate degree of hydration in a closed system must therefore correlate to the quantities of hydration products, and in turn relate to the original cement and water contents used in the mix, that is the original w/c. It is this relationship that Mills, 1965 found to exist when studying cement pastes cured at normal temperatures in his work conducted at University of the Witwatersrand. He concluded that ultimate degree of hydration, αult in cement pastes can be related empirically to w/c by the equation:-

α ult =

Data points Age (days) w/c αult (%) χ2

4

1.031.wo 0.194 + w0

Where wo represents the original water-cement ratio of the paste mixes. 3.2

Table 1. Chi-square test of w/c estimation (produced with data from Mills, 1965).

Accuracy

This equation was derived from using mathematical and geometric relationships based on simplified model assumptions. The credibility of this relationship is founded on strong correlations between experimental results and predictions tested on data by different authors referenced in Mills, 1965 ((Powers, 1947; Verbeck & Foster, 1950; Cernin, 1960), and his own data. These have been plotted for a selected practical range of w/c’s = 0.25 to 0.85 as seen in Figure 1.

All data

Mills (pc)

Mills (pc + ggbs)

29 Various Various Various 0.57

8 448 0.25–0.85 54.9–83.3 0.03

15 448 0.22–0.45 50.1–51.8 0.29

CONCLUSION

The aim of this paper was to examine the αult-w/c hydration equation as an approach for estimation of w/c in hardened concretes. Accuracy and precision of the method hasn’t been tested nor proved although good w/c estimations were determined within the confines of experimental data used. While the method does have useful advantages over current techniques, constants used in the original expression appear to be rather inappropriate for modern cementitious materials whose characteristics and chemistry changed remarkably over the past century. With modifications for robustness and appropriate material constants, it is believed that the core mathematical and experimental approach employed holds promise to an effective approach for w/c estimation in hardened concretes.

Figure 1. Correlation between original and estimated w/c values (produced with data from Mills, 1965).

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Selection of an optimal method for humidity elimination in masonry buildings I. Netinger Faculty of Civil Engineering, Osijek, Croatia

H. Anton Civil Engineering Institute of Croatia, Osijek, Croatia

D. Bjegovic´ Faculty of Civil Engineering, Zagreb, Croatia

ABSTRACT: The paper is about humidity problems in masonry buildings, which are the most frequent building style in East Croatia. The working principles of some known methods for humidity elimination in buildings including methods of vertical and horizontal sealing, and drainage are explained. Applying a multi-criterial analysis, the selection of an optimal method for humidity elimination is demonstrated.

1

INTRODUCTION

In this paper, there had been set forth the acknowledged methods of the masonry capillary humidity elimination, their advantages and disadvantages, as well as the efficiency of the same. Applying multi-criteria analysis, based upon criteria considered as being crucial, the authors used the approach to select best of the methods for rehabilitation of a house. 2

REHABILITATION METHODS OF THE MASONRY DE-HUMIDIFICATION

According to the mechanism of effect, the existing rehabilitation methods might be sorted into three groups: procedure of horizontal sealing, procedure of vertical sealing, and draining. 2.1

Rehabilitation methods of de-humidification by horizontal sealing

The method of horizontal sealing can be applied by several techniques, usually mechanical procedures, injection or electrical-chemical procedures, to disrupt the capillary humidity flow inside walls. Mechanical procedures, operate on very simple principles. The method comprises making openings or slots mechanically or by machine sawing along the wall, and might include sealing of capillaries. The sealing layer might be in form of bituminized metal foil, lead foil or foil made of artificial materials, usually laminates. The layer of water resistant mortar or sheet metal made of stainless steel is pressed into horizontal joints. Procedure of injecting appears to be a very efficient one, as far as elimination of humidity out of masonry

is concerned. Drilled holes inside a wall are injected with a substance which acts as sealant to pores, making them water resistant (hydrophobic). Applying a specially designed vessel equipped with a proportioning device, there should be poured the substance for blocking the capillary humidity penetration (water glass, potassium water glass, silicone micro-emulsions, paraffin, organic resins like acrylics, epoxy resin, bitumenized or microconcrete emulsions), by which there would be possible moisturizing the walls along the entire length of the drilled hole. Permanent moistening guarantees formation of continuous water resistant barrier. Electrochemical procedure of de-humidification is based on the principle of re-directing the natural polarity of capillary humidity flow (upwards) by installing the electrodes and applying current to block humidity flow and re-direct it to flow in the opposite direction, that is, into the ground. Within this procedure, the electrodes ensure direct current, where the current runs from the positive electrode across the wall, towards a negative electrode. 2.2

De-humidification by vertical sealing

Procedures of vertical sealing presume placement of vertical insulation across the masonry, like: water resistant mortars, water resistant suspensions based upon concrete, or bituminized strips. 2.3

De-humidification rehabilitation by draining

Draining is applied to prevent side penetration of surface waters formed from rain as well as limiting rise of high water levels. A draining pipe made of perforated concrete or plastic is placed at the depth of freezing or at base depth. It is surrounded with filling of coarse gravel.

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3

Table 1. Collected data on price, efficiency and durability of each of the procedures for a house rehabilitation.

APPLICATION OF MULTI CRITERIA ANALYSIS TO SELECTING THE OPTIMAL METHOD OF HUMIDITY REHABILITATION OF A HOUSE

Each of the described methods has its advantages and disadvantages which are evidenced in price, procedure efficiency, complexity of work, duration of work, procedure duration among others. Faced with the difficulty of selecting the best method of humidity rehabilitation of a house, it might be useful to use multi-criteria analysis. In this paper, the Analytical Hierarchy Process (AHP) method was applied to criteria essential to owner of facility and criteria important to both to the owner and contractor. 3.1

Input parameters

Relevant data presented in Table 1 was collected from enterprises dealing with de-humidification of structures and applied to multi-criteria approach to selection of an optimal method. Since vertical sealing and draining as separate procedures are not sufficient for efficient de-humidification rehabilitation, the methods had to be applied combined with some other procedures of humidity prevention. For evaluating the optimal rehabilitation method, criteria essential to the owner of the facility i.e. economical criteria, procedure efficiency, procedure durability, and also the procedure impact to stability of the facility within the work phase, as well as criteria essential from point of view of the contractor (complexity of the procedure and duration of the work) were considered. By subjective evaluation, the authors assigned weightings to the criteria as follows: In case of evaluating the optimal method per criteria essential to owner of the facility, 25% weighting was assigned to each criteria. In case of evaluating the optimal method per criteria essential both to owner of the facility and the contractor, the weightings were 20% to economy, 20% to efficiency, 20% to durability, 20% to procedure impact to stability of the facility, 10% to procedure complexity, and 10% to duration of work In Table 2 are shown the input parameters used for selection of optimal de-humidification rehabilitation method. 3.2

Method

Price (EUR/m′)

Efficiency (%)

Durability (years)

1

70

100

50

2 3 4 5

45 30 40 30

70 60 70 40

15 15 15 50

1 = Mechanical procedure—by pressing of sheet metal. 2 = Procedure of injecting. 3 = Electrochemical procedure. 4 = Vertical sealing—by bituminized strip. 5 = Draining. Table 2. Input parameters for selecting the optimal dehumidification rehabilitation method.

Method

Price (EUR/m′ )/Efficiency (%)/Durability (years)/Impact to stability of the facility/ Procedure complexity/Duration of the procedure

1 2 3 4 5

70 / 100 / 50 / *** / *** / *** 100 / 100 / 15 / **** / ***** / ***** 70 / 90 / 15 / *** / **** / **** 45 / 70 / 15 / ** / ** / ** 30 / 60 / 15 / * / * / *

1 = Mechanical procedure—by pressing of sheet metal. 2 = Draining + vertical and horizontal sealing by bituminized strip. 3 = Vertical and horizontal sealing by bituminized strip. 4 = Procedure of injecting. 5 = Electrochemical procedure. Less “*” markings mean less impact to the facility stability, lower complexity of the procedure and shorter duration of work-out.

low price, non-interference with facility’s stability, low complexity of work and shortest work duration. 4

The selection of an optimal method

From the owner’s point of view, the optimal selection results favour the mechanical procedure of humidity rehabilitation. By this procedure, such a result appears to have its justification in the entire efficiency and duration of problem solving. If criteria, essential from the owner’s point of view are accompanied by criteria presumed as being essential by the contractors too, the electrochemical method appears to be the best solution of de-humidification. Although having lower efficiency and durability compared to other methods, the selection of electrochemical method is justified by its relatively

CONCLUSIONS

This work deals with the problem of humidity present within masonry construction and familiar methods for its repair. Based upon collected data regarding the price prevailing within the market of the Republic of Croatia on efficiency, durability and impact of each of the methods to stability of the construction, multi-criteria analysis was used to select the optimal method of humidity rehabilitation. The mechanical procedure was favoured as optimal for criteria essential to owner. When criteria essential to both the owner and contractor were considered, the electrochemical method was selected as optimal.

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Repair of guard gallery columns, Robben Island Maximum Security Prison H.G. Smith Consultant Project Manager, Heritage Clerk of Works to Robben Island Prison Project

M.G. Alexander Department of Civil Engineering, University of Cape Town, South Africa

ABSTRACT: The paper describes the deterioration of stubby brickwork columns supporting the roof of the Guard Gallery of the Robben Island Maximum Security Prison. The columns had cracked severely through both brick joints and encasing plaster. Investigations revealed that the cracking was due to severe corrosion of an embedded vertical steel tie rod that acts to hold down the roof. The corrosion was caused by chloride contaminated mortar and the ingress of moisture. The remedial measures are also described. 1

INTRODUCTION: CONTEXT, AND THE PROBLEM

The Maximum Security Prison on Robben Island, in Table Bay, was built in 1964 to hold political prisoners. It was notorious world-wide as the prison where Nelson Mandela and others were held. After the negotiated end of apartheid in the early 1990’s, the prison was finally closed, and the buildings transferred to the Robben Island Museum to be developed as a tourist site. Robben Island was soon declared a World Heritage Site. In July 2007, three months in to a repair and renovation contract for the Maximum Security Prison, serious failure of a number of structural columns in the Guard Gallery was identified. Workmen opening up to repair what had been assumed to be plaster cracks, common throughout the 40-year old prison buildings, found that the masonry within the plastered columns was severely cracked, and in places individual bricks were split through and loose. Subsequent careful removal of areas of plaster from a number of the columns revealed the extent of the problem and indicated that the columns had most probably lost their integrity. Figure 1 shows cracking in a column of the Gallery. Identifying the cause of the failure required further detailed investigation and analysis. 2

Figure 1. Severe vertical cracking of a column. Column width approx. 27 cm.

PHYSICAL AND STRUCTURAL DESCRIPTION

The guard gallery is 31 m long and 2.5 m wide, with a reinforced concrete slab floor, flanked by plastered masonry walls 1.3 m high on which are built the plastered brick columns in question, carrying the pitched roof. There are eight exposed columns at 2.95 m centres on each long side of the gallery. The columns are

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stubby: 1.5 m high with a cross-section of approximately 380 mm wide × 270 mm deep (including the plaster). The columns comprise brick courses of 1½ bricks wide and 1 brick deep, with a coursing of 90 to 95 mm. The roof structure (timber beams, rafters and battens, covered by fibre cement sheets) is supported on the columns, and is tied down by steel metal tie rods that go down the centre of the brick columns. The main form of deterioration of the columns was severe vertical cracking, interspersed with a

certain amount of horizontal cracking. The vertical cracks varied in width, from fairly narrow (<1 mm) to alarmingly wide, ca. 20 mm. It was decided that further opening up of a typically deteriorated column would be necessary to understand the cause of the cracking. Exposed by this opening up was the roof holding-down bolt, a 16 mm steel rod continuous down the centre of the column and embedded in the mortar at the centre of and between the three bricks’ alternating bonding. Significantly this rod, lightly surface corroded over the first 250 mm of its length below the roof and eaves, was increasingly severely corroded over the remaining 900–1100 mm down to the parapet of the Gallery wall. At the worst areas of corrosion, up to 50% of the rod was corroded, with rusted fragments coming away with the mortar. It was clear that the corrosion products caused lateral strain (i.e. expansion) in the brickwork, reflected into the plaster cladding. Thus, the plaster was not the cause of the deterioration—an important fact to establish in regard to understanding the cause.

3

ANALYSES OF MATERIAL SAMPLES

Representative mortar and plaster samples (and a brick) were removed from the demolished column, and analysed for chloride content—see Table 1. Typical values that would put a steel rod embedded in a cementitious material at risk from a corrosion viewpoint would be ca. 0.05% of chlorides by mass of the material. The results indicate considerably higher values than this, of the order of 7 times larger in the mortar, and 5 times larger in the plaster. The bricks appeared to be relatively un-contaminated with chlorides. The cause of the corrosion of the steel bar in each column was clear: it was highly likely that a substantial amount of chloride (salt) had been included in the mortar and plaster mix. This was unlikely to have originated environmentally from wind-blown chlorides over the years, considering the high level of chlorides in the materials. It was more likely that the chlorides were introduced in the mortar and plaster mixes directly—possibly by way of use of unwashed beach sand, or even sea-water as mixing water.

Table 1.

Chloride analyses of materials. Percent chlorides (by mass of material)

Material

Sample no.

Indiv.

Average

Mortar

1 2 3 1 2 3 1

0.3802 0.3512 0.3554 0.2402 0.2353 0.2350 0.0360

0.363

Plaster Brick

4

0.237 –

REPAIR

The specification for the repair works includes adding in two extra rods on each side of an existing rod (spotwelded to the un-corroded length of the original at top and bottom), and additives to the mortar for both joints and plastering to eliminate future water penetration. Measures also include coating of new and original steel rods in three coats of bitumen tar to isolate them from any remnants of corrosive chemicals in the original mortar. The columns will be rebuilt one at a time, to preserve adequate anchorage of the roof during the works. 5

CONCLUSION

A quick response to uncovering an instance of unforeseen deterioration of the fabric of an iconic element of the Prison; thorough investigation, recording and analysis; the participation of appropriate specialist skills and techniques in the process; and teamwork on the part of all concerned have resulted in a full understanding of the nature of the failure and the selection of an optimal solution to a difficult problem. [Employer: Department of Public Works (Project Manager J I D Janse van Rensburg) for Robben Island Museum; Contractor: RUWACON (Pty) Ltd (Mr Rayno Schutte, Site Agent); Principal Agent: Trudi Groenewald P.Arch.(SA), RIBA, AAArb, of Groenewald & Preller, Architects; Heritage Clerk of Works: Howard Smith, BA(Arch) (Bristol), AADipl.(Planning), M.Phil (UCT); Specialist Consultant: Professor Mark Alexander, PhD, MSc(Eng), BSc(Eng); Robben Island Museum represented by Shaun Davis, Heritage Manager.]

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Theme 3: Concrete repair, rehabilitation and retrofitting Repair methods and techniques

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Concrete repair: Research and practice – The critical dimension A.M. Vaysburd Vaycon Consulting, Baltimore, USA

P.H. Emmons Structural Group, Inc., Hanover, USA

B. Bissonnette Laval University, Quebec, Canada

ABSTRACT: Creating a scientifically based concrete repair technology is an ultimate goal of the industry at the beginning of the 21st Century. Improved performance of repaired concrete structures will depend on whether this objective is achieved or not. It will depend on the willingness and the ability of all partners in the concrete repair field to cooperate in developing such a technology. Although significant research advancements have been made with regard to understanding (but not control of) several chemical and physical phenomena responsible for repair failures, the trend to the “low performance” concrete repairs has yet to be reversed. Research is not occupying its time-honored place well ahead of practice, and in some areas of concrete repair, it is lagging behind. With special attention to durability of repaired structures, the paper addresses some of the shortcomings in the scientific input in areas of repair durability such as condition evaluation methods, key electrochemical factors, service life prediction and performance testing techniques.

1

INTRODUCTION

There are several missing links in the present day knowledge of concrete repair. It must be admitted that too little is usually known about electrochemical processes that take place in repaired structures; condition evaluation techniques and methods are not reliable; corrosion processes and their influence on service life of a repaired structure are not fully understood; repair options cannot be rationally compared without modeling the future performance. With only the partial knowledge that is presently available, the desired performance of the repaired concrete structures cannot be assured confidently, nor can safety of the structures be assured and the maximum possible value be realized from the repair dollar. One of the greatest ills is considered the long lag between research and practice—the time elapsing between the findings of research and their acceptance and application by engineering practitioners. The engineer is really the person expected to bridge the gap between research and practice. The engineer in the concrete repair field, unfortunately, has very little to bridge. There is no gap to be bridged; on the best there are a few hairline cracks. The complex situation with repair failures clearly illustrates that, when dealing with repair durability problems, wrong assumptions, simplifications, generalizations, and approximations may be very unproductive, and even dangerous. A critical review of

some of the durability issues critical to repair, which is given next, will illustrate this point. Unfortunately, not just for the concrete technology, but for economics of world countries, the situation with corrosion and deterioration of concrete structures, and premature failure of expensive remedial actions has not led to accepted theories, questions answering and reliable practical actions. Following are just a few of such questions: − How chlorides depassivate the reinforcing steel? − What does “chloride threshold?” mean, if anything? − When corrosion occurs, what amount of “corrosion products” are necessary to crack the concrete? − Assuming continuous “good” bond between reinforcement and concrete, where the “critical amount” of corrosion products accumulates? − Does half-cell potential really represent corrosion activity? − Is rapid chloride permeability test reliable? − Can the corrosion inhibitor in concrete repair lead to “extra life” by prolonging the time until the next remedial action or until the replacement? Current different codes of practice and recommendations and research “discoveries” dial with issue of the chloride “threshold” as it is a given, a fact, a “rule of thumb.” And, because of this fiction everything, of course, is being unjustifiably blamed on the “threshold” disregarding the multiplicity of other critical

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factors involved in corrosion of steel in concrete. How many healthy structures were replaced, how many remediation projects were inadequately designed and failed because of the “chloride threshold” myth? But, time after time, research reports and papers are providing analysis and judgments based on “chloride threshold.” It is an unfortunate fact, that in condition evaluation practices, this test leads to conclusions that the corrosion occurs or does not occur at the time of measurement. The half-cell potential, however, often represents the chemistry of the solution in contact with steel, which may not relate to corrosion at all. The test may determine the thermodynamic risk of corrosion, but is not an indication of active corrosion taking place and does not evaluate the kinetics of the electrochemical reaction. The test is also not applicable for carbonation-induced corrosion where clearly defined anodic regions are absent. (Mackechnie & Alexander 2001) Presently, permeability is being evaluated according to the ASTM C 1202, “Rapid Chloride Permeability Test.” But how reliable is this test? There are some indications (Zhang et al. 1991) that the charge passed through the concrete is not directly related to the rate of chloride penetration. On the basis of preceding discussion, it is reasonable to question whether a direct relationship between electric conductivity (charge passed) and chloride permeability, exists. More research in this area is necessary to resolve controversy. A complete reliance in concrete repair on corrosion protection practices used in newly constructed structures, including the use of corrosion inhibiting admixtures, may, therefore, become a part of the problem instead of solution to the problem. What happens when corrosion inhibitor is added to the repair material? The local nature of the repair does not address the whole structure’s corrosion problem if chlorides or carbonation are widespread. Even if the local repair is

adequate with necessary concentration of inhibitor, it can become a clean (noncorroding) cathodic area that stimulates increased corrosion around it. When an existing concrete structure suffers from corrosion of reinforcement and concrete deterioration, it means that steel is depassivated and an essential cathodic reaction is taking place in the structure. Usually, only several areas of such a structure are exhibiting corrosion damage and are being repaired. Replacement of the most intensely anodic regions of the reinforcement with passive steel in the repair areas removes the adventitious form of sacrificial cathodic protection that was applied to the steel in the neighboring non-repaired regions prior to repair. The corrosion process in the structures is not stopped, it is ongoing. It is quite obvious that in repair service life models (which hopefully will be developed) it will be difficult to incorporate the synergetic effects of a number of aggressive elements and mass transport processes simultaneously present in the repaired structure, in addition to any temperature and humidity effects which can be very significant. However, to implement a mathematical model will allow us to learn so many things about the structure which we would never learn otherwise. The benefit of the modeling is the opportunity to gain an in-depth knowledge of the durability-related issues for a given structure.

2

CONCLUSIONS

The researcher has to have his fingers on the “pulse of practice” in order to be able to successfully help “sift the grain from the husk” and make research work more use-oriented. Making the right choices in concrete repair research and practice will help to ensure that the condition of the infrastructure is safe and it is fully operational.

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Performance based rehabilitation of reinforced concrete structures S.L. Matthews BRE Ltd., Garston, Watford, UK Co-ordinator of CONREPNET Thematic Network and Convenor FIB Commission 5: Service Life Aspects

J.R. Morlidge BRE Ltd., Garston, Watford, UK

ABSTRACT: Some informed owners have recognised that there is a requirement for more durable and effective measures for the protection and repair of concrete structures, with a better balance being struck between ensuring longevity (with the attendant impacts on cost, environmental and societal issues etc) and shorter-term financial considerations. Effective extension of life should improve overall sustainability and lower the through-life environmental impact of concrete structures and others assets. To help address these issues, a European funded thematic network CONREPNET was launched in 2003. The network looked at the performance of previously repaired concrete structures, reviewed current industry practices and research activities, examined the concept of a performance-based approach to the protection and repair of concrete structures, developing a vision and a methodology for its application. While the outcomes of the first two areas are of considerable interest by virtue of the findings from three questionnaire surveys, it is in the latter area that the greatest impact may arise and was the main focus of the network. Keywords: Concrete structures, maintenance and inspection, performance-based intervention, concrete protection and repair, EN1504, owner briefing, through-life performance, residual/remaining life cost

1

INTRODUCTION

2

Most concrete structures provide satisfactory performance over an acceptably long service life period. However deterioration processes affect all structures and materials so through-life management is needed. Premature failure of repairs and lack of certainty in the durability and performance of some repaired concrete structures affects Europe and many parts of the world. Accordingly there is a need to achieve more durable repaired concrete structures. To help address these issues, a European Commission sponsored thematic network, known as CONREPNET, was established looking at performance-based remediation of reinforced concrete structures. The network was launched in Madrid in February 2003. Its aim was to improve the performance of repaired concrete structures by looking at past, present and future circumstances. The network facilitated the transfer of information from research to practice, promoted a performancebased approach to the remediation of concrete structures and established related future research and development needs.

CONREPNET ACTIVITIES

The activities involved: • Participation of stakeholders throughout the remediation of concrete structures supply chain. • Cataloging the performance of previously rehabilitated concrete structures. • Examining current industry practice and the stateof-the-art in research in this area. • Mapping RTD activities concerned with deterioration and rehabilitation of concrete structures. • Developing means of evaluating the performance of rehabilitated concrete structures, together with related benchmarks. • Exploring performance-based concepts to improve the delivery of durable and effective preventive and remedial interventions. • Linking widely at national, European and international levels to aid progress on these issues and to disseminate the findings. Information about concrete durability and remediation issues was collected from industry and researchers. Problems and barriers to achieving durable rehabilitated concrete structures were identified,

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which has allowed new avenues for industry and researchers to pursue to be proposed.

3

REVIEW OF PERFORMANCE IN SERVICE AND CURRENT PRACTICE

Three questionnaire surveys were undertaken looking into: 1. The performance of past repairs and other forms of intervention, with information on about 230 structures being obtained and analysed. 2. Contemporary industry practices, with information being obtained from 54 respondents, and 3. Recent research and technical development (RTD) activity, with information on some 66 projects being obtained and analysed.

4

LIMITATIONS OF CURRENT PRACTICE & BARRIERS TO GOOD PRACTICE

Limitations of contemporary practice include difficulties associated with work organisation, diversity and limitations of technical standards relating to materials rather than to the actual/future performance of the ‘repaired’ structure, workmanship and control of execution—both of which are critical to the production of durable interventions. Barriers to achieving good practice and durable interventions and repaired structures include lack of effective inspection and assessment before the intervention method is chosen, a lack of clarity on what constitutes effective inspection and assessment, that many testing criteria are related to the performance of the repair product measured by laboratory tests, rather than on the basis of performance achieved on site, inappropriate specification of concrete protection and repair products and the change in performance of an intervention or repair material on ageing.

5

PERFORMANCE-BASED INTERVENTION

Previous work on performance-based concepts was used to develop strategies, techniques and processes which could deliver more durable and effective protection and repair of concrete structures. These approaches, which have been termed performancebased intervention (PBI). They are concerned with preventive or remedial activities taken to extended service life, using an approach which involves the practice of thinking and working in terms of the end goals rather than specifying the means by which the desired result can be

achieved. PBI implies a proactive approach to structure management and intervention strategies. As such decision-making associated with PBI would take into account not only basic economic considerations, such as whole-life cost issues, but also recognise the social, economic and environmental drivers which now form the wider framework of sustainability related matters. To be effective PBI requires definition of the functions that a concrete protection or repair product has to fulfil in the structure, the so called ‘final works’, together with methods to verify its suitability for the intended use. Identification of Performance Requirements and Performance Indicators will not be an easy task due the lack of precedents and previous experience. PBI methodologies for the evaluation/selection of intervention options could involve the following: • Confirmation of the management philosophy for the asset and the required performance standards. • Identifying appropriate performance indicators. • Establishing the threshold levels for the performance indicators to be used in the design and evaluation of the intervention. • Identifying intervention methods appropriate to the desired result. • Definition of the acceptance testing and verifications required. • Identification of means for monitoring through-life performance. • Comparison of the available intervention methods to establish the most appropriate. 6

CONREPNET PUBLICATIONS

The paper describes the main findings and outcomes of the network, which have been presented in detail in the following two recent publications: Matthews S.L., Sarkkinen M. and Morlidge J.R., 2007, Achieving durable repaired concrete structures: Adopting a performance-based intervention strategy, CONREPNET Project report, IHS BRE Press, Watford, UK. Tilly G.P. and Jacob J., 2007, Concrete repairs: Observations on performance in service and current practice, CONREPNET Project report, IHS BRE Press, Watford, UK. ACKNOWLEDGEMENTS The author, and other members of the project team, gratefully acknowledges the financial support provided by the European Commission for this work and for the valuable contributions made by the project partners and the wider project membership.

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Repair of bridge IB42 over the Limpopo river at Stockpoort border post E.J. Kruger The South African National Roads Agency Ltd., Pretoria, South Africa

W.S. Humphries Nyeleti Consulting (Pty) Ltd., Pretoria, South Africa

ABSTRACT: During February 2000 South Africa experienced devastating floods in the northern and eastern regions of South Africa. The Limpopo River, which forms the border between South Africa and Botswana, was one of the rivers that flooded. After the floods an inspection of the 60 year old bridge at the Stockpoort Border revealed severe structural cracks which clearly threatened the overall structural integrity of the bridge. For this reason the bridge was closed for road traffic. Due to there being no alternative route through the river the border post was also closed which had a negative impact on the local farming and business communities. An assessment of the bridge revealed severe structural cracks in deck beams, splitting pier tops, locked bearings and unstable abutments. A number of innovative cost effective and aesthetically acceptable repair solutions were then priced. The structure was successfully repaired and reopened after being closed for two years.

1 1.1

INTRODUCTION AND BACKGROUND

1.4

February 2000 floods

During February 2000 the Northern and Eastern parts of South Africa experienced an exceptionally wet cyclonic weather pattern that resulted in high rainfall over a large area which resulted in severe floods in most river catchments. The Limpopo River flooded for its full length. At the Stockpoort Border Post river crossing the catchment area of the Limpopo River is 82 570 km2. From available rainfall data a flood peak at the specific river site was determined at 7 900 m3/ sec which corresponded to a flood with an approximate return period of 30 years. 1.2

The team involved quickly came to the conclusion that heavy vehicles could definitely no longer be allowed to use the bridge. After much deliberation it was decided to totally close the bridge for all traffic.

2

The Stockpoort Bridge (IB42) crosses the Limpopo River which forms the northern border of South Africa with Botswana. The bridge was constructed in 1937 as a single lane structure and is located at the Stockpoort Border Post.

3.1

1.3

3.2

Bridge inspections after February 2000 floods

An inspection of the bridge revealed severe structural cracks in the super-structure. These cracks were so severe that there existed a real possibility of total failure of the bridge.

OBJECTIVE OF PAPER

The objective of this paper is to discuss the cost effective and innovative repair methods adopted for the repair of this bridge and further to highlight the lessons learnt through this project. 3

Location of bridge and countries involved

Decision to close the bridge

STRUCTURAL ASSESSMENT Type of structure & original design

The bridge consists of a combination of seven simply supported, reinforced concrete decks which are supported on solid mass concrete piers and reinforced concrete abutments. Structural deficiencies encountered

When the bridge was inspected after the February 2000 floods it was noticed that some flood debris was lodged between the deck and one of the pier tops. According to eyewitness accounts of border post personnel the flood had overtopped the bridge deck by 100 to 200 mm.

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A major structural crack was observed at the end of the main spine beam, directly above the bearing. On examination of the “as built” drawings of the bridge it was found that the longitudinal beam reinforcement did not extend to the end of the beams but was bent upwards at roughly the centre line of the bearings. This also confirmed the observations made on site during the subsequent repairs. In addition it was observed that the deck top slabs have butted together as certain roadway joints had totally closed and not reopened. The reason for this could also have been that the abutments rotated forward in the flood conditions and were in fact propped by the decks and prevented from overturning. Calculations confirmed that the abutments were in fact unstable under flood drawdown conditions. From the observations of the bearings it was evident that the sliding bearings had locked and their intended sliding function was no longer operative, or they required large forces to slide. This locking of the sliding surfaces caused the bridge to act as a continuous structure in the longitudinal direction. 3.3

Repair options considered and implemented

The following different structural deficiencies had to be addressed by the repair options: • inadequate sliding of bearings • the splitting of pier and abutment tops • the incorrect detailing of reinforcing anchorage at beam ends • the abutment instability. Two basic options were considered. Option 1: This first option was to demolish the beam ends and provide adequate reinforcing anchorage and to install new bearings. Option 2: This involved widening and strengthening the existing pier tops so as to allow new bearings to be placed further into the span so that the main bending reinforcing was anchored beyond the end of the new bearings. As this second option was the most innovative as well as the most cost effective the consulting engineers were instructed to proceed with this design.

Aesthetic appearance of repaired bridge

By innovation it was possible to economically repair the bridge in such a way as to restore it to its original intended structural behavior. This was done in such a way that also did not affect the overall aesthetic appearance. 4

LESSONS LEARNT

• One aspect that design engineers need to take into account is adequate anchorage of bending reinforcement beams beyond bearings. Inadequate anchorage can lead to the failure types observed on this bridge. • When bridge repairs are undertaken it is always necessary to totally understand the reasons for failure and to make sure any repair solutions address the actual structural problems that precipitated the failure of a structural system. • Often in projects the community is not adequately informed as to why certain decisions such as the closure of the bridge are taken. 5

Previous repairs undertaken

From the detail inspections it was obvious that similar failures had occurred in the past as it was observed that one of the beam ends had been repaired on a previous occasion. In addition, anti-splitting dowels had been inserted in the abutments and pier tops to counteract splitting of the un-reinforced mass concrete at some positions. 3.4

3.5

CONCLUSION

The following important points are highlighted from the paper: • The importance to act immediately when severe problems are observed, especially when the safety of the public is at stake. • One should always consult and endeavour to maximise communication with local communities. • Not to jump to early conclusions if failures are observed—the easiest solution would have been to demolish the total structure and rebuild it (there was certainly strong motivation for that). • Address the structure holistically. • To address all aspects of the structure, whether directly linked to the problem or not. • Accept the challenge to come up with innovative and cost effective solutions. Finally, when dealing with a historic old bridge structure, such as the Stockpoort Bridge, one should always endeavour to preserve their inherent character and beauty. This was achieved with great success in the repair of the Stockpoort Bridge. ACKNOWLEDGEMENTS The authors of this paper whish to acknowledge the positive part played by the governments of South Africa and Botswana in the funding of the repair of the bridge. The part played by South African National Roads Agency Limited as implementing agent is also acknowledged.

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The aging of concrete sewers: Replace or rehabilitate V.A. da Silva Afri-Coast Engineers SA (Pty) Ltd., Port Elizabeth, South Africa

A.M. Goyns PIPES CC, Tshwane, South Africa

ABSTRACT: Until fairly recently, 40 years was considered a realist design life for a sewer. The limitation of this concept is highlighted by the collapses or near collapses of several concrete sewers in the Nelson Mandela Bay Metropolitan area over the last few years. During 2004 the municipality undertook a study of the maintenance backlog of its sewers. The authors were commissioned to undertake the condition assessment of several sewers that were where known to be critical, predict their remaining life and propose and implement the necessary remedial action so that these sewers would provide the required service life. This paper describes technical and practical issues of assessment and rehabilitation that arose with two of these sewers in Govan Mbeki Avenue, Port Elizabeth. Keywords:

1

Biogenic acid attack, condition assessment, rehabilitation, replacement, sewer corrosion.

4

INTRODUCTION

Sewers are gravity pipelines that are usually placed at depth below all the other services in urban areas to ensure sufficient gradient to maintain self cleansing velocities. Any problems with a sewer will also disrupt the other services above it. The limitation of applying a 40 year design life for sewers has been highlighted over the past three years by the collapses or near collapses of several concrete sewers in the Nelson Mandela Bay Metropolitan (NMBM) area. 2

Concrete sewers are subjected to the same external conditions as any other buried structure. Under certain conditions the sewers may also be subjected to internal corrosion above the waterline from sulphuric acid (H2SO4) formed as a result of bacterial oxidation of hydrogen sulphide (H2S) that is generated below the waterline and is released into the sewer atmosphere, or acid attack below the waterline due to acidic effluent. The manifestation of these conditions and their causes are different.

MAINTENANCE BACKLOG STUDY 5

During 2004 the Nelson Mandela Bay Municipality (NMBM), undertook a maintenance backlog study of its sewers. The first step of this study was to develop a model to identify and prioritize the sewers that needed a condition assessment to determine maintenance or rehabilitation needs. 3

DURABILITY OF SEWERS

REQUIREMENTS FOR A SEWER

Pipe material durability is the main factor contributing to the service life of a sewer, but not the only one. Apart from product strength, water-tightness and robustness it should be possible to rehabilitate the original pipe at the end of its design life. Sewers must be analysed as part of a pipe/soil system. Leaks from the sewer to the adjacent soil must also be addressed.

CONDITION ASSESSMENT

Condition assessment is generally taken to mean the gathering of visual data from inside a pipeline using a CCTV camera, processing the data to list and classify the defects to determine the physical condition of the pipes and joints and the need, if any for maintenance and rehabilitation. It actually has a much wider scope and covers the gathering of data from as many sources as possible With this information the performance of pipes, joints and the surrounding soil can be determined both qualitatively and quantitatively. CCTV inspections do not provide a complete assessment. It is necessary to do physical measurements to check the output and in effect calibrate the output. The actual corrosion in a sewer can be determined by cutting a window into a section of sewer and taking measurements so that the actual material losses can be determined.

281

6

GOVAN MBEKI AVENUE SEWERS

The project involved the condition assessment, and the subsequent design and specification for the rehabilitation of 570 m of 450 mm and 560 m of 900 mm nominal diameter concrete sewers located a few meters apart under Govan Mbeki Avenue in the city of Port Elizabeth that was being redeveloped as a pedestrian mall leading to the City Hall. As these sewers were deep and run beneath the heavily trafficked section of Govan Mbeki Avenue, non-disruptive methods of sewer rehabilitation had to be considered. A further consideration was that the redevelopment project was being undertaken simultaneously with the sewer rehabilitation. These factors ruled out the excavation of open trenches. The condition of the two sewers was assessed by via a CCTV inspection, cutting a window through each and establishing certain effluent properties. Both sewers were in a poor condition, but the nature of deterioration was very different. The 900 mm sewer had been operating for about 25 years. The pipes were precast concrete and were made using dolomitic aggregate with a sacrificial layer. The CCTV footage showed that the pipes had a relatively smooth internal finish. However on close viewing a sill just above the average flow level could be seen. The sill can be clearly seen in Figure 1. The relatively smooth surface of the pipes shown in Figure 2 illustrates how both the binder and aggregate were corroded, typical of what happens in a sewer when dolomitic concrete is subjected to H2SO4 attack. Several trenchless rehabilitation techniques were investigated. When all factors were taken into considerations the final solution chosen was a spirally wound system consisting of a steel reinforced HDPE

Figure 1.

Photo of 900 mm sewer taken through window.

Figure 2.

Photo of 450 mm sewer taken through window.

profile that provides a continuous liner that fits closely to the host pipe bore. Once in place the annular space between the host pipe and liner are grouted to prevent water paths forming. The CCTV footage of the 450 mm sewer showed a different assessment from that inside the 900 mm diameter sewer, as shown in Figure 2. This was cast in place, in two sections with horizontal construction joints at about 60° either side of the crown. The sewer was made using a siliceous aggregate. This photo shows that the aggregate was exposed and the mortar eroded away, typical of acidic effluent attack. Remedial action was necessary to re-establish the water tightness of this sewer. In this case a spirally wound PVC liner that fitted tightly into the host pipe could be used as the sewer section had remained circular.

7

CONCLUDING REMARKS

These case studies high light the fact that there is no substitute for following sound design principles when designing new sewers or investigating the rehabilitation of existing ones. On the Govan Mbeki sewers the condition assessment was fully justified as it indicated the urgency of rehabilitating both sewers to ensure that there would be no future problems due to them leaking or collapsing and this then resulting in surface settlement or sink holes forming in this prestige inner city development. When considering the rehabilitation of sewers it is essential to establish the nature of the problems, their nature, extent, severity and their cause. If this is not done the symptom could be treated rather that the cause.

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Concrete retrofitment solutions utilized at the Van der Kloof Dam spillway bridge P.D. Ronné BKS (Pty) Ltd., Engineering and Management, South Africa

R.S. Maliehe Ibhayi Contracting CC, South Africa

ABSTRACT: Concrete was used as the material of choice to retrofit the bridge over the spillway of the Van der Kloof Dam. The 28-year old bridge had been closed to public traffic after reports were received of “rollercoaster” sensations experienced by road users. Retrofitment was required to remedy the structural concerns, a combination of excessive deflections and vibrations experienced under traffic loading due to low transverse stiffness of the structural system. Specialist concretes utilized in the remedial strategies included self-compacting concrete and the construction of a concrete pavement (or roadway). This paper details the initial structural concerns and records in detail the construction sequence, materials used, construction techniques employed in constructing the structural retrofitment actions utilizing the specialized concretes.

1

BACKGROUND

The Van der Kloof dam is South Africa’s second largest composite concrete arch dam and currently the highest dam wall at 107 m. The bridge over the spillway had been closed to traffic after reports of “rollercoaster” sensations by road users. A combination of specialized structural analyses and diagnostic testing techniques were employed and determined that the bridge deck exhibited significantly low transverse stiffness. Structural retrofitment with new reinforced concrete members, including the installation of internal transverse beams located at third points, as well as the construction of a fully-bonded concrete pavement, were constructed to significantly increase the transverse stiffness of the bridge superstructure.

2 2.1

RETROFITMENT SOLUTIONS Construction of new transverse beams

Two internal reinforced concrete transverse beams were installed at third points using self-compacting concrete placed through holes cored through the existing deck slab. This approach avoided extensive demolition actions and ensured that the existing deck could be used to support all temporary access measures. Bond to the existing longitudinal beams was ensured by roughening the bond surfaces and installing the compression steel fastened in holes drilled through the beams using rapid curing chemical injection mortar. The tensile reinforcement was placed at a level below the soffit line of the existing longitudinal

beams in order to preserve tensile continuity thereby avoiding large lap lengths and alignment issues. The shear reinforcement ensured that the compression and tensile steel were adequately connected in each bay. A highly adjustable, steel shutter system was developed to account for the variable alignment of precast beams, an effect of the original construction methodology. The formwork was fixed to the existing beam surfaces and double sealed to prevent potential spillage of flowable concrete into the river below. A viscosity-controlled, self-compacting concrete was developed to be introduced through a 125 mm diameter opening, fall 1,5 m and achieve suitable flow within the formwork system. The concrete was mixed using a pan mixer on site using commerciallysourced aggregates and routinely performance tested (for flow distance and time). It is believed that this is the first time in South Africa that viscosity-controlled self compacting concrete has been routinely mixed on site by a main contractor. The steel shutters were removed after four days and the beam surfaces were coated with a proprietary curing compound. Hereafter the superstructure was jacked and the elastomeric bearing pads replaced, as well as all patch repairs and coatings were completed, prior to the removal and relocation of the temporary access system. 2.2

Construction of concrete pavement

The bond interface was prepared using an asphalt milling machine to achieve a surface roughness with amplitude of 10 mm showing exposed coarse aggregate particles.

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The formwork and embedded dowels were installed to final design levels that allowed a uniform crossfall of 2% towards the downstream side and a guaranteed minimum 100 mm thickness for the concrete pavement. The dowels within the pavement panel were chemically anchored to the longitudinal beams while additional anti-curling dowels were chemically anchored to the deck slab around the edges of the panels. The concrete pavement was reinforced using 12 mm diameter high yield stress steel (450 MPa) bars at 100 mm centres each way, with the transversely oriented steel placed on the top layer. The close spacing of the mesh was chosen as a crack limitation measure. The concrete mix contained aggregates with proven abrasive resistance properties and was further designed to attain minimum flexural and compressive strengths. Owing to elevated temperatures (in excess of 30 degrees Celsius) and high local wind speeds, the contractor elected to place concrete in the early hours of the morning and have concrete placement operations complete by 10h00. The roughened concrete surface was cleaned using 300-bar water jetting and further sprayed with water (mainly overnight) to maintain a saturated state. Approximately 45 minutes before concrete operations started, all standing water was removed and the concrete surface attained the saturated surface dry condition. The mixed concrete was discharged and transported to the placement location using two dumpers. The concrete was moved within the panel using hand methods and compacted using poker vibrators. A single pass of an aluminium straight edge was followed by additional compaction using a motorized, vibrating screed beam. A double pass of an aluminium straight edge followed to attain a “smooth and shiny, yet dimpled” appearance of the concrete surface, in preparation for the final finishing techniques. Once the concrete had stiffened sufficiently, an aluminium bull float was dragged in the orthogonal direction to reduce any waves induced by the previous longitudinally-orientated operations. A final waiting period followed, which culminated in a judgement call when the concrete had stiffened sufficiently to allow surfacing texturing operations to be completed, but was still fresh enough to ensure that the resulting texturing depth of 3 mm could be attained. This judgement call to proceed is undoubtedly the key to achieve consistency in the final appearance of the concrete pavement and should only be attempted by a suitably experienced technically competent person. Wet hessian was dragged back and forth in the direction of traffic (longitudinally). Thereafter a grooving comb applied a particular pattern in the orthogonal,

transverse direction with the aid of a straight edge. The straight edge assured that the grooved pattern, which were induced by 3 mm wide spring steel elements spaced variably between 15 mm and 40 mm apart, had consistent depth and was correctly aligned. The burlap-dragged and grooved concrete surface formed the final trafficked surface of the bridge over the spillway, as well as reduced noise generated by vehicle traffic and aided surface drainage. Curing was achieved by the application of proprietary solvent-based curing compound by airless spray gun, followed by covering with a water saturated underfelt layer for 3 days. 2.3

Other rehabilitation activities

After the construction of the transverse beams for each span, the superstructure was jacked using a threestage displacement-controlled procedure to replace the elastomeric bearing pads, as well as supporting epoxy mortar plinths. Thereafter the temporary access platform was removed and relocated to the next available span. The rehabilitation process was repeated at the newly accessed location. The concrete pavement construction followed after the completion of several minor remedial activities to the top surfaces of the bridge deck. The replacement of bridge expansion joint system followed utilizing concrete nosings and silicone sealant realigned to the final levels of the concrete pavement. The asphalt surfacing was replaced for the remaining roadway portions over the dam wall and installation of new roadstuds and roadmarkings was completed before the bridge was reopened to public traffic. 3

CONCLUSIONS

This rehabilitation project presented a unique opportunity to utilize concrete in various specialized applications. Structural retrofitment was achieved by the construction of transverse beams, using selfcompacting concrete introduced through holes cored through the existing deck slab, without extensive demolition of the existing structural members. This encouraged the use of a simplified temporary access system that was fully supported by the existing superstructure. The access system also permitted suitable access to allow the replacement of elastomeric bridge bearings. Thereafter, a fully-bonded concrete pavement was constructed, along with a new expansion joint system. Adverse ambient environmental conditions were counteracted by the careful planning of the concreting operations and in particular pride in the work of the construction team.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Rehabilitation of major bridges over the spillways of three of South Africa’s largest dams P.D. Ronné & A.A. Newmark BKS (Pty) Ltd., Engineering and Management, South Africa

E.J. Viljoen Department of Water Affairs and Forestry, South Africa

ABSTRACT: A unique combination of concrete repair, rehabilitation and retrofitment techniques were utilized on three major bridges traversing the spillways of the Gariep, Van der Kloof and Pongolapoort Dams. State-of-the-art vibration-based dynamic tests were utilized together with visual inspections and standard structural analysis techniques in order to correctly identify the root cause(s) of the deterioration and post-retrofitment to assess the contribution of the remedial actions. The remedial work comprised of structural retrofitment to the superstructures of the Gariep and Van der Kloof dams, concrete repairs and the application of protective coatings to the superstructure of the Pongolapoort dam, the replacement of bridge bearings, the replacement of bridge expansion joints, the replacement of asphalt surfacing and ancillary repair items. Challenging structural retrofitment activities performed to the Van der Kloof bridge superstructures included the installation of new transverse beams and construction of a fully-bonded concrete pavement (or roadway), while an epoxy bonded steel plate strengthening system was used at the Gariep bridge. Physical constraints posed by the respective double curvature concrete arch dam walls and the envisaged scope of work at each structure resulted in the development of novel temporary access systems.

1

1.1

CONDITION ASSESSMENT OF STRUCTURES Observed defects

At Gariep, flexural crack patterns were observed on the longitudinal beams, particularly the outer two beams on each side of the superstructure. In addition the varying bearing support conditions, i.e. overstressed outer bearings and central bearings lifted clear of the plinths, were observed for the elastomeric bearings. These defects were attributed to transverse hogging induced by eccentrically prestressed transverse beams and insufficient distribution steel throughout the web of the longitudinal beams that exacerbated the cracking in the beams. At Van der Kloof, road users reported experiences of “rollercoaster” sensations and the road had been closed to public traffic. Significant longitudinallyorientated cracking was observed in the deck slab, midway between the longitudinal beams. The existing superstructure configuration utilized two end transverse beams in each span and relied on the deck slab to provide transverse stiffness. Inadequate transverse stiffness of the superstructure was determined as the root cause of the observed defects. At Pongolapoort, the superstructure elements were generally in a sound condition, while the quality of finishing was generally poor with spalling of cantilevered deck ends due to low cover over the reinforcement and

isolated portions of the original formwork were not removed. The bridge expansion joints were filled with accumulated debris resulting in the restriction of the movement capacity of the bridge and spalling of longitudinal beam ends. 2

2.1

DESIGN AND CONSTRUCTION STRATEGIES Provision of access

The greatest challenge facing the contractors was the provision of access to the bridges over the spillways. The physical constraints posed by the structures were amongst others the height above ground level, inability to found on the curved spillway profile below and occupational safety considerations in the presence of local valley wind systems. At the Gariep and Van der Kloof Dams, the platforms were constituted from conventional scaffold components, i.e. typical steel ladder beams and steel scaffold boards, supported by steel bearing plates attached via dual steel wire cables. The supporting cables were placed at each of the four corners and fed through cored holes in the concrete deck slab and anchored to purpose designed steel anchor frames above the bridge deck. The platforms were erected utilizing common rigging principles with the aid of a crane truck with a 6 ton capacity to lower the

285

platforms. Once lowered each corner of the individual segments were winched into place using proprietary cable puller devices located on the anchor frames. Thereafter a fixed length safety cable was secured, as well as independent safety lines, before general access to construction personnel was permitted. All construction personnel used safety harnesses attached to the independent safety lines. A rolling access gantry, similar to a crane girder, with an additional access tower and suspended walkway of limited width was used at Pongolapoort. The access frame was located on steel rails attached to reinforced concrete traffic barriers of the bridge and was sufficiently mobile to allow workmen to relocate the gantry structure by hand to the successive work areas. During relocation between spans, the unmanned walkway was rotated clear of the substructure elements, using water filled tanks as a counter weight. Importantly, since no alternative traffic route exists, this access system permitted the flow of public traffic over the bridge. High visibility, maximum height limitation signage and barriers were installed at each end of the bridge to protect the overhead portion of the access gantry.

including cleaning and treatment of exposed reinforcement. Since the areas of low cover to reinforcement were limited to the deck cantilever edges, a proprietary cementitious protective coating was applied to the deck cantilever ends and exposed beam surfaces to improve the resistance further weathering and to improve aesthetics.

2.2

A total of 825 m of bridge expansion joints were replaced at the three major bridges to provide sufficient capacity to accommodate temperature movements only. New concrete nosings were cast at Gariep and Van der Kloof. Thereafter the exposed joint faces were prepared by reaming and cutting of 10 mm by 10 mm chamfers to remove concrete laitance and as preparation for the installation of a fuel-resistant silicone sealant. At Pongolapoort, the expansion gaps between the precast beam ends were reinstated by clearing entrapped debris, while multi-directional core stitch drilling was required to create the required temperature movement capacity gap for the cantilevered deck ends. Asphaltic plug-type joints were installed in the roadway, while new concrete nosings with silicone sealant were constructed in the sidewalk.

Structural retrofitment at Gariep Dam

The outer two longitudinal beams on each side of the deck were strengthened using epoxy bonded steel plates, including shear straps bolted and tensioned into the compression flange of the T-beam configuration. Both exposed concrete surfaces and steel elements were painted with proprietary protective coating systems. Post-rehabilitation dynamic assessments indicated significant improvement in the flexural stiffness of the retrofitted structure. 2.3

Structural retrofitment at Van der Kloof Dam

The deck was retrofitted with two internal transverse beams installed at third points, using self-compacting concrete installed through core holes in the existing deck. Every batch of self-compacting concrete was performance tested for both flow distance and flow time to verify viscosity properties of the concrete. Thereafter an additional 100 mm thick concrete pavement, fully bonded to the existing scarified deck, as well as dowelled directly into the longitudinal beams and new transverse beams, was constructed in order to improve the load distribution characteristics of the deck slab. The reduction of free vibrations, improved load distribution and improved transverse stiffness were determined from post-rehabilitation dynamic structural assessments. 2.4

Concrete repairs at Pongolapoort Dam

The isolated areas of spalled concrete were repaired using proprietary cementitious patch repair system,

2.5

Bearing replacement operations

A total of 596 bridge jacking and bearing replacement operations were performed at Gariep and Van der Kloof bridges. A carefully displacement-controlled three phase jacking procedure was followed utilizing 60 ton capacity hydraulic jacks with lockable rings. Thereafter, 68 guide bearings, each with a 10 ton and 40 mm capacity, were installed centrally at both ends of each span of the Gariep and Van der Kloof superstructures.

2.6

2.7

Replacement of bridge expansion joints and reinstatement of joint gaps

Asphalt surfacing and miscellaneous drainage improvements

At Van der Kloof, the asphalt surfacing was replaced with medium continuously graded asphalt in the roadway over the dam wall, excluding concrete pavement constructed on the bridge over the spillway. At Pongolapoort, the asphalt surfacing over the entire length of the dam wall was replaced with SBS modified, medium continuously graded asphalt surfacing. For both dams, the block joints in the dam wall were reinforced using a fiberglass based proprietary asphalt reinforcement system. Local improvement to the superstructure drainage channel systems were also installed.

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Protection of concrete with water repellant agents – What is required to achieve a sufficient penetration depth? A. Johansson, M. Janz, J. Silfwerbrand & J. Trägårdh Swedish Cement and Concrete Research Institute, Stockholm, Sweden

ABSTRACT: Water repellent agents such as silanes have been shown to give a good protection against moisture and chlorides and thereby prolonging the service life for the construction. However, a requirement for a successful impregnation is a sufficient penetration depth. The effective penetration depth of different silanes was measured by cracking water repellent treated concrete plates, spraying water on the cracked surface and measuring the distance from surface to the sharp line between dry and wet concrete. The results are presented as a function of three parameters: time, degree of saturation and porosity. They show how these factors affect the effective penetration depth of the silane. It can be seen that by prolonging the treatment from one to five hours results in doubling the penetration depth something which easily can be achieved for example by using a paste instead of a liquid.

1

INTRODUCTION

Water repellent agents such as silanes and siloxanes have been shown to give a good protection against moisture and chlorides and thereby prolonging the service life for the concrete structure. However, a requirement for a successful impregnation is a sufficient penetration depth defined as the distance from the surface to the sharp line between dry and wet concrete after it has been sprayed with water, see Figure 1. The three most important factors for the penetration depth are time, porosity and degree of saturation. The time referred to is the duration of contact between the water repellent agent and the concrete surface. The porosity and degree of saturation refer to the concrete pore system and the amount of moisture inside the concrete at the time of the impregnation.

2

MATERIALS AND METHOD

The effective penetration depth of three different water repellent agents was investigated as a function of three parameters; time, degree of saturation and porosity. The degree of saturation is represented with the RH (relative humidity) in which the samples were conditioned before the treatment. 300 water repellent treated concrete plates were used in this experiment. In addition to these, four plates were used to measure the sorption coefficient in order to represent the porosity. The concretes were cast in 80 litre boxes and conditioned for three months in 100% RH and 20°C. A total of eighty 75 mm cores were then drilled out of the boxes and cut into 30 mm thick plates. To ensure a uni-dimensional flow when the samples were treated

Figure 1. The effective penetration dept is the distance from the surface (bottom) to the sharp line between wet (dark) and dry (light) concrete. The penetration depth in this picture is 21 mm.

with the water repellent agent the perimeter of the plate was sealed before they were placed inside climate boxes, leaving the up and downside open. The plates were then conditioned for six month in different RH and then treated with the silanes by capillarity from one side. When the desired time had passed the plates were removed from the box, wiped off with paper and then placed horizontal in open air. 3

RESULTS

Figure 2 shows the different penetration profiles of the three tested water repellent agents while Figure 3 illustrates the influence of RH. Based on the results presented an empirical equation is derived which describes how the efficient penetration depth depends on these different factors.

287

z = C1 ⋅ A ⋅ [1 − φ ]⋅ ln (C2 ⋅ t + 1)

(1)

[mm] 16 14 12 10 8 6 4 2 0 0

Table 1.

Molecule size - chemical reactivity w/c = 0.6, 59% RH triethoxy(isooctyl)silane triethoxy(isobutyl)silane triethoxy(propyl)silane

100

.1 .5 1

200 5

300

400

10

50

30

20 15

86.4⋅10−3

4.17

Triethoxy(isobutyl)silane

270⋅10−3

0.167

Triethoxy(isooctyl)silane

930⋅10−3

0.0133

Factor

Before

After

Result (z)

Time (t)

1 hour

5 hour

∼double*

Porosity (A)

0.009**

0.018***

∼double

RH (Φ)

90%

80%

∼double

* Only valid for triethoxy(isooctyl)silane. ** Corresponds to approximately w/c = 0.45. *** Corresponds to approximately w/c = 0.55.

4

5 0 100

200

5

10

30

400

50

CONCLUSIONS

It is important to emphasize that the experiment was conducted on non carbonated concrete and at a temperature of 20°C. The coefficients presented in Table 1 probably need to be revised if those conditions change. Nevertheless, based on this investigation and interpretation the following conclusions can be drawn:

10

.1 .5 1

Triethoxy(propyl)silane

Change

Degree of saturation w/c = 0.8, triethoxy(isooctyl)silane

0

C2 [s−1/2]

[h]

59% RH 75% RH 85% RH 93% RH 97% RH

25

C1 [m3s1/2/kg]

Table 2. Example of how Equation 1 can be used to evaluate the impact of different factors.

[s1/2]

Figure 2. The penetration profiles of three different water repellent agents. The time on the x-axis is the duration of contact for the treatment. The penetration depth is measured four months later.

[mm]

C1 and C2 for different water repellent agents.

[s1/2] [h]

Figure 3. The influence of the conditioning environment around the specimen. The time on the x-axis is the duration of contact for the treatment. The penetration depth is measured four months later.

where z [m] is the penetration depth; t [s] is the time; A [kg/m2s1/2] is the material property, called the sorption coefficient, used in this experiment to characterize the porosity of the concrete. C1 and C2 are empirically determined material parameters related to the water repellent agent used and they are presented in Table 1. The degree of saturation is represented numerically with the factor: [1 − φ] where φ = v/vs and v is the vapor content of air [kg/m3]. Index s means saturated. Concrete has a huge variation in material properties and the environmental conditions at the time for the treatment can change quickly. It is difficult to derive an equation that can predict the penetration depth of a water repellent agent at every situation. Equation 1 should therefore just be used to illustrate how different factors affect the result and use this as a tool when choosing a suitable surface treatment. Table 2 shows some examples on how this can be done.

1. Duration of contact, degree of saturation and porosity have all a major influence on the outcome of a water repellent treatment. 2. The equation presented describes qualitively how much these factors affect the effective penetration depth of the water repellent agent. A longer time for the treatment which can be achieved by using a paste instead of a liquid can increase the penetration depth significantly. 3. The chemical reactivity and the size of the water repellent agent affect the effective penetration depth. A silane with a small alkyl group enters the concrete faster than a silane with a big one. It thereby results in a higher effective penetration depth. A high chemical reactivity of the silane has the opposite effect. When the polymerisation starts the molecule size increases and the uptake is slowed down.

ACKNOWLEDGEMENT Financial support from the Swedish Research Council for Environment, Agriculture Sciences and Spatial Planning is gratefully acknowledged.

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Accelerated transport of corrosion inhibitors as complementary methodology for electrochemical chlorides extraction method M. Sánchez & C. Alonso Institute of Construction “Eduardo Torroja” (IETcc-CSIC), Madrid, Spain

ABSTRACT: In the last decades there has been an important development of non-conventional repairing techniques known as Electrochemical Maintenance Methods: Cathodic Protection (CP), Electrochemical Chloride Extraction (CE) and Electrochemical Realkalisation (RE). In contrast with conventional repair by removing the concrete that causes the corrosion, the electrochemical methods aim at taking away the aggressive conditions but maintaining the concrete cover. In this work, transport of corrosion inhibitors by migration simultaneously to a Chloride Extraction treatment is proposed. The inhibitor migration from the concrete surface to the reinforcement is expected together with the electrochemical removal of chloride. After the treatment the inhibitor has to be dissolved at the concrete pores and acts improving the repassivation of the reinforcement, increasing in this way the efficiency of the CE treatment. Two organic-based corrosion inhibitors have been tested: 1) mixed commercial inhibitor based on aminoalcohol compounds and 2) ethanolamine. It has been found that the migration of the mixed commercial inhibitor simultaneously to the connection of the rebar as cathode for the extraction of chlorides enhances the regeneration of the passive film that protects the rebars from corrosion.

1

RESUMEN

The present work has been focused on evaluating the viability of accelerating the penetration of an organicbased inhibitor through the concrete by migration. The analysis of synergic effects associated with the simultaneous application of an electrochemical chloride extraction treatment and transport by migration of the organic-based corrosion inhibitors has been considered. The studies have been performed on mortar samples made with: CEM I 42.5R/SR, resistant to sulphates and with low content of alkalis and aluminates (<4%), normalized CEN UNE-EN 196-1:1196 sand and deionized water. Water/cement ratio of 0.5 and sand/cement ratio of 3/1 have been employed. For the Electrochemical Chlorides Extraction treatment, sodium chloride was introduced in the mix, added to the water in 1% wt of cement. The application of two different corrosion inhibitors containing aminoalcohol compounds has been considered: 1. Ethanolamine (EA): a 0.2 M solution has been applied. 2. Mixed commercial inhibitor (CI) with an organic part (aminoalcohol-based) and an inorganic part (phosphate-based). A 9% vol. solution has been applied. The active group of each inhibitor which transport is expected to be accelerated by migration is NH3+. Two different has been developed: 1. Transport of organic-based corrosion inhibitors by migration.

These tests have been developed with a typical migration cell with two watertight compartments. The anode is placed into the inhibitor solution (anolyte), and the transport by migration of cations of the inhibitor is forced to the cathode, located in other compartment filled with water (catholyte). A voltage of 12 V is established between both electrodes. To evaluate the transport of the inhibitor, samples were taken periodically from the catholyte during the connection of the electric field. The identification of the active group of the inhibitors is made by Infrared Spectroscopy (IR) of the samples of the catholyte. 2. Cathodic connection of rebars: Migration of organic-based corrosion inhibitors and Electrochemical Chlorides Extraction. Mortar samples of 2 × 5.5 × 8 cm have been used for the application of the treatments. In each sample two steel rebars were located. A rod of graphite was embedded that acts as counter-electrode during the electrochemical measurements. An electric field of 12 V is connected between the rebar acting as cathode and an external electrode of titanium oxide acting as anode. The anode is placed into the inhibitor solution and thus, the transport of the cationic active group of the inhibitor to the rebar is enhanced. The connection is maintained for 5 days. The corrosion response of the rebars was analyzed before and after the application of the different treatments by measurements of the corrosion potential (Ecorr). Also the Polarization Resistance (RP) measurements were periodically obtained for the further estimation of the corrosion current density (icorr).

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At the end of the tests, the samples were broken and a visual observation of the rebars was made in order to confirm the superficial state of the rebar expected from the electrochemical measurements. Below are described the main results obtained in present work: 1. Organic-based corrosion inhibitors migration through hardened mortar pores. The transport by migration of the two organic-based corrosion inhibitors has been confirmed by IR analysis, identifying the characteristic transporter group at 1650 cm–1 corresponding with the amine group. In this way, the transport of amine group through the mortar pores until the catholyte should be confirmed by the detection of this characteristic band. When EA was transported, the identification of this band was after 2 days of test; however, when the commercial mixed inhibitor was transported this band appeared after 4 days of testing. These results are very interesting in relation with the possibility of applying corrosion inhibitors on hardened reinforced concrete, accelerating its penetration by migration under the action of an electric field. When the inhibitor reaches the rebar surface it could act to improve the resistance against corrosion. 2. Effect of the electric field on the superficial state of the reinforcements. The connection of a rebar as a cathode during the application of an electric field enhances the electrochemical reactions of reduction of water and iron oxides. Thus, the rebar surface changes as a consequence of the connection to the electric field. Then, it is important to study the electrochemical behaviour of the rebars after the treatment to guarantee that a protective situation against corrosion remains. The effect that the application of an electric field has on the superficial state of a rebar has been studied by application of migration tests with water on samples without chlorides. The arrangement employed has been similar to an extraction chlorides test: connection of the rebar as cathode and connection of an external anode of titanium oxide. After the electric field finished, very cathodic potentials (–0.8 V vs SCE) and quite high corrosion current values are registered that should indicate an active corrosion processes although no aggressive condition is present in the mortar. However, these corrosion current values should be associated with lost of material employed in the regeneration of the new passive film since the visual inspection of rebar after the test did not show rests of oxides. The new passive layer once developed protects the rebar.

3. Effect of the application of the inhibitors on the electrochemical response of the rebar The possibility of transporting an organic-based corrosion inhibitor by migration under the effect of the electric current has been considered. It is expected that the active group of the inhibitor reaches the rebar surface and enhances the formation of the passive film. The application of organic-based corrosion inhibitors by migration under the presence of an electric field promotes similar polarizations of the rebar to the case in which the electric field was connected without inhibitor. The cathodic polarizations of the rebars are related in all cases with an important increase of icorr values. Nevertheless, when the transport of the mixed inhibitor CI is considered, the time in which a new situation of protection against corrosion is reached is reduced. 4. Chloride extraction with simultaneous transport by migration of organic-based inhibitors The simultaneous application of an organic inhibitor has also been considered, expecting the transport of the cation of the inhibitor at the same time that the removal of chlorides from concrete. In this way, two positive actions should be promoted: decrease of the chloride content in the concrete and presence of the corrosion inhibitor at the level of the rebar, improving its protection against corrosion. In a similar way to the case of application of corrosion inhibitors by migration, the connection of the rebar as cathode to make a chloride extraction simultaneously with the inhibitor migration promotes important polarizations of rebars when the disconnection takes place. High values of icorr after the disconnection of the treatment for chlorides extraction simultaneously to inhibitor migration were registered in similar way than the observed in the case of migration of inhibitors in absence of chlorides. However, the effect of the inhibitor can be observed in the rate of recuperation of corrosion current density values. The application of the inhibitor IC promotes a faster reduction of icorr values that promotes less loses of material. The possibility to accelerate the transport of organic-based corrosion inhibitors by migration under the presence of an electric field can be concluded. The transport by migration of a commercial mixed inhibitor (IC) can be recommended as complementary technique for repair a chlorides-contaminated mortar. A reduction in time of regeneration of the passivity is obtained.

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Crack repair in concrete using biodeposition N. De Belie & W. De Muynck Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Gent, Belgium

ABSTRACT: The ability of certain bacteria to promote the precipitation of calcium carbonate, has been used advantageously for consolidation of concrete and stone or for the densification of sandy soils. Previous research has illustrated a reduction of the capillary permeable porosity and an increased resistance to damage processes such as chloride ingress and carbonation by this biodeposition procedure. In order to explore the crack healing potential of a biodeposition treatment, standardised cracks of 0.3 mm were produced in concrete specimens by introducing thin copper plates in fresh concrete and removing them after 1 day, or by performing splitting tensile tests on concrete cores wrapped in fibre reinforced polymer sheets. The use of pure bacteria cultures as in the previously developed biodeposition procedure, did not result in sufficient calcium carbonate precipitation to completely bridge the cracks. Therefore, the bacteria were protected in a silica sol, resulting in the formation of a bioceramic material (sol-gel or biocer) which was able to bridge the cracks completely. The crack healing potential was illustrated by microscopic evaluation, ultrasound transmission measurements and low pressure water permeability tests. The treatment of cracks with the bacteria incorporated in the sol-gel resulted in a large reduction of the water permeability.

1 1.1

INTRODUCTION General

In concrete, cracking is a common phenomenon due to the relatively low tensile strength. Without immediate and proper treatment, cracks tend to expand further and eventually require costly repair. If microcracks grow and reach the reinforcement, not only the concrete itself may be attacked. Also the reinforcement may corrode when it is exposed to water and oxygen, and possibly carbon dioxide and chlorides. For crack repair, a variety of techniques is available. Injection with epoxy resin or cement grout are the most popular ones. Traditional repair systems have a number of disadvantageous aspects. Therefore bacterial induced calcium carbonate deposition has been proposed as an alternative and environment friendly crack repair technique. 1.2

Principle of microbiologically induced carbonate precipitation

Ureolytic micro-organisms can induce extracellular precipitation of CaCO3 by degradation of urea into ammonia and carbon dioxide. This increases the pH at the cell surface and promotes the microbial deposition of carbon dioxide as calcium carbonate. In a project carried out in collaboration between the Magnel Laboratory for Concrete Research and the Laboratory for Microbial Ecology and Technology of Ghent University, first the criteria for the selection of calcium precipitating Bacillus sphaericus strains were established. Then, concrete and mortar samples of different

porosity were treated with bacteria in combination with different calcium sources and nutrient solutions. CaCO3 deposition resulted in a decrease of the water absorption ranging from 65 to 90%. As a consequence, the carbonation rate and chloride migration decreased. Similarly, an increased resistance towards freezing and thawing was noticed. The use of calcium precipitating micro-organisms for crack remediation has not yet been investigated in depth, but the possibility was proven by Ramachandran et al. (2001). However, curing for 28 days in a urea-CaCl2 solution was necessary. In the currently presented research the possibility of crack repair through application of micro-organisms, is explored further.

2

MATERIALS AND METHODS

Standardised cracks were made in concrete samples of 160 × 160 × 70 mm by introducing copper plates of 0.3 mm thickness up to a depth of 10 or 20 mm into the fresh concrete. More realistic cracks were obtained by splitting tests on FRP reinforced cylinders. The crack width varied between 0.01 mm and 0.60 mm (mean value: 0.20 mm). Cracks were repaired by traditional methods and by biodeposition treatments as listed below. 1. Traditional methods: epoxy and cement grout The two component epoxy resin was injected into the crack using an injection needle. The two component cement bound mortar (grout) was applied using a spatula. The area near the crack was protected with tape during the application.

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2. Biodeposition treatments Following treatments were carried out: – Medium The samples were immersed for 3 days in an equimolar solution of ureum (20 g/L) and CaCl2.2H2O (49 g/L) or Ca(NO3)2.4H2O (78 g/L).

1.8E-09 1.6E-09 1.4E-09 k (m/s)

1.2E-09

– BS + CaCl2 The samples were immersed for 24 h in a Bacillus sphaericus culture grown overnight. Then they were immersed for 3 days in an equimolar solution of ureum and CaCl2.2H2O.

– Sol-gel + BS + CaCl2 The samples were treated with sol-gel as described above, followed by the BS + CaCl2 treatment. Ultrasound transmission experiments were carried out on concrete prisms of 160 × 160 × 70 mm with a standardized crack of 10 or 20 mm deep. The efficiency of crack repair was investigated by monitoring the water permeability of the specimens. The test method was a slightly modified version of the low pressure water permeability test described by Wang et al. (1997). 3

RESULTS AND DISCUSSION

Almost all tested biodeposition treatments, except the medium and BS + CaCl2 treatments, were able to bridge cracks. With the ultrasound measurements the increase in transmission time due to the fissure could be determined: 0.22 ± 0.10 µs for the “10 mm” fissure and 0.43 ± 0.18 µs for the “20 mm” fissure. The effect of the epoxy, sol-gel and BS + CaCl2 treatments was also evaluated using ultrasound transmission. For the “10 mm” fissure, the epoxy injection and the sol-gel treatment reduced the transmission time with 0.43 ± 0.20 µs and 0.27 ± 0.11 µs respectively. The sol-gel treatment allowed to seal the fissure, whereas the epoxy treatment probably also had a consolidating

8.0E-10 6.0E-10 4.0E-10 2.0E-10 0.0E+00 0.0

– BS + sol-gel + CaCl2 or BS + sol-gel + Ca(NO3 )2 A suspension of bacteria in sol-gel was applied in the cracks using a syringe. This was repeated at 5 min intervals until the entire crack was sealed. Then, the samples were immersed for 3 days in an equimolar solution of ureum and CaCl2.2H2O or Ca(NO3)2.4H2O. – Sol-gel The samples were immersed for 20 min in the silica sol and dried for 24 h at room temperature. This treatment was repeated three times.

1.0E-09

0.1

0.2

0.3

0.4

0.5

0.6

0.7

Crack width (mm) untreated epoxy BS+sol-gel+Ca(NO3)2 sol-gel

BS + CaCl2 grout BS+sol-gel+CaCl2

Figure 1. Water permeability coefficient k vs. crack width for the different crack repair systems.

effect on the concrete near the crack. For the “20 mm” fissure the epoxy, sol-gel and BS + CaCl2 treatments, the decrease in transmission times indicated that neither the epoxy, nor the sol-gel treatment allowed to seal the “20 mm” fissure completely. The BS + CaCl2 treatment on its own was not at all able to bridge the crack. Fig. 1 shows the coefficient of permeability k of the different samples versus the crack width. All treatments resulted in a decrease of permeability when compared to the untreated cracked samples. The grout and BS + CaCl2 treatments were least efficient in improving the watertightness, whereas the epoxy and BS + sol-gel + CaCl2 or BS + solgel + Ca(NO3)2 treatments were most efficient. This shows that with the alternative biodeposition treatments, in which the bacteria are incorporated in a sol-gel and provided with a calcium source, a similar effect can be obtained as with a traditional epoxy injection.

4

CONCLUSION

It was shown that crack repair can be obtained through biodeposition treatments in which a Bacillus sphaericus culture is incorporated in sol-gel. A calcium source such as CaCl2 or Ca(NO3)2 should be provided.

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Cracked concrete repair with epoxy-resin infiltration R. Felicetti DIS, Politecnico di Milano, Milan, Italy

V.H. De Domenico COSIPI Ltd., Milan, Italy

ABSTRACT: A cracked concrete repair technique is presented, based on the infiltration with low-viscosity slow-setting epoxy-resins. In the reference application system, the infiltration is performed by sealing the crack mouth with a hot-melt glue and by injecting the resin under the seal with equally spaced syringes. An alternative resin formulation with modified surface tension has been also tested, which can be applied by simply brushing the open crack mouth. In order to check the effectiveness of this technique, a set of 16 plain concrete beams have been preliminarily cracked up to prescribed levels of residual crack-mouth opening in the range 0.1–0.8 mm. Then, the beams have been repaired by using both the cited application systems. In the second bending tests most of the specimen cracked in a different place and at a higher failure load. This confirms the effectiveness of this repair method in restoring the mechanical continuity of cracked concrete.

1

2 ADOPTED CRACK REPAIR TECHNIQUES

INTRODUCTION

In the last decades an increasing attention has been paid to the durability of concrete structures, to the point that the control of porosity, permeability and crack opening is getting a more stringent requirement than the strength in the selection of the material grade and reinforcement detailing. Nonetheless, in case of failure to comply to crack opening provisions, different methods are available to restore the required durability of a structure. Among them, the injection of cracks with epoxy resin is an interesting option, which allows to “freeze” the inelastic strain localized in the open fractures and to recover the material continuity and tensile strength. In fact, no significant enhancement of the strain capacity can be expected from these very thin layers of relatively deformable polymer. Then, a successful implementation relies on the preliminary stabilization of the factors originating the crack pattern. The basic condition for the practicability of the method is the ability of the repair compound to penetrate deep in the cracks, a feature that can be hardly obtained with the common highly-viscous adhesives. Special polymer formulations are needed, with lowviscosity and slow-setting in the temperature range of interest for practical applications. In this paper the viability of crack infiltration with this kind of products is checked on a set of pre-fractured plain concrete beams. The bending response of the repaired specimens is then regarded as an indicator of the overall performance of the system.

According to normal practice, active injection and gravity-feed are the reference methods for filling a crack with resin in order to provide a structural bond between the adjacent sides and a sealing plug against external agents. Both of them have been applied in the present investigation, though the former technique has been considered as a reference. Active injection of a polymer compound requires a series of entry ports to be glued at regular intervals along the visible cracks and the intermediate stretches to be sealed for preventing any leakage. In the present investigation a fluid resin (230 cps @ 25°C) has then been injected at low-pressure (0.5 ÷ 2.0 bar) by means of an array of syringes actuated by an adjustable number of rubber bands. Beyond the low pressure exerted during the injection process, capillary suction is expected to allow the resin to permeate cracks as thin as a few hundredths of a millimetre. An alternative resin formulation with modified surface tension and improved penetration performance (190 cps @ 25°C) has been also tested on a smaller number of samples. This can be applied by simply brushing the open crack mouth, as in the gravity feed repair technique. It has to be noted that the gravity force actually plays a secondary role in the infiltration process, though repairing a crack extending from a floor or a deck has the significant advantage of the much simpler delivery of the resin to the crack mouth.

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3

EXPERIMENTAL PROGRAMME

In order to check the effectiveness of the two aforementioned repair techniques, 16 plain concrete beams have been preliminarily fractured up to different residual crack mouth openings in a relevant range for practical applications (0.1, 0.2, 0.4 and 0.8 mm, a set of four specimens for each case). The specimens are 80 × 160 × 480 mm concrete prisms (fcm,cube = 60 N/mm2), simply supported on a 400 mm span and subjected to a central load (Fig. 1). The tests have been conducted beyond the load peak by controlling the average strain at the lower cord so to reach the target residual opening after the load removal. The fairly good concrete grade lead to relatively smooth fractured surface, with mostly split coarse aggregate (rounded siliceous gravel—size = 12 mm) and a reduced surface available for bonding. The following step consisted in repairing the cracked specimens by means of the two cited resin infiltration techniques. After the required curing

Figure 1. Experimental setup for strain-controlled threepoint bending tests (span = 400 mm).

30

load (kN)

test 0.4b 20

repaired beam

(b)

10

preliminary cracking 0 0.0

0.1

0.2

0.3

crack opening (mm) Figure 2. Couples of load-crack opening curves confirming the recovery of stiffness and the general improvement of load capacity of the repaired beams.

period (at least a couple of days), the beams are tested for a second time and their response and crack path are compared to the former results. 4

CONCLUSIONS AND OUTLOOK

In this paper the viability of cracked concrete repair by means of two resin infiltration techniques has been checked (low-pressure injection and crack mouth brushing). The results obtained in two consecutive series of bending tests on pristine and repaired plain-concrete beams allowed to draw the following conclusions: − Despite the relatively high concrete grade, in almost all of the tests, the repaired specimens exhibited an improved load capacity and the new crack started propagating from a different point than the former one. − The results have been partly influenced by some surface shrinkage cracks, which may have reduced the bending response of the specimens in the preliminary cracking test. Once the weakest cross section has been repaired with resin infiltration, further cracks are likely to switch to other parts of the beam intrados, revealing an enhanced bending capacity. − The resin proved to be able to penetrate cracks as thin as 0.01 mm, leaving unglued just the tip zone. This un-bonded portion showed to have a sizeable influence for relatively thin repaired cracks (say up to 0.4 mm maximum aperture), though this effect may have been emphasized by the load removal and crack closure prior to infiltration. − Based on the performed tests, the low-pressure resin injection and the gravity feed technique seam to lead to comparable results. These first encouraging results call for a thorough investigation of a series of still open issues concerning the quality control during the application and the actual improvement of the structural durability. Different Non Destructive Testing techniques are available to monitor the crack infiltration and the following polymerization process. Ultrasonic methods are presently the most promising support for quality control during crack repair operations. Another important aspect to be deepened concerns the rebar protection provided by the repaired cover concrete, and then its permeability to water and gases. The effect of sustained or repeated service loads is also to be taken into account with regard to this latter point. Finally, the influence of the injected resin on the bond of the rebars crossing the repaired crack is of interest as far as the serviceability behaviour of the refurbished structure is at issue. The cited aspects are the object of the future research activity which is being planned by the authors on the topic of cracked concrete repair with resin infiltration.

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Concrete repair with ultra ductile micro-mesh reinforced abrasive resistant and impervious mortar C. Flohrer & M. Tschötschel HOCHTIEF Construction AG, Mörfelden-Walldorf, Germany

S. Hauser DUCON GmbH, Mörfelden-Walldorf, Germany

ABSTRACT: For rehabilitation of industrial floors or bridge deck overlays in most cases cost effective and high potential repair materials are needed. The application of ultra ductile micro-mesh reinforced abrasive resistant and impervious high strength mortar has the advantage to work with thin layers (2–5 cm) of this repair material. The high ductility of the repair system will be achieved by using micro-reinforcement mats with slurry infiltration. The damaged surface will be left in place and doesn’t have to be demolished and disposed because of crack bridging properties of the micro-mesh reinforced mortar. The ultra ductile material is also used for the repair of waste areas or extremely loaded floors with impervious overlays. Besides having high compressive and flexural strength, it is an extremely ductile material, expansion joints are not necessary. Also for strengthening of columns, increasing resistance against blast or earthquake, the micro-mesh reinforced mortar will be applicable.

1 1.1

Table 1.

MATERIAL Composition and material characteristics

DUCON® stands for DUctile CONcrete and represents the combination of a high-performance or ultra-highperformance concrete and a microreinforcement from steel wire meshes (Fig. 1), developed by Hauser. Table 1 shows the most important material characteristics. 1.2

Production

Production is based on the placement of the prefabricated micro-reinforcement and the infiltration of the concrete slurry. Between 1 Vol.-% and 10 Vol.-% of microreinforcement are embedded, typically, 6 Vol.-%. Due to the fine mesh size of the microreinforcement, the crack width can be reduced to values of w < 0,1 mm which qualifies DUCON as impervious material. Thus, DUCON can be executed with a small concrete cover of a few millimetres compared to 20 to 50 mm of standard reinforced concrete.

Figure 1. Self-compacting high-strength concrete + Microreinforcement.

Material characteristics.

Type

Value

Compressive strength Flexural strength Tensile strength Shear strength Modulus of elasticity Thickness Ductility factor

90–180 N/mm2 (cube) 16–75 N/mm2 9–20 N/mm2 3–16 N/mm2 >35.000 N/mm² ≥10 mm–500 mm >10

1.3

Flexural strength and ductility

One major benefit of DUCON is that the material performance is programmable. Various set-ups and qualities of the micro-reinforcement (e.g. steel characteristics) allow the adaption of the material performance to the specific application. Figure 2 shows a stress-deflection diagram from flexural bending strength tests (Fig. 2) with three different possible adjustments A, B, C in comparison with standard concrete and typical fiberreinforced concrete. Note that the curves for standard concrete (grey) and fiber-reinforced concrete (blue) are shifted horizontally for a better legibility. A, B, C of DUCON in comparison with standard concrete and a typical fiber-reinforced concrete. Besides having high compressive and flexuralstrength, the properties can thus be adjusted to achieve an extremely ductile material. The ductility and its high strength are the key characteristics for high energy absorption of high speed dynamics and dynamics in combination with large deformations.

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2 2.1

slab could be quickly repaired by a 25 mm DUCON overlay and a time- and cost consuming demolition and reconstruction of the concrete could be avoided.

REPAIR APPLICATIONS Blast protection columns

Here, two types of production have been applied; a retro-fitting of existing columns (Fig. 3) or tube columns made of DUCON used as integrated formwork. They have been applied for earthquake protection (Zekaria 2001) and for blast protection (Schuler & Mayrhofer 2007). 2.2

Impervious and abrasive resistant overlays

Based on the high performance material characteristics of DUCON, impermeability, durability, freezethaw resistance and corrosion resistance in combination with crack control have been tested for the International Code Council (ICC) approval which was obtained for the applications on the US market. These characteristics are well suited for applications of thin impervious overlays on top of existing or damaged flooring or concrete structures. Figure 4 shows an application at a chemical plant where 7.500 m² of damaged and cracked concrete

2.3

DUCON as structural overlay

The existing concrete slab of the Main Train Station of Frankfurt has been damaged and a structural retrofit of the slab has been necessary. A reinforcement of the slab from below has been impossible, due to mechanical installations. The existing floor fill (60 mm) has been replaced by 60 mm DUCON. DUCON now performs as retrofit of the existing concrete slab and carries the structural loads (Fig. 5 and 6). The advantage is that due to the replacement of the gypcrete the granite stone cover could be executed 2 days after the installation of DUCON and the DUCON overlay doesn’t change the height of the former floor set-up. 2.4

DUCON as high performance overlay

Heavy duty off-set printing machines are sensible to tilting and deformation of the foundation, which leads to loss of print quality. In addition a crack propagation of the surface has to be avoided. For this specific project DUCON replaced the existing floor fill and performed as retrofit of the existing concrete slab, as energy absorption for dynamic loads and as crack controlled foundation of the printing machine.

Figure 2. Stress-deflection diagram for three different adjustments.

Figure 4. DUCON overlay, 25 mm, applied on an existing cracked concrete slab of a chemical plant.

Figure 3. Retro-fitting of existing columns for earthquake protection and blast protection.

Figure 5. Repair and strengthening of concrete slab—Train station Frankfurt.

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Electrochemical treatments of corroded reinforcement in concrete C. Christodoulou Faber Maunsell, Birmingham, UK

ABSTRACT: Electrochemical treatment methods have been used to arrest corrosion and protect the serviceable life of a structure. This paper analyses the corrosion initiation mechanisms, the chemical reactions taking place at the steel interface and the transportation process. The available options are briefly discussed and their advantages and disadvantages examined. The conclusions drawn, focus on the current practice and how the discussed electrochemical treatments can be used to optimize performance.

Reinforced concrete structures are widespread throughout the world and exposed to different climate conditions. Corrosion may be induced due to chloride ingress, cast in chloride and carbonation. Steel in concrete is normally passive and stable, and under a high alkaline environment it develops a passive oxide protective film. The diagram shown in Figure 1 illustrates the thermodynamic stability of the film in alkalinity. Even in the presence of chlorides, the oxides making up the protective film remain the most stable products and a significant reduction in pH needs to occur to de-stabilise the steel-concrete system. When break out of this film occurs, corrosion will initiate and propagation will be subject to concrete resistivity, humidity, chloride ingress, defects, cracks etc. The chloride threshold level may be defined as the critical level after which the concentration of chlorides is sufficient to sustain local breakdown of the passive film and therefore initiate micro-corrosion cells. It is expressed as a ratio of the total chloride to cement content of concrete (i.e. a weight percentage). Based on a variety of researches and literature,

Figure 1.

Corrosion initiation arrest mechanism.

typical threshold levels range between 0.2 and 2.5% by weight of cement. Although chloride levels are easily measured, the cement content can only be estimated as laboratory verification is difficult. Although bound chlorides will contribute to corrosion development, still there is no sufficient evidence to prove a correlation between the chloride binding and chloride threshold levels. During the electrochemical treatment process, it has been observed that sodium silicate gel and calcium hydroxide are by-products of the treatment which will fill the pores and the interface voids of the concrete. These will provide a reservoir of hydroxyl ions at the steel which suggests that electrochemical treatments not only arrest corrosion but they also increase the reservoir of OH- which then increases the chloride threshold levels. Corrosion rates are usually expressed as a current density, a rate of weight loss or a rate of section loss. A corrosion rate of 1 mA/m2 is approximately a loss of 10 g/m2/year or otherwise a loss of section of 1µm/year. As indicated from Figure 2, rates below 1 or 2 mA/m2 are considered negligible and corrosion development highly unlikely. Any rate higher than the aforementioned ones may result in the disruption of the concrete cover and passive film breakdown. Local fall in pH as low as 4 has been observed and the local corrosion rates may exceed 100 mA/m2 or in some extreme cases exceed 1000 mA/m2. A range of control measures exist to influence corrosion kinetics and the electrons movement. However, the reinforcement has a very low resistivity to the electrons movement and hence a successful arrest mechanism will have to control either the kinetics on the anodic and cathodic sites or by influencing the resistance of electrolyte (i.e. concrete) between those sites. It has been established that moisture levels and temperature will also affect the corrosion rates. At very dry environments and high temperatures chloride ingress is a very slow process, similarly to the dissolved oxygen availability in water saturated

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Figure 2. Corrosion rate plotted as a function of potential shift and current density together with an example of its interpretation.

environments. Thus, the interior areas of buildings are hardly ever affected by corrosion and it is the exterior that usually needs to be protected. Electrochemical treatments can offer excellent results when applied properly. The contaminated but otherwise sound structure will be treated and continue to function as intended, increasing the probabilities reaching its design life. Patch repairs have been widely used in the past, but in some cases with disappointing results. In cases of cast-in chlorides, the fresh alkaline mortar will probably worsen the overall corrosion stability. Where carbonation is the main cause of the problem and it has not reached a depth behind the reinforcement, it probably still remains the most cost effective solution. Impressed current cathodic protection has been the main component of the repair and maintenance strategy on bridge structures. It provides a long term protection and the current output can be regulated to meet the performance of the system. However, this

method requires availability of a constant power supply, significant electrical wiring and periodic monitoring of the system. Galvanic anodes instead have low installation and maintenance costs and provide an alternative strategy for remote areas where a DC supply is not readily available. Uncontrolled anode-steel shorts present no risk to system function and stray current corrosion of discontinuous steel is limited. While the technology currently available is generally less powerful than impressed current, sacrificial anodes can be applied to areas of need. Additionally, they can provide a cost efficient solution, when the specified anodes have a long serviceable life and the corrosion extent is not great. Generally, the technology is better used when it forms part of an overall corrosion arrest strategy. A hybrid technique uses galvanic anodes both in an impressed current and a sacrificial role. Any future corrosion risk can be managed by applying a brief impressed current treatment to arrest the corrosion process and thereafter continue protection by means of sacrificial corrosion. Realkalization and chloride extraction are temporarily techniques part of an overall management strategy and they do not require the extensive permanent installation of a traditional impressed current cathodic protection. However, they require the application of high current densities contrary to the impressed current cathodic protection. In general, design and detailing of electrochemical treatment systems needs to consider the risk of failure in current distribution and output to arrest ongoing corrosion. Reinforcement corrosion in high resistivity concrete presents a particularly challenging problem. Concluding, all the aforementioned corrosion arrest mechanisms are aimed to arrest corrosion and prevent further deterioration of the structure and by no means will they replace the already lost section.

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Electrochemical impedance monitoring of carbon fiber as an anode material in cathodic protection M. Chini & B. Arntsen Norut Teknologi, Narvik, Norway

Ø. Vennesland Department of Structural Engineering, Trondheim, Norway

J.H. Mork Maxit Group AB, Oslo, Norway

ABSTRACT: In impressed current CP systems, an insoluble anode is used on the surface or within the concrete structure. The most frequently and widely used anode material is titanium covered with noble metal oxides. Meanwhile much work is carried out to replace the anode material or improve the application technique such as using conductive coatings and paintings. In this regard, carbon fiber meshes have been introduced as an alternative for the last decade. Unfortunately, scientific work and publications concerning electrochemical properties of carbon embedded in concrete are scarce. Hence, a series of experiments have been designed and carried out in order to improve knowledge of its behavior as anode. One of these tests is to monitor the behavior of carbon by means of Electrochemical Impedance Spectroscopy (EIS) measurements. After carbon was polarized in different potential values, EIS have been done. Also effect of addition of 1 and 3 percent chloride ions has been investigated.

1

INTRODUCTION

Reinforced concrete is widely used in construction. The corrosion of the steel reinforcing bars in the concrete limits the lifespan of concrete structures (Chang 2000). The direct cause of destruction is increase of volume of steel reinforcement as the result of corrosion product formation (Darowicki 2003). Therefore the registered need for rehabilitation and repair and the associated costs is a matter of great concern for those responsible for assessment and maintenance of affected structures. A corrosion control method for steel reinforced concrete involved cathodic protection (Chang 1995). Cathodic protection (CP) is a widely used and accepted method of corrosion control for reinforced concrete structures (Mork et al. 2006). Cathodic protection (CP) of steel reinforcement in concrete structures is obtained by applying a direct current through the concrete from an anode system which is usually applied on the concrete surface. The anode material is a critical component of a CP system, because it serves to distribute the protective current across the structure and provides the locations for anodic reactions to take place in lieu of the reinforcing steel. There are some materials which are used as anode. Although the most frequently used anode material is titanium mesh covered by activated metal oxides, as a more recent anode material the carbon

fiber meshes have been used in last decade in some countries. Since carbon fiber used as anode material is a new CP system, the published papers about this anode material are scarce (Chini 2007) and very few researchers have work on this theme. Hence, a series of experiments have been designed and applied in order to improve knowledge of its behavior as anode. One of these designed test programs is investigation of electrochemical impedance behavior of carbon fibers in electrolyte which is simulated of pore solution of concrete. In this paper obtained data on electrochemical impedance spectroscopy (EIS) of carbon fibers after polarizing in different potential values in electrolyte would be presented. Also effect of existence of chloride in electrolyte was investigated and obtained data in different amount of chloride ions would be illustrated. 2

EXPERIMENTAL PROGRAM

To investigate impedance on carbon in order to clarify its behavior as anode in CP system, EIS test has been done on carbon after polarizing it in different potential values. In this experiment, after carbon fibers were polarized anodically from open circuit corrosion potential (OC potential) as 20, 40, 60, 100, 150, 300, 500, 750, 1000, 1500, 2000, 2500, 3500, 5000 mV,

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impedance measurements were applied. Afterwards the impedance amplitude of 10 mV was applied to carbon electrode while the system kept on the polarization potential. These measurements were operated by Impedance analyzer 1260 from Solartron Company. Also this was cooperated with Electrochemical Interface 1287; by using Solartron 1287 the potentials were continuously applied and the data was stored. A titanium MMO was used as counter electrode and the reference electrode was SCE. Since it is believed that existence of chloride ion plays an important role in corrosion process, therefore measurements were also done in chloride contents of 0, 1 and 3 percent of solution volume in the same polarization potentials as above.

3 TEST PREPARATION In CP system the carbon is used as mesh. The applied anode material is a continuous, highly flexible mesh woven of carbon fibers. The threads may consist of thousands of filaments with a thickness of 7 µm of each filament. Therefore it was very difficult to measure the corrosion current density among longitude of fibers. Although some techniques have been introduced (Mork et al. 2006), there are still uncertainties on measurement of surface area. Hence, in order to have more accurate assumption of active surface area, after various experiments, the cross section corrosion monitoring was preferred. In this method 10 bundles of 24000-filament thread were cast in a non-conductive transparent epoxy. After molding, the specimen was cut and ground to have a smooth and clear cross section. This method gives a surface area of around 10 mm2. All testing process was monitored and saved by Solartron 1287. After polarization, electrochemical impedance spectroscopy test was applied in frequency range of 0.1 Hz to 10 MHz and monitored by Solartron 1260. To simulate the concrete pore solution, a mixture of sodium, potassium and calcium hydroxide was used. pH of the solution was 13.2 and the measurements made in room temperature between 22 and 25 centigrade. Chloride ions were added as NaCl to make 1% and 3% chloride solutions.

4

CONCLUSION

The aim of this research is to clarify some of electrochemical behavior of carbon fiber in anodic polarization. In this regard, several test setups has been done by authors. In this paper oxidation of carbon is monitored by means of EIS measurements. Results indicate that carbon has not affected a lot by anodic polarization up to 500 mV and it shows a diffusion control behavior. In contrary, above 500 mV, corrosion of carbon starts and more than 1500 mV polarization one could observe mass transfer control (Warburg line) in its corrosion mechanism. More than 2 volt the system seems to change and formation of new products disturb the monitoring, as well as reach to a stable system is difficult. This behavior is more apparent in higher polarization potentials. Unfortunately above 2000 mV polarization, the results were so disturbed that even positive readings came for imaginary part which is meaningless. This phenomenon could be due to that the system and carbon was not stable anymore or some other oxidation products covered the working electrode. In addition, in those cases which charge transfer or combination of kinetic and diffusion were observed, the center of semicircles were displaced below the real axis. This tilting of center could be because of overlapping of different circles or the dielectric relaxation time of basic material. As mentioned before, although carbon follows the same concept behavior (oxidation mechanism) in different amount of chloride, existence of chloride ions has an important role in kinetic of the reactions. It is clearer on results that the addition of chloride increases the rate of oxidation (lower polarization resistance or smaller circle-diameter); however the amount of chloride does not affect very on the results. So adding chloride into solution does not affect carbon behavior but increase the kinetic of reactions. Also disturbing of data started in lower potentials in 1% chloride compared with 3%; and in the Nyquist plot for 3% chloride content shows multi time constant system. Authors think that it could be due to formation of more stable oxidation products and accumulation of them on the surface of working electrode. Reaching a better explanation needs more tests on by-products of oxidation of carbon in alkaline solution.

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CP of the rear reinforcement in RC-structures – Numerical modelling of the current distribution M. Bruns & M. Raupach Institute of Building Materials Research (ibac), RWTH Aachen University, Germany

ABSTRACT: In the last decades Cathodic Protection (CP) has become a well known and accepted rehabilitation method for reinforced concrete structures especially in case of chloride induced reinforcement corrosion. Thereby the anode is usually attached at the concrete surface closest to the corroding reinforcement in order to minimise the voltage required between anode and reinforcement. In specific circumstances an anode installation at the concrete surface next to the corroding reinforcement is not practicable, i.e. due to difficult access. In such cases it would be helpful to know whether it is possible to protect the reinforcement by an anode being installed at the opposite surface of the structural element. The question to be answered is: How does the protection current distribute to the reinforcement layers, and does the rear reinforcement layer receive sufficient current to be effectively protected, without getting over protection of the reinforcement layer next to the anode at the same time. Since the distribution of protection current depends on several parameters like the geometry and amount of the reinforcement, concrete resistivities, the polarisation behavior and geometry of corroding and passive zones, a general answer to this question is not possible. In order to investigate the influence of these parameters on the possibilities and limits to protect the rear reinforcement, a 3 D-FEM simulation of the current and potential distribution was developed and its accuracy was verified by laboratory results. This paper presents the numerical approach as well as the results of parameter studies carried out to show the impact of the reinforcement distribution, the concrete resistivity and the polarisation behaviour of the reinforcement on the current distribution within the reinforcement.

1

CATHODIC PROTECTION (CP) OF THE REAR REINFORCEMENT LAYER

When surface applied anodes are used for CP of the reinforcement typically the anode system is placed on the concrete surface next to the corroding reinforcement layer in order to assure that the mayor part of the applied current polarises the reinforcement layer which shall be protected. In specific circumstances this typical anode installation next to the corroding reinforcement is not practicable, i.e. due to difficult access to this surface or due to unacceptable traffic blocking that would be needed during the installation (see Fig.1). In such cases also conventional repair is difficult or even impossible and it would be helpful to know whether it is possible to protect the reinforcement by an anode being installed on the opposite surface of the structural element. The question to be answered is: How does the protection current distribute to the reinforcement layers and, does the rear reinforcement layer obtain sufficient current to be effectively protected. The distribution of protection current depends on several parameters like the geometry and amount of the reinforcement, concrete resistivities, the polarisation behaviour and geometry of corroding and passive zones. In order to investigate the influence of these

Figure 1. Potential applications for CP of the rear reinforcement. Left: i.e. bridge deck or ramp in a parking garage. Right: tunnel in a ground containing chlorides.

parameters on the possibilities and limits of the CP of the rear reinforcement, a 3 D-FEM approach of the current and potential distribution was used and its accuracy was verified by laboratory results.

2

EXPERIMENTAL INVESTIGATION AND NUMERICAL SIMULATION OF SIMPLE GEOMETRIES

To develop and verify the simulation model laboratory tests on concrete specimens containing two

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Figure 2. Specimen set-ups for laboratory tests to determine the distribution of the CP-current for simple geometries (left: specimen type V, right: specimen type H).

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Figure 4. Fraction of the total CP-current that is consumed at the rear reinforcement layer.

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PARAMETR STUDY

Figure 3. Measured and calculated rebar currents to the rebars (top) and rebar potentials (bottom) of specimen V for different protection currents.

The numerical approach was additionally used to carry out parameter studies on the influence of the reinforcement content, the reinforcement geometry and the concrete resistivity on the distribution of the protection current and on the reinforcement potential during CP with an anode applied at the rear surface. Three reinforcement geometries within a square slab of 0.60 m width and 0.15 m height were investigated. In order to investigate the influence of the concrete resistivity on the current distribution, the calculations were carried out with different concrete resistivities. The results of the parameter study showed, that the fraction of the protection current that is achieved at the rear reinforcement is strongly depending on the one hand on the reinforcement content, on the other hand on the concrete resistivity and the fraction decreases with increasing anode current density (see Fig. 4). Further the following conclusions were drawn from the results of the parameter study:

embedded reinforcement bars were carried out. The specimen setups are given in Figure 2. Details of the numerical approach used for the calculation of the current and potential distribution are given in the full paper. A comparison of the measurements from the laboratory tests and the calculated currents and the potentials of each rebar for different protection currents is given in Figure 3 for specimen geometry V.

• Macro cell corrosion at the rear reinforcement can already be suppressed at anode current densities below 10 mA/m2. • At higher concrete resistivities the concrete resistance between the anode and the reinforcement layers predominantly defines how the CP-current distributes to the reinforcement. • At lower concrete resistivities and lower protection current densities the polarisation behaviour of the reinforcement significantly contributes to the current distribution which leads to higher current fractions at the rear reinforcement.

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rebar potential mV vs. SCE

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4% chloride measured 4% chloride calculated 0% chloride measured 0% chloride calculated

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The application of electrochemical chloride extraction to reinforced concrete bridge members P.E. Streicher & G.E. Hoppe HHO Africa, Cape Town, South Africa

V.A. da Silva Afri-Coast Engineers SA (Pty) Ltd., Port Elizabeth, South Africa

E.J. Kruger South African National Roads Agency Ltd., Pretoria, South Africa

ABSTRACT: Some of the bridges of the Burman Road interchange in Port Elizabeth are located in a very high risk location and are in direct exposure to tidal and sea spray action. Chloride ions are derived from seawater and sea spray and enter the concrete by absorption and permeation of chloride containing water and by diffusion through saturated or partly saturated concrete. Concrete members of the three most exposed structures exhibited such high chloride contents that there remained only 2 options, i.e. demolish and rebuild or remove some of the chlorides from the concrete. After detailed cost/benefit analysis, it was decided to demolish the most affected bridge and to electrochemically remove some of the chloride ions from the other two bridges. Electrochemical Chloride Extraction (ECE) involves placing a temporary electrolyte on the concrete surface, and electrically driving the negatively charged chloride ions out of the concrete by negatively charging the reinforcing steel, and placing a positively charged anode in the external electrolyte. The chloride extraction system is commercially available and this is the first time it has been used in South Africa on this scale. Keywords:

1

Rehabilitation, desalination, repair.

INTRODUCTION

The Burman Road bridges were built in 1968. By 2000 the condition of these bridges had deteriorated as a result of durability problems associated with extreme marine exposure conditions. Poor concrete quality, ASR cracking and low cover to the reinforcing steel had exacerbated the corrosion problems in many instances. The bridges exhibited a wide range of durability problems. This included cracking and spalling of concrete due to chloride induced corrosion and cracking of concrete due to Alkali Silica Reaction (ASR). The focus of this paper is on the application of electrochemical chloride extraction (ECE) torehabilitate certain members of the affected bridges.

Figure 1. Locality Map of the N2 National Road at Port Elizabeth.

2

DURABILITY ASSESSMENT OF BRIDGES

are shown in Table 1. The rehabilitation methods considered included:

Chloride corrosion of the reinforcing steel was diagnosed as the cause of deterioration which was widespread with severe spalling evident. Chloride levels were determined in all the major members and these

1. Do nothing 2 Apply silane coatings 3. Apply a combination of silane coating and corrosion inhibitor (Sike Ferrogard, 1996)

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4. Apply ECE 5. Demolish and rebuild. It was decided for chloride levels above 1% with respect to cement mass, options 4 or 5 would be selected as option 3 becomes less effective at high chloride levels. Certain portions of bridges B1292, B1293 and B1294 fell in this category. This included the soffit edges and cross beams on bridges B1292 and B1293, and the canal column on bridge B1294. A life cycle cost analysis was performed to select between options 4 and 5. It was decided to apply ECE to selected portions of bridge B1292 and B1294 and to rebuild B1293. 3

an electronic systems are required on site, and hence no monitoring or maintenance is required (Berkley and Pathmanaban, 1990). This is providing the concrete surface is sealed against the reintroduction of moisture and chlorides.

4 APPLICATION ECE ON THE BURMAN ROAD BRIDGES A total of 900 m2 of bridge surfaces were treated. This was the first large scale use of this system in the RSA. ECE was applied at the following locations: • On bridge B1292, the bridge beams on the bridge deck soffit. • On bridge B1294, the canal pier column (round) This column was ideally suited to ECE, as the concrete cover was very low (−23 mm).

ELECTROCHEMICAL CHLORIDE EXTRACTION (ECE)

ECE involves electrochemically reducing chloride ion levels within the cover concrete and around the reinforcing steel. This process is illustrated in figure 2 (SHRP-S-347, 1993). The process requires the placing of a temporary anode on the concrete surface and negatively charging the reinforcing steel. This process is illustrated in Figure 2 (SHRP-S-347, 1993). The process requires the placing of a temporary anode on the concrete surface and negatively charging the reinforcing steel. The negatively charged chloride ions are repelled from the negatively charged reinforcing steel and migrate through to concrete to the external anode. At the same time, positively charged alkali ions from the electrolyte and within the concrete (Li,Na,K,Ca) migrate towards the reinforcing steel, increasing the pH level at the steel. After treatment, the steel is in a re-passivated state, and no corrosion will take place unless new chlorides enter the concrete. This re-passivated state is achieved by a combination of lowering the chloride level and increasing the alkali level at the reinforcement. By coating the treated areas, future chloride ingress can be delayed or even prevented. Unlike most coating systems that effectively only reduce the rate of corrosion ECE and cathodic protection are the only rehabilitation techniques that halt corrosion. As ECE is relatively expensive process it only becomes viable if demolition can be avoided by treating limited portions of a structure. The major advantage of ECE over impressed current cathodic protection is that no permanent electrical

5

SUCCESS OF TREATMENT

The performance of the treated areas after 4 years has been found to be very good with no spalling or rust staining evident.

6

CONCLUSION

ECE was successfully undertaken at the Burman road bridges. Treatment periods were within the time estimated for the specified chloride levels at level of reinforcing steel to be achieved. This being the first large scale ECE project in RSA, it took the contractor several months to perfect the system, but the final system worked very well. The use of the ECE treatment allows existing bridges, that otherwise would have to be demolished, to be rehabilitated. The demolishing and rebuilding of most bridges, specially under urban conditions, has a major impact on traffic.

ACKNOWLEDGEMENTS This paper is published by kind permission of the client, the South African National Roads Agency Limited.

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The corrosion protection of embedded steel reinforcement in reinforced concrete structures using galvanic anodes Heinz Bänziger BU Contractors, Sika Services AG

Jörg Vogelsang Surface Science Laboratory, Sika Technology AG

Georg Schulze Steel Protection Laboratory, Sika Corrosion Protection GmbH

ABSTRACT: Corrosion protection of steel structures using galvanic anodes is well established and a popular process. It was also introduced some years ago for the protection of steel reinforcement embedded in reinforced concrete structures. The currently available systems operate through prefabricated zinc anodes which are electrically connected to the reinforcement, so that the zinc anodes are sacrificially corroded as an alternative to corrosion of the steel. These systems have proved practical to apply in the field and where selected and installed correctly they can provide long lasting corrosion protection. The new system presented in this paper consists of a mixture of zinc powder and epoxy resin which is applied like a repair mortar directly onto the reinforcement and the concrete. A patent has also been filed for this process. (Note:- The information given in this paper represents the standard of knowledge as at May 2008. It is perfectly possible that additional information will be obtained during the approval process.)

1 1.1

INTRODUCTION Idea

To develop a galvanic anode that can be used as a component of the replacement concrete or repair mortar. One advantage of this method is that foreign bodies which might disturb the structure of the concrete are not incorporated. This type of anode is also capable of supporting loads and transmitting them to the surrounding mortar or concrete. An additional important advantage is that the anode material can be modelled to almost any shape or size and can therefore be adapted to the requirements.

Figure 1.

Cross section anode.

2 ANTICORROSIVE EFFECT 1.2

Initial experiments

Initial experiments were conducted to check whether it was at all possible to formulate an anode of this type. An anode with sufficient stability to be applied on the reinforcement bars, even in both vertical and overhead locations, plus with a suitable layer thickness. Our years of experience with epoxy mortars and systems served us very well and soon we had the first formulations ready for the basic preliminary tests.

2.1

The optimum epoxy resin quantity

Further experiments had to be carried out to find the maximum level to which epoxy resins can be filled with zinc particles, so that their electrical conductivity is retained and the galvanic protection is guaranteed. In the practical tests, we also verified the combinations which met the defined requirements by using simple dispersions in salt water.

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3

Optimum workability combined with optimum galvanic protection

In the next stage of development the two main requirements were considered together. The purpose of this stage was to combine conductivity, miscibility and application characteristics in one formulation. These were essential requirements for a successful product. It did not always prove easy to mix the material. In the first few seconds of mixing, a very fine zinc particle dust developed and was deposited in the immediate environment. This problem was overcome with a special sealing system, a lid with a hole in the middle through which the mixing spindle passes. This allows the material to be mixed without the local dust pollution.

3.1

Tests on realistic samples

The effect on reinforcing steel was verified with specially designed samples. In these tests it also became clear that not all the current methods of analysis are suitable for testing galvanic corrosion protection. The methods normally used in steel corrosion protection such as the salt spray test according to EN ISO 9227 do not simulate actual behaviour and can lead to misinterpretation of the results. For these reasons a test method was developed which is realistic and plausible and gives reproducible results. 2.4

Durability of anode, life expectancy

Dependent on the severity of the corrosion, the condition of the steel surfaces, environmental exposure and the nature of the concrete, an average life expectancy of 25 years can be expected. 2.5

Dosage

The kg/m2 dosage or consumption of the zinc anode is influenced by various factors. From experience we recommend a consumption which gives a life of 20 to 30 years. For average corrosion rates we recommend a consumption of 400 to 600 g anode material per m2 of steel surface. Anode behaviour over the operating period

Due to the sacrificial function of the zinc anode, the zinc oxidises and an increase in volume occurs. To prevent this causing the mortar covering the anode to spall, the repair mortar has a defined pore structure into which these oxidation products can expand.

Field applications

In principle this process can be used wherever steel reinforcement corrosion exists or is anticipated. Practical experience shows that galvanic protection systems operating with or without an external power source are particularly well suited for reinforcement protection where there is the presence or risk of chloride induced corrosion. The advantage of a system operating without an external power source is that they are almost entirely maintenance free! The necessary functional checks can be carried out as part of the regular structural inspections. 3.3

Functional checks on the structure

The check on the function of the anode is carried out by taking the potential measurement between the anode and an external reinforcement bar which is electrically connected to the exposed anode. By connecting this external steel reinforcement to the anode via an ammeter, it is possible to measure the protective current directly which then provides the data on the status and activity of the anode.

4 4.1

2.6

Initial trials on the Sika test wall

It is only by direct application on real structures that a definitive conclusion can be obtained on the functionality, usability, applicability and durability of a system of this nature. We conducted the initial trials on an inhouse test wall which had been specifically designed for this type of full scale test application. Corroded steel bars were exposed, derusted with a wire brush and protected with galvanic anodes weighing 80 to 100 g. A dosage of approximately 500 g per m2 of steel surface was used as this represents an average operating period of 25 years. 3.2

2.3

FIELD TRIALS

FURTHER PROCEDURES Field trials and product approval

It is scheduled to launch the Sika “FerroShield” process on the market in the spring of 2009. We intend to concentrate initially on selected markets and the most suitable structures. We want to work with customers who are interested in long-term monitoring of the process, leading to relevant approvals.

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Re-alkalisation technology applied to corrosion damaged concrete G.K. Glass FaberMaunsell, Birmingham, UK

A.C. Roberts & N. Davison Concrete Preservation Technologies, Notts, UK

ABSTRACT: This work examines the processes of steel corrosion initiation and arrest in chloride contaminated concrete. It is noted that the local production of acid is an essential feature in chloride induced corrosion damage. An acidification—realkalisation model of chloride induced corrosion has been developed to improve the explanation of some experimental observations. A simplified electrochemical treatment consisting of a hybrid of a pit re-alkalisation process to arrest corrosion followed by supplementary galvanic protection to maintain a high pH and steel passivity has been applied to concrete structures. Both the pit-realkalisation and supplementary galvanic treatments are delivered from a permanently installed sacrificial anode system. Risk management includes monitoring and a strategy to deal with any future risk of corrosion. An identified risk may be treated using a 2 week pit re-alkalisation treatment from a permanently installed sacrificial anode system with an impressed current connection.

Current per anode (mA)

The work looks at a two stage electrochemical treatment to arrest the chloride induced corrosion process and subsequently prevent corrosion initiation. A brief high current treatment is used to rapidly arrest corrosion by inducing conditions at the steel promoting steel passivity. It is shown that the main protective effect is the generation of inhibitive hydroxyl ions at the steel. A reservoir of hydroxyl ions stabilises the steel passive film by preventing the local pH reduction that would otherwise accompany and accelerate the chloride induced corrosion process. This temporary treatment is termed pit re-alkalisation and typically lasts 1 to 2 weeks. Following this initial treatment, the high pH at the steel is maintained using a low maintenance galvanic treatment. The treatment is applied using a single hybrid sacrificial/impressed current anode system. Sacrificial anode oxidation reactions occur relatively easily and facilitate the delivery of very high

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Figure 1. The current delivered by one anode in a hybrid treatment applied to 0.25 m2 of steel in concrete containing 4% chloride by weight of cement.

Figure 2. The installation of sacrificial anodes for use in a hybrid electrochemical treatment on a bridge beam.

impressed current densities off the anode surface at low safe DC voltages while embedded in a porous material in cavities formed in reinforced concrete. This is used to rapidly restore the high pH at otherwise acidic corrosion sites (pits) on the steel. The corroding sites are moved from locations on the reinforcing steel to the installed sacrificial anodes by a brief impressed current treatment. At the end of the brief impressed current treatment the remaining sacrificial anode is connected to the steel to maintain the high pH at the steel. Relatively little charge is require to restore steel passivity in the pit re-alkalisation treatment and the galvanic preventative treatment is applied at a low current density. Thus very long lives are in theory possible from the sacrificial anode system. This work reviews data from laboratory studies, site trials and full system installations with particular focus

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Figure 3. Steel potential in a laboratory specimen containing 4% Chloride (wt% cement) prior to the hybrid treatment and the potential decay on interrupting the treatment after 60 days.

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Figure 5. Measured potentials and corrosion rates in a bridge structure.

i appl ⎛ ⎛ ⎞ ⎛ ⎞⎞ ⎜ exp ⎜ 2.3∆E ⎟ − exp⎜ − 2.3∆E ⎟ ⎟ ⎜ β ⎟ ⎜ ⎜ βa ⎟⎠ ⎟⎠ c ⎝ ⎠ ⎝ ⎝

Figure 6. Potential map 1 month after switching a hybrid system to galvanic mode in a sheltered concrete column.

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Figure 4. The theoretical corrosion rate plotted as a function of potential shift and current density together with an example of its interpretation.

being placed on monitoring and acceptance criteria. Laboratory data shows that steel passivity is readily induced and maintained in chloride contaminated concrete. The galvanic protection applied is strongly dependent on the level of chloride contamination existing in the concrete structure and also responds positively to temperature and moisture changes. Risk management includes monitoring and a strategy to deal with any future risk of corrosion. Corrosion risk may be assessed non-destructively using corrosion potential and corrosion rate measurements. Steel corrosion rates are related to the applied current density and the steel potential shift achieved by the applied current. A conservative estimate of the potential shift is given by the potential decay and the applied steel current density can be estimated from the local current density delivered off an anode segment.

Passive steel is indicate by positive steel potentials and low corrosion rates. Corrosion rates have been determined on several field structures subject to this treatment using remote monitoring techniques. Another method of monitoring uses potential mapping. The presence of strong sacrificial anodes is indicated by strong peaks in the potential map and this indicates that the anodes are functioning. If the steel fall within the anode field it receives some protection. The absence of peaks between the installed anodes indicates that there are no anodes and the steel and therefore the corrosion risk is negligible. The use of such a non-invasive potential mapping technique is more compatible with low maintenance electrochemical treatments. An identified risk may be treated by re-applying the temporary pit realkalisation treatment from the permanently installed sacrificial anode system. Access to the anode is not required, only an accessible connection to the anode is required. This is because the anode is left in place after the temporary treatment. In some practical cases the application of the pit re-alkalisation treatment has been delayed until after scaffold access has been removed because access has been restricted. This paper challenges existing theory and proposed a new model for steel corrosion initiation and arrest in chloride contaminated concrete. While the theory may be debated further, a significant practical benefit has been achieved in the form of a new powerful but simple treatment for corrosion damaged concrete.

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On pathology and rehabilitation teaching of concrete structures: A case study N.G. Maldonado, R.J. Michelini & N.F. Pizarro National Technological University. Regional Mendoza Faculty. CEREDETEC, Mendoza, Argentina

M.E. Tornello National Technological University, Regional Mendoza Faculty, Civil Engineering Department, Mendoza, Argentina

ABSTRACT: The appearance of damages in reinforced concrete structures is recent due to material aging but the pathology of buildings implies a field of multidisciplinary knowledge. The challenge of education is how to incorporate this subject: in higher education or in postgraduate courses. If the knowledge must be in higher education, it seems that it is necessary a change in teaching about materials and performance of structures. Postgraduate courses may include research and development activities. The results of international networks give validation to these subjects for the practitioners through handbooks which treat of to mitigate the lack of professional formation in material durability. When natural phenomena appear, such as earthquakes or tsunamis, in order to retrofit structures, no traditional engineering solutions are required. Different academic sources are analyzed and judged and how the teaching related to codes and laws for professional practice must be improved.

1

INTRODUCTION

The appearance of damages in reinforced concrete structures is recent because of the use of new materials present at the beginning of century XX, which revolutionized urban growth worldwide but they were only specified by mechanical requirements. Due to aging of material in the structures and the problems derived from use and environmental conditions, studies about causes, repair techniques and retrofitting measures and prevention actions began. Relying on type of detected problem, more appropriate intervention alternatives based on available materials, workmanship, time and related costs can be chosen. Nevertheless, the pathology of buildings implies a field of multidisciplinary evaluations, the causes of their activation mechanisms and appearances, and their concurrent causes make lab simulation difficult under the same environment conditions and the appropriate evolution of construction problems. The published handbooks due to networks sponsored by CYTED such as Red DURAR (1997) or Red Rehabilitar (2003) validate these subjects for the professional graduates, especially referred to main cause of deterioration in reinforced concrete all over the world: the corrosion of reinforcing steel. These handbooks are a mitigation to lack of professional formation in material durability. The problem is increased in the presence of natural phenomena such as earthquakes. Nowadays, in different countries there are preliminary standards, some of them in discussion such as: COST 509, ACI Committee 365 and 201, GEHO,

RILEM 130-CLS. But, for performance evaluation, we need material studies and solutions of no traditional modeling. The proper characteristics of civil engineering, with different environments, technical practices, workmanship and materials generate difficulties for standardization. Seismic retrofit requires engineering approaches dramatically different from the traditional construction of structures. The challenge of education is how to incorporate the new subject. The first question is when it must be taught, in higher education or in postgraduate courses. If the knowledge must be placed in higher education, there must be a change in teaching about materials or teaching must be adapted to the performance of structures with the inclusion of time on deformations and modeling. If the knowledge must be in postgraduate courses, there are two important aspects: the incorporation of research and development activities and the formation and the practitioner’s experience in order to integrate a different knowledge for solutions of pathology diagnosis. The National Technological University has in its institutional program, postgraduate courses in relation to engineering of structures in different headquarters in the country. The Regional Mendoza Faculty approved in 1997 the first curricula of Master in Seismic-Resistant Structures (MS IES) with accreditation of project by CONEAU (National Evaluation and Accreditation Board), Resolution 398/99 (1999) and it was actualized in 2005. The six courses dictate during 1999–2008 introduce students to pathology aspects, impacts of construction

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materials (steel, concrete, masonry), laws, codes and standards for seismic-resistant construction, strategies for managing and reducing risks and impacts as well as development of rehabilitation techniques and strategies. 2

COURSE DEVELOPMENT

At the end of the courses the student will: − be introduced to scientific concepts of damage, the problems involved in the decision to upgrade existing buildings and the formulation of guidelines for a comprehensive rehabilitation of structure with efficient strategies and techniques according to seismic risk. − be able to discuss future directions in the area of risk earthquake reduction. − be introduced to tools for quantifying economic issues due to durability of materials and vulnerability of buildings and facilities in urban area prone to earthquakes. − be able to complete an independent research project (a search of the engineering literature, the writing of concise report and its presentation). During these courses, laboratories were available for students and they prepared their seminar experiences and they are integrated in research work-teams. 3

Thus, in the field of retrofitting and strengthening of structures, it is necessary to build bridges between basic research of materials and applied engineering.

COURSE ELEMENTS

Each course has a complete description of subjects, with objectives, schedule of seminars and examinations. It also includes recommended textbooks and the complementary papers to study. The activity includes a technical report to identify one line of research projects to evaluate and apply experimental works or modeling. All proposals should address the ways in which education and training are integrated within the research programs. Efforts to incorporate interdisciplinary educational activities and student teamwork are also encouraged. 4

failures during a long time, not specifically determined. This aspect of safety, in term of loss of human lives, contributes to a conservative approach in codes and standards. Most regulations are based on one level of performance, but the state of the art shows that it is necessary to evaluate other levels in relation to costs (FEMA 547 2006). As population and facilities grow, risks of seismic problems increase. Solutions are not only unique but they require a great number of specialists such as geoscientists, geotechnical engineers, structural engineers, architects, contractors, government officials and politicians. After 100 years of service life of the first reinforced concrete building in Paris (Helene 2007), aspects of maintenance and service life have been recently considered in few codes; therefore the concept of probability of failure is not incorporated in structural pathology. − On the other hand, the use of cementitious base of materials is extensively diffused by lower costs in comparison with new materials (steel, plastic, ceramic), but the limitation in the performance of physical and chemical processes of these materials became important research laboratory at micro and nano scale as well as natural scale for interaction with material and built environment. This subject was not considered in the study of materials and the planning of higher education reduced the time of formation about materials science.

COURSE OBSERVATIONS AND FEEDBACK

Two barriers are being detected during the courses: − In the teaching process of structural safety we hope that the buildings perform without catastrophic

5

CONCLUSIONS

− The subjects of courses point out the necessity to restate the teaching of materials in relation to service life, interaction with its environment and deterioration. − We believe that it is necessary to maintain the subject of pathology and rehabilitation as postgraduate formation. − It appears that the role of specifically international networks for new techniques is valuable but each country must adapt the best schedule of teaching according to academic formation of practitioners and graduates. − We consider promissory to include the subject in research programs and to share the examples in international base data.

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Comparative study of cleaning techniques to be used on concrete indoors M. Bouichou Cercle des Partenaires du Patrimoine, Champs-sur-Marne, France

E. Marie-Victoire & D. Brissaud Laboratoire de Recherche des Monuments Historiques, Champs-sur-Marne, France

ABSTRACT: Concrete in France is now a major concern in the field of conservation of cultural heritage. A lot of churches are listed and often show high levels of indoors dirtiness due to candle soot. But indoors cleaning is a hard task. The techniques which have shown their efficiency outdoors, are awkward to perform indoors, water and abrasive spreading and seeping in. Recently, new techniques such as a water injection-extraction system or new latex pads were developed and appeared interesting for this specific application. Their performances and potential side effects were evaluated by testing them inside a church dating back to the beginning of the 1930ies, covered in black soot deposits. If the water injection-extraction appeared very efficient, without deleterious impact, the cleaning level obtained with the latex pads varied from poor to excellent depending on the pad; some dissolution facies, as salt crystallization, appearing with some of tested products.

Outdoors cleaning on concrete has been quite studied and therefore very efficient techniques are available depending on the dirt deposit (thin or thick) and on the surface to be cleaned (façade or sculpture, decayed or not …). Thus, wet or dry abrasive projection, pads or even laser cleaning are commonly used. But the problematics indoors is slightly different. Effectively, indoors dirt deposits, which are mainly due to candle soot, are generally darker and less hardened than the black crusts frequently encountered outdoors. Another constraint lies in the confinement, specifically in churches. Actually, abrasive projection or water-based techniques spread water and abrasives that implies protection and recycling measures that can be tricky to install indoors. Considering a church interior, the use of such techniques would mean to protect any sculpture, painting or stained glass window. Therefore two types of techniques recently developed: water injection-extraction and new latex pads. The water injection-extraction system (G) is based on a vacuum-washing technology. Low pressure water is sent on the surface through a sucking head under low vacuum, which also recovers the dirty water by a suction action. Thus water is confined. Two pressures were tested: 2 bars and 20 bars. The latex pad technique consists in the application (brushing or airless spraying) of a natural latex dispersed paste on the surface to be treated. When a polymerized film is formed, it is peeled off. The dirt deposit is then trapped in the film, which is removed by the mechanical peeling action. Depending on the temperature and on the relative humidity, the time

before peeling can vary from 24 h to 48 h. The major compound of the pads is hevea sap, which was initially stabilized in a liquid phase by adding ammonia. As the ammoniac emanations were problematical either for the applicator or for metals sensitive to ammoniac …, new stabilizers were recently developed. Some of the producers also introduced mineral additions or chemical agents (Diamine Tetra Acetic Ethylene …). As a consequence, three families of products (E, F, R) were tested with twelve products in total, containing or not ammonia or EDTA (which is generally used to dissolve calcium). As the testing areas were small, all the products were hand-brushed. Tests were performed in a church made of concrete built in 1934 and listed in 1979. The inner walls that conserved the wooden prints of their forms were covered with a thin and homogeneous very dark dirt deposit, probably linked to candle fumes. But as the Saint-Esprit Church is richly decorated with wallpaintings and mosaics, it seemed hard to introduce cleaning techniques using water or abrasives inside the church. Therefore it appeared to be an interesting site to test alternative techniques. For each technique, 0.16 m² surface had to be cleaned. The aim of the tests was to eliminate the dirt deposit, preserving the wood-looking skin of the concrete. Three points were examined: the implement ability, the efficiency and the potential side effects. A maximum duration of 5 minutes was necessary either to clean the 0.16 m² testing area with the G technique or to apply the latex pads. After a short time of installation (water and electricity supply) the G technique appeared easy to apply.

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Concerning the latex pads, it is to be noticed that the operators applied much higher quantities of products than the consumption recommended by the producers. Some of the pads were also clearly too fluid (E1, E3, E4 and E5), when others were too pasty (F3, E2), both being difficult to apply. Ammoniac emanations were encountered for all the tested products indicating that ammonia was probably reduced but not totally replaced (ex: important ammoniac emanations with R2 and R4). Dealing with the polymerization duration, unexpected 8 days were necessary for 2 of the F latex pads. Finally, some of the pads were too adhesive and therefore hard to peel-off (F2, F3, E3), when others almost did not stick to the surface (E1, E2, E5). The cleaning efficiency varied from good to insufficient. The best cleaning was obtained with the G technique, with an homogeneous and luminous result. In that case, no difference in efficiency was noticed between the 2 tested pressures. The performances of the latex pads are more uneven. The F latex pads (F1, F2 and F3) lead to an unsatisfactory cleaning, dark stains being still visible (Fig. 2). On the contrary, the R pads (R1, R2 and R3) lead to an homogenous cleaning, but darker than the one obtained with the E pads. To quantify the efficiency, color measurements (L, a, b system) were performed before and after cleaning, and on a control area. The color measurements confirmed the naked eyes observations, the performances achieved with the G technique being the best, but very close to the E latex pads. The worse cleaning efficiency was obtained with the F latex pads. On the most representative areas, cores were sampled at the dirty/cleaned interface. It is with the G technique that the least dirt residues were observed, thus confirming the color measurements and the visual observations. On the contrary, the interface dirty/cleaned of the area treated with the F3 latex pad was almost not distinguishable, so numerous the dirt residues were in the cleaned area. On the areas treated with the E1, E2 and E5 pads (which were the less adhesive), some dirt residues were also observed. A visual observation of the latex films revealed the presence of particles pulled out from the concrete skin (small aggregates …). It is interesting to note that the F pads (F1, F2, and F3) which were the most adhesive caused the most important loss. Finally it is important to precise that may be some particles were also pulled-out with the G technique, but as they were sucked with the dirty water, the phenomenon could be hardly quantified. Binocular observation of the surfaces of the cores treated with any of the latex pads did not reveal the

presence of any latex residue. But close to the interface dirty/cleaned, some white crystals were observed on the areas treated with the E1, E3, E5 pads. SEM observations indicated the presence of a few latex particle on areas treated with the R1, E1 and E3 pads, lots of residues where encountered on the surfaces treated with F1 and F3 pads. EDS analysis evidenced the presence of calcium, potassium and traces of sodium and chlorine on the surfaces treated with the E1 and E3 pads, when calcium, sodium, chlorine, potassium, and traces of nitrogen were encountered on the surfaces treated with the R1, F1 and F3 pads. On the surface treated with the F3 pad, EDS maps were realized. They revealed a clear sodium pollution, associated with calcium and sulfur dissolution on the cleaned surface. Finally, the observation of the E1 and E3 cores confirmed the presence of crystals at the interface dirty/cleaned. These crystals are potassium and calcium-based. FTIR analysis performed on the reference sample and on the surfaces treated with G, E3, F1 and F3 indicated the presence of calcite and quartz. On the spectrum obtained on the surface cleaned with F3, new bands appeared compared to the non-cleaned area. The same bands were observed on the F3-film spectrum, which indicates a latex pad pollution on the treated surface. On the spectrum obtained for the E3cleaned zone, new bands were also evidenced compared to the non-cleaned area, but these bands do not correspond to the E3 latex film. An analysis carried out on a powder precisely sampled at the interface dirty/cleaned on the E3 core, where the potassium and calcium-based neo-crystals were observed, revealed the presence of identical peaks, with a higher intensity. Those bands are probably a signature of these neo-crystals. In conclusion, the best efficiency was obtained with the water injection/extraction system, which was quite easy to use and without any side effect identified. Concerning the latex pads, some of them were too fluid (most of the E pads) and difficult to apply. Some pads (F2 and F3) were also too adhesive and hard to peel-off. Dealing with efficiency, depending on the pads, it varied from good (E and R pads) to insufficient (F pads). Finally, a clear impact on the concrete was identified with the pads the most adhesive, that induced losses of the concrete skin (F pads). A chemical pollution was also evidenced with the majority of the latex pads, sometimes generating neocrystallization (E1 and E3 pads). New analysis are ongoing to better understand these chemical interactions. A second series of tests was also performed on another type of concrete (bush-hammered).

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Repair materials and systems

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Material Data Sheet Protocol – From anarchy to order Fred R. Goodwin BASF Construction Chemicals, Beachwood, Ohio, USA

Alexander M. Vaysburd Vaycon Consulting, Baltimore, Maryland, USA

Peter H. Emmons Structural Group, Baltimore, Maryland, USA

ABSTRACT: The Material Data Sheet Protocol was developed and accepted by the International Concrete Repair Institute (ICRI) to address the current state of anarchy in the concrete repair industry, and establish order of information by providing a consistent, logical and systematic methodology for testing and reporting of information for cement based repair materials. Rather than develop multiple specifications to provide acceptance limits for material performance in a multitude of repair situations, the Material Data Sheet Protocol allows the specifier to choose verifiable properties optimized for the requirements of a particular repair situation. The applicator can obtain useful information about yield, working time, surface preparation, application temperature range, curing and compatibility, as well as verify the material performance. The material producer can optimize products based on market needs and technology improvements, rather than concentrating on closely passing acceptance levels of an existing specification in a commodity based market. This paper presents an overview of the ICRI Repair Material Data Sheet Protocol.

1

INTRODUCTION

The Webster’s Dictionary definition of the anarchy as “disorder in any sphere of activity” fully reflects the present situation with material data sheets. Selecting and specifying the most appropriate concrete repair materials based on product literature can be a daunting task because of the variety of test methods and material properties used to characterize these materials. The engineer has very limited, and sometimes misleading information on which to base selection and specification of materials for a particular repair project. Typically, only data on properties favorable to the particular material are reported. Also, test procedures used to determine material properties vary widely and modifications are often poorly documented. Such information does not provide user confidence in the given properties, and is not a credible basis for selection of materials that will result in durable repairs. The Material Data Sheet Protocol was developed by the International Concrete Repair Institute (ICRI) to address the current state of anarchy in the concrete repair industry for cement based repair materials. Rather than develop multiple specifications for the different material performance needed in varied repair situations, the Material Data Sheet Protocol allows the choice of verifiable properties optimized for the requirements of a particular repair situation. A discussion of the rel-

evance, interpretation, and suggested limiting values of many types concrete repair materials has been published in a related document, ACI 546.3 R-06 Guide for the Selection of Materials for the Repair of Concrete.

2

DESCRIPTION OF THE PROTOCOL

The Protocol is divided into five sections. A brief introduction of the contents of each section will be discussed below. 2.1

Section 1, Repair material description

Section 1, Repair Material Description, is divided into three subsections: Recommended Use; Benefits; and Limitations. Benefits include claims such as shrinkage compensated, colored, or rapid hardening. Limitations require reporting the minimum and maximum application thickness, the minimum and maximum application temperature, any material modifications (i.e., aggregate extension) and the recommended curing regimen. 2.2

Section 2, Compositional data

Section 2, Compositional Data, provides a means of classifying the binder chemistry, defines the number of components, and requires determination of possible

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deleterious components within the proprietary composition. The reporting of these components is not intended to disclose proprietary information, but is based on references to compositional limits from ASTM C 150 and ACI 222. Reporting the pH is necessary to determine if steel passivation can occur when the repair material is applied onto reinforcing steel in the concrete. The pH of the repair material when freshly mixed with water is to be reported as well as the pH of the hardened repair material. The aggregate characteristics in the repair material are described in accordance with the sections of ASTM C 33 using the material retained on a 0.1-mm (170- mesh) sieve. 2.3

The package label must contain the volume yield of the product as cubic meters (or cubic feet) per package, the shelf life listed as a “use-by” date, and the minimum and maximum storage temperatures and conditions. 2.5

Section 5, How to use the material

Section 5, How to Use the Material, includes reporting the aggregate extension requirements. Concrete surface preparation is in accordance with the ICRI Concrete Surface Profile. The type and amount of mixing liquid, instructions for material application and temperatures, finishing guidelines, curing regimen, application thickness, and cleanup recommendations are listed in this section.

Section 3, Material properties

Section 3, Material Properties, typically specifies different tests for mortar and concrete materials. The test method used for the reported result must also be reported, as some results may be determined by specified alternate methods. Plastic properties are reported first and include:

3

⎯Density and Yield (ASTM C 185 [mortar] or C 138 [concrete]). ⎯Setting Time (ASTM C 266 or C 191) at both the minimum and maximum stated application temperatures (the mortar fraction should be sieved from concrete materials for setting time). ⎯Air Content (ASTM C 185 [mortar] or C 231 [concrete]).

⎯“one size fits all” for a type of material (certainly not the best case for concrete repair), ⎯the acceptance that minimum performance limits represent the true level of performance needed for a material, rather than the lowest common denominator that was agreeable among the industry specification developers (frequently material suppliers), and ⎯the commoditization of a material type which limits further development of technology to address the application need (i.e., once a specification has been developed, the products complying tend to compete upon price with limited further investment in product differentiation).

The hardened properties are also reported in Section 3 of the Protocol. A different demolding and curing regimen is specified based upon the speed of hardening and polymer modification. Different tests are typically specified for mortar and concrete mixtures and include direct and splitting tensile, flexural, compressive strengths, modulus of elasticity, bond strength, length change, CTE, resistance to freezing and thawing, deicer scaling resistance, compressive creep, RCP, chloride permeability, sulfate resistance, chemical resistance, and cracking resistance. 2.4

Section 4, Packaging and storage

Section 4, Packaging and Storage, requires labelling of the packages in accordance with the “Product Marking” section of either ASTM C928 or C 1107.

THE FUTURE OF PROTOCOLS

Protocols are a new type of document to the concrete repair industry that can help resolve several of the obstacles that exist regarding the development of specifications:

4

CONCLUSIONS

The ICRI Inorganic Material Data Sheet Protocol provides useful information to the material purchaser, applicator, and specifier. Advantages of the Protocol approach to material property characterization are that standardization organizations can more easily reach consensus, specifiers can select the important material properties and performance based on their experience, and the properties can be verified due to the transparency of the test methods.

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Utilization of high performance fiber-reinforced micro-concrete as a repair material M. Skazlic Materials Department, Faculty of Civil Engineering, University of Zagreb, Croatia

D. Bjegovic´ Materials Department, Faculty of Civil Engineering, University of Zagreb, Croatia Civil Engineering Institute of Croatia, Zagreb, Croatia

M. Serdar Materials Department, Faculty of Civil Engineering, University of Zagreb, Croatia

ABSTRACT: Problems of early degradation, reduced service life and repair costs that overhead the price of structure itself, have resulted in development of new materials and technologies for repair of concrete structures. Recent results show that High Performance Fiber-Reinforced Micro Concrete (HPFRMC), used as repair material, has very good properties concerning stability, serviceability and durability. For that reason there has been a growing interest in the use of HPFRMC for repair and rehabilitation of concrete structures. The aim of this research was to evaluate the efficiency of high performance fiber-reinforced micro-concrete as a repair material when applied on concrete beam. During the research testing of compressive strength, flexural strength, modulus of elasticity and toughness has been performed on concrete beams repaired with high performance fiber-reinforced micro-concrete and cement mortar. Results of this research have proven that the use of high performance fiber-reinforced micro-concrete as repair material has both economical and technical advantages. 1

INTRODUCTION

HPFRMC is material with excellent mechanical properties and with durability properties better than those of normal concrete, which makes its utilization as repair material very interesting. The aim of this research was to evaluate the efficiency of HPFRMC applied on existing concrete structural elements with insufficient bearing capacity and/or durability in order to reinforce the tension zone. 2 2.1

EXPERIMENTAL PROGRAM Materials

Mix designs of concrete (C25/30) used to prepare 150 × 150 × 600 mm beams and HPFRMC used to reinforce the tension zone of concrete beams are presented in Table 1. Properties of concrete and HPFRMC used in this research, in fresh (density, slump and porosity) and hardened (compressive and tensile strength and modulus of elasticity) state are presented in Table 2. As it can be seen in Table 2, due to the careful mix designing, HPFRMC has high compressive and tensile strength, while similar modulus of elasticity as base concrete. High values of compressive and tensile strength are preferable for the use of this material for reinforcing concrete in tension zone, while similar values of modulus of elasticity as those of the base concrete are necessary in order to achieve the compatibility of two materials. 2.2

Specimens preparation

Specimens were demoulded after 24 h and part of the concrete cross section was crushed and removed in order to prepare

the surface before placing HPFRMC. At 28 days age surface of the specimens was coated with emulsion in order to have better adhesion between base concrete beam and repair material. Specimens were then placed back to the mould in which HPFRMC was poured. Reinforcing procedure of the concrete beams was performed in three different ways, Figure 1: A. Cross section of concrete beam dimensions 120 × 150 mm was reinforced in tensile zone with 30 × 150 mm layer of HPFRMC, B. Cross section of concrete beam dimensions 90 × 150 mm was reinforced in tensile zone with 60 × 150 mm layer of HPFRMC, C. Cross section of concrete beam dimensions 120 × 90 mm was reinforced on the sides and in tensile zone with 30 mm thick layer of HPFRMC. 2.3

Testing procedure

In order to evaluate the efficiency of the repair procedure and material, bending behaviour of the specimens was tested. Testing was performed according to standard ASTM C 1609, with strain rate 0.05 mm/min up to the displacement of 4 mm. 3

RESULTS

The comparison of the results of testing bending behaviour of all four specimens at 3 days is given in Figure 2 and at 28 days in Figure 3. Mechanical properties of concrete repaired with HPFRMC are shown in Tables 2 and 3.

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Table 1. Mix design of concrete and HPFRMC used in this research.

Table 3. Testing results of bending behaviour.

Material

C25/30

After 3 days

Peak force (kN)

Displacement at peak force (mm)

Cement, kg Water, kg Aggregate, kg 0–4 mm 4–8 mm 8–16 mm Superplasticizer, kg Silica fume, kg Steel fibers, kg Polypropylene fibers, kg Water/binder ratio

350 175

HPFRMC A HPFRMC B HPFRMC C

63.9 66.4 64.3

0.09 0.21 0.09

After 28 days

Peak force (kN)

Displacement at peak force (mm)

R HPFRMC A HPFRMC B HPFRMC C

33.4 67.4 77.6 73.5

0.05 0.08 0.51 0.25

978 391 587 – – – – 0.50

HPFRMC 925 193 1150 – – 35 148 200 2 0.18

Table 2. Properties of concrete and HPFRMC used in this research. C25/30

Properties in fresh state Density (kg/m3) Temperature (°C) Slump (cm) Porosity Properties in hardened state Compressive strength (MPa) Tensile strength (MPa) Modulus of elasticity (GPa)

A

HPFRMC

2380 25 7 1.2

2580 30 23 3.8

35.6 4.7 31.8

116.5 32.2 32.0

B

C

R

60 50 force (kN)

Material

70

C

40 30

A 20

B

R

10 0 0

0,5

1

1,5

2

2,5

3

3,5

4

displacem ent (mm )

Figure 2. Results of flexural loading test of concretes at 3 days. 90 A

80 70

B

C

R

B

force (kN)

60

C

50 40

A

30 20

R

10 0 0

0,5

1

1,5

2

2,5

3

3,5

4

displacement (mm)

Figure 3. Results of flexural loading test of concretes at 28 days. Figure 1.

Cross section of specimens.

From the results presented in Table 2 and 3 it can be seen that compared to control concrete (R) that was strengthened with HPFRMC, all HPFRMC reinforced concretes showed an appreciable increase in flexural strength. Among all types of reinforcing, specimen HPFRMC B showed the best behaviour under flexural loading. 4

CONCLUSION

The results presented in this paper show that high performance fiber-reinforced micro-concrete (HPFRMC) can have

larger compressive and tensile strength but the same modulus of elasticity as normal concrete, which makes this material very interesting for the use in repair and strengthening of concrete structures. Results presented in this paper show that concrete beams that were reinforced with 60 mm thick layer of HPFRMC have the highest tensile strength and the best behaviour under flexural loading. It can be concluded that this modus of reinforcing is the optimal one. From the comparison of results of flexural loading performed at 3 days and 28 days it can be concluded that HPFRMC can bear designed flexural loading even at 3 days, which is another advantage for the utilization of HPFRMC as repair material.

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Properties of modern rendering systems based on mineral binders modified by organic admixtures J.C.-M. Capener Division of Building Technology, Chalmers University of Technology, Göteborg, Sweden

ABSTRACT: Composite mineral-organic materials and in particular cement-organic materials are increasingly used in today’s construction applications and civil engineering projects. Polymer-modified render is a material that possesses several properties superior to conventional renders, such as better workability, crack resistance, adhesion to substrate and flexural strength, all of these important properties to renovation and rehabilitation projects, as they improve the durability of the material and ensure proper performance. The importance of understanding the mechanisms behind the materials behaviour is crucial when designing new and advanced products for the industry. In this paper, the findings of a doctorate project studying the influences of different organic admixtures on the early age and hardened properties of mortars, such as water retention, plastic shrinkage and drying shrinkage are presented. The mechanisms of the hardening process, from plastic to hardened state, moisture transfer and obtained structure are examined. The results show that the cellulose water retainer not only improves water retention, but also significantly introduce air content in both cement and limecement binder systems. Styrene-Butylacrylate-based copolymer significantly increased the flexural strength as expected, but revealed a high dry shrinkage. The combined effect of the air-entrainer and the cellulose ether reduce the plastic shrinkage of all compositions, but increase the drying shrinkage, due to their high air contents, but also the lower elastic modulus created by the entrained air. Styrene-Butylacrylate copolymer and vinyl acetate/ethylene copolymer also contribute to improving water retention, but have a significant effect on moisture transport properties.

1

Table 1. Admixture types investigated in this paper and their properties as specified by the producers.

INTRODUCTION

The main part of the doctoral project consisted of studies on the interactions between composition, hardening process, microstructure and transport properties. The influence of different admixtures, especially organic polymers and air entraining agents, was studied. In the plastic state the consistency and plastic shrinkage as well as water retention are studied. Among the decisive properties of the hardened mortars the focus was kept on vapour transport, moisture transport properties and drying shrinkage.

2

Admixture 1: Air entraining agent—C14/C16-alpha olefin sulphonate. An anionic surfactant used for air entraining and modification of consistency. Admixture 2: Water retainer—Methyl-hydroxyethyl cellulose (MHEC). A non-ionic cellulose derivative. Admixture 3: Vinyl Acetate/Ethylene Copolymer (Et/VAc). Thermoplastic latex of anionic type. Admixture 4: Styrene-Butylacrylate Copolymer (SBA). Synthetic rubber latex. Redispersible polymer powder of anionic character.

EXPERIMENTAL STUDY

In this study four common and representative types of admixtures were investigated. Their properties are described in Table 1. The cement and lime/cement based mortar compositions are described in Table 2 with an increasing amount of Admixture 3, Et/VAc. There is a similar set-up for Admixture 4, SBA, but with a different notation. Here lower case “d” denotes Admixture 4 as opposed to “c” for Admixture 3 in Table 2.

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3

MAIN RESULTS AND DISCUSSION

In this paper, the influence of different admixtures with the special focus on organic polymers and air entraining agents was investigated. A certain admixture, or combinations of two or more, will influence not only one property but affect the behaviour of the mortar in both the fresh state and hardened material.

0,5

Table 2. Mortar compositions based on constant consistency and grouped by binder. Increasing amount of Admixture 3. Thermal power [mW/g]

C

Constant consistency Group: C-c Group: LC-c Reference Ref. + Adm. 1 Ref. + Adm. 2 Ref. + Adm. 1 & 2 Ref. + Adm. 1 & 2 + Adm. 3 (3%) Ref. + Adm. 1 & 2 + Adm. 3 (6%) Ref. + Adm. 1 & 2 + Adm. 3 (12%) Ref. + Adm. 1 & 2 + Adm. 3 (24%)

3.1

C C-a C-b C-a-b C-a-b-c3

LC LC-a LC-b LC-a-b LC-a-b-c3

C-a-b-c6

LC-a-b-c6

C-a-b-c12

LC-a-b-c12

C-a-b-c24

LC-a-b-c24

C-a-b-d12

0,4

C-a-b-d24 0,3 0,2 0,1 0 0

20

40

60

80

100

Time [hours]

Figure 1. Thermal power evolved during hydration for Group C-d. The retarding effect of Admixture 4 is clearly seen.

Calorimetric measurements

A small retardation effect is seen on the introduction of Admixture 3 in the cement-based compositions. This effect, however, is much greater for Admixture 4 where a significant retardation is imposed on the hydration, both for the cement- and the lime/cement-based compositions with increasing dosages of polymer, as seen in Figure 1 for the cement based compositions. 3.2

Drying shrinkage

Air entrainers and cellulose ethers give the same increase in shrinkage compared to the reference mortar in cement-based compositions. However, in lime/cementbased systems, cellulose ethers yield a somewhat higher drying shrinkage compared to the air entrainers. Admixture 4 gives a significant increase in drying shrinkage for both cement- and lime/cement-based compositions, with rising polymer/binder ratio. The high air content can explain part of this increase induced by Admixture 1 and Admixture 2, but the retained water held in compositions with Admixture 2 also contribute to the shrinkage as it can dry out at later ages. This does not explain the further increase in drying shrinkage found with the rise in polymer/binder ratio of Admixture 4, see Figure 2. Here, the explanation probably lies in the microstructure and pore size distribution created by the film formation. For organic materials and in this case mineral-organic materials, there is a large amount of adsorbed water in the organic material microstructure. Here, drying does not cause capillary pressures by forming menisci, but rather a volume reduction as water evaporates and the molecules of the organic material contracts.

Figure 2. Drying shrinkage for Groups LC-c and LC-d, RH = 50%, (x) indicating (c) or (d) depending on admixture type used.

4

CONCLUDING REMARKS

For the specific types of admixtures investigated in this study, the following conclusions can be drawn: • The results show that the cellulose water retainer not only improve water retention, but also significantly introduce air content in both the cement and the cement-lime binder systems. • The combined effect of the air entrainer and the cellulose ether reduces the plastic shrinkage of all compositions, but increases the drying shrinkage, due to their high air contents, but also the lower elastic modulus created by the entrained air. • Vinyl Acetate/Ethylene co-polymer tends to reduce the plastic shrinkage of lime/cement based materials whereas an increase was found for cementbased compositions.

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Low shrinking self-compacting concretes for concrete repair C. Pistolesi, C. Maltese & M. Bovassi Mapei S.p.A., Milan, Italy

ABSTRACT: Self compacting concrete repair using cementitious materials often causes the formation of many cracks caused by the drying shrinkage in the hydration phase of the hydraulic binders applied on the rigid underling layer. A carefully studied mix—design together with expanding agents are normally used to limit the effect of this shrinkage. In this paper we studied special self-compacting concretes with high dimensional stability, by using a combination of expanding and shrinkage reducing admixtures. In particular, we studied the synergistic effect due to the interaction between calcium oxide as expanding agent—which works during hardening phase—and a special shrinkage reducing admixture in order to compensate long term concrete shrinkage and this synergistic effect was also studied during a test performed on a job site.

1

INTRODUCTION

Self-compacting concretes are used more and more often to build new structures (Seto et al., 1997), as they can be applied very easily, they are very fluid and have a high compaction level in situ. Self-compacting concretes (SCC)—as normal cementitious material—undergo drying due to hydration of the hydraulic binders (Person & Creep, 1999). The drying shrinkage of new cementitious materials, which are applied on rigid and shrinkage-free concretes, could cause the formation of many cracks that compromise the durability of the repair (Freidin, 2000). On the basis of scientific knowledge, the drying shrinkage can be limited if the composition and particular concrete conditioning are carefully studied (Neville, 1996). In the scientific literature the use of expanding agents based on calcium oxide (CaO) and calcium sulphoaluminate (C4 A3S) is well-known (Kokubu, 1972). Concrete admixed with such expansive agents expand on the first few days and a form of prestress is obtained by this restrained expansion with steel reinforcement. The efficacy of the expanding agents, used in mortars and/or concretes to prevent cracks formation, is much influenced by the conditioning of cementitious materials immediately after casting. In fact, when humid curing is not ensured, expanding effect does not occur and shrinkage can not be efficiently compensated by the expanding action. Some researches have concentrated their interest in the use of specific organic admixtures, which reduce the superficial stress of the water contained in the capillary pores (Hua & Francis Young, 1997; Cerulli et al., 2001).

The combined effect of an expanding agent (EA) and shrinkage reducing admixture (SRA) has been investigated, in the experimental program, to get a SCC with high dimensional stability even in absence of humid curing, as it usually happens in concrete repairing works. 2

EXPERIMENTAL

The laboratory testing has regarded the preparation of mortars and concretes, on which such fresh and hardened properties like fluidity, mechanical strengths, shrinkage and restrained expansion have been measured. The tests on mortar have been carried out using different dosages of SRA (0,4 ÷ 2% on cement weight) and also in combination with the EA (dosed at 7% on cement). Tests on concrete have been carried out on a combination of 5 kg/m3 of SRA and 30 kg/m3 of EA, on the basis of the preliminary results that have been obtained by the mortars. 2.1

Materials

For the preparation of the mortars the following materials have been used: cement type CE I 52,5R (in compliance with EN197-1 norm); SRA based on propylenglycol ether; EA based on calcium oxide; siliceous normalized sand according to EN 196-1. Concretes have been prepared by using Stabilcem SCC/F as binder (produced by Mapei S.p.A.), which is suitable to prepare standard self compacting concretes (STABILCEM SCC/F is a binder with pozzolanic reaction, containing a limestone filler, superplasticizers, anti-segregation agent). It has been modified adding the same admixtures used in the mortars and a

321

selected siliceous aggregate (according to UNI 8520) having a maximum diameter of 12 mm. 2.2

Tests for fresh and hardened mortars

The mortars have been mixed according to EN 196-1: SRA has been added to water before mixing, while, EA to the cement powder. Density (according to EN 1015-6) and consistency (according to EN 13395) have been tested on fresh mortars. Drying shrinkage (according to UNI 6687) has been measured on the mortars containing SRA, while, restrained expansion (according to UNI 8147) has been measured only for the mortars containing both SRA and EA admixtures. Restrained expansion has been determined by measuring the dimensional variation of a mortar specimen reinforced by a threaded metal. Compressive strengths have been measured according to EN 196-1 after 24 hours, 7 days and 28 days. 2.3

Restrained expansion tests with prism conditioned under water show that the presence of SRA does not modify maximum expansion, which remains rather constant. On the contrary, in the same tests—where the prism are cured at 20°C and 50% RH.- the action of SRA is evident: it increases expanding agent effectiveness. A negative effect on the compressive strength has been found in the mortars containing SRA: the strength reduction is higher at 1 day, while is less evident after long curing times. The positive effect of SRA on the initial concrete expansion is confirmed. The effectiveness of both admixtures SRA+EA is more evident on the final restrained shrinkage after 3 months. The tests carried out on concretes have shown a small negative effect on the mechanical strengths, due to the presence of SRA.

3.3

Tests for fresh and hardened self-compacting concretes

SCC have been prepared with the same water/binder ratio and controlled in fluidity (UNI 11041), slump loss retention and density (EN 12350/6). The tests on the hardened SCC have been carried out to determine the restrained shrinkage according to UNI 8148. Mechanical strengths development has been evaluated up to 3 months according EN 12390-3.

Special Self Compacting Concrete was used for the repair of large piles of a high way viaduct damaged because of the reinforcements corrosion. Concrete was prepared by using Stabilcem SCC/F as binder, aggregate 0–12 mm, Expancrete and Mapecure SRA (produced by Mapei S.p.A.) as expanding agent and shrinkage reducing admixture.

4 3 3.1

RESULTS AND DISCUSSION

CONCLUSIONS

Based on the results obtained in this paper, the following conclusion can be drawn:

Properties of fresh mortars and concretes

The fluidity of mortar improves, when the dosage of SRA increases. When the same mortars are admixed with SRA in combination with EA, a reduction of the initial consistency occurs due to the presence of the EA. The samples of SCC have shown similar behavior to the mortars. Density remains almost constant for all concretes and mortars. 3.2

Job site test results

Properties of hardened mortars and concretes

SRA has a good behavior as shrinkage reducer. Increasing the dosage of SRA, the drying shrinkage lowers.

− increasing the dosage of SRA a significant drying shrinkage reduction can be observed; − SRA reduces early compressive strengths compared to reference mixture (without admixture), nevertheless the difference becomes less then 10% at longer ageing times; − EA based on calcium oxide works very well when the mixture is conditioned in highly humid environment, while in dry conditioning (50% H.R.) it does not work enough; − a combination of SRA and EA in SCC leads to a higher early restrained expansion and a very low final drying shrinkage; − the use of SCC and admixtures like SRA and EA to repair very large concrete structure gives good results without any crack-formation.

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Effects of fiber and silica fume reinforcement on abrasion resistance of hydraulic repair concrete Yu-Wen Liu Department of Civil and Water Resource Engineering, National ChiaYi University, Taiwan

Chin-Chun Lee Hoping Construction Office of Taipower Company, Taiwan

K.S. Pann Civil Engineering Department, Cheng Shiu University, Kaohsiung, Taiwan

ABSTRACT: Concrete of hydraulic structure is easily eroded when hit by waterborne sand flow. This experiment was an attempt to increase the abrasion resistance of hydraulic repair concrete by adding fibers and silica fume. Three different types of fibers, containing steel fiber, carbon fiber, and Polypropylene fiber, were added to repair concrete, also the abrasion resistance was measured with waterborne sand flow testing method and compared with plane silica fume concrete. Test results show that optimized fiber—silica fume combinations can better improve the abrasion resistance of repair concrete. At the silica fume—cement ratio of 20%, the fiber concrete can remarkably enhance the abrasion and impact resistance. When hit by waterborne sand flow, the abrasion resistance was better for silica fume concrete combine with carbon fibers, steel fiber and glass fiber than plain silica fume concrete. In addition, the carbon fiber and glass fiber concrete have rather high impact resistance than silica fume concrete.

The flexural strength of four types concrete as shown in Figure 2. It can be found that the flexural strength of GFC is the highest in the four types of concrete at 28-days and 56-days. In this study, the main purpose of using fibers is to improve the abrasion and impact resistance of concrete. Figure 3 represents the abrasion loss rate by water borne sand flow of each concrete. It can clearly be found, the fibers addition to silica fume concrete can improve the abrasion resistance. The feature points to a significant change in the abrasion behavior if fiber inclusions are added to the concrete. Because the fibers increase the

Compressive Strength (MPa)

In Taiwan, the most significant erosion problems are due to the abrasive effect on concrete surface of hydraulic structures. The purpose of this study is therefore aiming at investigating the performances of silica fume concrete, steel fiber, carbon fiber and glass fiber reinforced silica fume concrete. The test conducted included determining the properties of fresh concrete, compressive strength, impact resistance, and abrasion resistance. The four types concrete can obtain well workability with adding superplasticizer. The slump of silica fume concrete (SC) is 160 mm and 95 mm at 0 min and 45 mins, respectively. For CFSC, GFSC and SFSC, the slump is 170–140 mm, 155–120 mm and 135–125 mm, at 0 and 45 mins, respectively. Moreover, the slump loss (45 mins.) of three types fiber concrete are small, except the silica fume concrete (SC). Figure 1 shows the compressive strength of four types concrete. For the concrete made with w/cm of 0.28, and addition 1.0 vol.% of carbon fiber shows the compressive strength of 10–15% higher than the three other types of concrete at 7-days. For SC, CFSC, GFSC and SFSC, at 7 days age, the percentages are 86, 95, 85 and 80, respectively, of 28-days compressive strength. But the steel fiber reinforced concrete has the highest compressive strength than the three other types concrete at 28-days and 56-days. The above-mentioned later age improvement can be attributed to the enhancement of steel mortar interface related with frictional forces.

120 110

7 day 56 day

28 day

100 90 80 70 60 50 SC

CFSC

GFSC

SFSC

Figure 1. The compressive strength of four types concrete.

323

16

56 day

12 8 4

500 382 400

382

Fracture

408 408 271

300 200 100

64 15 15

0

0 SC

CFSC

GFSC

SIC

SFSC

CFC

GFC

SFC

Figure 5. The number of impact at beginning crack and fracture of four types concrete, respectively.

Figure 2. The flexural strength of four types concrete.

Abrasion loss rate (cm3/hr)

Beginning crack

600 28 day

Number of impact

Flexure Strength (MPa)

20

30 25

28 day

56 day

20 15 10 5 0 SC

CFSC

GFSC

SFSC

Figure 3. The abrasion loss rate of four types concrete.

bond stress between cement matrix and aggregate, the cement matrix and aggregate aren’t be removed easily by abrade. The usage of carbon, glass and steel fibers decrease the abrasion loss rate by 14%, 10% and 39% as compared to silica fume concrete, respectively. The surface layer of concrete specimens had been wear out and the coarse aggregate becomes exposed. But the surface layer of steel and carbon fiber concrete appear to be smoother than the silica fume concrete (SC). On the other hand, from figure 4, it can be found that the fibers don’t be pulled out from concrete specimens. Then the efficacy of fibers in concrete as same as the aggregate, as the strength or hardness of fibers are stronger which result in the higher abrasion resistance of concrete specimens. The number of impact at beginning crack and fracture and the failure modes of four types concrete specimens, respectively, after impact test as shown in figure 5 and figure 6. Because the addition of fibers, the specimens of CFSC, GFSC and SFSC have a rather high impact resistance compared with the silica fume concrete. It is noted that, under the impact action, the number of impact at beginning crack is the same as fracture for silica fume, carbon fiber and glass fiber

concrete specimen. However, the number of impact at fracture is 210 higher than at beginning crack for steel fiber concrete specimens. Due to the dimension of carbon fiber and glass fiber which are 2.8% and 6.4% of steel fiber, respectively, and the amount of carbon fibers and glass fibers are rather more than the steel fibers in unit volume of concrete. As a result, when the micro-crack just begins, the carbon and glass fibers can hold up the micro-crack extending. As the numbers of impact achieve the fracture energy of concrete specimens, the internal cracks breaking through the bind of fiber simultaneously result in the cracks which extend quickly and the fracture. However, the steel fiber concrete doesn’t fracture immediately till the fibers are pulled out. CONCLUSION As the repair material for hydraulic structures must have the performance of high abrasion and impact resistance in Taiwan, then major experimental results show that the three types fibers concrete can obtain good workability, and the compressive strength increases. At the early age, the compressive strength of carbon fiber concrete is the highest one among four types concrete, but the steel fiber concrete’s compressive strength is the highest at the later age. In addition, fiber—silica fume combinations can better improve the abrasion resistance of repair concrete. At the silica fume—cement ratio of 20%, the fiber concrete can remarkably enhance the abrasion and impact resistance. When hit by waterborne sand flow, the abrasion resistance was better for silica fume concrete combining with carbon fibers, steel fiber and glass fiber than plain silica fume concrete. Moreover, the carbon fiber and glass fiber concrete have rather high impact resistance than silica fume concrete.

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Use of calcium sulfoaluminate cement to improve strength of mortars at low temperature J. Ambroise & J. Péra Laboratoire Génie Civil et Ingéniérie Environnementale, Institut National des Sciences Appliquées de Lyon, Villeurbanne, France

ABSTRACT: One of the main properties required by mortars for concrete repair is rapid strength development regardless of the temperature. The strength development of mortars based either on Portland cement (OPC) or Calcium Aluminate Cement (CAC) is easily managed at 20°C, utilizing usual accelerators. At 5°C, it is more complicated. Calcium Sulfoaluminate Cement (CSA) introduced in mortars either as binder or as CAC accelerator represents a solution to solve this problem. This paper presents two examples of CSA utilization. The first example deals with the development of sealing mortar for road works using CSA as binder. Such mortar presents 2-hr strength higher than 9 MPa at 5°C and time of workability higher than 20 minutes at 20°C. The second utilization concerns the acceleration of self-levelling topping mortar based on CAC. When 25% CAC is replaced by CSA, the 1-day strength shifts from 1 to 8 MPa at 5°C without modifying the time of workability at 20°C.

1

INTRODUCTION

2

The efficient repair and replacement of concrete pavements and bridge decks often requires a rapid setting material that can be placed, cured, and opened to traffic in a relatively short period of time. Several repair materials are marketed for the repair of deteriorating concrete structures. These repair materials are classified into different types, such as cement, epoxy resins, polyester resins, polymer latex and polyvinyl acetate. Cement-based materials and polymer/epoxy resins are the most widely used among the repair materials Calcium aluminate cement (CAC) has several important characteristics including early strength gain, chemical resistance, and excellent refractory properties. Accelerated OPC and CAC work very well at 20°C, but their performances decrease at 5°C. In the present study, calcium sulfoaluminate cement (CSA) has been introduced in mortars either as binder or as CAC accelerator to get good performances at 5°C in two cases: − development of sealing mortar for road works; − acceleration of self-levelling topping mortar based on CAC. Yeelimite or Klein’s compound (anhydrous calcium sulfoaluminate, 4CaO.3 Al2O3.SO3) is the key ingredient of CSA cement. Its importance derives from the fact that on hydration in the presence of lime and calcium sulphate, it rapidly hydrates to form ettringite, leading to high early strength development.

PROPERTIES OF CSA USED IN THE STUDY

A Chinese commercially produced calcium sulfoaluminate clinker (CK) has been used in the present study and its amount of yeelimite was very high: 73.5%. To prepare CSA cement, CK was mixed with re-crystallized gypsum (RG), a by-product of the manufacture of phosphoric acid by the Prayon PH2 process, with two hemihydrate stages followed by a dihydrate process producing co-crystallized gypsum with low P2O5 content. The pure gypsum content of RG determined by DTA-TGA was 89.4%. 3

SEALING MORTAR FOR ROAD WORKS

The following requirements had to be fulfilled: − the mortar is self-compacting; − the time of workability is higher than10 minutes at 30°C, and 20 minutes at 5°C and 20°C; − the 2-hr compressive strength is higher than 7.5 Mpa, regardless of the temperature; − the drying shrinkage is minimum. The workability was measured by means of the static spread of a truncated cone presenting the following dimensions: φinf = 80 mm; φsup = 70 mm; h = 40 mm. Mortar was self-consolidating when the spread was higher than 120 mm. Two points were optimized: − the particle size distribution of the sand; − the borax content.

325

Three particle size distributions were investigated: − S1 = 50% 0/1 mm + 50% 2/5 mm; − S2 = 25% 0/1 mm + 25% 1/2 mm + 50% 2/5 mm; − S3 = 33% 0/1 mm + 33% 1/2 mm + 33% 2/5 mm. S3 led to the best performances regarding both the workability and the 2-hr compressive strength at 5°C. The recommended borax dosage is 0.051%: the mortar presents the best workability and the higher strength between 2 and 5 hours after casting. After 5 hours, the strength at 5°C is equivalent to that recorded at 20°C. The difference between the strengths recorded at 20°C and 5°C decreases versus time and more rapidly with a dosage of 0.061% borax. The long-term strength of sealing mortars stored at 5°C and 70% RH increased between 1 day and 28 days, regardless of the borax content: +30%. This study shows that it is possible to develop a sealing mortar based on calcium sulfoaluminate cement presenting high strength after 2 hours at 5°C. The recipe is robust and usable between 5 and 30°C without major changes: only the retarder content has to be adjusted. 4 ACCELERATION OF SELF-LEVELLING TOPPING MORTAR BASED ON CAC The aim of this study was to investigate the accelerating effect of CSA on a topping mortar based on CAC. The initial composition of the dry matter contained in the mortar was as follows: − sand + filler + admixtures: 71.7%; − CAC + anhydrite: 26.8%; − OPC: 1.5%. The water/solids ratio was maintained at 0.24. The following requirements had to be fulfilled: − sufficient fluidity, assessed by the measurement of the time taken by 80 mL of mortar to flow through an aperture of 8 mm (9 sec. ≤ t ≤ 15 sec.); − self-levelness, measured by means of the static spread (SS) of a truncated cone presenting the following dimensions: φinf = 80 mm; φsup = 70 mm; h = 40 mm. The following values had to be reached: − 145 mm ≤ SS ≤ 165 mm after 5 minutes, − 135 mm ≤ SS ≤ 160 mm after 20 minutes; − 1-day compressive strength at 5°C ≥ 5 MPa.

In a first step, CSA cement was composed of 75% CK and 25% RG (CSA25). CSA25 replaced the initial binder (CAC + anhydrite + OPC) at different contents varying from 5 to 25%. In a second step, the influence of CSA composition was investigated. The results can be summarized as follows: − the flowing time and the static spread are not influenced by the percentage of CSA25 introduced in the mortar; − a strength of 5 MPa is reached when the amount of CSA25 is higher than 15%; − the performances at 7 and 28 days do not depend on the quantity of CSA25 present in the mixture; − the best accelerating effect is obtained with 25% of CSA25. Two other CSAs were investigated: − CSA 15: 85% CK + 15% RG; − CSA 30: 70% CK + 30% RG; Their behavior was compared to that of control mortar (0% CSA) and CSA25. The results independent of the CSA composition.

5

CONCLUSIONS

The use of calcium sulfoaluminate cement is interesting to solve the problem of strength development at low temperature (<5°C). When mixed with re-crystallized gypsum in the following proportions (75% clinker + 25% gypsum), it can be used as either a binder in the design of sealing mortar or an accelerator in topping mortar based on calcium aluminate cement. In sealing mortar, the strength reaches 9 MPa, 2 hours after casting at 5°C. The workability at 20°C is not modified. The recipe is robust and usable between 5 and 30°C without major changes: only the retarder content has to be adjusted. In topping mortar, it is possible to replace 25% of the initial binder without affecting the workability. The 1-day strength is 8 times higher at 5°C. The composition of CSA has not any influence on the properties of the mortar.

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Development of Strain-Hardening Cement-based Composites for the strengthening of masonry A.-E. Bruedern, D. Abecasis & V. Mechtcherine Institute of Construction Materials, TU Dresden, Germany

ABSTRACT: The low tensile strength and high brittleness of masonry structures can lead to collapse in the event of an earthquake. Hence there is a need to improve the mechanical performance and consequently the safety in a great number of buildings in countries where such risks exist. In this paper the applicability of Strain-Hardening Cementbased Composites (SHCC) for strengthening masonry is investigated. Several SHCC compositions suitable for this purpose were developed based on the previous work by the authors. Apart from the high ductility and shear resistance of the hardened material, the optimisation parameters were workability and sprayability of SHCC in the fresh state as well as good adhesiveness to the masonry. In the first step, a series of simple rheological and mechanical tests were performed on fresh and hardened SHCC in order to estimate the major parameter combinations for the main tests. Subjected to advanced testing methods, the material compositions chosen, all containing 1.8% of PVA fibre, were subsequently investigated with regard to their sprayability, changes in their rheological properties over time, and their mechanical performance. Finally, triplet specimens representing a piece of masonry were reinforced with the developed SHCC and tested for shear behaviour. The tests showed that the static load bearing capacity of masonry, as well as the energy absorption capacity increased manifold due to such reinforcement.

1

Table 1.

DEVELOPMENT OF SHCC COMPOSITIONS

The starting point for developing SHCC compositions suitable for the strengthening of masonry was a composition developed by authors in an earlier investigation. In order to achieve the desired behaviour of concrete in both the fresh and hardened states, several changes were undertaken in the reference mixture. From this point on the mixture S1.8 as shown in Table 1 will be referred to as the “base mixture”. The description of further compositions investigated is given in following chapters when presenting the results obtained for them. Simple slump flow tests using were performed to test the quality of mixtures in the fresh state and to sort out the improper compositions. Only if the results of these tests were promising was the sprayability of SHCC tested. The fine-tuning of compositions was performed using a rheometer. Furthermore, three-point bend tests were performed on hardened SHCC for all sprayed mixtures. Finally, few representative mixtures were chosen for the testing of further mechanical properties as well as mechanical performance with regard to strengthening of masonry. 2

SPRAYING TESTS

The wet spraying method was applied without the addition of admixtures via the nozzle. SHCC was spayed into wooden boxes of dimensions 800 mm × 500 mm × 60 mm. The adhesion of the mixture to the wall and the form stability (either keeping its shape or running down the wall) was assessed. After the material had hardened,

Composition of the reference SHCC.

SHCC

S1.8

S1.8-CA3 [kg per m3]

S1.8-LP10

Cement CEM I 42.5 HS CA-Cement Fly Ash Quartz Sand 0.06/0.2 mm Water SP (polycarboxylate ether) SP (sulfonate naphthalene) VA Underwater compound PVA Fibre Air-entraining agent

327

315

358

– 766 546

10 766 546

– 572 598

342 3.2

342 3.2

291 –

–

–

8.9

3.2

3.2

23.4 –

23.4 –

3.5 21.1 0.46

small specimens were cut from the plate to perform mechanical tests. The results of the spraying tests showed that the best sprayability was observed for SHCC mixtures for which slump flow values between 150 mm and 160 mm had been measured in the tests with small cones. The maximum possible fibre content in SHCC depends on the type and the configuration of the spraying machine. The maximum fibre content which did not affect the sprayability of the mixtures negatively was 1.8% by volume for the small machine used in this study.

327

3

RHEOMETER TESTS

5

Rheometer tests were used to monitor the development of the viscosity of mixtures continually over time, which is crucial for the applicability of SHCC by spraying. The mixtures were tested in the System HAAKE MARS over a span of 45 minutes. The mixtures were prepared without fibre to avoid possible uncertainties related to the separation of fibre in the process of testing. Fresh SHCC should have a low viscosity during the first minutes after the mixing to make spraying possible without congesting the hose or the pump, but then after spraying onto a wall it should stiffen rapidly to avoid the fresh mix’ running off the wall. In the first series the influence of superplasticiser was studied on mixes of otherwise identical composition (the base matrix S1.8 without fibre) and equal initial workability (slump flow value). The effect of using a partial replacement (3%) of Portland cement by calcium aluminate cement (S1.8-CA3) was investigated in the next test series on mixtures containing Glenium Ace 30 and Woermann FM 375, respectively. Furthermore, the influence of air entrainment was studied. Fig. 1 shows the development of the viscosity for these mixtures over time. The mixture with addition of calcium aluminate cement and superplasticiser Woermann FM 375 showed the most favourable behaviour, i.e. a sufficiently long “dormant” phase followed by a rapid increase of viscosity.

4

TRIPLET TESTS

To estimate the effect of an SHCC reinforcement layer on a masonry wall, triplet specimens were used, with the outside bricks standing on supports and the middle brick pressed down until the connections collapse (cf. Figure 2). Figure 2 shows representative loaddeformation curves obtained from the triplet tests with and without strengthening layer, respectively. Table 3 gives the average values and the standard deviations of the maximum load and the deformation at failure obtained from the triplet tests. Both parameters increased by an order of magnitude as a result of the strengthening using SHCC.

Table 2.

Mechanical properties of the selected mixtures.

Mixture

Compressive Strength MPa

Tensile Strength MPa

Strain Capacity %

S1.8 S1.8-CA03 S1.8-LP10

48.4 (1.84) 47.7 (4.60) 43.7 (2.79)

3.02 (0.02) 2.93 (0.32) 3.11 (0.22)

1.3 (0.10) 0.8 (0.21) 1.2 (0.44)

3 tests per series, standard deviations are given in parentheses.

20

MECHANICAL TESTS

F

10

15 Load F [kN]

Three-point bend tests on small beams were performed for each material composition showing favourable rheological behaviour in the slump flow and spraying tests. Uniaxial tension tests on prisms (290 mm × 40 mm × 40 mm) were performed to test the strain-hardening behaviour of sprayed mixtures. Additionally, the compressive strength was tested using small cubes 40 mm × 40 mm × 40 mm, which were cut from the sprayed slabs as well. Mechanical characteristics of SHCC as obtained from the compression and tension tests are given in Table 2. The data are presented only for three SHCC compositions used subsequently in the triplet tests (see Chapter 5).

10 with strengthening layer 5 without strengthening layer 0 0.0

0.5 1.0 1.5 2.0 Displacement δ [mm]

2.5

Figure 2. Typical load-deformation curves obtained from the triplet tests with and without strengthening layer, respectively.

S1.8-ACE30

9

Viscosity [Pas]

S1.8-CA03-ACE30

8

S1.8-CA03-FM375

7

S1.8-LP10-FM630

Table 3. Average results of the triplet test (standard deviation is given in parentheses).

6 5

Specimen type and mixture

Maximum load [kN]

Deformation at failure [mm]

Without strengthening layer With 10 mm S1.8 With 10 mm S1.8-CA03 With 10 mm S1.8-LP10

2.2 (0.6) 18.4 (1.5) 16.5 (2.0) 13.1 (2.0)

0.25 (0.04) 3.50 (0.66) 2.20 (0.60) 2.10 (0.50)

4 3 2 1 0

0

10

20

30

40

Time [min]

Figure 1. Viscosity development of the selected mixtures.

3 tests per series, standard deviations are given in parentheses.

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Air permeability of hardener-free epoxy-modified mortars as repair materials M.A.R. Bhutta Research & Development Center, Takamura Holdings Co., Ltd., Japan

K. Imamoto University of Science and Technology, Japan

Y. Ohama Research Center for Advanced Mineral Aggregate Composite Products, Kangwoon National University, Korea

ABSTRACT: The purpose of this study is to investigate the air permeability of hardener-free epoxy-modified mortars by applying a non-destructive testing method developed by RILEM Technical Committee 189-NEC. Considering intelligent patch materials for the repair work of deteriorated reinforced concrete structures, hardener-free epoxy-modified mortars using a bisphenol A-type epoxy resin without any hardener are prepared with various polymer-cement ratios, and tested for moisture content, air permeability and strength. As a result, although there is no significant strength improvement, the coefficient of air permeability of the hardener-free epoxy-modified mortars is markedly decreased with an increase in the polymer-cement ratio.

1

2.2

INTRODUCTION

The air permeability of patch materials or cover concrete plays an important role in the durability of repair work for deteriorated reinforced concrete structures because it controls the rate of penetration of air with moisture. The purpose of this study is to investigate the air permeability of the hardener-free epoxy-modified mortars by applying a non-destructive in-situ testing method specified in the State-of-the-Art Report of RILEM Technical Committee on Non-Destructive Evaluation of the Penetrability and Thickness of Concrete Cover. Hardener-free epoxy-modified mortars using a bisphenol A-type epoxy resin without any hardener are prepared with polymer-cement ratios (epoxy-cement ratios) of 0, 5, 10, 15 and 20%, and tested for moisture content, air permeability and strength after a 28-d dry curing. 2 2.1

Moisture content determination

The moisture content of cylindrical specimens was determined by using Concrete and Mortar Moisture Tester (HI-520, manufacturer: Kett Electric Laboratory) before air permeability test. 2.3

Air permeability test

The air permeability test of cylindrical specimens was carried out by using the “Torrent Permeability Tester”, developed at “Holderbank” Management & Consulting Ltd. (currently Holcim Support Ltd.) in Switzerland. Figure 1 demonstrates the air permeability test of a cylindrical specimen. 2.4

Strength test

Beams specimens 40 × 40 × 160 mm for flexural strength and broken half specimens 40 × 40 × 80 mm

TESTING PROCEDURES Preparation of specimens

According to JIS A 1171 (Test Methods for Polymer-Modified Mortar), hardener-free epoxy-modified mortars were mixed with a mass ratio of cement to fine aggregate 1:3, polymer-cement ratios of 0, 5, 10, 15 and 20% and a water-cement ratio of 50.0% at the flow range of 170 ± 5. Cylindrical specimens φ 150 × 300 mm for air permeability test and beam specimens 40 × 40 × 160 mm for strength test were molded, and subjected to a 20°C-60%(RH)-dry curing for 28 d.

Figure 1. A view of air permeability test.

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for compressive strength were tested in accordance with JIS A 1171.

60.0

5.0 Compressive

Flexural

Flexural strength (MPa)

20.0 10.0

0.0

5

10

15

20

Polymer-cement ratio (%)

Figure 4. Polymer-cement ratio vs. compressive and flexural strengths of hardener-free epoxy-modified mortars.

15.0

60.0

0.20

10.0

40.0 30.0 20.0

Compressive

5.0

Flexural

10.0

0.0 0.012

0.15

Flexural strength (MPa)

50.0

0.0

2

m )

-16

30.0

0

0.25

(×10

10.0

0.0

0.30 Coefficient of air permeability

40.0

Compressive strength (MPa)

Figure 2 illustrates the coefficient of air permeability (kT) of hardener-free epoxy modified mortars with different polymer-cement ratios. The air permeability of unmodified mortar is quite higher than that of epoxy-modified mortars. The air permeability of the epoxy-modified mortars is decreased with increasing polymer-cement ratio. Figure 3 represents the coefficient of air permeability (kT) vs. moisture content (M) of hardener-free epoxy modified mortars. It can be seen that there is the reasonably good correlation between the coefficient of air permeability and moisture content of the hardener-free epoxy-modified mortars. Figure 4 exhibits the compressive and flexural strengths of hardener-free epoxy-modified mortars with different polymer-cement ratios. No significant

50.0

Compressive strength (MPa)

3 TEST RESULTS AND DISCUSSION

15.0

0.014

0.022

0.068

0.247

Coefficient of air permeability (×10-16 m2)

0.10

Figure 5. Relationship between coefficient of air permeability (kT) and compressive and flexural strengths of hardener-free epoxy-modified mortars.

0.05 0.00

0

5

10

15

20

Polymer-cement ratio (%)

Figure 2. Polymer-cement ratio vs. coefficient of air permeability (kT) of hardener-free epoxy-modified mortars.

kT = 1/(29.982 M–158.88) Correlation coefficient = 0.89

0.20

-16

2

m )

0.25

(×10

Coefficient of air permeability, kT

0.30

0.15

strength improvement is observed in the epoxy-modified mortars with polymer-cement ratios of 5, 10, 15 and 20% after a 28-d-20°-60%(RH)-dry curing. Figure 5 shows the relationship between the coefficient of air permeability (kT) and compressive and flexural strengths of hardener-free epoxy-modified mortars. There is no significant relationship between the compressive and flexural strengths and the coefficient of air permeability (kT) after a 28-d-20°-60% (RH)-dry curing.

0.10

4

0.05 0.00

5.0

6.0

7.0

8.0

9.0

10.0

Moisture content, M (%)

Figure 3. Relationship between coefficient of air permeability (kT) and moisture content (M) of hardener-free epoxymodified mortars.

CONCLUSIONS

The air permeability of hardener-free epoxy-modified mortars is markedly decreased with increasing polymer-cement ratio. There is the reasonably good correlation between their coefficient of air permeability and moisture content, but no correlation is found between their coefficient of air permeability (kT) and strength after a 28-d-20°C-60%(RH)-dry curing.

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Performance of corrosion inhibitors in concrete exposed to marine environment I.N. Robertson University of Hawaii at Manoa, Honolulu, Hawaii, USA

C. Newtson New Mexico State University, Las Cruces, New Mexico, USA

ABSTRACT: Corrosion of reinforcing steel is a major contributor to early deterioration of concrete structures in a coastal environment. This paper presents the results of a five year field monitoring study involving 25 reinforced concrete panels exposed to a marine tidal zone. Seven different corrosion inhibiting admixtures were used in the panels to compare their effectiveness at preventing or delaying onset of corrosion. This study was funded by the Hawaii Department of Transportation, Harbors Division, to evaluate corrosion inhibiting admixtures when used in concretes made with basalt aggregates common to Hawaii and other Pacific Islands. The corrosion-inhibiting admixtures included in this project were categorized into two types. Type 1 admixtures attempt to reduce the concrete permeability—Xypex Admix C-2000, latex modifier, Kryton KIM, fly ash, and silica fume were used in this study. Type 2 admixtures attempt to raise the chloride concentration threshold at which corrosion of the reinforcing steel is initiated—Darex Corrosion Inhibitor (DCI), Rheocrete CNI, Rheocrete 222+, and FerroGard 901 were used in this study. Based on half-cell measurements taken over a five-year period, it was concluded that the control panel with lower water-cementitious material ratio (0.35) performed significantly better than the panel with higher watercementitious material ratio (0.40). It was also concluded that panels with Type 1 admixtures recorded significantly lower half-cell potentials than the corresponding control panel. Panels with Type 2 admixtures showed varying degrees of corrosion probability based on half-cell measurements.

1

INTRODUCTION

Reinforced concrete structures along Hawaii’s shoreline are exposed to a marine environment that promotes corrosion of the reinforcing steel. In its advanced stages, corrosion of the reinforcing steel requires that the structure be rehabilitated to remain in service. Currently, rehabilitation of corroded structures is requiring substantial resources from the Hawaii Department of Transportation construction and maintenance program. The study reported in this paper was intended to identify effective and reliable corrosion inhibiting measures that will delay the onset of corrosion, prolong the service lives of future structures, and reduce maintenance costs for deteriorating structures. The study was conducted in three phases. The first part of the study was a field investigation aimed at evaluating the effectiveness of corrosion inhibiting measures in existing structures along Hawaii’s shoreline. Results of this phase are not discussed here. Phase II of the research included a laboratory study to develop practical and reliable guidelines for the production of concrete mixtures with low permeability. The laboratory study also investigated the corrosion inhibiting effects of mixture proportions such as pozzolanic materials and corrosion inhibiting admixtures.

Phase III of the investigation was the initiation of a long-term monitoring program to assess the performance of slabs produced using the most promising inhibiting methods identified in the laboratory study. 2

LABORATORY ACCELERATED TESTS

An accelerated laboratory corrosion study was performed based on ASTM G 109-92. The focus of this portion of the project was to develop practical and reliable guidelines for the production of concrete mixtures with low permeability and increased resistance to corrosion. Figure 1 shows the laboratory specimens stored in the University of Hawaii (UH) structures laboratory basement. A single ASTM G 109 cycle consists of ponding for two weeks with a 3% NaCl solution, and drying for two weeks. A total of 656 individual specimens representing 100 different concrete mixtures were included in the accelerated laboratory corrosion study. Ponding cycles initiated between 1999 and 2001. At the time of writing this paper, over 200 of these specimens have not shown signs of corrosion and are still being subjected to ponding cycles in the laboratory.

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Figure 1.

Lab specimens at University of Hawaii. Figure 2.

Nine types of concrete mixtures using Hawaiian sources of aggregates were produced using eight types of admixtures intended to inhibit corrosion. Compressive strength, elastic modulus, Poisson’s ratio, and slump tests were performed for all of the mixtures. Concrete permeability, chloride concentration, pH, and air content were also measured.

3

FIELD EXPOSURE PANELS

The focus of this portion of the project was to initiate a long-term monitoring program for slabs produced using the most promising inhibiting methods identified in the laboratory study. A total of 25 reinforced concrete panels were fabricated in the structures laboratory at UH. The panels were installed at the field site at Pier 38 in Honolulu Harbor one week after concrete placement (Figure 2). Each panel was suspended in the seawater such that the bottom of the panel was always submerged, the middle of the panel was in the tidal zone, while the top of the panel was above the highwater level. Each mixture used to produce a reinforced concrete panel was tested for compressive strength, elastic modulus, Poisson’s ratio and air content, and each panel was tested annually for chloride concentration, pH and half-cell potential.

4

CONCLUSIONS

Based on the observations from the accelerated lab tests, the following conclusions were drawn: 1. Corrosion was observed in all specimens that had recorded a macro-cell current exceeding 10 µA.

Installation of panels at Honolulu Harbor.

This current measurement is therefore an accurate predictor of the presence of corrosion. 2. Based on the results of this study, it would appear that the potential limits for the half-cell measurements have shifted. More accurate results were obtained if the half-cell limits were modified from –200 mV and –350 mV to –100 mV and –200 mV, respectively, when using a copper sulfate electrode (CSE). 3. The chloride concentration at the level of the reinforcing steel did not show a strong correlation with observed corrosion. Chloride levels ranged from 0.5% to 6.5% by mass of cement. 4. The pH level at the top reinforcement did not show any correlation with the severity of corrosion of the reinforcement. All specimens in this study had pH levels between 12 and 13. Based on the observations during the field exposure study, the following conclusions were drawn: 5. The control panel using Kapaa aggregates with water/cement ratio of 0.35 had chloride concentration levels that were significantly lower than the control panel with 0.40 water/cement ratio. 6. Half cell readings on the Kapaa control panel with 0.35 water/cement ratio show a low probability of corrosion. Half cell readings on the Kapaa control panel with 0.40 water/cement ratio are significantly higher indicating a 90% probability of corrosion after 40 months field exposure. 7. Panels with 20 liters per cubic meter (4 gallon per cubic yard) doses of DCI or Rheocrete CNI showed significant reduction in half-cell readings compared with the control specimen. 8. Panels with silica fume or fly ash showed significant reduction in the half-cell readings compared with the control specimen.

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Hydrophobic treatments on concrete – Evaluation of the durability and non-destructive testing M. Raupach & T. Büttner Institute of Building Materials Research, Technical University of Aachen, Germany

ABSTRACT: For many years hydrophobic agents have been used to reduce the water ingress of building materials. Since the applications of traditional materials like waxes or oils in the Mayan culture the chemical composition of water repellent materials has been further developed. However, the question of the durability of hydrophobic treatments especially used as a surface treatment on concrete surfaces is still not fully answered. The central aim of this work was the evaluation of possible deterioration mechanisms that might lead to a loss of effectiveness of water repellent treatments and hence their durability. In order to investigate a spectrum of commercially available products, six different hydrophobic agents with different organic residues as well as different solvent types were investigated. These were applied to two different types of concrete. The specimens were aged artificially by accelerated carbonation, exposure to UV-light and high-alkaline environment. After the artificial ageing different performance parameters were measured, such as contact angle, water absorption and impregnation depth. Furthermore, some samples were analyzed with the new NMR-Mouse® technology. This technology reveals to be a promising possibility of measuring the impregnation depth of hydrophobic treatments non-destructively.

1

INTRODUCTION

2

The main damage mechanism of steel reinforced concrete structures exposed to outdoor weathering is the ingress of water and water pollutants, such as chlorides and sulfates. Both result in the corrosion of the concrete or even the steel reinforcement. Therefore the reduction of the water uptake of the porous material can effectively be used to prevent such damages. Especially in the field of conservation and repair of concrete structures hydrophobic treatments are widely used because the application can be fairly easy done and the appearance of the structure is usually not altered. Despite the relative long time “modern” hydrophobic treatments have been used, the mechanisms which may lead to the degradation of their performance have not been clarified yet. Investigations of the durability are mostly based on the performance of treatments applied to outdoor structures, but in general, the important technical details about these applications are only known to a very limited extend. The study had two main goals: the identification of different possible degradation mechanisms; and, reproducing these in a specific laboratory setup. Then the influence of these artificial ageing conditions on the performance of hydrophobic treatments could be evaluated. Figure 1 shows the interaction of hydrophobic treatments with the concrete substrate and the possible factors influencing the durability.

INVESTIGATION

In the current research two different types of concrete were used. A rather porous substrate was represented by a C20/25 according to DIN EN 206-1:2001. A rather dense material was represented by a C040 according to DIN EN 1766:2000. In order to avoid the influence of any organic substance, no fly ash, silica fume, super-plasticizer or release agents were added to the concrete.

Exposure by: Concrete: - UV-Radiation - Carbonation - Alkalinity - Intrusion of dirt by gases - Humidity - Temperature

- Porosity - Hydration stage - Carbonation depth - Alkalinity

DURABILITY

Hydrophobic Agent: - Active agent - Active content - Organic residue - Type of solvent

Concrete combined with the hydrophobic agent - Impregnation depth - Distribution of active content over depth

Figure 1. Interaction of hydrophobic treatments and concrete concerning the durability.

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3.1

hydrophobicagent

Storagein poresolution: -pH-value13.3 -23°C; Closedbox -t = 28; 35 d

AcceleratedCarbonation 2.0 ± 0.4 Vol.-% CO2 t = 28; 35 d

The idea behind the use of a NMR technology in combination with hydrophobic treatments was to establish a measuring system, which is able to detect the impregnation depth non-destructively. Conventional NMR devices used for computed tomography or materials testing are rather large and immobile due to its weight and size. The NMR-Mouse® is a handheld device with two small permanent magnets, which are mounted on an iron yoke with anti-parallel polarization to form the classical horseshoe geometry. The radio frequency field is generated by a surface coil which is placed in the gap. The measurement area of this specific NMR-Mouse® has a size of approx. 4.0 cm² and a depth of 200 µm. The first step of the study was to determine water distributions in a sample. A concrete core was put into water with the treated side upwards, so water penetrated through the non-treated surface from the bottom side into the concrete. The NMR-Mouse® was then applied to the treated side and the intensity of the relaxation signal was measured in 0.5 mm steps. The measurements of two specimens show an impregnation depth of 2–3 mm. This impregnation depth could also be obtained with the commonly know method by splitting the specimens in two halves and spraying water on the freshly broken surface.

UV(A)-Radiation (290 – 450 nm) t = 28 d

Figure 2. Overview over the different types of specimens and artificial ageing methods.

Figure 3. Standardized NMR intensity over depth indicating water saturated and hydrophobic areas in the concrete.

4 The hydrophobic agents were delivered by a manufacturer of pure silane/siloxane products and were diluted according to the technical specifications. The selected active agents represent commonly used products. The originally produced concrete slabs were divided into three different types as shown in figure 2. The different storage conditions are also indicated in this figure. After the various storages the change of performance of the water repellents was quantified using for different methods: • • • •

3

NMR-Mouse® Measurements

Contact angle measurements, Capillary water absorption, Impregnation depth and NMR-Mouse® measurements.

RESULTS AND DISCUSSION

The following paragraph will only present selected results of the NMR-Mouse® measurements, the other results are shown in the extended version of the paper.

CONCLUSION AND OUTLOOK

The results of the present study can be summarized as follows: • Two hydrophobic agents showed an increase in effectiveness upon exposure to high alkaline solution; • The influence of the UV-radiation is limited to the surface only, but in two cases a loss of the hydrophobic performance could be measured. The reasons for this behavior are being investigated in ongoing tests; • A change of the pore structure due to carbonation seems to affect the hydrophobic performance of water repellents substantially. The reasons for this alteration are not yet clarified; • The NMR-Mouse® technology seems to be a suitable way to investigate the impregnation depth indirectly in a non-destructive manner. Future research will focus on further developments of the NMR-Mouse® technology for on-site measurements. It also will be used in tests aimed to answer the questions raised in this study regarding the durability of the hydrophobic effect.

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Application of the NMR-technique to concrete-coatings J. Orlowsky & M. Raupach Institute of Building Materials Research (ibac) of RWTH Aachen University, Germany

M. Baias & B. Blümich Institute of Technical and Macromolecular Chemistry (ITMC) of RWTH Aachen University, Germany

ABSTRACT: The application of the NMR-MOUSE® (Nuclear Magnetic Resonance Mobile Universal Surface Explorer, registered trademark of RWTH Aachen University) for performance tests on concrete-coatings is studied. In combination with a special lift, measurement depths from 0 to 5 mm with a minimum step size of better than 10 µm are possible. The measurement is non-destructive, so that not only geometrical properties such as the thickness of different coating layers can be determined but also functional properties can be investigated like the ingress of water into the coating as well as ageing processes.

1

NMR-TECHNIQUE

NMR is a method to interrogate molecular properties of matter by interrogating atomic nuclei with magnetic fields and radio-frequency irradiation. This method allows accurate determinations of the chemical structures, molecular order and dynamics, morphology of semicrystalline polymers, material heterogeneities, and spatial structures in terms of magnetic resonance images. Instead of using highly homogeneous magnetic fields as in conventional NMR spectroscopy, one can perform relaxation measurements in inhomogeneous fields. Such measurements are of great value in well logging and material science. Inhomogeneous magnetic fields can easily be realized by one-sided permanent magnets and surface coils. An NMR device of this kind is the NMR-MOUSE®. With the profile NMR-MOUSE (Figure 1), high resolution profiles can be recorded by placing the sensor on a stepper-motor driven, high-precision automatic lift which changes the distance between the sensor and the sample. The lateral dimensions of the sensitive slice are 20 mm by 20 mm. The most important contrast parameters employed by the NMR-MOUSE are the signal amplitude corresponding to the proton density, and the transverse relaxation times T2 which relates to the molecular mobility and the effective self-diffusion coefficients of liquids within the sensitive volume. This information is obtained using the Carr-Purcell-Meiboom-Gill (CPMG) pulse sequence.

Figure 1. Set-up consisting of the NMR-MOUSE and the lift used for NMR measurements.

mortar specimens were investigated. Their compositons are specified in Tab. 1. From the systems OS5a-2 and OS5a-3 also films were tested.

3 3.1

2

TEST PROGRAM

Two classes (OS5 and OS11) of surface protection systems according to a German guideline applied to

RESULTS Thickness of the surface protection systems

Measurements of the proton density in steps of 50 µm allow conclusions about the composition and layer thickness of the surface protection system. Figure 2 shows the sequence of layers of a coating by the

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Table 1. Composition of the investigated coatings.

10 Shortcut Base coat

Inner layer

OS 5-4 after 10 h in water at 23°C OS 5-4 at 23°C and 50% rel. humidity

Top layer

OS5a-2 AcrylateAcrylatedispersion dispersion filled (solid content: ∼ with sand 18 Vol.-%) OS5a-3 Polymer-cementsand-composite (acrylic acid ester-copolymer) OS5a-4 Copolymerdispersion (solid content: ∼ 23 Vol.-%)

OS11-3 Epoxy

1,4

Amplitude

Acrylatedispersion (solid content:∼ 53 Vol.-%) 2 layers of acrylatedispersion (solid content:∼ 55 Vol.-%) – 2 layers of acrylatedispersion (solid content: ∼ 55 Vol.-%) Polyure- Polyure- Polyurethane thane thane with sand

8 6 4 2 0 0

1

2

3

4

5 6 7 Echo time in ms

Figure 3. T2 decay curves for OS5a-4 showing water ingress after 10 hours in water at 23°C.

Storage in water leads to a softening of the dispersion, which is noted by a lengthening of the transverse relaxation time (black curve in Fig. 3).

Amplitude Top layer

OS5a-2

1,2

4

1,0 Inner layer

0,8

Base coat

0,6 0,4 Mortar

0,2 Surface 0,0 0

500

1000

1500

2000

2500 3000 Depth in µm

Figure 2. Proton density as function of depth through OS5a-2 (50 µm resolution).

corresponding proton density profile. The different coating layers can clearly be identified. 3.2

Water ingress

Figure 3 shows two decay curves of OS5a-4, measured inside the top layer. Water ingress after storage in water for 10 hours can be clearly detected. The water could only pass the OS5a-4 system because the four sides of the mortar were coated with epoxy.

CONCLUSION AND OUTLOOK

The reported analysis is part of a research project (project partners: German Polymer Institute (DKI), Instiute of Technical and Macromolecular Chemistry (ITMC) and Institute of Building Materials Research (ibac)) where different test methods are used to characterize weathered concrete-coatings for comparison with data obtained by the NMR-MOUSE. The current results demonstrate that the NMR-MOUSE is a promising tool for non destructive testing of coatings on concrete. The thickness of the coating and its constituting layers can be measured. Furthermore, water ingress through the surface protection system can be detected as well as changes of the material due to weathering. The currently employed lift of the NMR-MOUSE is designed for indoor measurements on specimens. Future measurements with the NMR-MOUSE will be conducted on buildings outside. For routine applications, the influence of temperature and steel reinforcement in the concrete structure on the NMR signals has to be determined. Actually we plan to build a robot carrying the sensor that moves in x, y, and z-directions as we believe, that use of the NMR technology will result in a powerful new test method for on-site measurements.

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Durability of adhesion of epoxy coatings on concrete; causes of delamination and blistering M. Raupach & L. Wolff Institute of Building Materials Research of Aachen University, Germany

ABSTRACT: Epoxies have already been used for impregnation and coating of concrete parts since more than 40 years. Nevertheless delamination and blisters sometimes occur when the coating, especially the epoxy base coat, is exposed to humidity and alkaline environment from the concrete. The phenomenon of blistering of epoxy coatings has been reported for the first time about 35 years ago: Some weeks or month after application usually liquid filled blisters occur within or under the coating. In recent research the mechanisms of blistering were mainly described as osmotic transport processes in the contact zone between concrete and base coat. In a research project at the Institute of Building Materials Research of Aachen University, ibac, funded by the German Research foundation DFG it could be shown that osmotic transport processes are not involved in the effect of blistering. In fact interactions between the epoxy base coat and the capillary system of the concrete, depending on the composition of the epoxy and the capillarity of the concrete, lead to flaws in the base coat and variances of crosslinking of the epoxy. Change of pressure in the partially water filled pore system of the concrete depending on humidity variation then lead to blistering in the area of these flaws and destruction of the coating.

1

INTRODUCTION

The damage pattern of the so-called “osmotic blistering” is nearly as old as the application of concrete coatings itself. Nowadays this damage pattern is held responsible for about 3% of all coating damages. Basically, two kinds of blisters can be discerned at epoxy coatings on concrete: − Blisters occurring immediately during or after application: These are often due to air entrapped during manufacturing in the coating material or to air escaping from the substrate. − Blisters occurring later: These blisters appear clearly after the application of the coating material, i.e. a few weeks up to months, sometimes even years later. In most cases, these blisters are filled with liquid. Figure 1 exemplarily shows these two different kinds of blisters. The left picture shows an opened blister in a PUR-coating as a result of entrapped air in the coating material. The right picture displays a cross-section of a liquid-filled blister of a crack bridgeable coating with delamination between base coat and main surface protection layer. The first-mentioned blisters are mostly due to faulty manufacturing during application. Therefore, in this paper, only the mechanisms of the delayed blisters are considered. Within the framework of a research project funded by the Deutsche Forschungsgemeinschaft DFG (Ger-

man Research Foundation), extensive investigations regarding the mechanisms of blistering of reactive resin coatings on concrete were conducted at the Institute of Building Materials Research (ibac), RWTH Aachen University during recent years. As opposed to the hitherto understanding of the mechanisms of blistering, the investigations of this research work furnished basically new findings regarding their causes. − Osmotic transport processes are irrelevant at the development of the blisters because there is no semi-permeable membrane for the relevant substances of the blister liquids. − An epoxy base coat applied to concrete is not homogeneous. Moreover, due to physical interactions with the mineral substrate, separations within

Figure 1. Application-related blister resulting from entrapped air in the coating material (left); delayed liquidfilled blister with delamination between base coat and main surface protection layer (right).

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the liquid epoxy may occur which can exert a major influence on the adhesion to the mineral substrate. − A chromatography effect of the investigated epoxies at the application on a mineral substrate could only be proven for benzyl alcohol in individual cases. − The pressure within a capillary pore structure of the concrete which is necessary for a delamination of a base coat or coating clearly ranges below the existing pull-off strength, in particular cases by more than 90%. − Local imperfections within the crosslinked epoxy base coat are the main source for the development of a blister above the base coat. Thus, these imperfections represent the necessary link between the capillary pore structure of the concrete and the inside of the blister.

The reason for this difference is the incomparable stress of the coating bond by a uniaxial tensile load at the determination of the pull-off strength in contrast to a peel stress between coating and substrate at the development of a blister. Both these basically different stress states at the determination of the pull-off strength as well as at the delamination of a coating are schematically presented below in figure 2. At a developing blister, the illustrated peel stresses do not occur only in one direction, they appear, however, in an axially symmetric way as a result of the circular shape of the blister. In contrast to previous findings, capillary pressures measured inside the capillary pore system of concrete of up to about 0.3 N/mm², can thus already lead to a local adhesion failure of a coating.

Whereas in previous works regarding the mechanisms of blistering the tensile bond strength of the respective coating layer was put in relation to the pressure necessary for a delamination of the coating, the laboratory tests and FE-calculations. showed that the stress on the bond at the edge area of the blister cannot be correlated directly with pull-off strength of the coating. The pressure of at least 1.5 N/mm² which has up to now been assumed to be necessary for a delamination was thus too high by more than a decade. The laboratory tests as well as the FE-calculations proved that a local adhesion failure can occur already at pressures of about 0.1 N/mm².

2

Figure 2. Comparison of tensile load (left) and peel stress (right) in a bond joint.

SUMMARY

In the presented paper different typical damage patterns occurring at trafficable coating systems were specified. The focus was directed on damage patterns whose cause was often described with the so-called “osmotic” blistering in the past. Results of a research project conducted at ibac impressively pointed out that osmotic transport processes at the development of liquid-filled blisters within the polymer coating as well as of large-scale delaminations of single coating layers are not relevant. So the main explanation of osmosis as cause of the damage was assumed to be a pressure as high as the respective pull-off strength of the coating necessary for its delamination. Laboratory tests as well as FE-simulations of the bond between coating and concrete, however, indicated that a local delamination of the coating may already occur at significantly lower maximum pressures of about 10% of the respective pull-off strength. To avoid a blistering by a careful application of the coating is only possible to a limited extent. Moreover, a modification of the existing epoxy formulations is necessary to eliminate the development of local flaws or large-scale separations at the application of the base coat on the concrete. The development of suitable test methods is planned to targetedly examine the “susceptibility to blistering” of epoxy base coats in the laboratory.

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Protection and rehabilitation of chemically high stressed concrete surfaces by thin glass K. Schubert Hochschule Karlsruhe Technik und Wirtschaft, Germany

B. Hillemeier & P. Osselmann Technische Universität Berlin, Germany

ABSTRACT: The German wastewater net is approximately 1 Million km long. The distance from earth to the moon is 385.000 km. Approximately 9% [1] of this network is in need of immediate repair. Another 30% will need middle or long term repair. Due to water use reduction the amount of damaged piping will tend to increase. The total amount of wastewater facilities in need of repair in the EU has not been sufficiently estimated. The Technical University of Berlin has developed a new surface protection system for chemically high stressed concrete surfaces and components. The system consists of a flexible thin glass layer and mineral polymer silicate grout. Due to its thinness and flexibility the glass can be applied overlapping without any butted joints. Uses for this system are in new construction, refurbishing and retrofitting of wastewater plants, and as a long lasting protection for facades in conjunction with the use of transparent adhesives.

1

INTRODUCTION

Recent research and development in the area of mineral construction materials has shown that it is possible to use high performance concrete effectively to produce elements and structures which are able to resist chemical attack in sewage systems and aggressive ground water or soils. Pipes which are made from acid-resistant, high performance concrete are monolithic and satisfy all the stipulated chemical and physical requirements [1]. The “densest packing” of the aggregate mix and grading it down to the finest range minimizes the amount of cement to 50% of that required by a conventional high-performance concrete [2]. Optimizing the binder made from reactive powders and special cements also produce an acid-resistant concrete which is suitable for sewage systems, cooling towers, and foundation structures in aggressive groundwater. This concrete is resistant to dissolving and expansion attack and is highly impermeable to the diffusion of chlorides, gases, and water. The dense packing and low cement content reduces the concrete shrinkage to approximately 40% less than that of a conventional concrete. Only biogenic acid corrosion which cannot be prevented by structural design measures requires additional protection for pH values below 3.5 (continuous exposure) and 2.0 (periodic exposure). For these conditions glass provides the best Protection. The bonding of glass and concrete to make a composite material is one innovation aim for the near future. Glass is flexible and can be adjusted to the geometric conditions on the construction site. Thin glass is flexible because the flexibility is a function of the moment of inertia. At the Technische Universität

Figure 1. Sewage system collector after biogenic sulphuric acid corrosion.

Berlin a mortar creating a high-strength chemical bond between glass and concrete has been developed. In most cases, the two materials were only stuck to each other via weak Van der Waals forces. However, glass can also be bonded to concrete chemically using a special polymer silicate with high alkalinity. The bonds between the glass and the concrete using polymer silicate involve covalent, Si-O-Si bridging bonds to form a chemical compound type bond. With both innovations, a special glass lining system has been developed using thin glass and silica adhesive. The glass is supplied by Schott AG and the adhesive by MC-Bauchemie. Figure 2 shows a concrete pipe clad with thin glass at the Institute of Construction Engineering at the Technische Universität Berlin.

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in Gelsenkirchen-Heßler have shown that pores are opened but the directly adjacent glass remains undamaged at 300 bar water pressure. The opened areas are then covered with glass and mortar patches. The immediate objective is to optimize the time needed to apply the glass. Exterior water pressure is resisted by the compound system without problem. A corresponding longevity test is taking place at the Institute for Civil Engineering at the TU Berlin with a water pressure of 5 bar from the backside, equal to 50 m water column since October 2004. 3 Figure 2. 1600 mm concrete pip, clad with thin glass, overlapped and with labyrinthine joints at the Institute of Construction Engineering at the Technische Universität Berlin.

The current research work and test installations on construction sites are expected to reveal the possibilities and limitations of this novel glass-lining system. Because it is thin, the glass sheets are applied to the background overlapping one another so that the glass joints are labyrinthine. The corrosion resistance of glass is exemplary in comparison to other materials, particularly because the thin glass is a borosilicate glass.

2

QUALITY ASSURANCE

As in all technically sophisticated solutions, the application of the Glass-Lining system requires a quality assurance adapted to the task in the planning, controlling, and execution. Special care has to be dedicated to the substrate. The reconstruction of the surface follows the power washing of the damaged surface and the restoration with shotcrete and fine filler. The surface to be coated with glass is curved, but should be as smooth as possible. Before the coating with thin glass the surface has to be coated with the KONUTECT adhesive with approximately 2 mm thickness and allowed to harden one day. Afterwards the glass is to be applied only by trained staff. It is a learned skill to apply the thin, brittle glass to the adhesive applied with a grooved trowel free of pores. After the restoration covered pores are opened during a power wash test with 300 bar pressure to simulate high pressure cleaning. Tests in the Schwarzbach sewer pipe

DISCUSSION, CONCLUSIONS AND ACKNOWLEDGEMENTS

The advantages of the glass lining compared to conventional polymer based protection systems are: − Extremely high chemical resistance. − Flexible glass can be applied on curves and concavities without any pre-forming. − Simple Repairs of surface damage are made with the same components. − Polymer/plastic protection systems retain osmotic pressures because the polymer/plastic layer is not closed against the diffusion processes. − Glass does not creep, remains undeformed over defective substrates, and therefore will not peel. Groundwater and hydrostatic pressure can lead to bumps and deformities in plastic linings. − The facility which is to be protected or repaired can be reopened and reutilized within hours. The glass lining protects the polymer silicate grout which can harden sheltered from an aggressive environment. Uses for this system are in new construction, the refurbishing and retrofitting of wastewater plants, as well as long lasting protection for facades in conjunction with the use of a transparent adhesive. REFERENCES [1] Berger C., Lohaus J. “Zustand der Kanalisation in Deutschland” Deutsche Vereinigung für Wasserwirtschaft, Abwasser und Abfall e.V., Hennef, 2004. [2] Hillemeier B., “Säureresistenter Beton für Abwasserrohre—Möglichkeiten und Grenzen”, BWI Betonwerk International, Vol. 6, Dec. 2001, pp. 52–55. [3] Hillemeier B., Hüttl R., “Hochleistungsbeton— Beispiel Säureresistenz, Betonwerk und Ferstigteiltechnik”, Vol. 1, 2000, pp 52–60.

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Bonded concrete overlays and patch repairs

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Adhesion – A challenge for concrete repair L. Czarnecki Building Materials Engineering Department, Faculty of Civil Engineering, Warsaw University of Technology, Warsaw, Poland

ABSTRACT: The adhesion between overlays and concrete substrate is one of the most important factors that affects the reliability and durability of repair. Achievement of maximum bond strength is basic recommendation for various types of repair joints. In this paper, the necessity of the above recommendation has been discussed with regards to: repair material type, strength of concrete substrate and environmental conditions of bond service. The possibilities of creation higher level of adhesion as well as economical and technical limitations of the adhesion improvement have been stressed. The need for new generation materials—High Adhesion Repair Materials, HARM for repair of the High Strength Concrete, HSC is formulated.

The European Standard EN 1504-10 defined the bond as the adhesion of the applied product or sys-tem to the concrete substrate. Adhesion is defined as adherence of two materials in contact or in a more colloquial way “of being stuck together”. The adhesion can be considered ass a fundamental issue in repair of concrete structure. From a scientific point of view, fundamentals of adhesion belong to the nano-technology and surface physic and chemistry in particular. The corresponding engineering research area will include mechanical interlocking on the rough surface of a substrate on the micro-scale. The adhesion between overlays and concrete substrate is one of the most important factors that affects the reliability and durability of repair. Achievement of maximum bond strength is basic recommendation for various types of repair joints. The question arises how high bond strength is high enough? Some consideration on that has been presented in the paper with regards to: repair material type, strength of concrete substrate and environmental conditions of bond service. Adhesion is a decisive factor of repair effectiveness. In EN 1504 the adhesive bond level are formulated in two ways of expression of results:

In rough estimation, various oriented cracks occur regarding to the combination of values: concrete tensile strength, ftC; repair material tensile strength, ftR and adhesion strength, fA in relation to internal stresses, σt (Tab.1). There is a need for the compromise between economy and ageing sensibility (durability), compatibility level and reliability requirement. It is necessary to remember that ageing is more dangerous in case of polymer composite than mineral one, and external protection coating makes a difference too. According to the author’s suggestion: fA = (1.15 ÷ 1.30)ftk depends on the various repair materials and on different system: with- and without protection coating.

− threshold value in MPa; towards structural repair (EN 1504-3): >2,0MPa (Class R4), >1,5 MPa (Class R3); − pass/fail criteria towards structural bonding (EN 1504-4) e.g. hardened concrete-to-hardened concrete or fresh concrete-to-hardened concrete: the test shall result in fracture in the concrete. The adhesion of repair joint is effective if it enables the load transfer and ensures even distribution of stresses. It can be reached if bonding materials satisfy the conditions of physico-chemical compatibility and guarantee tightness of the joint.

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Table 1. Various kind of cracks in repaired system. Diagram

Relation

ft R < σ t < ft C ≤ f A f A < σ t < ft C ≤ ft R ft C < σ t < ft R ≤ f A ft R < σ t < ft C ≤ f A

σ t < ft C ≤ ft R ≤ f A σt—internal shrinkage stress, ftR—repair material tensile strength, ftC—concrete tensile strength, fA—adhesion strength.

Figure 1. Required adhesion strength, fA vs. concrete compressive strength, fCm adequately to the concrete strength classes. Areas of “adhesion usability” adequate to various kind of repair materials: MCC, PCC, PC are described.

Above considerations, give us a set of required adhesion strength ascribed to the given mechanical strength of concrete substrate. If we put together the concrete strength classes according to the EN 206-1 and required adhesion values than compare them with features of various repair materials, the computation diagram for repair usability can be built up (Fig.1). It is obvious that existing repair materials are good enough for ordinary concretes. Categorization “top down” for usability of repair materials for concrete substrate of various classes will be as follow: − PC—(Polymer Concrete and Mortars) below C60/75. − PCC—(Polymer Cement Concrete and Mortars) below C40/50. − MCC—(Modified Polymer Cement Concrete and Mortars) below C25/30. There is no effective (adhesive) enough repair materials for High Strength Concrete, HSC—classes C70/85, C80/95, C90/105, C100/115 and more. Of course, when we think about concrete repair, we usually have an old and weak concrete in mind, e.g.—below C20/25. However, an old concrete is not necessarily a weak one, it could be quite healthy and strong. Moreover, a damage caused by accidental impact (e.g. due to the traffic), overloading, settlement and explosion could also happen to new high-strength concrete elements. There is need for new generation

of repair materials—High Adhesive Repair Materials, HARM. The work on High Adhesive-Polymer Cement Concrete, HA-PCC is already under way. The PCC with adhesion bond (pull of test) around 5 MPa has been received till now. It is worth to stress that High Strength Concrete need—by nature—more aggressive surface preparation, which could result—in turn—in the increase of microcracks in concrete substrate. In such case “bonding layer” on concrete substrate below the repair materials will be justified or even needed. The repair materials—as the rule—should contain microfibers for better “microcrack bridging”. There is a need not only for new generation of repair materials but also for new generation of repair systems and methods. There are lots of factors affecting adhesives strength involved with; − concrete substrate: mechanical strength, surface roughness, microcracks, porosity, dampness, impurities, etc., − repair materials: viscosity, wetting (surface tension), setting shrinkage, thermal expansion, elastic modulus, creep, etc., − environmental impact: transportation phenomena (diffusion, osmosis, capillary suction), temperature level and change of temperature, humidity level and change of humidity, mechanical loading, degradation (ageing, carbonation, corrosion). Some of them could increase adhesion level, but— in general—are the reasons of adhesive destruction; in case of environmental usually gradually due to the time. Considering the factors affecting adhesion bond the High Adhesive Repair Materials need is still obvious.

ACKNOWLEDGEMENT In the paper has been used some ideas and figures formerly published in “Adhesion in Interfaces of Building Materials: a Multi-scale Approach (L. Czarnecki, A. Garbacz, eds), Advances in Materials Science and Restoration (AMSR), 2007. Author would like express acknowledgement to AMSR Editor prof. dr FH Wittmann and Aedificatio Publishers. This work has been prepared in the framework of the Warsaw University of Technology grant nr 504 G 1080 7007.

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Concrete repair and interfacial bond: Influence of surface preparation B. Bissonnette, A. Nuta, M. Morency & J. Marchand CRIB—Laval University, Department of Civil Engineering, Québec, Canada

A.M. Vaysburd Vaycon Consulting, Baltimore, Maryland, USA

ABSTRACT: The aim of concrete repairs is to prolong the useful service life of an existing structure, to restore its load-carrying capacity and stiffness, and/or to strengthen its members. A prerequisite to achieve adequate composite action is lasting bond between the existing substrate and the new-cast material. In this respect, concrete surface preparation prior to repair material application is of critical importance. In fact, regardless of the repair material and application method employed, the quality of the surface preparation prior to repair will often determine whether a repair project is a success or a failure, and whether or not a repaired structure is durable. As part of a research project intended to lead to the development of performance criteria for surface preparation of concrete prior to repair, the experimental program reported herein focuses on the influence of mechanical integrity and roughness of the substrate upon bond strength development. The results indicate that both parameters can vary quite significantly depending on the removal technique and that their combined effects control to some extent the bond strength level that can be achieved. Still, the most important parameter seems to remain the mechanical integrity of the substrate. In that regard, the use of impacting methods such as jack hammering leaves significant damage at the surface, which can easily outweigh the benefits of an increased roughness.

1

INTRODUCTION

The general objectives of the research work reported in this paper are to provide the industry with guidelines for selecting the surface preparation method that will yield the best results with regard to bond strength and durability. In addition, the identification of performance criteria will enable a field assessment of the quality of surface preparation prior to repair works. The specific objectives of the project are to evaluate the respective influence of roughness and mechanical integrity of the prepared surface on bond strength development.

2

METHODOLOGY

In order to cover a sufficiently large spectrum in terms of roughness and, at the same time, to address most usual surface preparation techniques, the following methods were selected for investigation: sandblasting (SaB), shotblasting (ShB), scarifying (Sc), 15,000-psi handheld hydro-jetting (HJ), and 7-kg jack hammering (JH). In addition, to avoid the presence of induced damage and isolate the effect of roughness upon bond strength, artificially profiled test specimens were cast. Two series of 16 concrete slabs (625 × 1250 × 150 mm) were manufactured for the test program.

The first series was made with a 35-MPa ready-mix concrete, while the second series was prepared using a 25-MPa concrete. After surface preparation, evaluation of surface integrity and characterization of surface roughness were performed. The slabs were then repaired (75-mm overlay) with a 45-MPa repair concrete. For a wide and objective characterization of surface roughness, optical profilometry has been investigated in this study. The parameter used to describe surface roughness quantitatively is the average halfamplitude (Ra) of the profile. Surface integrity of the prepared test slab was evaluated through pull-off experiments and, on an exploratory basis, Schmidt hammer soundings. All 32 repaired test slabs were characterized exhaustively for bond strength with a combination of 16 pull-off tests, four direct tensile tests and four torsional shear tests on each of them. 3 3.1

RESULTS AND ANALYSIS Surface roughness

The largest half-amplitude values (1.50–3.75 mm) were obtained with the jack hammer and hydrojetting, while the lowest values were recorded respectively for the scarified, the shotblasted and the sandblasted surfaces (<1.00 mm). It could also be observed that for

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3.2

Substrate cohesion

For a given substrate quality, the average cohesion values obtained with sandblasting, shotblasting and scarifying are all close from the corresponding base concrete tensile strength. These substrates were virtually undamaged by the surface preparation operations. In comparison, the average cohesion values recorded for the jack hammered slabs are significantly lower, especially in the 35-MPa series. This is assumed to be a consequence of surface defects induced by the hammer tip, as reflected by the prevalent number of pull-off specimen failure occurrences near the surface. In the case of hydro-jetting, the lower recorded cohesion values are most likely not due to damage, as the bond strength test results subsequently show, but rather to a pull-off test bias for that given type of surface profile. The trends exhibited by the average Schmidt hammer results for the various investigated surface preparation techniques are somehow similar to those observed for the cohesion test results. 3.3

Avg. pull-off bond strength (MPa)

all slabs and CSP templates, surface roughness was uniform.

Bond strength

Except for the slabs prepared by jack hammering, the pull-off test results (Figure 1) are close to the corresponding substrate tensile strength values for both slab series. On jack hammered slabs, even though lightweight hammers (7-kg) were used, the recorded pull-off strength values are significantly lower and most of the time (>90%), failure occurred in the interface area. As far as the relationship between pull-off strength and substrate roughness is concerned, it appears that pull-off values slightly increase with the value of Ra, provided that no or limited damage is induced. Where the extent of damage becomes significant, as in the case here of jack hammered slabs, the positive influence of increased roughness is completely offset by the adverse effects of bruising. The bond strength tests performed in direct tension yielded similar trends, except for the effect of the substrate strength. The results actually suggest that the latter has little influence on the magnitude of the bonding forces developed at the interface. The torsional shear bond test results also show similarity with the pull-off data, in that case both in terms of magnitude and observed trends. Again the substrate strength and the presence of damage are

3.0 SaB

2.5 Sc

2.0 1.5

HJ

ShB HJ

SaB Sc

ShB JH

JH

1.0 0.5 0.0 0.0

25-MPa substrate 35-MPa substrate

0.5

1.0

1.5

2.0

2.5

3.0

3.5

Roughness half-amplitude - Ra (mm)

Figure 1. Results of pull-off tests (ASTM C1583) performed after repair on the prepared slabs.

influential parameters. Contrarily to what could have been inferred, roughness does not appear to play a more important role in shear than in tension.

4

CONCLUSION

The investigation has generated useful information for the evaluation and characterization of concrete surface preparation prior to repair. For field surface roughness characterization, by taking advantage of the optical tools, further development should yield a wider CSP value range to cover the majority of surface preparation techniques. Besides, the pull-off test is a convenient field method for evaluating both the mechanical integrity of the concrete surface prior to repair and the repair bond strength. Bond strength of concrete repairs depends on a number of parameters. It has been shown that in absence of substrate-induced damage, tensile bond strength increases with the substrate coarseness. Still, the most important parameter apparently remains the mechanical integrity of the substrate. In that regard, it must be stressed that the use of impacting methods such as jack hammering leaves significant damage at the surface, which can easily outweigh the benefits of an increased roughness.

ACKNOWLEDGMENTS This project has been financially supported by NSERC, the Québec FQRNT Research Fund and industrial partners (complete list provided in the paper).

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Correlation between the roughness of the substrate surface and the debonding risk F. Perez, M. Morency & B. Bissonnette Civil Engineering Department, Laval University, Québec, Canada

L. Courard GEOMAC Department, University of Liège, Belgium

ABSTRACT: This paper presents the influence of substrate surface preparation on the adhesion strength of repaired beams system. Bond between new and old concrete has been the subject of a number of investigations, but in most cases, only adhesion strength was addressed. This parameter was used to estimate the durability and/ or the debonding risk for repaired structure and it’s a generally accepted auditing standards. Usually, surface preparation of the substrate concrete is considered essential to achieve a durable repair because of its influence on the bond strength. To better understand debonding mechanism, in particular these related to surface preparation, roughness parameters were calculated to quantify the influence of surface preparation on the structural behaviour. Using this approach, repair beams prepared by way of four (4) concrete surface preparations were characterized. Results obtained show that repaired beams presenting a substrate with a rough surface permit to achieve a monolithic behaviour of the repaired system. Opposite structural behaviour, with large debonding, was recorded for those having smooth surface. However, all surface preparations used have promoted the same bond strength regardless the roughness of the substrate. The resulting analysis highlights the relation between roughness parameter αrough and the debonding mechanism of repaired beams. Such results will be useful to better predict the performance of concrete repairs.

1

INTRODUCTION

Many existing concrete structures are deteriorated and need to be replaced or rehabilitated. Deteriorations are often localized at the surface due to exposition to severe weathering conditions (de-icing salts and freeze-thaw cycles). Thin bonded concrete repair overlays are among promising rehabilitation approaches to extend the service life of bridge desks. A number of debonding mechanisms have been developed based on field observations and on mechanical considerations [Saiidi et al 90, Granju 04]. A literature survey showed that debonding is initiated in tension perpendicular to the interface [Hilsdorf et al 92, Bijen et al 94]. Therefore, the tensile strength is considered as a critical design parameter and the tensile stress perpendicular to the interface controls the debonding mechanism. Initiated by tension stress, debonding propagation in mode I is due to a combination of tension and shear stresses [Hilsdorf et al 92, Bijen et al 94, Granju 04]. The adhesion, and consequently the durability of a repair, depends on various phenomenons taking place in the interfacial zone [Courard 99]: the wettability of concrete substrate by the repair materials, concrete surface geometry, rheology of the repair concrete, roughness of the substrate surface, etc… The last one is often presented as the most important factor

to achieve a good bond. This improvement is mainly assigned to the increase of the contact surface [Talbot et al 94, Santos et al 07]. But other results suggested that roughness has no relevant effect on the bond strength [Silfwerbrand 90, Courard 99, Perez 05, Belair 06]. This misleading on the influence of roughness come from that roughness is often approached or evaluated with non effective method which give qualitative or not precise enough information.

2 2.1

ANALYSIS Relation bond strength and roughness

The relation between surface roughness and bond strength is plotted in figure 1. In first approximation, it seems that there is no relationship between roughness and adhesion. The results confirmed those obtained by Belair [Belair 2006]. 2.2

Relation between debonding and bond strength

Based on debonding mechanism, it seems to be impossible to observe a debonding until stress stays under the bond strength. This description highlights the prevalence of the tensile strength on the existence and the propagation of debonding. Figure 2 presents the length of debonding measured during loading tests

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Figure 1. Relation between adhesion and the roughness parameters Ra. Figure 3. Relation between debonding length and the roughness parameters αrough.

It exists a roughness threshold above which the risk of debonding is reduced. 3

Figure 2. Relation between adhesion and the debonding length.

in function of the adhesion evaluation. The debonding obtained on SCA-beams indicates that an adherence of 2,60 MPa does not guarantee the monolithic behaviour whereas an adherence of 1,80 MPa (R-JP7S) seems to be sufficient. Adhesion appeared to be a non sufficient parameter to prevent debonding. 2.3

Relation between debonding and roughness

Results below indicated that roughness is a good parameter to guarantee monolithic behaviour. Figure 3 present the length of debonding measured during loading tests in function of the roughness evaluation.

CONCLUSIONS

The aim of the research was to better understand the mechanisms involved in the cracking behaviour of bonded overlays used on reinforced concrete beams. This works have focus on the influence of the surface preparation. A complete evaluation of the bond developed between repair material and substrate has been made. Experimental study, as presented in this paper, emphasizes the influence of surface roughness on the structural behavior of repaired beams and on the debonding mechanism. Although tensile bond has been proposed by many investigators as a tool to evaluate the structural behavior of repaired beams, it is, regard to results obtained, that this parameter alone is not sufficient enough to assess the monolithic behavior, especially for cases of relative flat surface preparation. Results indicate that increase the roughness does not enhance the bond strength. Then, to insure monolithic behaviour, surface treatment must produce a minimal adhesion and must induce a certain level of roughness. While the roughness is higher than a certain threshold, the debonding risk decrease rapidly and monolithic behaviour is reached. By taking the roughness angle αrough as design parameter for repaired structure, it makes possible to evaluate the potential durability of structure even before the casting of the repair material.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Performance of spall repair materials D.W. Fowler & D.P. Whitney The University of Texas, Austin, TX, USA

D. Zollinger Texas A&M University, College Station, TX, USA

ABSTRACT: Spalling of concrete pavements has been commonplace in pavements. In Texas a common failure mode consists of shallow spalls in the transverse directions that occur along transverse cracks. Many repair materials have been used with varying success. A research study was conducted by The University of Texas at Austin and Texas A&M University to characterize the properties of the repair materials. Field observations were made of some of these materials that had been in place for several years, and a field trial was conducted on a busy interstate highway. The best performing materials were those that had very low moduli of elasticity, even though the coefficients of thermal expansion were high. The product of coefficient of thermal expansion and modulus of elasticity equal or less than 125 kPa/ºC was the best predictor of success for the materials tested.

1

INTRODUCTION

Previous studies have confirmed that spalling in concrete pavements is a consequence of the early age delamination cracks that form at essentially the same time that early age transverse cracks develop in continuously reinforced concrete pavements (CRCP). The delamination cracks are at a shallow depth below the surface of the pavement. The delaminations occur at an early age due to large evaporation-induced stress gradients that result in shear stresses near the pavement surface. The moisture- or evaporation-induced gradient is affected by wind speed and curing method during and after concrete placement. The evaporation results in differential drying shrinkage near the pavement surface, and the shrinkage produces shear stresses in the concrete near the surface that can cause the delamination. Once delamination has formed, it may develop later into spalls as a result of incompressibles, wetting and drying cycles, traffic loading, and other effects. A survey was performed to determine the repair materials that have been found to work successfully throughout Texas and in other state departments of transportation. Minimum requirements for repair materials included: • • • • •

Ready for traffic within four hours Resistance to weathering and abrasion Skid resistant surface Placement temperature of 10°C or above Resistant to deicers, motor oil, sodium chloride solution and brake fluid • Working times of 5 to 60 minutes. • Wet bond strength of at least 100 psi • Compressive strengths of at least 200 psi.

The materials that were tested in the laboratory included one hydraulic rapid setting concrete, three magnesium phosphates (MP), and one bituminousbased material, as well as five polymer concretes (one vinyl ester, one epoxy or EPC, and three polyurethanes or PUPC).

2

PROPERTIES OF MATERIALS

The flexible materials and some semi-rigid materials were so soft that they did not lend themselves to testing using the usual ASTM test methods for compressive strength (ASTM C 575) and flexural strength (ASTM C 580) in rigid concrete-repair materials. For those materials that could be tested the flexural, compressive, and tensile bond (ASTM C 1583) strengths were determined at 4°, 21° and 38°C. Curing shrinkage was also determined at the same three temperatures. The coefficient of thermal expansion, COTE (Texas DOT method), and the modulus of elasticity, E (ASTM C 469), were both determined for the materials at 21°C. Tensile bond strengths were measured for repair materials that had been placed on small concrete blocks with simulated spalls, which were thermally cycled between −7 and 49°C. Tensile bond tests were performed at the end of 75 cycles by coring through the repairs into the substrate and pulling the cores. No delamination of the repair materials was observed in the simulated repairs. In general the polymer-based materials had low compressive strength and higher flexural strengths compared to the rigid materials. Shrinkage was higher for the polymer materials, and the tensile bond strength varied. The modulus of elasticity was very

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low for the polymer materials, but the COTE was much higher compared to the rigid materials. The results of the materials tests would suggest that the polymer-based materials would not perform satisfactorily in pavement repairs due to the low compressive strength and E and the very high COTE. However, inspections of repairs made in high traffic areas in Houston showed that the polymer based materials, particularly one PUPC, had performed extremely well over a period of at least six years. Spalls with transverse cracks in the bottom in most cases had not developed reflective cracks in the repair material. In order to further investigate the performance of the materials a field test was conducted. 2.1

Field test

The field test was conducted on I-35 south of Fort Worth, Texas. The three-lane north bound concrete highway has heavy truck traffic. When the inside lane was constructed about 15 years ago, a car drove into the fresh concrete for a distance of about 40 m. The dual tracks were 50 to 75 mm deep. They were repaired with an unknown material, but the repairs eventually deteriorated. The Texas Department of Transportation permitted the use of the tracks to be used for repair using the repair materials used in the research study. Five manufacturers participated, and the materials were the three MPs, one PUPC, and the EPC. The previous repair materials were chipped out, the surface was blasted with abrasive, and the repairs were made by the manufacturers’ representatives. After about one year of service, the repairs were visually inspected. The MPs performed the most poorly. The EPC developed a continuous crack along the boundary, but only one transverse crack was observed in a length of about 30 m. The PUPC performed the best of all the materials. For the PUPC there was no observable cracking along the boundary or in the interior, including reflective cracking. 3 WHAT WAS LEARNED The study identified properties that relate to successful repair performance. The most critical properties appear to be the stresses and strains produced by

thermal expansion and shrinkage. A useful tool for evaluating the impact of these strains and stresses on durability is the product of COTE and E (E*COTE), the units of which are MPa/°C, or stress per unit temperature change. The materials that performed the best are those that have an E*COTE of 125 kPa/°C or less. To illustrate, one PUPC has an E*COTE of 0.51 × 117 × 10−6 = 60 kPa/°C. For a 30°C drop in temperature, the stress produced would be proportional to the difference in COTE for the concrete substrate and the repair material. For example, if the concrete’s COTE is 10 x 10−6 and the PUPC’s COTE is 117 × 10−6, the tensile stress for a 30 deg C change = (117–10) × 10−6 × 30 × 0.51, or 164 kPa. Since the material has tensile bond strength of 454 to 700 kPa, a measure of the system tensile strength, the repair would be likely to remain uncracked. This assumes that there is no delamination and that both materials exhibit a uniform change in temperature. If one of the MPs is used, the COTExE stress is calculated to be 400 kPa for an application with a 40°C drop. Since the tensile bond strength is ∼400 kPa at cold temperature, the repair could fail. The high COTE is more than offset by the much lower E. The lower E also implies a much more ductile material that did not exhibit excessive wear or rutting, and field repairs in Houston have performed very well for many years. Does this mean that materials with high modulus will not perform satisfactorily? Not necessarily. There have been many high modulus materials that have performed satisfactorily, but low modulus materials would appear to have a higher probability of success. More information must be collected on material properties and associated performance of these materials in the field.

4

CONCLUSIONS

Spall repair materials with a wide range of properties were included in a study to determine the performance. The properties included a wide range of coefficients of thermal expansion and moduli of elasticity, the properties that provided the best prediction of distress. Evaluating the product of these two properties in each material showed that those with the lowest COTExE values gave the best performance in the field.

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Innovative concrete overlays for bridge-deck rehabilitation in Montréal R. Gagné CRIB—University of Sherbrooke, Department of Civil Engineering, Sherbrooke, Canada

B. Bissonnette CRIB—Laval University, Department of Civil Engineering, Québec, Canada

R. Morin & M. Thibault Ville de Montréal, Montréal, Canada

ABSTRACT: Over the past four years, three repair projects using concrete overlays were conducted in collaboration with the City of Montreal: the Cosmos Bridge (2002), the Girouard Overpass (2005), and the Bonaventure Highway (2007). The Cosmos Bridge deck was repaired with five types of fiber-reinforced-concrete overlays: A high-performance concrete, a normal concrete, two shrinkage-compensating concretes, and a concrete containing a shrinkage reducing agent. The Girouard Overpass deck was overlaid with two types of air-entrained fiber-reinforced-concrete overlays containing a Shrinkage Reducing Admixture (SRA) or an expansive agent. The Bonaventure Highway reinforced-concrete deck was repaired using two types of fiber-reinforced overlays made with either an expansive admixture or fast-setting latex-modified concrete (15%). Performance follow-up has shown that using a shrinkage reducing admixture decreases the amount of shrinkage cracking by a factor of 2.5 to 3. Both type K expansive cement and powder expansive admixture generated enough initial expansion to avoid overlay cracking from restrained shrinkage.

1

INTRODUCTION

Over the past four years, three repair projects were conducted in collaboration with the City of Montreal: the Cosmos Bridge (2002), the Girouard Overpass (2005), and the Bonaventure Highway (2007).

2 2.1

cement is a silica-fume (8%) blended cement. The CNSC overlay is an ordinary concrete reinforced with 5.2 kg/m3 of structural synthetic fibers (L/D = 70/1.4). The C-SRC overlay is a concrete containing a liquid shrinkage reducing agent reinforced with 4.0 kg/m3 of structural synthetic fibers (L/D = 90/2.3). 2.2

REPAIRED BRIDGE DECKS Cosmos bridge

The Cosmos Bridge, built in 1966, consists of a continuous deck resting on two abutments and six piles. The experimental panel consists of five thin overlays 2.5 m wide and 9.2 m long oriented transversally with respect to the deck’s long axis. Each overlay is 80 mm thick and positioned side by side. All five air-entrained concrete overlays were made with different repair concretes, each containing fibers. The C-ECK overlay is a Type K expansive cement concrete reinforced with 38 kg/m3 of hooked steel fibers (length/diameter = L/D = 80/1.3 [mm]). The C-ECA overlay is a concrete containing 24 kg/m3 of an inorganic powder expansive agent classified under ACI-223 as a Type-G component. The concrete is made with a silica-fume (8%) blended cement, reinforced with 42 kg/ m3 of hooked steel fibers (L/D = 80/1.3). The C-HPC overlay is a high-performance concrete reinforced with 42 kg/m3 of hooked steel fibers (L/D = 80/1.3). The

Girouard overpass

The Girouard Overpass, built in 1960, is a conventional reinforced-concrete structure that includes two parallel bridge decks supporting a roadway and a railway. The entire roadway surface was repaired using two types of air-entrained concrete overlays. One half of the roadway surface (G-ECA) was overlaid with a concrete containing the same inorganic expansive agent used for the Cosmos Bridge. This overlay was reinforced with 39 kg/m3 of hooked steel fibers (L/D = 80/1.3). The second half of the roadway was overlaid with a concrete containing a shrinkagereducing agent (G-SRC) reinforced with 4.6 kg/m3 of structural synthetic fibers (L/D = 90/2.3). 2.3

Bonaventure highway

The Bonaventure Highway is a 6-lane reinforcedconcrete structure built in 1966. The test site includes two 4 × 27-m concrete overlays made with two types of fiber-reinforced concrete. The first strip is an airentrained concrete (B-ECA) with the same inorganic

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expansive agent used in the Cosmos Bridge, reinforced with 40 kg/m3 of hooked steel fibers. The second strip is a latex-modified fast-setting concrete (B-RSC) reinforced with 15 kg/m3 of high-flexibility ribbon-shaped metallic fibers (L = 20, W = 1, T = 0.024 mm). 3 3.1

RESULTS Cosmos bridge

3.1.1 Concrete properties Force-deflection curves were obtained from flexural testing of prisms according to ASTM C1018. The curves for C-ECK and C-ECA show high residual strength after the first crack. These high tenacities indicate that the fibers can limit the maximal crack opening and can produce concrete with multicrack behavior. The high tenacity of these concretes probably results from internal expansion, inducing pretensioning of the steel fibers before external loads are applied. 3.1.2 In situ performance Crack density is expressed in terms of linear meters of crack per square meter of overlay (m.l/m2). There is no data for C-ECK and C-ECA (expansive agent), since these overlays have yet to exhibit cracking. C-HPC and C-NSC, evidence surface cracking that increased linearly during the first 75 weeks following construction. The crack density development stops completely after approximately 100 weeks, reaching from 1.0 to 1.4 m.l/m2. Using a shrinkage-reducing admixture (C-SRC) decreases the crack density by a factor of 2.5 to 3. 3.2

Girouard overpass

3.2.1 Concrete Properties Force-deflection curves were obtained from flexural testing of prisms according to ASTM C1018. As observed with C-ECK and C-ECA, the high tenacity of G-ECA probably results from internal expansion, inducing pretensioning of the steel fibers before external loads are applied. Internal expansion occurred in G-ECA during the first 48 h in water. After 4 months of drying, shrinkage was approximately 500 µm/m, but some small residual expansion persists. These results suggest that the initial expansion is high enough to compensate most of the autogenous and drying shrinkage occurring during the first months of drying. 3.2.2 In situ performance The evolution of the surface crack density for overlays G-SRC was monitored. There is no data for

G-ECA, since it exhibited no cracking after 2 years. The evolution of crack density in G-SRC is similar to that of C-SRC, confirming that the use of a shrinkage-reducing admixture decreases the crack density by a factor of 2.5 to 3. The evolution of the in situ deformation measured in the Girouard overlays was monitored. These curves were obtained using vibrating wire gages placed vertically near the surface of the old concrete, just before placement of the new concrete. The expansive agent in the G-ECA overlay generated unrestrained expansion of approximately 300 µm/m during the first day after placement. The initial expansion still remains after more than 2 years. 3.3

Bonaventure highway

3.3.1 Concrete properties Force-deflection curves were obtained from flexural tests on prisms according to ASTM C1018. The results indicate that metallic fibers used in B-RSC yield higher performance with small crack openings. The expansive agent in G-ECA produced an internal expansion of approximately 500 µm/m during the first 48 h in water. After 4 months, some small residual expansion remains. As with G-ECA, these results suggest that the initial expansion is high enough to compensate for most of the autogenous and drying shrinkage occurring during the first months of drying. 3.3.2 In situ performance Visual inspections performed 2 months after construction indicate no surface cracking with either type of overlay. Further inspection will be performed to better evaluate the long-term in situ performance.

4

CONCLUSION

Construction and performance follow-up of the concrete overlay test sites support the following conclusions: – Shrinkage-reducing admixtures decrease the amount of restrained shrinkage cracking by a factor of 2.5 to 3. – Both type K expansive cement and inorganic powder expansive admixture generate enough initial expansion (0–48 h) to prevent overlay cracking due to restrained shrinkage. – Fiber reinforcement limits crack opening and crack-penetration depth. After two years of service, the crack penetration depth did not reach the substrate-overlay interface. This mechanism has probably helped to eliminate overlay debonding.

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Mortar mix proportions and free shrinkage effect on bond strength between substrate and repair concrete R. Abbasnia, M. Khanzadi & J. Ahmadi College of Engineering, Iran University of Science and Technology, Tehran, Iran

ABSTRACT: Bond strength between concrete substrate and repair materials is one of the important parameters, which affects the performance of patching repair in concrete structures. In this paper, to study the mix proportion’s and free shrinkage effect on bond strength, the effect of several factors (including w/c and c/s ratio, aggregate type and grading in cementitious and modified cementitious material with silica fume and latex epoxy) are investigated. Also, the obtained results from the slant shear test under natural ambient condition are used to the measurement of mixture proportions and free shrinkage effects on bond strength. Based on the results obtained, increasing the w/c ratio or using fine aggregate with smaller size will reduce the bond strength, while increasing the c/s ratio, up to 1/2.5, and using silica sand aggregate, will increase bond strength in both ordinary and modified cementitious materials. Furthermore the bond strength was found to be independent of repair material’s tensile strength. Keywords:

1

Bond Strength, Repair Materials, Shrinkage, Substrate Concrete, Surface Shrinkage.

INTRODUCTIONS

A Large number of existing concrete structures worldwide are in urgent need of effective and durable repair. However, it has been estimated that almost half of all concrete repairs fail due to lack of reliable and perfect bond (Cleland & Naderi, 1986 ). Good bonding between repair materials and existing concrete substrate is of vital importance in the concrete repairs (Guangjing & Jinwei, 2002). The strength and integrity of the bond depends not only on substrate concrete properties and the interface factors (such as surface roughness and soundness, bond adhesive, humidity conditions, …), but also physical and chemical characteristics of repair materials. Although, many research efforts have been carried out on different factors, such as interface surface preparation method (Austin et al. 1995), modulus mismatch and specimen size effect (Cleland & Naderi, 1986), but little information is available on material mix proportions and resulting bond strength. Therefore to achieve more hindsight on bond strength, the effect of repair materials mix proportions as one of the important factor in patching repair application has been studied in this paper. 2

TEST METHOD

Slant shear test (ASTMC882) was performed to determine the bond strength. In addition, for each mix

design free shrinkage, modulus of elasticity and the compression and tensile strength were measured. According to the ASTM standard, cylindrical mold with 76.2 mm diameter and 152.4 mm height which is divided along the diagonal axis has been used for slant shear test.

2.1

Specimen preparation method

After filling the bottom part of cylinders with substrate concrete (ASTMC109), samples were removed from the molds and then kept in water for 28 days. The specimens were then left to desiccation in laboratory for an additional 30 days. Then the interface surface was cleaned of any extra dust or loose concrete and prepared by using epoxy primer and increasing the roughness under the desired humidity condition. As the next step, the repair materials were casted (ASTMC33) and then repair portions for all of specimens were wet-cured by covering with wet burlap for 28 days. At last, both side of the specimens were capped (ASTMC617-76) and compression force was applied. The value of the ultimate load was taken as the bond strength. The probable modes of failure are substrate failure, interface failure, and mortar failure. However the best case would be the state that all part of spacemen behaves uniformly.

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3 3.1

RESULTS AND DISCUSSION Mechanical property and free shrinkage effect

According to this research, no significant dependence between tensile and bond strength was observed. This fact could be related to the type of stress state that acts on interface zone (compression-shear stress instead of tensile stress). Also, the bond strength in repair materials with high value of free shrinkage decreases strongly due to micro crack spreading in interface zone. 3.2

Silica fume and polymer admixture effect

According to this study, the bond strength increases with the silica fume content (because of interface zone microstructure improvement). However, the optimum value of silica fume appears at 8%, and any additional silica fume content dose not cause considerable increases in bond strength. The improvement of the interface zone microstructure by using silica fume in repair materials is not only the consequence of its pozzolanicity, but also of the ability of very small fly particles which fit in between cement particles. This noticeable improvement in the microstructure leads to a significant increase of the intermolecular force and mechanical interlocking. However, the main part of this significant increases, is related to partially results from the chemical reaction between the active silicon dioxide of silica fume and the Ca(OH)2 to form C-S-H. This leads to an even denser interface zone with higher bond strength. Considerable increase was observed in bond strength when latex epoxy was used as an admixture. This is probably due to epoxy accumulated at some lower parts of rough surface, which leads to the creation of a thicker polymer layer at interface zone. 3.3

Aggregate grading and type effect

In this research, two different type of aggregate and grading have been used, and the resulting bond strength was measured. Based on the results obtained, the mechanical properties and bond strength of cementituse and modified cementituse materials (with silica fume and polymer admixtures) reduce, when smaller size of aggregate is used. Furthermore, these repair materials regardless of admixture type, have a higher value of free shrinkage than mixes with bigger one. In this case, usually failure surface was tending to form in interface zone. Also, the mixes with silica sand aggregate have higher bond strength than the one with limestone, during the first days. However, after 56 days, the bond strength of these two different sand types tends to reach an equal value. Also, results show that, in mixes with silica fume as an admixture, the compression and tensile strength did not change considerably

when silica aggregate was used instead of limestone. In this case, modulus of elasticity increases while the free shrinkage decreases. This can be attributed to the fact that the silica aggregate has a greater elastic modulus than the limestone, hence there is higher constraint against any volume changes. This case dose not have an important effect on bond strength except in the early ages (this issue is very important by considering the vulnerability of young concrete). 3.4

W/C and C/S ratio effect

To investigate the w/c and c/s ratio effect on bond strength, three different ratios were used for each case. Any changes in c/s ratio in mixes with silica fume as an admixture, have a considerable effect on the bond strength. In this case, the bond strength increases to a maximum value with increasing the c/s ratio to an optimum value (1/2.5) and starts to drop behind this point. By increasing the w/c ratio the bond strength drops quickly in all mixes, especially in polymer modified mixes. In this case, the failure surface tends to pass across the repair materials. Bond strength decrease, can be attributed to reduction of mechanical properties with increase of w/c ratio. 4

CONCLUSIONS

In this research, to study the mix proportion’s effect on bond strength, the effect of w/c and c/s ratio, aggregate type and grading in ordinary and modified repair material were investigated and the results can be concluded as follow. The optimum value of silica fume content is 8%, and any additional silica fume dose not increase the bond strength. Also, considerable increase was observed in bond strength when latex epoxy was used as an admixture. The mechanical properties and bond strength of mixes with silica fume and polymer admixtures reduce when smaller size of aggregate is used. Also, regardless of admixture type, these mixes have a higher value of free shrinkage. If the compression—shear stress state is dominant on the interface surface, among the repair material’s properties, the ones that increase the surface shrinkage (e.g. c/s, w/c) and those which reduce compression strength have more influence on the bond strength. The mixes with silica sand aggregate have higher bond strength than the one with limestone during the first days. However, after 56 days, the bond strength of these two types tends to reach an equal value. Over-all, mix properties have a smaller effect on bond strength in comparison to the interface surface properties.

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Evaluation of saturation and microcracking of the superficial zone of concrete: New developments L. Courard & J.-F. Lenaers University of Liège, Belgium

The aim of the concrete repair or overlay is to prolong the useful service life of the deteriorated/distressed structure or its element, to restore the load-carrying capacity and the stiffness, and to strengthen the structure or its member. Consequently, monolithic action in the composite repair structure is the final aim. A prerequisite for monolithic action is sufficient lasting bond between the existing substrate and the new-cast material. In this respect, of critical importance to the efficiency of the composite repair system is the concrete surface preparation prior to application of the repair material. Some surface preparation techniques are inducing superficial microcracking (Bissonnette and Courard, 2006). With regard to the type and the energy of the preparation, the number and the length of microcraks may vary. On the other side, water and vapor can ingress through the porosity and the microcrak network (Courard and Degeimbre, 2003). These two effects could be theoretically analysed by evaluating the quantity of water or air that could be introduced into capillaries and cracks by means of permeation or capillary absorption. Two test methods have been performed: an “Initial Surface Absorption Test (ISAT )”, developed at Queen’s University of Belfast and a modified capillary suction (MCS) test, recently developed at the University of Liège. The first proposed device offers several improvements over other existing devices and makes it an attractive alternative for non-destructive field testing: it is compact, costs less, and the test duration is short (approximately 10 minutes). The MCS device was used at the GeMMe laboratory of University of Liège to conduct test series using cylindrical cores base slabs made with three (3) different concrete mixtures (different W/C, compressive strengths and air contents). After the curing period (moist curing for 28 days followed by air drying for 60 days at 50% R.H.), each of those slabs has been superficially prepared (virgin (NT), sandblasting (SB) and hydro-demolition (HD)) and exposed to different moisture conditions, ranging from dry state to SSD conditioning state (control). The samples have been finally tested with the MCS device to determine the absorption-time curves. Air permeability results (ISAT ) show a relationship between the decreasing of the pressure and the porosity, the compressive strength and the degree

of saturation. However, results were not taken into account for the analysis, due to many problems of waterproofing between metallic ring and substrate. Water absorption permeability indexes (Fig. A) give relatively high correlations with degree of saturation (Fig. B). Statistically speaking however, there is no influence of the quality of concrete, with regard to its compressive strength. On the contrary, for a specific resistance, a light influence of the surface preparation seems to be possibly pointed out: absorption rate is higher for surfaces prepared by hydro-jetting than sandblasting. The same type of correlation can be observed with the degree of saturation: the higher the saturation, the higher the permeability index (Fig. B). Water capillary absorption tests (MCS) give clearer and less dispersive information. The parameters influencing the test procedure are of course more easily controllable. The same tendencies are observed for the two methods. (Fig. C).

Figure A. Permeability index for the different conditions (concretes, atmospheres and surface preparation).

Figure B. (ISAT).

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Permeability index versus degree of saturation

ing or to the roughness. Moreover, it doesn’t allow distinguishing concretes with different compressive strength. Other devices (sometimes easier and more precise) are already used. The Autoclam doesn’t really bring an interesting alternative to determine the real state of saturation of a concrete substrate. However, the advantage of the autoclam is that it can be used in situ and is a non destructive test. Moeover, the results obtained by the Autoclam are in accordance with these obtained with a capillary absorption test.

Log (Permeability index (mg/min^0.5).

3,0 2,5 2,0 1,5 1,0 0,5 0,0 0,0

0,5

1,0

1,5

2,0

2,5

Log (Capillary absorption (mg/s^0.5))

Figure C. Comparison between permeability index and capillary absorption.

The results finally obtained show that the Autoclam can be used as a quantitative test which is really sensible to the saturation of the substrate. The indexes measured are also influenced by the state of the surface but we don’t know if it’s actually due to the crack-

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Spalling of sprayed concrete under tunnels fire conditions C. Féron, C. Larive & G. Chatenoud CETU, Bron, France

ABSTRACT: Sprayed concrete is widely used in tunnel linings. A special technique based on this material has been developed at CETU for the waterproofing of tunnels. It results in a self-supporting shell of sprayed concrete, which is directly sprayed on insulating foam and waterproofing membrane. In order to qualify this technique regarding the fire risk, tests have been performed under the most severe fire curve that has been especially designed for tunnel fires. Most concrete spall under this thermal load, and the aim of the test was to check if the sprayed concrete was as likely to spall as poured concrete, and if polypropylene fibres were as efficient as in poured concrete. Only sprayed concrete was tested in this preliminary study. The first result is the successful mixing and spraying of the concrete, which was not guaranteed with high amounts of polypropylene fibres (2 kg/m3). Then a fire test has been performed on four slabs including two thicknesses and two mixes: with and without fibres. High values of spalling were obtained for ordinary sprayed concrete whereas fibres avoided or greatly decreased the spalling. It has also been confirmed that spalling increases with the thickness of the slab and that reinforcement doesn’t stop it.

1 1.1

INTRODUCTION Background

Spalling of concrete has been an issue for many years now and many mixes have been studied. Yet studies seem to focus on high-strength concretes which are more likely to spall, and self-compacting concretes that need to be compared to ordinary concrete regarding all their properties. In tunnels, those concretes are not so widely used, and fires are much more severe than the standard fire curves that are used to design buildings. Therefore, a lot of testing is still needed on normal strength concretes under severe fire loads. In particular, sprayed concrete is commonly used in tunnels, and its fire behaviour needed to be evaluated. That was the aim of the experimental study presented here. 1.2

Fire resistance of tunnel structures

In tunnels, critical structures are designed under a specific fire curve, which may lead any type of concrete to spall. Indeed, the temperature increase is very steep in the first minutes, and that induces a very high risk of spalling. The only way to assess and quantify spalling under that thermal load is testing. No general rules considering the type of concrete, the water content or special reinforcement can replace testing to predict the spalling of a specific mix design. 2 2.1

EXPERIMENTAL STUDY

developed for qualification of passive protection materials. We have adapted it for quantifying the spalling of unprotected structures. We modified the dimensions of the samples and used the normative referential for the testing procedure only, which allowed us to test 4 slabs at the same time (3 m × 1 m each) in a representative and reproducible way. The 4 slabs constituted the roof of the 4 m × 3 m oven. 2.2

We chose to compare 2 thicknesses and 2 concrete mixes: a current sprayed concrete mix design (Ordinary Sprayed Concrete, OSC) and the same one with 2 kg/m3 of monofilament polypropylene fibres (PPfSC). The specifications on the fibres were chosen according to the EN 1992-1-2 and its French Appendix, adding a diameter below 50 µm. The 2 thicknesses have been chosen in the common range of the lining shells made of sprayed concrete: 160 mm and 200 mm. Each slab was reinforced just like the structure would be. The main difference remains that the slabs are flat and that the shells are vaulted. In order to restrain the strains of the slabs, they were bonded on their cold face on metallic H profiles. 3

3.1

Existing testing methods

Fire resistance of structures can be evaluated by testing. The European Normative setup has been initially

Chosen parameters

EXPERIMENTAL RESULTS AND DISCUSSION Spraying of concrete with polypropylene fibres

There are two techniques that can be used to spray concrete: the wet- or dry process. The dry process is the one we wanted to test because it is used to build self-supporting sprayed concrete

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shells as it ensures a better quality of concrete through the dense re-bars network (high compaction strength allows to avoid shadow effects). Adding metallic fibres in concrete sprayed through the dry-mix process is possible, but we were wondering if polypropylene fibres could be used as well. Indeed, they are very light and we feared that most of them would fly away during the spraying of the dry mix. Moreover, polypropylene fibres usually dissociate in wet concrete thanks to repulsive effects into water. With the dry-mix process, we worried about their ability to mix. Fortunately, we were pleased with the results of the spraying. Although we could see fibres flying in the air, the sprayed mix was very homogeneous and the density of fibres was quite high (Fig. 5). We regret that no standard test exists to quantify the amount of polypropylene fibres in fresh concrete, because we would have liked to know the exact amount of fibres in the in-place sprayed concrete, being started with 2 kg/m3. Considering the amount of fibres in the air, it seems necessary that the spraying is being done outside, although no toxicity is revealed on these fibres. Moreover, as a precaution, the spraying operator wore a mask. The spraying procedure of the concrete was not modified due to the fibres; the concrete was as easy to spray as the one without fibres. The workability and the adherence on the formwork were similar with the two concretes. No blockage occurred in both cases. 3.2

Spalling measurements

The day after the test, the slabs were removed from the oven and the spalling measured. At first sight, when the slabs were taken out of the oven and even before, the efficiency of the polypropylene fibres was obvious. The 160 mm thick slab with fibres did not spall at all, whereas the one without fibres had an average spalling depth around 40 mm with maximum values reaching 93 mm. The 200 mm slab with fibres had a very homogeneous superficial spalling depth of 15 mm, whereas the slab without fibres spalled up to 110 mm, with an average of 57 mm. 3.4

Effect of reinforcement

The spalling is not modified, even locally, by the reinforcement. Neither is it stopped, which confirms literature results. 3.6

Comparison with poured concrete

The amount of spalling obtained during this test is the deepest we have ever observed on poured normal strength concrete: average spalling is about 50% higher and local deepest spalling is nearly twice. But this experiment had not been conceived to have a direct comparison between sprayed and poured concrete. Therefore the only results it can be compared to come from experiments on concrete that does not have the same composition, water content, compressive strength or thickness. All these parameters are of great influence on spalling. No general statement about spalling of non protected sprayed concrete can be concluded, all the more as this is one of the very first published result on that subject. The similar and noticeable result in poured and sprayed concrete is the remarkable efficiency of the monofilament polypropylene fibres (32 µm in diameter), here with a content of less than 2 kg/m3 due to the inevitable loss of fibres during spraying.

4

Description of the fire test

About 6 months after casting, the 4 slabs were tested in a qualified fire resistance laboratory. The fire load was the HCinc curve. 3.3

3.5

CONCLUSION

Thanks to this experiment, we learned a bit more about the fire behaviour of sprayed concrete. Polypropylene fibres can be used in this material, even in dry-way spraying, in about the same amounts as recommended for poured concrete (2 kg/m3). Even if some of the fibres fly away during spraying, a sufficient amount remains in the concrete and spreads uniformly. This results in a very good efficiency regarding spalling under HCinc fire load, reducing it to zero or 15 mm depending on the thickness of the slab. Moreover, the experiment confirmed that spalling is increasing with the thickness of the slab and that the reinforcement does not stop it. It also confirmed that precautions must be taken in case of use of self-supporting sprayed concrete shell in tunnels submitted to fire regulation (longer than 300 m in France). One of these precautions can be to use monofilament polypropylene fibres in sufficient amount.

Effect of thickness

It can be observed that the spalling depth, with or without fibres, increases with the thickness of the slab. That result is logical and was expectable.

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Repair of slab surface with thin SFRC overlay J. Šušteršicˇ, I. Leskovar, A. Zajc & V. Dobnikar IRMA Institute for Research in Materials and Applications, Ljubljana, Slovenia

ABSTRACT: In the paper, repair of slab surfaces with thin SFRC overlays are discussed on the base of results of permanent control test of SFRC properties during the construction of 4 selected overlays and findings of permanent observations of those overlays during their uses of three years. Fundamental principle of slab surface preparation is shown as well. This research is about repair of concrete slabs for industrial or commercial floors, not for highways. SFRC placed in thin overlay can have different values of compressive strength and ultimate splitting strength. But, portions of first crack strength and equivalent strength up to the crack width of 0,2 mm remain equal values, approximately over the whole range of ultimate splitting strength values.

1

INTRODUCTION

In practice, there are many surfaces of concrete slabs which are disintegrated due to different actions and because those damages have near surface character, underlies remain undamaged and they are still proper for further use. Therefore, damaged layer has to be removed and replace by new overlay with the same thickness. High quality bond between old and new layers has to be achieved. In the paper, fundamental principles of repair of slab surface with thin SFRC overlay with thickness of up 3 cm, approximately are briefly shown and some properties of used SFRC are discussed.

2

PREPARATION OF SLAB SURFACE BEFORE SFRC OVERLAY CONSTRUCTION

Careful milling was carried out over the whole area of repaired surface in depth of 1 cm, approximately. The rough surface was cleaned by water under high pressure. Excessive moisture in all area was removed and the slab surface was dried before grout was placed. The grout made from cement, polymer, sand and water was spread out by brush very careful all over the appointed area a little while before the fresh SFRC was placed.

3

MATERIALS AND MIX PROPORTIONS

Following materials and their proportions were used for mixing of SFRC: (1) cement CEM II 42,5 S, in quantity of over 400 kg/m3, (2) effective water to cement ratio (w/c)eff ≤ 0,40, (3) high-range superplasticizer, up to 1 mass % of cement content, (4) shrinkage reducer, up to 2 mass % of cement content, (5) silica fume, 5 mass % of cement content, (6) hooked steel fibers with diameter of 0,4 and 0,5 mm, and with length of

16 mm, in content of 0,25 and 0,5 vol. %, (7) aggregate with maximum grain size Dmax = 4 or 5 mm, when thickness of overlay (tov) was up to 2 cm, (8) aggregate with Dmax = 8 mm, when tov ≤ 3 cm.

4

TEST METHODS

Properties of SFRC were tested during the construction of several overlays. Following test methods were used: (1) slump test was carried out in accordance with EN 12350-2; (2) compressive strength were tested on the cubes with edge of 15 cm, at the SFRC ages of 1, 7, 28 and 33 days, the test method is given in EN 12390-3; (3) ultimate splitting strength, strength at the first crack and equivalent strength up to the crack width of 0,2 mm were calculated, taken into account parameters of load—CMOD curves obtain by wedge splitting test (WST) method; WST was carried out on the cubes with edge of 15 cm, at the SFRC age of 28 days.

5

TEST RESULTS AND THEIR DISCUSSION

Results of permanent control test of SFRC properties during the construction of 4 selected overlays are discussed. Those overlays were constructed in the period of last three years. 5.1

Workability of fresh SFRC

Fresh SFRC was mixed in the ready-mixed concrete plant and then it was delivered with a truck mixer to the site. Individual results of slump measurements carried out on the site range from 160 mm to 230 mm. Such workability of fresh SFRC was proper for pumping and placement into the overlay.

359

Figure 1. Results of compressive strength of different SFRC obtained during the construction of 4 overlays.

5.2

Figure 2. Maximal load—CMOD curve of SFRC of 1st overlay, and minimal load—CMOD curve of SFRC of 4th overlay.

Compressive strength

While workability of fresh SFRC was similar approximately in all 4 applications, compressive strength of 28-day-old SFRC specimens differentiated between them, as it can be seen in Figure 1. Those differences resulted from different loadings and environment conditions which SFRC placed into the several overlays have to resist during their use. 5.3

Ultimate splitting strength, strength at first crack and equivalent strength up to the crack width of 0,2 mm

The same differences had been obtained, when SFRC of several ovelays were tested in accordance with WST method. Two extreme load—CMOD curves are shown in Figure 2. If ultimate splitting strengths (fsu) derived from load—CMOD curves of SFRC of all four overlays are compared with compressive strengths, good correlation between those strengths is achieved. Strength at the first crack (fFC) and equivalent strength up to the crack width of 0,2 mm (f0,2) were calculated, as well during the construction of overlays, taken into account parameters of load—CMOD curves. If both strengths fFC and f0,2 are compared with fsu of SFRC of all four overlays, very good correlations are achieved. It means, that portions of fFC and f0,2 in regard to fsu remain more or less equal values. It can be seen in Figure 3. 6

CONCLUSIONS

On the base of results of permanent control test of SFRC properties during the construction of 4

Figure 3. Correlations between fFC/fsu and f0,2/fsu ratios and ultimate splitting strength fsu.

selected overlays and findings of permanent observations of those overlays during their uses of three years, following conclusions can be made: (1) fresh SFRC has to have proper workability (much the same as workability of self compacted concrete) to be achieved dense SFRC by easy placement without use of high energy of vibration; (2) class of compressive strength as well as ultimate splitting strength fsu of SFRC depend on levels of loadings and environment conditions which SFRC placed into the several overlays have to resist during their use; (3) proportions of first crack strength fFC ≈ 0,80 × fsu and equivalent strength up to the crack width of 0,2 mm f0,2 ≈ 0,70 × fsu are required to be taken into account in calculation of thin SFRC overlay, irrespective of class of fsu. Quality repared slab surface with thin SFRC ovelay would be achieved not only by abovementioned requirements of SFRC properties, but also by proper prepared slab surface before SFRC placement.

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SHCC repair overlays for RC: Interfacial bond characterization and modelling G.P.A.G. van Zijl & H. Stander Institute of Structural Engineering, University of Stellenbosch, South Africa

ABSTRACT: Bonded overlays are increasingly used in concrete and reinforced concrete repair and rehabilitation applications. Fibre-Reinforced Concrete (FRC) and various classes resorting under FRC, ranging from Ultra-High Performance Fibre-Reinforced Cements (UHPFRC) to fibre-reinforced Strain-Hardening Cement Composites (SHCC) present interesting options for such overlay repairs. The modelling and characterization of SHCC and its bond with concrete or Reinforced Concrete (RC) as substrate, are presented here. An interface model based on multi-surface plasticity is proposed for Finite Element (FE) modelling of the interface region of an overlay repair system. It enables realistic modelling of the physical processes of interfacial delamination and shear-slipping along a SHCC-RC interface. The model incorporates a Coulomb-friction material law, bounded by a tension cut-off, as well as a compressive cap. After the initial thresholds have been reached in these different stress regions, softening is activated, leading to reduced limit surfaces in tension and Coulomb-friction. To capture other sources of non-linearity, a material model based on continuum damage, which captures the nonlinear behaviour of SHCC, is employed. In this paper also physical laboratory experiments to characterize the parameters of the interface and the SHCC material models are described. Case studies are performed by laboratory experiments and comparative FE analyses to verify and validate these models and the characterized parameters.

1

INTRODUCTION

Overlay repair by fibre-reinforced Strain Hardening Cement-Based Composites (SHCC) should exploit the multiple cracking feature of SHCC, whereby significant ductility is achieved in loading actions including tension, flexure and shear. The multiple cracks are controlled to fine widths, whereby gas, moisture and chloride ingress is retarded significantly, affording structural durability to the repaired system. To allow such deformability and multiple cracking of an SHCC overlay, a smooth substrate surface has been reported to be preferable, in contrast to typical approaches in concrete repair. By a smooth interface the mechanism of combined de-lamination and overlay cracking, termed kink-crack trapping is mobilised. Complete delamination is prevented by the relatively low matrix crack toughness of SHCC, allowing crack initiation and growth in the overlay, rather than further de-lamination along the substrate-overlay interface. Such cracks are eventually “trapped” by effective fibre bridging in the SHCC overlay, whereby the resistance to further cracking becomes larger than the resistance to further de-lamination. By an appropriate balance between bond and overlay properties, multiple cracks may arise, with associated ductility and durability. To engineer such a balance, characterisation of interface properties in terms of resistance to de-lamination and shear-slipping is required. In addition to mechanical action, drying shrinkage is a major action to be considered in overlay repair.

Volume change due to moisture loss of the exposed overlay is restrained by the substrate at the level of the interface, whereby the interfacial shearing bond may be exceeded. SHCC has been reported to undergo significant shrinkage, in the region of 0,2%. Note, however, that this shrinkage is more that one order lower than the tensile strain capacity of SHCC, with full restraint. In contrast, the free shrinkage of plain concrete exceeds its fracture strain. This means that unreinforced concrete overlays may fracture under full restraint to its free shrinkage, while SHCC will have fine cracks while retaining full tensile resistance. In this paper, the focus is on interface testing, using various surface preparation methods, and modelling of the repair layer strategy to verify the parameters and modelling approach. Appropriate bond tests for SHCC-concrete overlay systems are described. The results of interface characterisation tests with these test procedures, as well as larger tests on composite beams (SHCC overlays on concrete substrate) are reported and discussed. The interface test data is used to calibrate an appropriate constitutive model for the interface. This model is used in combination with a recent computational constitutive model for SHCC to analyse the interface tests. The characterised interface models and parameters are subsequently used in computational analysis, to demonstrate the model and parameter validity. Note that shrinkage is excluded here, but will be considered in subsequent computational studies with the models calibrated here.

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2

OVERLAY MATERIALS EMPLOYED

In this paper a particular concrete substrate was used throughout, with compressive strength in the range 35 MPa and E-modulus 29 GPa, using a constant w/b ratio = 0,5 and 13 mm coarse aggregate. For the overlay, a standard SHCC was developed at Stellenbosch University.

3

5

INTERFACE CHARACTERISATION TESTS

In order to characterise the different modes of interface behaviour objectively, care is taken to clearly distinguish between these modes. Here, de-lamination, or separation of the overlay and substrate normal to the interface, is distinguished from relative shearing of the overlay and the substrate along their interface. For shear, a push-off test was adopted in this study and refined by finite element (FE) analyses incorporating assumed, realistic interface parameters and the concrete an SHCC material behaviours as characterised in separate tests. A direct tension test was chosen for measuring the adhesive (tensile) strength and fracture energy of the interface. A semi-rigid upper platen, and rigid lower platen was selected as the most appropriate test setup for objective parameter characterisation, based on FE analysis results. A comprehensive series of shear and tensile tests were performed with the above two tests. Various substrate (concrete) surface preparations were used, including no preparation, scraping, sandblasting. In addition, two moisture preparations were used, namely soaking for 24 h, as opposed to 10 min moistening before casting of the SHCC overlay.

4

Subsequently, an overlay repair system simulation was analysed, in the form of four point bending of a beam prepared by centrally separated concrete substrate beams, bound together by a SHCC overlay. The superior flexural behaviour of SHCC overlay to a concrete overlay was demonstrated computationally, and shown to be in good agreement with experimental test results. CONCLUSIONS

Appropriate shear and tensile tests have been proposed and refined through finite element analyses, to characterise the SHCC-concrete substrate interface. A series of interface characterisation tests led to the following conclusions: − An increased level of substrate surface roughness is associated with an increased shearing and tensile strength at the interface. − A lower water content level of the substrate leads to larger shear and tensile bond strength than higher water content. It is believed that a higher level of capillary absorption occurs in the substrate at lower moisture content, leading to higher interlock across the interface boundary. − The interface shear and tensile strength increases with age in the range tested here (7 to 24 days). Note however that shrinkage was limited in these experiments and not considered in the FE analyses. The interface test data has been used to calibrate an interface constitutive model, which was subsequently validated by larger SHCC-concrete overlay beam experiments and analyses. This modelling strategy and parameters can be used in subsequent optimisation of other SHCC-concrete overlay design strategies, including consideration of the influence of drying shrinkage.

ANALYSIS OF OVERLAY APPLICATION ACKNOWLEDGEMENT

The measured interface parameters were verified by re-analysis of the tests with the Coulomb-friction model for the interface, and incorporating substrate and overlay nonlinearity with a continuum damage model and smeared cracking for the SHCC overlay and the concrete substrate respectively.

This research was jointly sponsored by the Department of Trade and Industry through a THRIP grant to project SAPERCS, and several South African industry partners. The support by Infraset Infrastructure Products is gratefully acknowledged.

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Modelling the performance of ECC repair systems under differential volume changes Jian Zhou Microlab, Faculty of Civil Engineering and Geosciences, Delft University of Technology, Delft, The Netherlands

Mo Li Advanced Civil Engineering Materials Research Laboratory (ACE-MRL), Department of Civil and Environmental Engineering, University of Michigan, USA

Guang Ye, Erik Schlangen & Klaas van Breugel Microlab, Faculty of Civil Engineering and Geosciences, Delft University of Technology, Delft, The Netherlands

Victor C. Li Advanced Civil Engineering Materials Research Laboratory (ACE-MRL), Department of Civil and Environmental Engineering, University of Michigan, USA

ABSTRACT: An ultra ductile material, known as Engineered Cementitious Composite (ECC), has been proposed to be one of most promising repair materials. An analytical model was developed to calculate the stresses in ECC repair system under differential volume changes. With this model, the performance of ECC repair system was investigated. In an ECC repair system, although the shrinkage of ECC increases dramatically after the first cracking of ECC, the stress increases only by a small amount. It can be expected that the use of ECC can reduce the potential of repair material failure in tension and interface delamination and therefore enhance the durability of repair systems.

1

INTRODUCTION

Most materials used in concrete repairs have a tendency to deform due to shrinkage, heat release at early age and ambient temperature change. The restraint of these deformations by substrate concrete induces stresses in repair systems. The stresses can lead to vertical cracking through the thickness of the repair material, peeling of the repair material from the substrate concrete and/ or delamination of the interface. ECC has been proposed to be one of the most promising repair materials. Unlike common cement-based materials ECC shows tensile strain-hardening behaviour with strain capacity in the range of 3–7%, which is hundreds of times of the strain capacity of common cement-based materials. Figure 1 shows the typical tensile stress-strain curve and the average crack width of ECC. The high ductility of ECC is achieved by multiple cracking. When ECC is used as repair material, the multiple cracking can release stresses in repair systems induced by differential volume changes. The risk of repair material in tension and interface delamination is therefore reduced. It can be expect that the use of ECC can enhance the durability of concrete repairs. This phenomenon has been demonstrated both in the laboratory and in the field. An analytical model was developed to calculate stresses in repair systems subjected to

Figure 1. The tensile stress-strain curve of ECC.

differential volume changes. In this paper, this model will be further developed to estimate the performance of ECC repair systems under differential volume changes. The modelling results, with comparisons to experimental observations, are reported here. 2

MODEL DEVELOPMENT

An analytical model was developed based on the classical plate theory and the assumption of the linear relation between shear stress and slip at the interface. It has been successfully used to calculate the stress and strain

363

6

Stresses (MPa)

5

σ xx

4 3 2

σ yy

1

σ xy

0 0

500

1000

1500

2000

Shrinkage (µstrain)

Figure 2. The stresses in ECC repair system at 28 days calculated with Equations 4.

60 50 Delamination (µm)

in the concrete repairs subjected to differential volume changes. This model is further developed to estimate the performance of ECC repair systems. ECC, under tensile load, behaves differently before and after the first cracking which corresponds to the bend-over point in the tensile stress-strain curve. Accordingly, the tensile stress-strain curve of ECC can be divided into two parts. When the stress is lower than the first cracking strength, ECC behaves like common cement-based materials. When the stress is between the first cracking strength and the ultimate strength, ECC shows strainhardening behaviour. In terms of structural response, the unique behaviour of ECC can be reflected by using the different elastic moduli before and after first cracking. The calculation of stresses in the ECC repair system was divided into two stages. In the first stage, i.e. when the tensile stress in ECC is smaller than the first cracking strength of ECC, the stresses can be calculated by using the model with the elastic modulus of ECC before first cracking. In the second stage, the stresses can be calculated by using the model with the stress-strain gradient after first cracking. By superimposing the stresses calculated in these two stages, the stresses in ECC repair system induced by the differential volume change can be determined.

40 30 20 10 0

3

0

SIMULATION

5

10

15

20

25

30

Age (days)

Layered ECC repair systems were experimentally investigated. In the layered repair systems, a layer of ECC repair material with the thickness of 50 mm was bonded on the top concrete substrate with the thickness of 100 mm. The length of repair system was 1560 mm, and the width of repair system was 100 mm. The free shrinkage and uniaxial tensile tests of ECC were carried out at 28 days. The tensile strain capacity of ECC was higher than 2.5%. The measured first cracking of ECC was averaged to be 5.0 MPa, and the measured ultimate strength was averaged to be 6.0 MPa. The modulus of elasticity before the first cracking was 20,000 MPa and the stress-strain gradient after the first cracking was 40.4 MPa. The material properties measured from experiments were used as input in the calculation of stresses in ECC repair systems. Figure 2 shows the calculated results. When the shrinkage of ECC increases to 392 µstrain, the tensile stress in ECC reaches the first cracking stress of 5.0 MPa. After the first cracking, although the shrinkage increases around 3.5 times, the stresses in ECC repair system only increases a little bit. This is because of the low stress-strain gradient in strainhardening stage. The experiments show the same phenomena. As the shrinkage increases 85% after 4 days, the interface delamination opening increases 24% as

Figure 3. Interface delamination opening at the ends of ECC repair system.

shown in Figure 3. The final tensile strain of ECC was calculated to be 0.18%, which is much smaller than the tensile strain capacity of ECC of 2.5%. It means that ECC cracks but does not fail in tension under the relatively high differential shrinkage. 4

CONCLUSIONS

An analytical model was developed to calculate the stresses in ECC repair systems according to the bilinear behaviour of ECC under uniaxial tensile load. With this model, the performance of ECC repair systems was investigated. In ECC repair systems, the increase in the shrinkage after the cracking of ECC results in only a small increase in stresses. Although ECC has relatively high drying shrinkage compare to normal concrete, the tensile strain remains much smaller than its tensile strain capacity. It can be expect that the use of ECC can reduce the potential of the repair material failure in tension and interface delamination and therefore enhance the durability of repair systems.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Prevention of damages in industrial floors with screed layers R. Breitenbücher & B. Siebert Institute for Building Materials, Ruhr—Universität Bochum, Germany

ABSTRACT: Industrial floors usually are subjected to high abrasion. To resist such wears different measures referring a surface-improvement are applied in practice. Nevertheless, due to various reasons damages can arise in such members which impact the utility significantly and shorten the real service life time. In many cases an improvement is performed by a screed layer applied wet-on-wet onto the casted concrete. Although this construction has often been successfully executed, occasionally damages in terms of superficial delaminations occur. Such effects are mostly raised since the concrete has already stiffened or set too far at the time of the screed mortar application. In order to assure on one hand a wet-on-wet application and simultaneously on the other hand a sufficient load-bearing capacity for the smoothing machineries on the surface, a penetration test to check the state of the concrete setting was developed at the Institute for Building Materials of the Ruhr-Universität Bochum. With this tool the appropriate time slot for the final screed application can be determined in situ.

1

INTRODUCTION

3

As in other sectors also industrial floors have their specifics which must be considered in the design, construction and maintenance. Otherwise the susceptibility referring defects and damages may increase significantly. Such negative phenomenas usually appear in form of more or less intense cracking, breakout of edges or delaminations in multilayer systems. The development of industrial floors is mainly influenced by ideas of craft and empiricism. Thus, varieties in the design as well as in the execution have been developed over years. The general experiences are documented in a few leaflets with informative character only. Not least the lack of systematic investigations has led to a greater damaging potential for industrial floors.

2

SERVICEABILITY OF INDUSTRIAL FLOORS

Industrial floors mostly are subjected to high abrasion for example by heavy traffic loads and frequent impacts or strokes. Often a combination of such loads arises. Incomplete information during the design about the later utilization or possible changes in use during the lifetime can lead to inappropriate matching between loadings and resistance of an industrial floor. Consequently, in such cases the serviceability can be limited to an end quite before the scheduled service life. For intensively stressed industrial floors special measures must be taken already during the construction phase in order to achieve a durable high abrasion resistance. One possible measure is the application of a specially produced screed wet-on-wet on the concrete slab before its setting or hardening.

DELAMINATION OF THIN LAYERS

Industrial floors are often successfully applied with separate screed layers in a monolithic way. However, from time to time some damages in terms of bond failure have been observed. Frequently, these locations of bond failures are identical with areas of simultaneous surface cracking. After opening such zones, segregations directly in the bond zone between the screed layer and the concrete slab can be detected clearly in most cases. The surface of such separations can appear smooth or rough. Usually there are two distinct characteristics: − Type 1: Accumulation of fine mortar directly in the separation plane, − Type 2: Separation/loss of bond. In case of the first type an accumulation of fine mortar can be observed below the screed layer, i. e. in the zone of bond failure. The separation plane has a relatively smooth surface and obviously an adhesive bond between concrete and mortar has never been developed. In contrast to this type there are defects without any "physical separation layer" (type 2). Rather more the development of a significant bond between the concrete slab and the screed layer was impaired in this case from the beginning, although a wet-on-wet application of the screed-mortar was planned. Obviously the fresh concrete had already stiffened/set to a large extent before the screed mortar has been applied. Thus a real monolithic bond—as gained by a optimal wet-on-wet application—could not be obtained. Such a poor bond is caused significantly by a tightrope walk between two more or less conflicting boundary requirements. On the one hand side the concrete must be stiffened to such an extent that a sufficient load-bearing capacity for the mechanical

365

surface treatment is already obtained. On the other hand the setting must not have proceeded so far yet that the wet-on-wet bond between the two layers cannot be ensured anymore. Therefore in practice only a comparatively small time slot remains for a technically adequate application of the screed mortar. At present this decision is based on empirical criterions which only consider whether the concrete is accessible or not. Since this criterion is very subjective, it certainly can result in quite different assessments. Up to now no generally accepted rules for a proper definition of this period have been developed. Thus the construction process of the described weton-wet application is often linked with certain risks that may impair the bond between the screed layer and the concrete slab.

4

Figure 1. Penetration test with force gauge and penetrator (circular cone).

DEVELOPMENT OF A TEST METHOD FOR THE ADEQUATE WET-ON-WET PRODUCTION OF INDUSTRIAL FLOORS WITH SCREED LAYERS

Within the scope of a research project a test device has been developed at the Institute for Building Materials at the Ruhr-Universität Bochum based on the determination of the continuous increasing penetration resistance of setting concrete (Fig. 1). With this equipment the reactive force is determined which is necessary to push a metallic cone of defined geometry into the fresh concrete. By means of this penetration test method investigations were carried out to check the interactions between the stiffening/setting state of the concrete at the time of the screed application and the attainable bond between these two layers. For this purpose a typical screed mortar was applied wet-on-wet on different concrete slabs in the lab after different waiting periods. Before each mortar application the stiffening/setting state in the concrete surface was determined by the penetration test. Comparing the average values of the penetration forces with the average bond strengths determined later on, a relatively clear correlation was found. Exemplarily the results of a repeated test series are shown in Figure 2. As expected decreasing bond strengths of the screed layer are associated with increasing penetration forces determined shortly before the screed application. A sufficient monolithic bond with strengths exceeding 2.0 N/mm² could be proved for all the samples, on

Figure 2. Average bond strength after 7 days in dependence from the average penetration resistance of the concrete determined directly before applying the screed layer (penetration with 20°-cone; penetration depth d = 2.0 cm).

which the screed layer had been applied before the penetration force exceeded about 250 N. On the other hand it also could be determined that penetration forces of at least about 150 N are sufficient for an adequate surface treatment with power trowels. These specified penetration resistances are only valid for the concrete compositions used here and under the specific boundary conditions of these laboratory tests. In practice they have to be defined in individual checks considering the project-specific conditions. Nevertheless, the mentioned limit values can be taken for a rough orientation.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Evaluation of the effect of load eccentricity on pull-off strength L. Courard University of Liège, Belgium

A. Garbacz & G. Moczulski Warsaw University of Technology, Poland

1

INTRODUCTION

Concrete repair works involve most of the time the removal of the damaged or contaminated surface layer. It is necessary to assess the effect of that operation (tool, procedure, etc.) upon the integrity of the residual concrete substrate, prior to the placement of the repair material, since it can affect significantly the adhesion of the repair system (Courard & Bissonnette, 2004). A parameter which was very rarely experimentally studied is load eccentricity. The principle of pull-off test is to set a load perpendicularly to investigated surface. Two main reasons may however induce eccentricity: the first is angle of inclination α, caused by inaccurate core drilling. The second is really load eccentricity. It is caused by inaccurate positioning the steel dolly on the top of the core. The main objective of this program was to investigate the influence of load eccentricity on pull-off test, for angles of inclination up to 5°. 2

RESEARCH PROGRAM

The first part of research program was devoted to numerical analysis. Firstly, the analysis was performed in order to know if the load eccentricity and the angle of inclination can induce the same effect on pull-off test. Secondly, the stress distribution around the core has been analyzed. Thirdly, percentage of stress variation in peripheries of the core bottom was investigated. The second part was devoted to laboratory experiments. Experiments were performed to measure the influence of angle of inclination on the tensile pull off strength and compare results to modeling. 3

Finite elements were used for modeling the three partial cores by means of Lagaprogs software. According to the assumptions of the project, the models had different eccentricity angles—0°, 2° and 4°—and core depths—15 mm and 30 mm -, respectively. The specimens were designed in two dimensions and analysis was defined as a plane strain problem. Figure 1 shows the mesh of finite elements. The load was assumed to be uniformly applied to the material surface; The analysis was focused on cross section and its remarkable points (A and B) as presented on Figure 1. 3.1

Causes of eccentricity

Firstly, the analysis was performed in order to know if the two causes of eccentricity could induce the same effect on pull-off test. The analysis was realized for angle of inclination 4° and load eccentricity 4°, and a core depth of 30 mm. The two “ways” of eccentricity give results so similar as it can be said that effects are really the same. The only difference is in values of σx stress distribution. However, the differences in values remain negligible. From this analysis, it appears that σy stresses are the most important in values, which means that main influence of angle of inclination is on σy stresses.

·A

NUMERICAL MODELING

B

·

y

Numerical modeling was based on the elastic constitutive law for solid elements at constant temperature. The numerical analysis was conducted with program Lagaprogs, developed at the University of Liege, Belgium.

x Figure 1. Mesh of finite elements for model with 30 mm core depth and 4° of angle of inclination.

367

Core depth and angle pull-off strength [MPa]

3.2

Secondly, σx and σy stress distributions were calculated for variable core depths and angles of inclination. For angle of inclination of 0°, σx and σy stresses are symmetric. The symmetry does not exist for cores with angle of inclination of 2° and 4°. For 0° angle of inclination, at the edge of the bottom of the core, stresses σx are very small. Stresses increase when angle of inclination increases. Especially at the bottom of the core, there is bigger increase of stresses, either σx or σy, near point B (it can indicate that, in this place, failure begins) and bigger decrease near point A.

0

2

4 o

angle of inclination [ ]

Figure 2.

4

Core depth 15mm Core depth 30mm

4.5 4.0 3.5 3.0 2.5 2.0 1.5 1.0 0.5 0.0

Pull-off strength for concrete slabs C30/37.

EXPERIMENTAL INVESTIGATIONS

Experiments were performed on three types of concrete: C30/37, C40/50, C50/60. The concrete slabs surfaces were prepared and treated by sandblasting. The roughness of surface was checked by Sand Patch Test (EN 13036) and the Average Texture Depth was equal to 0.90 mm. According to the scope of research program, the tensile pull-off test was conducted with two core depths: 15 mm and 30 mm. For each depth, the effect of 0°, 2° and 4° deviation was analyzed. Results show very small differences between each concrete type. Maximum difference is about 13% (Table 1). The value for C40/50 at angle of inclination of 4° and core depth of 30 mm is an exception in which pull-off strength was higher than for core without angle of inclination. A significant influence of load eccentricity on tensile strength was observed mainly for concrete C50/60 at 4° of angle of inclination and 15 mm core depth. The decrease was 13% for pull-off strength. The investigation on surface failure allowed setting specific observations. A global observation, for all types of concrete used in tests, is that a higher percentage of aggregate failure than aggregate/paste interface failure in total surface failure, can be observed for core depth of 15 mm than for 30 mm: it probably explains the higher tensile strengths and better behavior of superficial concrete (Fig. 2). The percentage of failure of aggregate/paste doesn’t seem to change when angle of inclination increases (the values differ no more than 3%). In such situation, we cannot say that there is a real

Table 1.

Percentage change of pull-off strength (%). Core depth 15 mm

Core depth 30 mm

Concrete type

0°

2°

4°

0°

2°

C30/37 C40/50 C50/60

0 0 0

8 8 3

5 10 13

0 0 0

6 3 6

4° 6 −6 12

influence of load eccentricity (2° and 4°) on failure of aggregate/paste interface versus failure surface. 5

CONCLUSIONS

On the basis of the results the next conclusions can be drawn: • the influence of load eccentricity and angle of inclination can be treated in the same way. • according to numerical analysis, the influence of inaccurate coring is observed. For core depth of 30 mm, an increase of maximum stress is observed in the periphery of the bottom of the core, up to 9% for 2° and 19% for 4°, respectively. For core depth of 15 mm, the maximum stress increase goes up to 6% for 2° and 14% for 4°, respectively. • with regards to results of laboratory experiments, it can be observed that, if an influence of angle of inclination is observed on pull-off strength, it is however not statistically significant. • regarding investigation of failure mode, it can be observed that the load eccentricity does not significantly change the shape of failure mode for angles of inclination not higher than 2° and 4°. REFERENCES Austin, S. et al, 1995. Tensile Bond testing of concrete repairs, Materials and Structures, 28, 249–259. Bungey, J.H. and Madandoust R. 1992. Factors influencing pull-off tests on concrete, Mag. Conc. 44, No. 158, 21–30. Cleland, D.J. and Long, A.E. 1997. The pull-off test for concrete patch repairs, Proc. Instn Civ. Engrs Structs & Bldgs, 122, Nov., 451–460. Courard, L. and Bissonnette, B. 2003. Essai derive de l’essai d’adhérence pour la caractérisation de la cohésion superficielle des supports en béton dans les travaux de réparation: analyse des paramètres d’essai, Mater. Struct., 37(269), 342–350. Czarnecki, L. and Emmons, P.H. 2002. Repair and protection of concrete structures, Kraków, Polski Cement, (PL).

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Proposal of an experimental programme for determining tensile relaxation in bonded concrete overlays C. Masuku, H.D. Beushausen & P. Moyo University of Cape Town, Department of Civil Engineering, South Africa

ABSTRACT: Restrained shrinkage in bonded concrete overlays can cause stress build up, cracking and even debonding. Tensile relaxation is the main mechanism of stress relief in concrete overlays resulting in less possibility of cracking and debonding. The research described in this paper aims at presenting an analytical solution based on tests performed in order to assess the tensile relaxation in concretes subjected to restrained shrinkage. Although only limited test results are available to date this paper serves as an introduction to the topic and presents a review of existing literature. Concrete specimens were subjected to constant strain proportional to shrinkage restraint on real specimens as obtained from related studies. Test results will be developed into an analytical model describing tensile stress relaxation and its variation with shrinkage strains and the effects of material composition and maturity. These results aim at predicting the likelihood of failure in bonded concrete overlays. Based on this, cracking and debonding can be mitigated in the design phase.

1

INTRODUCTION

Presently, some of the concrete structures in South Africa are approaching the end of their service life whilst others need repair from time to time. The bonded concrete overlay technique is one of the main repair techniques used in repairing deteriorated concrete structures. This technique involves the removal of a distressed surface layer on a concrete base (substrate) and replacement with a fresh layer of concrete i.e. the overlay. Due to thermal and hygral deformations the fresh overlay will tend to contract more than the already matured substrate, leading to differential deformation between the two composites. Tensile stresses are set up in the overlay whilst the top section of the substrate is subjected to compression. The tensile stresses in the overlay may lead to cracking and or debonding of the concrete overlay. Cracking and debonding may be prevented if the stress due to restrained shrinkage can be reduced to levels below the tensile strength of the overlay. If the bonded concrete overlay has sufficient elastic strain capacity as well as tensile relaxation to counteract the effects of shrinkage restraint, the restraint stresses are reduced considerably. In normal concretes the free shrinkage strain is approximately 6 times the tensile strain capacity, which lies in the range of 100–200 microstrain. Therefore the main mechanism of stress relief in concrete is tensile relaxation. Tensile relaxation reduces the stress induced in overlays, in some cases preventing cracking and debonding. Although studies have been carried out in this field, there is still little experimental evidence with regards to the magnitude or modelling of tensile relaxation in bonded concrete overlays. Furthermore the effects of

time, the water binder ratio, the loading conditions and the maturity at loading are not adequately covered. Therefore the study described in this paper will attempt to show the effect of varying input parameters such as the w/c, the strain imposed and the maturity at loading on tensile relaxation. The aforesaid input parameters were chosen based on common overlay properties and studies carried out previously. Although experiments are still ongoing, preliminary results, obtained from experiments, seem to be consistent with evidence presented in existing literature. 2 2.1

EXPERIMENTAL PROGRAM Introduction

Experiments were ongoing when this paper was drafted. 2.2

Preparation and curing of specimens

Tensile relaxation test specimens similar to the dog bone specimens were prepared. The dog-bone shape allows for easy gripping during testing. The dimensions were 270 mm × 40 mm × 40 mm (Fig. 1). Mixes with water binder ratio of 0.45 and 0.60 were used as presented in Table 1. Coarse aggregates were not included in the mix to minimize the effects of dilution and restraint from shrinkage. Free shrinkage strain, tensile strength and tensile relaxation dog bone test specimens were cast, wetcured and tested. Wet-curing was done by placing wet burlap and plastic sheets on the samples until testing at respective ages. Relaxation tests were done at ages

369

7 days

100

Detail

2 days

80

t/ 0

60

[%] 40 20

[hrs] 0 0

10

20

30

40

50

60

70

80

100

7 days

90

2 days

80 t/

0

[%] 70 60 50

[hrs]

40 1

Figure 1.

4

5

6

7

8

9

10

Stress relaxation at ages 2 days and 7 days.

Concrete mix proportions.

Constituent [kg/m3]

0.45

0.60

Cement Water Klipheuwel sand Superplasticiser

556 250 1490 8

417 250 1605 –

of 2 days and 7 days in order to determine the effect of maturity on tensile relaxation.

The major difference in relaxation occurs during early ages (roughly within the first 10 hrs). Thereafter not much relaxation occurs. The profiles of the curves flatten out and become somewhat similar. The preliminary results may agree with observations from previous research however it is still early to draw dependable conclusions as tests are still underway.

4 3

3

Experimental set-up. Figure 2:

Table 1.

2

CONCLUDING REMARKS

PRELIMINARY RESULTS

Fig. 2 shows the comparison between two 0.60 w/c mix specimens tested at ages of 2 days and 7 days. Both samples were subjected to the same degree of restraint i.e. 60% of the total free shrinkage. The specimen tested at 2 days had approximately 65% relaxation whilst the 7 day specimen reached approximately 50% stress relaxation after 72 hrs. Hence preliminary results indicate that relaxation decreases with age. This is in agreement with findings. From Fig. 2 it appears that relaxation develops rapidly after loading. Early development was observed within the first 10 hrs as shown on the detail (circled). The early development of relaxation was also reported in other studies.

Tensile relaxation in bonded concrete overlays may be influenced by w/c as well as the maturity at loading. In this respect, this study shall yield an analytical model of relaxation of tensile stresses induced due to restrained shrinkage of the overlay. Moreover the relaxation shall be expressed according to mix parameters (w/c ratio) and maturity of specimen at loading. Furthermore, results obtained from this study can be used to reduce the likelihood of failure in bonded concrete overlays by means of predicting the development of relaxation. Consequently this information can be used on site to optimize mixes and in design in order to mitigate the occurrence of cracks, debonding and subsequently the failure in bonded concrete overlays.

370

Structural repairs and strengthening

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Analytical modelling of retrofitted reinforced concrete members with flexible bond E. Raue, H.-G. Timmler & H. Schröter Bauhaus-Universität Weimar, Germany

ABSTRACT: The model for physically and geometrically non-linear analysis of retrofitted cross-sections and elements is a consequent continuation of the authors’ former works. Instead of the then assumed rigid bond between the different segments of the cross-section, the modelling includes flexible bond also. The mathematical model is based on a kinematic formulation of the mechanical problem by the Lagrange principle of Minimum of total potential energy. Stresses and strain energies are functions of the strains. The strain energy of a cross-section is determined by a contour integral related to the generalised function F to describe the material behaviour in an integral form. Thus a discretisation of the cross-section is unnecessary. Minimising the total potential energy of the entire member as a function of the displacements u, v und w enables the computation of deformations and stresses of composite structural elements, whereas the compatibility of these displacements and deformations must be taken into account. According to Lagrange principle flexible bond between the sub-sections will be considered in the model by means of energy expressions. Different bond effects like friction and adhesion as well as special bond components e.g. headed studs are included into model by separate shear force—slip relations.

1

INTRODUCTION

Conventional design of retrofitted elements often assumes rigid bond between the layers, so that the results are imprecise. This paper presents an alternative modelling which considers the physical nonlinear behaviour of the material, the geometric nonlinear behaviour of the elements and the flexible bond between the sub-sections also. 2

Using the theory of moderate deformations the compatibility condition for a geometrical nonlinear analysis is given by ε 0 = u′ ( x ) +

1 1 (v ″ ( x )) 2 + (w ″ ( x ))2 . 2 2

(6)

The material law is usually described by the stressstrain relation σ = σ (ε ).

BASICS OF MODELLING

(7)

The model is based on a kinematic formulation of the mechanical problem especially on the Lagrange principle of Minimum of total potential energy

An integral formulation of the material law by the functions W(ε), F(ε) and Φ(ε) are very effective. These functions are defined as the particular integrals

Π = Π i + Π a → Minimum.

W = W (ε ) = ∫ σ (ε ) d ε

ε

(1)

ε x (y , z ) = ε 0 + κ y y + κ z z

(8)

0

According to the Bernoulli hypothesis the strain at an arbitrary point of the cross-section is defined by the linear function

F = F (ε ) =

ε

∫ W (ε ) d ε

(9)

0

(2) ε

The deformations ε0(x), κy(x) and κz(x) depend on the displacements u(x), v(x) and w(x). In the case of geometrical linearity the equations of compatibility are defined by ε 0 = u ′ (x )

(3)

κ y = v ″ (x )

(4)

κ z = w ″ ( x ).c

(5)

Φ = Φ (ε ) = ∫ F (ε ) d ε.

(10)

0

The strain energy ΠiC of a cross-section with the region B is determined by the integral Π iC =

∫∫ W ( y , z ) dydz = ∫∫ W [ε ( y , z )]dydz , B

B

where W(ε) is the specific strain energy.

373

(11)

By means of the Gauss’ theorem the double integral is transformed into a contour integral along the boundary L of the region B Π iC =

1

∫∫ W ( y , z ) dydz = − κ ∫○Fdη.

(12)

B

The strain energy ΠiC of a composite cross-section consisting of m partitions with different material properties is defined as the sum of the strain energy Πj,iC of all partitions j Π iC =

segments. We assume that vertical displacements in the contact area cannot occur so that all segments of the cross-section have the same vertical deflections. In the case of a symmetrical and uniaxially loaded composite cross-section with only one flexible joint the variables of the problems are the two displacements in direction of the element u(x)1 and u(x)2 and the deflection w(x). The shear-stress transmitted in the joint due to adhesion, friction and special connectors etc. depends on the relative displacements s(x): s ( x ) = u1 ( x ) − u2 ( x ).

m

∑Π

C j ,i

(13)

.

j=1

The strain energy ΠiE of an element is obtained when ΠiC is integrated over the length l of the element

The behaviour due to adhesion is approximated by a bi-linear function ⎧0 ⎪ ⎪ s −s τ ad = ⎨ ad 2 τ ad 0 ⎪ s ad 2 − s ad 1 ⎪⎩τ ad 0

l

Π iE =

∫

(14)

Π iC dx .

0

For the potential energy of external loads for elements with the length l we obtain

0

3

(15)

Concrete cross-sections are usually polygonal. The geometry of a polygon is defined by the coordinates yi and zi or ηi and ζi of the corners Pi (i = 1, …, n). The strain energy related to the boundary between the points Pi and Pi+1 can be calculated by the formulas

Π iC,,i

(16)

C i ,i

⎧ ⎪T (sCo 2 < s ) ⎪⎪ Co 2 = ⎨ TCo1 + (TCo 2 − TCo1 )f (s ) (sCo1 ≤ s ≤ sCo 2 ) , ⎪T ⎪ Co1 s (0 < s < sCb1 ) ⎪⎩ sCo1

⎛ f ( s ) = ⎜1 − ⎝

(17)

.

(21)

(22)

where TCo,1, TCo,2, sCo,1 and sCo,2 are limit values of the corresponding functions and nCo is the order of the exponential function

n

∑Π

(0 ≤ s ≤ s ad 1 )

where µ is the friction coefficient depending on the surface roughness and σsq is the stress due to compression rectangular to the joint. This compression can be induced by external loads as well as by internal forces e.g. shear-forces. The behaviour of local connectors can be described by linear and non-linear functions. In the case of headed studs or reinforcement stirrups the formula is proposed for a single connector k

TCo ,k

whereby the strain energy of the whole polygonal crosssection is Π iC =

(20)

(s ad 1 ≤ s ≤ s ad 2 )

τ fr = µ σ sq

APPLICATION TO COMPOSITE CROSSSECTIONS AND COMPOSITE ELEMENTS

∆Φ i ⎧ 1 (κ ≠ 0, ∆ε i ≠ 0) ⎪ − κ ∆ηi ∆ε i ⎪ ⎪ 1 = ⎨ − ∆ (η F )i (κ ≠ 0, ∆ε i = 0) ⎪ κ ⎪ W im ⎪ 2 ( y i z i+1 − y i+1z i ) (κ = 0). ⎩

(s ad 2 < s )

The shear-stress transmitted by friction is defined by

l

Π aE = − ∫ [ p x ( x ) u ( x ) + p y ( x ) v ( x ) + p z ( x ) w ( x ) ] dx

(19)

⎛ sCo 2 − s ⎞ ⎜⎝ 1− s − s ⎟⎠ Co 2 Co1

nCo

⎞ ⎟ . ⎠

(23)

i=1

Rectangular cross-sections where the thickness compared to its length is small can be described by the coordinates of the end points yi0, zi0 and yi1, zi1 and the thickness. Assuming constant strains and stresses across the thickness of the element eq. (16) is transformed into

The total shear-forces T(x) in the joint is the sum of this described three components T ( x ) = Tad ( x ) + Tfr ( x ) + TCo ( x ) x

= ∫ τ ad b jo ( x ) dx + 0

⎧ ∆Fi ⎪ Π = ⎨ ∆ε i ⎪W A ⎩ im i C ii

( ∆ε i ≠ 0)

(18)

( ∆ε i = 0).

Composite cross-sections with flexible bond of the joint are characterized by relative displacements of the

4

x

∫

0

τ frb jo ( x ) dx +

∑T k ∈x

Co , k

.

(24)

EXAMPLES

In the paper the application of the presented method is demonstrated by a retrofitted column and by a precast beam element with in-situ concrete.

374

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Experimental and analytical investigation of concrete confined by external pre-stressed strips H. Moghaddam, S. Mohebbi & M. Samadi Department of Civil Engineering, Sharif University of Technology, Tehran, Iran

ABSTRACT: Strengthening is often a necessary measure to overcome an unsatisfactory deficient situation or where a new code requires the structure or a member of it to be modified to achieve new requirements. In the engineering practice such restraint to lateral dilation, indicated by the name of confinement, has been traditionally provided to compression members through steel transverse reinforcement in the form of spirals, circular hoops or rectangular ties. Steel and concrete jackets are other techniques for providing additional confinement for compression members, too. This paper presents the results of experimental and analytical study on the application of strapping technique for retrofitting of concrete columns. Experimental program included axial compressive tests on cylindrical and prismatic small-scale columns which were actively confined by prestressed metal strips. Test Results showed significant increase in strength and ductility of columns due to active confinement with metal strips. The effect of various parameters such as pretensioning force in the strip, number of strip layers wrapped around the specimens and spacing of confining strips on strength of concrete is studied. Nonlinear Finite element models of tested specimens were also made and analyzed. The observed stress-strain behaviors of columns with different levels of confinement are compared to those obtained from finite element method.

1

INTRODUCTION

efficient for use as a repair and strengthening technique for RC structural members.

In this paper, an easy technique of retrofit of concrete is presented. The main aim of this research was quantification of the enhancement of concrete strength and ductility by the application of the technique. The results of experimental and analytical studies on performance of the technique are discussed. This study focused on application of this technique for high strength concrete. The technique used for strengthening concrete columns in this study, involves post-tensioning highstrength packaging straps around the column (by using standard strapping machines used in the packaging industry) and subsequently locking their ends in metal clip. Commercially available strapping tensioners and sealers make it easy to pretension the strip and fix the strip ends in the clamps. The available straps have widths of 10 to 50 mm and thicknesses of 0.5 to 1.12 mm. In terms of strength, high strength strips in excess of 10000 kg/cm2, are available in the market. The strips are tensioned to 25 percent of their yield stress. Hence, an effective lateral stress is applied on the column prior to loading. This has many benefits such as full utilization of the strip capacity and prevention from premature crushing of the confined concrete, as would be the case with not properly tightened strips. The low cost of strip and speed and ease of application of the strapping technique make this method

2

EXPERIMENTAL PROGRAM AND OBSERVATION

The concrete specimens were fabricated in the structure and concrete Laboratory at the building and housing research center (BHRC). Experiments included 25 cylindrical and 15 prismatic concrete specimens. The column models were made of a relatively high-strength concrete with no air-entrainment. The concrete reached an average uniaxial compressive strength of about 50 MPa. The specimens were removed from the forms after 2 days and put into water to be moist cured. The cylindrical and prismatic specimens were tested after 428 days after casting. Two sizes of metal strips were used for strengthening of the specimens. The specimens were strengthened by using 16×0.5 mm and 32×0.8 mm strips. In addition to the difference in width and thickness, the material behavior of the strips was also dissimilar. Although both strips have similar strengths, the elongation of the strips, that is an important characteristic of the confining elements, is quite different. The 32 mm wide strip has larger ductility making it more suitable for application as a confining element for concrete. One of the important parameters in this study was to compare the active and passive external lateral con-

375

finements by this technique. In order to do so, some of the cylindrical specimens were tensioned only to 40 kg (which will be called passively confined specimens hereafter) while a tensioning force of 250 kg was applied in pretensioning the other specimens (which will be called actively confined specimens hereafter). In fact, the metal strips of the latter specimens are tensioned to 0.31 of their yield strain. The results obtained from strain gauges as well as displacement transducers were analyzed. It was observed that the axial stress and the confining pressure kept increasing until the value of lateral strain reached the yield strain of the strips in a circumferential direction. The specimens reached their maximum strengths when one or more of the strips yielded. After the peak stress, the strips ruptured one by one resulting in the loss of axial stress. Column specimens with two layers of the metal strips gained larger strengths as well as a larger ultimate axial strain as compared with column specimens with one layer of the metal strip. The observed stress-strain behavior of some of the test specimens are drawn in the following figures. In these figures, the results of specimens that were actively confined with only one layer of 16 mm strip confined with strips with different clear spacing (s’) are shown. It can be concluded from the figures that: 1. This technique has been able to increase the strength of concrete up to 2.3. 2. An increase in the spacing between the strips, has always led to increase in strength of confined concrete. 3. The concrete confined with double layer metal strips has generally shown better enhancement in concrete strength than confinement with single layer. 4. Active confinement resulted in more increase in concrete strength than the passive one this is mainly because whilst the ordinary passive confinement is mainly utilized after the core concrete has dilated (which means that some cracks have occurred in it); the active confinement influences the core concrete even before load application.

3

NONLINEAR FINITE ELEMENT MODELING

Finite elemnt models of the tested specimens were made by using eight node solid elements in ABAQUS

program. The concrete damaged plasticity model of the program was used for modeling the nonlinear behavior of concrete. This model uses concepts of isotropic damaged elasticity in combination with isotropic tensile and compressive plasticity to represent the inelastic behavior of concrete. It consists of the combination of non-associated multi-hardening plasticity and scalar (isotropic) damaged elasticity to describe the irreversible damage that occurs during the fracturing process. Concrete damaged plasticity model requires that the elastic behavior of the material be isotropic and linear. Model of cylinderical and prismatic specimens consisted of solid elements for concrete with the abovementioned plasticity model and shell elements for modeling metal strips with stress-strain relationship as observed in tensile tests of strips. The bottom surface of models are restrained and the load was applied by incrementally increasing the displacement of nodes of the top surface. There is relatively good agreement between analytical and experimental results. However, the NFEM has underestimated the experimental results. It should be mentioned that the pretensioning force in the strips in actively confined specimens have been applied in the model before application of the incremental displacement. Also the nonlinear finite element method has underestimated the post-peak part of stress-strain behavior of confined specimens.

4

CONCLUSIONS

The applied technique for strengthening of concrete columns could increase strength, ductility of concrete considerably. The technique was able to increase the peak strength and its corresponding strain of concrete up to 230 percent. The gain in ductility of confined concrete was very sensitive to the ductility of the metal strip used. Active confinement resulted in better enhancement of strength and ductility of confined concrete than passive confinement. The efficiency of confinement in cylindrical specimens, i.e. the gain in strength and ductility, was greater than that of prismatic ones. The damaged plasticity model was capable to estimate the behavior of confined concrete with reasonable accuracy. However it underestimates the results of both actively and passively confined concrete.

376

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Flexure and shear behavior of RC beams strengthened by external reinforcement F. Minelli & G.A. Plizzari University of Brescia, Italy

J. Cairns Heriot-Watt University, Edinburgh, UK

ABSTRACT: A novel strengthening technique for Reinforced Concrete (RC) beams based on non-prestressed external reinforcing bars anchored at the supports of simply supported beams or slabs is herein considered. External reinforcement can be easily adopted to retrofit existing reinforced concrete structures. The simplicity of the proposed technique offers advantages in ease of installation, quality control on site and future inspections if comparison to other methods. An analytical and numerical evaluation of the available experimental results as well as a parametric study allows investigation of the effect of different percentages of external reinforcement and of different collapse mechanisms (shear and flexure), in order to optimize the amount of external reinforcement that could provide enough bearing capacity and sufficient ductility prior to collapse. The numerical analyses enable to highlight the collapse phenomena, with emphasis to both the arch and the beam action, the ductility and the stiffness in the cracked stage.

1

INTRODUCTION AND BACKGROUND

The technique is illustrated diagrammatically in Figure 1. High yield threaded bars are applied to both sides of a RC beam, close to the soffit level of the beam. On all but short spans there are benefits from use of deflectors to avoid a reduction in effective depth as the beam deflects. External bars can thus easily be installed by hand. The use of external unbonded reinforcement offers the potential of providing a more cost effective and less disruptive solution to the problem of strengthening simply supported RC beams. The use of external unbonded reinforcement retains many of the merits of unbonded post tensioning but dispenses with the need for specialist stressing operations and expensive tendons and fittings.

Figure 1. Description of the technique herein investigated.

where k = (1.6-d ) with d (effective depth) expressed in meters, and.

τ rd = 0.25 ⋅ VRd ,3 =

2 ANALITYCAL AND NUMERICAL MODELS Beams having both internal bonded and external unbonded reinforcement show a hybrid behavior involving both beam and arch action. Concerning beam action, there are many well established models, such as those included in the international standards. In this study, the model implemented in the previous version of EC2 (1993) is considered, consisting in a stirrup contribution (VRd,3) to which a concrete contribution has to be summed up (VRd,1). The expressions for the concrete and stirrup resisting contributions are the following: VRd ,1 = τ rd ⋅ k ⋅ ( 40 + 1.2ρs ) ⋅ bw ⋅ d

(1)

f ctk , 0.05

γc

.

Asw ⋅ 0.9 ⋅ d ⋅ f sd p

(2)

Concerning the arch action, a number of interesting analytical models were developed starting from the pioneer study of Kani (1966). Three different models will be analytically compared against the experiments and numerical results. They are: 1- Kani’s model (1966); 4 d 1 vu , arch = M fl ⋅ ⋅ ⋅ l a 0.9

(3)

being a the distance between the support and the loading; 2- Kim et al.’s model (1999);

377

⎛ d⎞ vu , arch = 204.8 ⋅ ρ 0 ,9 ⋅ (1 − ρ ) ⎜ ⎟ ⎝ a⎠

r + 0 ,6

2.5

⋅

Deep Beam b=300 mm, d = 960 mm,

(4)

fck = 25 MPa, fsy = 374 MPa

rs = 0,5%

⋅ ( ρ)

−0 ,1

Vr2/Vmax2

⎛ d⎞ being r = ⎜ ⎟ ⎝ a⎠

3- Russo et al.’s model (1991). 5 ⎛ a⎞ vu , arch = 206.9 ⋅ ξ ⋅ ρ 6 ⋅ ⎜ ⎟ ⎝ d⎠

where ξ =

Kim Kim White

1000

2.0 0,6

Kani Russo Zingone Puleri

1.5

Numerical Analysis

300

1.0

− 52

(5) 0.5 0.000

1 1 + d (25 ⋅ d ) a

Figure 2.

The parametric study was conducted on RC beams of various cross sections, all having a concrete with a strength ( fck) of 25 MPa (C25/30) and steel rebars with a yield stress ( fyk) of 374 MPa. A simply supported beam with a span of 5 m and a uniformly distributed load was chosen. The parametric study was performed by considering the following variables for all four cross sections: 1. Internal reinforcement ratio (ρs), chosen equal to 0.5%, 0.7%, 1.0%, 1.5% and 2%. The latter case was then skipped as it turned out not to be feasible for external retrofitting. 2. External reinforcement ratio (ρs,ext), chosen equal to 0, 0.5%, 0.7%, 1.0%, 1.5% and 2%. A brief summary of the parametric study is herein summarized in form of step-by-step procedure: 1. Design for flexure and shear of the beam without external reinforcement. The required shear resistance was set to be 10% higher than the shear corresponding to the flexure failure of the member (with the classical stress-block coefficients included in the current EC2). Stirrups were calculated by using the EC2 requirements. 2. Calculation of the extra shear contribution ∆V (∆V=bw d νu,arch) due to addition of the external reinforcement by considering the three abovementioned analytical models. Calculation of the increase in flexure resistance (due to the external reinforcement) and, consequently, of the corresponding shear force (once again, 10% higher of that corresponding to flexure failure); 3. Calculation of the ratio between the shear resistance of the retrofitted beam Vr2 (sum of the beam and arch action, Vr2=VRd1+VRd3+∆V ) and the shear force corresponding to flexure failure of the strengthened beam Vmax,2. If Vr2 /Vmax,2 is lower than the unity, the member would experience a brittle undesirable shear failure, whereas a desirable flexure failure will occur in the opposite case.

0.004

0.008

rs ext

0.012

0.016

0.020

Vr2/Vmax,2 vs. ρs,ext plot for the deep beam.

Numerical analyses were performed using program VecTor, based on the MCFT, which is a well known model for representing the nonlinear behavior of RC structures. Figure 2 show the Vr2 /Vmax,2 ratio vs. the external reinforcement percentage ρs,ext for the three analytical models (Eqs. 3–5) and for the numerical analyses, in the case of the beam 1 m deep. One can notice that the three analytical models considered give a similar answer with a small external percentage of reinforcement. In most of cases, the shear collapse originates when the external reinforcement percentage is similar to the internal one. This preliminary result should be considered a rather safe but, at the same time, a very easy-to-use design condition for the retro-fitting of beams with this technique. With increasing effective depth, more external reinforcement can be used (up to a ratio ρs,ext/ρs,int=2), given that a proper shear design was performed with the suitable scale effect considerations. In all cases, the model proposed by Russo et al. (1991) seems to be the most conservative , even with respect to the numerical analyses. The latter show that all analytical models are often too safe, especially while dealing with very high total longitudinal reinforcement ratios. 3

CONCLUDING REMARKS

The amount of external reinforcement should be accurately designed in order to have a ductile behavior with the flexural strength slightly lower than the shear strength. Both analytical tools or numerical FE programs can be utilized for a proper design of external reinforcement, even though, if the existing member is properly designed for shear, adding a percentage of external reinforcement approximately equal to the one of the bonded (existing) reinforcement, the failure mode should not change.

378

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Simplified verification of the bond resistance of externally bonded reinforcement at the area of the support moment Christian Muehlbauer, Roland Niedermeier & Konrad Zilch Department of Concrete Structures, Technische Universitaet Muenchen, Germany

ABSTRACT: The resistance of reinforced concrete members, retrofitted with externally bonded reinforcement, i.e. CFRP strips, steel strips or CFRP sheets, highly depends on the load bearing capacity of the bond. Compared to the bond between reinforcing steel and concrete the bond between externally bonded reinforcement and concrete shows a very brittle behavior due to the limitation of bond fracture energy. Thus the transmittable force can only be increased to a certain limit with increasing the bond length. In order to transmit high forces cracks along the bond length have to form in the concrete with increasing the load to the retrofitted concrete member. The forming of the tensile force in the externally bonded reinforcement occurs at the single sections, which are located between the cracks. Thus the bond resistance of the externally bonded reinforcement has to be proofed over the whole area of the flexural member, which is subjected to shear. Some standards on retrofitting concrete members with externally bonded reinforcement only provide a verification of the anchorage at the end of the bond length similar to the verification of reinforcing steel and a global limitation of the strain of the externally bonded reinforcement. Other design concepts recommend a limitation of mean bond stress neglecting bond fracture limitation and unpropitious crack pattern. These procedures may lead to an uncertain design. At the department of concrete structures at Technische Universitaet Muenchen a lot of research work was done to proof the bond resistance of the externally bonded reinforcement over the whole area of a flexural member, which is subjected to shear. For the most critical case of a beam at the area of the support moment a simplified verification of bond capacity is presented.

1

INTRODUCTION

At the Department of Concrete Structures at Technische Universitaet Muenchen a lot of research work was done to proof the bond resistance of externally bonded reinforcement over the whole area of a flexural concrete member, which is subjected to shear. From this work a simplified verification of bond capacity for the most critical case of a beam, which is retrofitted with externally bonded reinforcement, at the area of the support moment was derived. The procedure of this simplified verification is as follows: First of all the decisive single section, which is located between two cracks, has to be determined. In the following article this single section is called decisive crack element. At the area of the support moment this decisive crack element is located at the point where the value of the flexural moment and the shear stress has its maximum. Thus the decisive crack element is located at a distance of al, which is the shift measure, from the intermediate support (see Figure 1). The decisive crack element can be determined with a maximum crack moment of the concrete Mcr,max and a maximum crack spacing ar,max. For this crack element the existing stress increase exist∆σ1L in the externally bonded reinforcement has to be determined. The

stresses at the cracks ∆σ1L and the existing stress increase exist∆σ1L in the externally bonded reinforcement have to be determined according to a strain situation supposing a plane strain distribution. Furthermore the resisting stress increase res∆σ1L at the decisive crack element has to be determined. This can be done using an equation whereas the resisting stress increase res∆σ1L mainly depends on the fracture energy of the concrete and the stress in the external reinforcement at the lower stressed crack. Of course the stiffness and the geometry of the reinforcement have a stake in the resisting stress increase, too. The verification of resisting stress increase is provided if the existing stress increase exist∆σ1L is less than the resisting stress increase res∆σ1L.

2

DETERMINATION OF CRACK FORMATION

The decisive crack element can be supposed to be located at the point where the maximum values of flexural moment and shear force meet themselves. That is at a distance al, which is the shift measure, from the intermediate bearing with a maximum crack spacing ar,max (see Figure 1).

379

Figure 2. Maximum possible strength increase in the externally reinforcement res∆σiL(σiL) at crack element. Figure 1.

3

Location of decisive crack element.

res∆σ 1L , k =

DETERMINATION OF EXISTING STRESS INCREASE EXIST∆σ1 L

The bond load is defined as stress difference exist∆σ1 Λ between two cracks. Therefore the stresses ∆σ1L(x1) and ∆σ1L(x2) at the decisive crack element have to be determined at ultimate limit state. The existing stress increase exist∆σ1 Λ in the externally bonded reinforcement can then be calculated as follows: exist ∆σ 1L = σ L2 − σ 1L

4

+ (σ 1L , k )2 − σ 1L , k

(2)

The verification of bond capacity is provided, if the design value of resisting stress increase res∆σ1 Λ,δ at the decisive crack element is greater than the design value of existing stress increase exist∆σ1 Λ,δ: exist ∆σ 1L , d ≤ res∆σ 1L , d

6

The bond resistance is defined as maximum possible stress increase res∆σιΛ in the externally bonded reinforcement between two cracks. According to Niedermeier 2001 the maximum possible stress increase res∆σιΛ(σιΛ) is a function of the stress ∆σ1L at the lower stressed crack edge at the location xi, see Figure 2. For the simplified verification of the bond resistance of externally bonded reinforcement at the area of the support moment the resisting stress increase at the decisive crack element (see Figure 1) can be determined using equation (2):

tL

5 VERIFICATION OF BOND CAPACITY

(1)

DETERMINATION OF BOND RESISTANCE

2 ⋅ GF , k ⋅ E L , m

(3)

SUMMARY

A simple and safe design concept to proof bond capacity of flexural concrete members strengthened with externally bonded reinforcement is presented. Based on this design model more factors that increase the bond capacity can be considered and integrated to this design model. That is subject of current experimental and theoretic research work that is done at the Department of Concrete Structures at Technische Universitaet Muenchen.

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Loading test on a retrofitted pretensioned concrete girder after fire L. Taerwe & E. Annerel Magnel Laboratory for Concrete Research, Department of Structural Engineering, Ghent University, Ghent, Belgium

ABSTRACT: This paper describes a real scale fire test with details about the temperature development in the building and in some concrete members. An industrial hall erected with precast concrete elements was constructed to serve this purpose. A roof girder which was seriously damaged during the fire, was repaired by shotcreting and submitted to a static loading test up to failure. It was found that the girder showed a load bearing capacity comparable to a reference girder. Combination of simple calculation in combination with the finite element program DIANA, used for the temperature development in the girder, confirms this result.

1

INTRODUCTION

In 1970’s, the regulations in Belgium required 2 hours of fire resistance for structural elements in high-rise buildings. This requirement was current for both the precast and the cast-in-place concrete. However, only the precast industry had to prove this fire resistance with fire tests, while for the cast-in-place concrete analytical methods were available. The problem was that these analytical methods over-estimate the fire resistance, putting the precast industry in an disadvantageous position. To understand the behaviour of buildings composed of precast concrete elements during a fire, a full scale experiment was performed. An industrial hall was built for this purpose, composed of commercially available elements, with their actual connections, realistic loads and submitted to a completely controlled fire. This experiment had to yield information concerning the assumption that concrete structures have a better fire resistance than separate elements, but should also study the effect of thermal deformations on the structure. Furthermore, the fire itself was studied, namely, the maximum temperature, the spreading of the temperature in the different parts of the building and the influence of the air supply. After the fire, a structural assessment of a repaired pretensioned roof girder was done. 2

DESCRIPTION OF THE BUILDING

The building was 12 m × 18 m in plan, had a free height of 6 m under the roof girders and consisted completely of precast concrete elements. The load bearing structure consisted of three portal frames, each consisting of two columns (cross section 400 mm × 400 mm) with a girder that supported the roof. The columns were anchored in the foundations. Wood was selected as fuel. Two preliminary tests were carried out to determine the amount of wood, the

geometry of the timber and the cribs, that would give the high temperatures required by ISO Standard 834, during a sufficiently long time. The total amount of wood used was 27 tons, meaning 125 kg/m², sawn in small beams, loosely stacked and dried. 3

TEMPERATURE DEVELOPMENT DURING THE FIRE

The ignition of the wood stacks took about 2 minutes, but it was only after 34 minutes, when the lightdomes failed, that the fire fully developed (flames escaping through the roof) and the inside temperature reached the level of the ISO-curve. From that moment on, the fire rate was roughly controlled by closing and opening of the doors. It was found that despite the very high amount of fuel that was used and the ventilation of the building, the heating curve of the ISO standard 834 was followed during only 1 h 15 min. Reliable measurements were not available anymore at that moment, because thermocouples had fallen down in the layer of burning charcoal, while others were mechanically interrupted. The fire extinguished after 120 minutes by lack of fuel. 4

PRETENSIONED ROOF GIRDER

The pretensioned roof girder with a total length of 18 m, was removed from the test site for further tests and observations to the Magnel Laboratory for Concrete Research. The I-shaped cross-section had a total depth of 1.2 m at midspan and 0.75 m at the supports. Both flanges had a width of 0.4 m and the thickness of the web equalled 0.12 m. Twelve 7-wire prestressing strands (nominal diameter 12.7 mm) were located in the lower flange and two strands in the upper flange. After the fire test, the girder showed quite severe damage. At most places surface spalling

381

could be observed. In some parts of the web severe spalling to depths of several centimetres occurred, even resulting in a hole through the full thickness of the web. At several places stirrups and longitudinal rebars were visible, while at some places the concrete cover of some strands also had disappeared. After a shotcreting repair, the girder was submitted to a static loading test with a span of 17.6 m. The service load of the girder was 15 kN/m (exclusive dead weight) which corresponds to four equivalent point loads F of 66 kN. During the static loading test, the four point loads were raised in steps of 5 kN up to the individual service load of 66 kN. After an unloading cycle, the four point loads were raised up to 66 kN again, after which they were further raised in steps of 10 kN up to failure. At each step the deflection and the deformations of the concrete at midspan and at points at one fourth of the spanwidth were carefully followed, as well as the rotations at the supports. Failure occurred at a load level of 162 kN per point load, which was 2.45 times higher than the service load. Failure occurred by crushing of the concrete in the upper flange after yielding of the prestressing strands in a cross-section where a broken strand could be observed after the fire test. This example shows that although the girder appeared to be quite seriously damaged after the fire, still a sufficient safety margin could be achieved by applying an appropriate repair technique. The residual compressive strength was determined on six cores (Ø 50 mm, height 56 mm), taken from the upper part of the web. The residual prestressing force is calculated as the difference in deformation of a reference length before and after cutting the strands. This measurement was done in zones close to an end block, where the concrete cover was removed and reference points were glued on the strands. To determine the residual mechanical properties of the strands, a tensile test was carried out on specimens from each strand after the static loading test. Recording of the stress-strain diagram provides the tensile strength, Young’s modulus and the 0.2 % proof stress. The strands at the corners of the cross section suffered the most from the fire load. 5

CONCLUSIONS

− The structural behaviour of the building was judged satisfactory as no major collapse occured.

− The experimental fire has given results which were considered to be more favourable than the tests carried out in laboratory conditions on separate elements. − The severe fire load resulted in a fully developed fire during about 1 hour and 15 minutes after an initial heating period of about 20 minutes. Hence, it can be concluded that in reality it is almost impossible that a fire with the usually encountered fire loads will last for two hours in the same compartment. However, during a fire test in laboratory conditions, the fire is maintained during two hours under the most unfavourable loading conditions. Indeed, in the 1970's the regulations required to apply the full service load during a fire test. − A pretensioned roof girder with a span of 17.6 m, which was severely damaged during the fire, was repaired by shotcreting and submitted to a static loading test. A sufficient safety margin was reached and demonstrates that concrete members can be reused after a fire by applying an appropriate repair technique. − Calculation of the residual resisting moment, based on experimentally determined residual properties, agrees adequate with the moment obtained during the static loading test. − FE analysis simulates after 90 minutes a temperature profile in the girder corresponding to a exposure of 85 minutes ISO curve.

ACKNOWLEDGEMENT The authors gratefully acknowledge the financial support from the former “Institute for Encouragement of Scientific Research in Industry and Agriculture (IWONL/IRSIA)” and the Belgian Federation of the Concrete Industry (FeBe). The project was co-ordinated by a working group consisting of representatives from the Laboratory for Fuel Technology and Heat Transfer (Ghent University), the Magnel Laboratory for Concrete Research (Ghent University), the Ministry of Public Works (Building Division), the Belgian Research Centre of the Cement Industry (CRIC), the Belgian Building Research Institute (WTCB/CSTC), the National Society for Fire Safety, the Fire Brigade Department of the city of Ghent and the Federation of the Concrete Industry.

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Anchorage failure of RC beams strengthened with FRP at the bottom face Fedja Arifovic & Björn Täljsten Department of Civil Engineering, Technical University of Denmark, Denmark

ABSTRACT: When RC beams are strengthened by FRP plates bonded to the bottom face of the beam they fail due to the lack of the mechanical anchorage of the plate. This paper shows that it is possible to relate the anchorage failure of tension face FRP-plated beams to the crack sliding capacity of the beams. The method suggested gives the highly desirable occurrence of the critical diagonal shear crack. The occurrence of this crack is suggested as the anchorage failure mechanism of the plate. The failure load that produces the sliding in such crack is shown to be a lower bound to the anchorage failure load of tension face plated beams. The assumption is that the FRP-plate must be extended at least one critical anchorage length beyond the theoretical occurrence of the critical diagonal crack.

1

INTRODUCTION

When an RC beam is strengthened by FRP plates bonded to the bottom face of the beam and subjected to bending it fails due to the anchorage failure of the plate. This means that the full composite action of the plate can not be expected. This paper suggests the occurrence of the critical diagonal shear crack as the governing failure mechanism of the anchorage failure. The occurrence of this crack is found by the means of the crack sliding theory. Thus, this solution suggests the shear resistance of an unplated RC beam without internal shear reinforcement as the lower bound to the load that produces the anchorage failure of tension plated RC beams. This requires a simplification of the dominant failure modes of tension face plated RC beams to the diagonal cracking failure (the diagonal cracking is shown in Figure 1). The crack sliding theory is developed for the beams without shear reinforcement by defining the load that produces diagonal cracking in such beams. These beams fail in shear by a major diagonal crack, and by demanding the sliding in such a crack the shear carrying capacity of the beam may be found. The reason for using the crack sliding theory is that it is based on the upper bound solutions of the theory of plasticity. The upper bound solutions are based on the geometrically possible strain fields. Thus, on the basis of a predicted failure mechanism the maximum energy dissipation in the structure may be found. By means of the work equation this leads to the carrying capacity of the structure. Since the upper bound solutions do not require any knowledge on the difficult stress state in the structure, they seem obvious for producing the simple design equa-

Figure 1. Debonding failures of a tension plated RC beam.

tions at the limit state for RC beams strengthened by bonded FRP. 2 2.1

METHODS Crack sliding theory

The crack sliding theory takes into account the cracking that influences the failure. It is the last formed diagonal crack that initiates at the bottom face and propagates to the load point that is of the interest. The theory is for beams without shear reinforcement. By the means of the crack sliding theory it is possible to predict a theoretical occurrence of the critical diagonal shear crack. It is well known that this crack is the failure mechanism that governs the failure of beams without shear reinforcement. The failure occurs

383

Figure 3. Anchorage failure mechanism.

Figure 2.

Crack sliding theory.

by sliding in this critical crack. Figure 2 illustrates this. The crack sliding criterion may be written as ⎛a− x⎞ 1+⎜ 2 ⎡ ⎝ h ⎟⎠ a− x⎞ a−x⎤ ⎛ ⎥ = f t . ef − fc ⎢ 1 + ⎜ ⎟ a 1c ⎝ h ⎠ h ⎥ ⎢⎣ + ⎦ h 2h

2

(1)

here x is the distance from the face of the support plate to the intersection of the critical diagonal crack with the bottom face. fc and ft:ef are the concrete compressive strength and the effective concrete tensile strength, respectively. Sliding failure in such a crack involves the use of the effectiveness factor ν for the compressive strength of cracked concrete. ν is a product of ν0 and νs. νs is the crack sliding factor and is set to 0.5 and ν0 is a function of the concrete compressive strength fc, beam height. ν0 is thus written as

ν 0 = λ f1 ( f c ) f 2 ( h) f 3 (φ )

(2)

λ is the load factor: 1.6 for point load. f-function are given to f1 ( f c ) =

3.5 fc

[MPa ]

(3a)

1 ⎞ ⎛ f 2 ( h ) = 0.27 ⎜ 1 + ⎟ [m ] ⎝ h⎠

(3b)

f 3 (φ ) = 15φ + 0.58 [% ]

(3c)

2.2

at the critical anchorage length (case 3) and outside the critical anchorage length (case 4). The occurrence of the crack is approximately found by the crack sliding criterion in which the beams are assumed overreinforced (tensile reinforcement does not yield at the failure). It is suggested to approximate the critical anchorage length of the FRP-plate to the ones found by simple anchorage strength models. Thus, when calculating the anchorage failure load the approach is to define the theoretical occurrence of the critical diagonal crack by satisfying the crack sliding criterion. If the FRP-plate is extended at least one anchorage length beyond the intersection between the crack and the bottom face (i.e. if positions 3 or 4 from Figure 3 are achieved) the crack sliding solution is on the safe side. It is thus assumed that the load carrying capacity is increased by increasing the anchorage length beyond the theoretical occurrence of the critical diagonal crack. If there is no sufficient anchorage length a mechanical anchorage must be provided. 3

The experimental failure loads that are due to a test database in which all tests exhibited concrete cover separation failure are compared to those using the crack sliding solution. In the calculation it is assumed that all beams were provided with FRP-plates with sufficient anchorage length (i.e. at least one anchorage length beyond the theoretical occurrence of the critical crack). The comparison showed that the crack sliding solution renders safe failure loads with a mean value of 0.8. The coefficient of variation is 20% and the correlation is 0.94. 4

Theory for anchorage failure of tension face plated RC beams

The anchorage failure of tension face plated beams is assumed governed by the critical diagonal shear crack. In Figure 3 it is illustrated that the crack may intersect the bottom face outside the range of the FRP (case 1), inside the critical anchorage length (case 2),

RESULTS

DISCUSSIONS AND CONCLUSIONS

It is shown the crack sliding failure predicts a load carrying capacity that is a lower bound to the load that produces the anchorage failure of a tension face FRPplated beam. The demand is that the FRP-plate extends at least one critical anchorage length beyond the theoretical occurrence of the critical diagonal crack.

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Strengthening and verification of the prestressed road bridge using external prestressing M. Moravcik & I. Dreveny University of Zilina, Zilina, Slovak republic

ABSTRACT: Many bridge structures were designed across the Váh river. The rehabilitation method one of such road prestressed a concrete bridge is presented in this paper. New condition of channel operation, diagnostic inspection of bridge, long term measurement of deflections and control Load Carrying Capacity (LCC) calculation constitute the major background for design of bridge rehabilitation. Strengthening of superstructure by external prestressing was finally selected as a way to increase the load carrying capacity.

1

THE BRIDGE STRUCTURE TECHNICAL CONDITION

LONGITUDINAL SECTION

HVOZDNICA

STAROVEC

A lot of new channels were built-up on the Vah river in Slovakia during 50th–60th years. There was realised more types of bridge structures as steel truss girders, reinforced concrete continuous beams, precast prestressed bridges and frame monolithic structures. The Hvozdnica bridge was built near the Bytca town at 1959 on the road that is connecting two little towns on both sides of the channel. It is one span direct structure where the structural system is composed from prestressed two-hinge frame with the ties in ending parts, see the Figure1. The span is 63,40 m and frame girder has parabolic profile from 3,42 m over the abutment to 1,26 m in the centre of the span, see the Fig. 1. The typical deterioration and failures were observed on the bridge structure. Some of them are caused by insufficient drainage system on the bridge. Waterproofing membrane was damaged so the water passes trough the cracks in the slab. The main failure on the superstructure is visible large deflection around 187 mm, due to supposed crucial influence of the concrete creep. This deformation has negative affected to dynamic structural behaviour too. The cracks approximately 0,7 mm width were observed just in bottom slab in middle cross section along the middle 10 m length. Concrete strength corresponds with proposed value. The concrete carbonation achieved the level 18–54 mm till to the reinforcement. The chloride content in concrete approximately 4-time exceeded the value compare to allowable standard value for prestressed bridges. Local corrosion appearance was found out due to the insufficient concrete cover. The following goals were formulated. To eliminate all defects and their causes on bridge structure, to fulfil the required conditions of channel regime, and to increase present load carrying capacity that

VAH CHANNEL

CROSS-SECTION

Figure 1. The Hvozdnica Bridge. Longitudinal and Cross Section.

are: normal load carrying capacity from 12,0 tonnes to 26,0 tonnes at least and exclusive load carrying capacity from 17,0 tonnes to 50,0 tonnes at least.

2

STRENGTHENING OF THE BRIDGE

The main goal of superstructure strengthening has been increase the load carrying capacity (LCC) to requested values. LCC analysis was based on allowable stresses calculation according to Slovak standard, that means to achieve the decompression state or the limit values of compression in concrete along the cross section. Two alternatives were considered: 1) to change the bridge structural acting to the new stay cable system with the new pylons creation, 2) applying the external unbonded prestressing led out of the box girder using the lifting effect of the cables, see the Fig. 2. Finally the second one was chosen. It has appeared to be more advantageous compare to first method

385

from its simplicity, economy and the short time of rehabilitation works point of view. The new external cables consisting from 6 tendons 7 φ Lp15,5–1800 MPa were designed to the strengthening of structure, see the Fig. 3. The rigid steel strut frames were designed as deviators with 1,80 m depth.

3

CONCLUSION

The following results and experiences of the bridge rehabilitation can be discussed: − The accurate diagnostic inspection, analysing of the failure reasons and LCC calculation compose the general background for the determining technical condition of the bridge. − Using external unbonded prestressing seems to be as a simple and effective method of the structure strengthening, see the Fig. 4. In this way such heavy exploited bridge structures can increase the live load corresponding to the present traffic requirements. External prestressing imagines fast and economy way of such types box girder bridges rehabilitation with. Also the residual negative deformation was eliminated by upper slab concreting. − Using external prestressing composed from 6 tendons 7 φ Lp15,5 achieved required values of load carrying capacity to the level Vn = 31 tonnes (Normal LCC), Vr = 96 tonnes (Exclusive LCC) and Ve = 143 tonnes (Exceptional LCC). Sections

Stresses [MPa]

Figure 2. The external prestressing effect.

Resulting stresses - bottom fibres

3,00

span [m] <0; L/2>

0,00

Figure 4. The final status of the bridge.

Figure 5. The in situ load test.

over the support, in mid-span and around L/4 span are exposed as a critical sections. As the decisive load combination was analysed complex load combination—with changing temperature effect (getting warm or cool) with load combination factor 0,7. − The in situ load test, see the Fig. 5, has approved the desired effects of the superstructure strengthening as followed: the lifting effect of prestressing, all critical cracks closing and the elastic behaviour of structures can be considered.

-3,00 before strength.

-6,00

inter. cable line

-9,00

lim. stresses

-12,00

ACKNOWLEDGEMENT

lim. stresses dev.=1,8m

-15,00 0

2

4

6

8

10 12 14 16 18 20 22 24 26 28 30 32 34 36 38 40

Figure 3. The internal and external cables effect.

The author gratefully acknowledges the support of Grant Agency of Ministry of Education of Slovak republic, VEGA 1/332/07.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Numerical simulation of continuous RC slabs strengthened using NSM technique J.A.O. Barros & G. Dalfré University of Minho, Guimarães, Portugal

J.P. Dias Civitest Company, Braga, Portugal

ABSTRACT: The effectiveness of the Near Surface Mounted (NSM) technique for the increase of the flexural resistance of Reinforced Concrete (RC) beams and slabs was already well proved. The NSM technique is especially adapted to increase the negative bending moments of continuous (two or more spans) RC slabs. However, the influence of the NSM strengthening on the moment redistribution capability of RC structures should be investigated. Recently, an exploratory experimental program was conducted to assess the level of moment redistribution that can be obtained in two span RC slabs strengthened with NSM strips for negative bending moments. To help the preparation of an extensive experimental program in this domain, the values of the parameters of a constitutive model, implemented into a FEM-based computer program, were calibrated from the numerical simulation of these tests. The main aspects of the experimental program are presented, the numerical model is briefly described and the numerical simulations are presented and analyzed.

1

INTRODUCTION

When an existing continuous RC structure is strengthened with Fiber Reinforced Polymer (FRP) materials, its ductility and “plastic” rotation capacity may be restricted or even extinct, due to, principally, the linear-elastic stress-strain response of the FRP up to its brittle failure. Tests on simply supported RC members strengthened with Near Surface Mounted (NSM) strips have shown that NSM strengthening elements debond or fail at much higher relative strain levels than Externally Bonded Reinforcing (EBR) strengthening systems. In general, NSM strengthened members are expected to be much more ductile than EBR strengthened members. Therefore, NSM technique is recommended for strengthening negative moment regions, since for this application the NSM technique can be restricted to open slits and fix, with an adhesive, the FRP strips to the concrete substrate. On the topic of moment redistribution of statically indeterminate RC members strengthened with NSM technique, a significant amount of moment redistribution was attained with NSM, when compared with EBR technique. With the final purpose of establishing design guidelines for the NSM flexural strengthening of continuous RC structures, an exploratory experimental program was recently conducted and the level of moment redistribution that can be obtained in two span RC slabs strengthened with NSM strips for negative moments was assessed. The values of the parameters of a constitutive model, implemented into

the FEMIX software, a FEM-based computer program, were calibrated from the numerical simulation of these tests. This model is able of simulating the concrete crack initiation and crack propagation, the nonlinear concrete compression behavior, the elastoplastic behavior of steel reinforcements and the elastic-brittle failure behavior of FRP elements. 2

EXPERIMENTAL PROGRAM

To assess the influence of NSM CFRP strips on the moment redistribution ability of continuous RC slabs, an experimental program composed of nine 120 × 375 × 5875 mm3 RC two-way slabs (Figure 1) was carried out, three of them were unstrengthened RC slabs forming a control set (SL15, SL30 and SL45), and six slabs were strengthened with CFRP strips according to the NSM technique (SL15s25, SL15s50, SL30s25, SL30s50, SL45s25 and SL45s50). The notation adopted to identify each slab specimen is SLxsy, where SL is the slab strip base, x is the moment redistribution percentage, MR, (15%, 30% or 45%), s means that the slab is strengthened, and y is the increase of negative moment of the slab cross section at its intermediate support (25% or 50%). 3

NUMERICAL SIMULATION

The concrete slab is considered as a plane shell formulated under the Reissner-Mindlin theory. In order to simulate the progressive damage induced by con-

387

F

F

120 110 100 90 80 70 60 50 40 30 20 10 0

120 1400

1400

125

2800

2800 5850 Figure 1. Test configuration.

130 120 110 100 90 80 70 60 50 40 30 20 10 0

Total load F (kN)

0

Figure 2.

5

10

120 110 100 90 80 70 60 50 40 30 20 10 0

( LVDT 18897)

15 20 25 30 35 Displacement (mm)

F

SG11 SG12

SG13 SG14

-1000 -2000 -3000 -4000 -5000 -6000 Strain (µm/m)

Experimental Numerical SG17 SG17 SG18 SG18/SG19 SG19

Total load F (kN)

( LVDT 60541)

F

Figure 3. Force—Concrete strain relationships at slab loaded sections.

SL15 SL15s25 SL15s50 Numerical LVDT 18897 LVDT 60541 F F

0

40

Force-loaded section deflection relationship.

crete cracking and concrete compression nonlinear behavior, the shell element is discretized in layers. Each layer is considered in a state of plane stress. The constitutive laws for the concrete, steel and CFRP strips and the values of the parameters adopted for the numerical simulations are described in the full-length paper. Due to the structural symmetry, only half of the slab was considered in the numerical simulations.

4

Experimental Numerical SG14 SG12/SG14 SG13 SG11/SG13 SG12 SG11

Total load F (kN)

1400

1400

125

TOP VIEW CONCRETE SGSs

F Laminate strips F

TOP VIEW LAMINATE STRIP SGSs

0 Figure 4.

4000

SG18 SG17 SG19

8000 12000 16000 Strain (µm/m)

20000

Force-CFRP strain relationship for the strips.

strengthened slabs had moment redistribution capability similar of the reference slab. ACKNOWLEDGEMENTS

RESULTS

Figures 2 to 4 present some of the main results of the simulations and show that the numerical model is able to capture with good accuracy the behavior of the constituent materials of this structural system during the loading process of the tested slabs. The comparison between the strains registered experimentally in the CFRP laminates and the strains determined by the numerical model shows that it can be assumed a perfect bond between NSM laminates and surrounding concrete. The contribution of the NSM laminates for the slab’s load carrying capacity was limited by the concrete crushing. The NSM

The authors wish to acknowledge the support provided by the “Empreiteiros Casais”, S&P®, Secil (Unibetão, Braga) Companies. The second author would like to acknowledge the National Council for Scientific and Technological Development (CNPq)—Brazil for financial support for scholarship (GDE/CNPq). The study reported in this paper forms a part of the research program “CUTINEMO—Carbon fiber laminates applied according to the near surface mounted technique to increase the flexural resistance to negative moments of continuous reinforced concrete structures” supported by FCT, PTDC/ECM/73099/2006.

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Non-linear finite element modeling of epoxy bonded joints between steel plates and concrete using joint elements L. Hariche Department of Civil Engineering, Djelfa University, Algeria

M. Bouhicha Civil Engineering Research Laboratory, Laghouat University, Algeria

ABSTRACT: The issue of maintenance and repair of existing structures has become a major issue, particularly extending the service lifespan of reinforced concrete structures. Amongst various methods developed for strengthening and rehabilitation of Reinforced Concrete (RC) structures, external bonding of steel or Fibre Reinforced Plastic (FRP) strips to the beam has been widely accepted as an effective and convenient method. However, the development of a design method that can properly describe and predict the behaviour of a strengthened RC beam is an extremely difficult task because of the relatively large number of likely failure modes that have been reported by researchers. One of the most critical failure modes encountered in plated beams is the premature and brittle rupture of the plate to concrete bond. This debonding of the plates tends to occur when the bond strength is reached locally. Consequently, the integrity of the strengthening system does not depend solely on the plate material but also on the properties of the interfaces involved in the joint, namely, the plate-adhesive and the adhesive-concrete interfaces. In this paper, a non-linear finite element model is developed for studying the behavior of epoxy adhesive bonded concrete-steel joints. The predictions of the model have been validated with experimental test data found in the literature. The study was conducted for varying plate and adhesive thicknesses, and the behaviour of the joint was assessed in terms of load carrying capacity and the shear stress distribution at the interfaces and in the adhesive. The results show that a non-linear finite element model may reliably predict the behaviour of epoxy adhesive joints.

P

Experimental Setup: Strain gauges

Steel Plates

P 6

4

Glue

3

40

60

5 30

4 30

3 30

2 1

10 10

30

Figure 2. Strain gauge positions. Concrete

60 150

Figure 1.

1

Experimental set-up for the joint test.

CONCLUSIONS

This study has shown that nonlinear finite element analysis may reliably predict the behaviour of epoxy bonded steel plate to concrete joints. Detailed information on the behaviour of such joints can be obtained from this analysis. The following results were observed:

1. Failure of the joint was always initiated in the concrete surface. Therefore, the load capacity of the joint is strongly correlated to the shear strength of concrete. The failure of the joint is gradual and propagates along the joint with the increase in loading. 2. The distribution of shear stresses along the bonded joint follows an exponential trend. The peak value of this stress tends to be concentrated in a very small region of the joint. 3. The increase of the glue thickness does not seem to have a very significant effect on the load carrying capacity of the joint [Muravljov 1994, Teng 2002]. 4. The existence of an anchorage length of the joint is confirmed during this study. It was found to be equal to 40 times the plate thickness.

389

20.00

Local force (kN)

5KN

Shear stress (N/mm2)

10KN 6.00

15KN 20KN

4.00 2.00

10KN 15.00

15KN 20KN

10.00 5.00

0.00

0.00 50.00

50.00

0.00

150.00

0.00

Figure 3. joint.

5KN

100.00

Experimental shear stress distributions along the

Figure 4. joint.

150.00

100.00

Experimental local force distribution along the

Finite Element Modeling of the Joint:

2

Symmetry

ftj2

Blocked nodes 12

x

Interface element

Compression (+)

Tensile (-) Kn Kt 72

Figure 5.

105 Rmax

-ftj

x

Finite element mesh used.

Figure 6.

Failure criterion for the concrete surface.

1.00

2.0

2. 0

0.00

0

40

80

120

160 0. 0

Shear Stress (N/mm2)

-1.00

Figure 7.

-2.00

-300

0

40

80

120

160

-2. 0

10KN 15KN 17,5KN 20KN

0.4fcj

0.0

0

40

80

120

160

-2.0

-4. 0

10KN 15KN 17,5KN 20KN

-6 .0

-4.0

-60

10KN 15KN 17,5KN 20KN

Shear stresses distribution in the concrete, the adhesive, and the steel along the joint.

5. The load capacity of the joint studied herein may be estimated by: P = 4.851 tp + 2.283 ta + 5.386 ft—8.936 Where: tp = plate thickness ta = adhesive thickness

ft = tensile strength of concrete. 6. The above results confirm and complement the available experimental observations and show the importance of using numerical models to analyse the behaviour of epoxy joints.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Increase in the torsional resistance of reinforced concrete members using Textile Reinforced Concrete (TRC) F. Schladitz & M. Curbach Institute of Concrete Structures, Technische Universität Dresden, Dresden, Germany

ABSTRACT: There are various methods by which to improve the torsional resistance of existing reinforced concrete members. One method entails the use of Textile Reinforced Concrete (TRC). Test results will be used to demonstrate that TRC combined with Alkaline-Resistant glass (AR-glass) can be used to strengthen the torsional resistance of quadratic reinforced concrete members. Both, an increase of improved resistance as well as improved serviceability, will be presented. Reduced torque and decreased crack widths are specific results that will be emphasized. An initial approach to determine the torsional resistance of square strengthened concrete components reinforced by TRC will also be introduced.

1

INTRODUCTION

The necessity to improve the structural behaviour of existing reinforced concrete members occurs more and more frequently as a result of building conversions, increases in load scenarios and corrective structural measures (e.g., shoring up walls, etc.). To date, these improvements are largely provided by strengthening through the use of reinforced concrete. The consequence of using such strengthening measures requires reinforcement dimensions and concrete cover (e.g., corrosion protection) of several centimetres, where by the cross-section and the dead load are drastically enlarged. In numerous studies different researchers have investigated whether or not textile reinforced concrete can be used to strengthen members, and if so, precisely how this works (Brameshuber 2006, Hegger et al. 2006). Textile reinforced concrete is a composite material with layers that consist of high tensile, fine-grained concrete and textile reinforcing. The small maximum aggregate size of 1 mm for the finegrained concrete matrix allows for minimal layer sizes of about 2 mm. AR-glass or carbon fibers, which were previously converted to textile fabric, are used as fiber material (Fig. 1). Depending on the anticipated load capacity, fabrics with up to four reinforcement directions can be produced. Thus, the reinforcing layer can be systematically strengthened while simultaneously optimizing the use of the carrying capacity. The qualification of the use of TRC in the retrofit of flexural components was already proved by several tests (Brückner et al. 2004, Weiland et al. 2006) and practical applications (Weiland et al.). The reinforcing effect regarding shear force resistance was demonstrated with beams and T-beams (Brückner et al. 2007, Schladitz et al. 2008).

Figure 1. Textile fabric of the reinforcement.

The improvements of the torsion-bearing strength as well as the mathematical determination of the carrying capacity of TRC strengthened members that have been achieved within the first tests will be shown in the following.

2

EXPERIMENTAL INVESTIGATIONS AND CONCLUTIONS

The experimental investigations included un-strengthened reference members as well as members with varying numbers of layers of textile reinforcement (Fig. 2 and Tab. 1). A fabric of AR-glass with a mass per unit area of 263 g/m² was used as textile reinforcement. The rovings used provided a fineness of 1200 tex (that means 1200 g/km length) and were arranged in at a ±45°configuration with 10.8 mm spacing. Components were examined in an experimental rig in which it was possible to measure the torsional loads. The two presses in opposite directions with

391

Figure 2. Arrangement of the strengthening. Table 1. Tested members.

Figure 3. Torque-twist curve.

Members

strengthening

1.1 and 1.2 2.1 and 2.2 3.1 and 3.2 4.1 and 4.2

Reference without strengthening 2 layer AR-glass fabric* 4 layer AR-glass fabric* 6 layer AR-glass fabric*

* Fabric name: NWM-3-003-07-p2 (15%).

equal loads can almost entirely eliminate shear forces and bending moments. As the experiments showed, the torsional resistance of the reinforced members was higher than that of the unstrenghtened members. The resistance increased proportionally as the number of textile reinforcement layers grew (Fig. 3). It can be seen that the strengthened members had a noticeably smaller twist

at the same load level as that of the unstrengthened members. The test results have shown the possibility of strengthening reinforced concrete members with TRC. Both, the torsional resistance and the serviceability, can be improved considerably by the use of textile reinforcement. Calculation show that the steel reinforcement as well as the textile reinforcement share in the load transfer. In addition, the torsional resistance can be investigated by existing strut and tie models. Future analyses will emphasize members which have traction strut angles other than 45° as well as units of differing geometric configurations. This way, a model for calculations is to be established to facilitate later practical application.

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Rejuvenating our aging structures: Cumberland street parking garage rehabilitation Philip Sarvinis Read Jones Christoffersen Ltd., Toronto, Canada

ABSTRACT: Built in 1968, this multi-level concrete parking facility has been subjected to years of moisture and chloride ion attack causing it to experience corrosion-related deterioration to the point that its structural integrity was being compromised. Over the years, moisture and chloride contamination has caused both the embedded mild steel and paper wrapped “Button-Head” post-tensioning to experience corrosion-related deterioration, which in turn, has resulted in the failure of the post-tensioning system. After years of monitoring and structural capacity calculations, sections of the structure were deemed to be no longer able to safely support the applied loading. This paper will illustrate the procedures implemented in the design of the new post-tensioning system, which enabled us to rejuvenate this 40-year old structure and extend its overall effective service life. This paper will also outline how the new system was installed while the parking facility remained operational.

1 1.1

BRIEF DESCRIPTION OF THE FACILITY General description

The Cumberland Street Parking Garage is located in the heart of Downtown Toronto’s High Fashion District in an area where vehicle parking is at a premium. As the area continues to develop with the introduction of high rise condominiums, parking is becoming an even bigger problem for local shoppers and workers in the Yorkville/Cumberland corridor. The parking structure, which was built in 1968, is typically above ground and consists of five parking levels, four of which are suspended parking levels. 3 ½ of the parking levels are above ground and 1 ½ are below ground. The structure has a rectangular footprint with approximate plan dimensions of 260 feet in the north/south direction and 100 feet in the east/west direction. The garage has the capacity to hold approximately 300 vehicles with approximately 120,000 sq. ft. of parking (excluding occupied space, mechanical rooms, parking offices, stairwells, etc…) of which 24,000 sq. ft. is on grade and 96,000 sq. ft. is suspended slab. 1.2

Structural description

The suspended slabs of this parking garage are typically 8″ thick cast-in-place post-tensioned concrete one-way spanning slabs. The slabs are supported by 30″ deep by 18″ wide post-tensioned concrete beams spaced at 28 foot centres and are also typically one-way spanning. The post-tensioning system within the suspended slabs is an unbonded post-tensioning system known as a “Button-Head System”. The post-tensioning system in the beams is also the “Button-Head System” however it is of the bonded type (i.e. placed in solid grout filled

ducts). The suspended slabs have a nominal amount of mild reinforcing steel embedded within them (i.e. # 3 bars at 16″ centres each way) to control cracking due to temperature fluctuations. 2

2.1

BUTTON-HEAD POST-TENSIONING SYSTEM Post-Tensioning basics

Post-tensioning is a method of introducing internal forces into a concrete element after the concrete element has been cast in order to counteract the external loads that will be applied to the element when it is put into use. When the concrete element is posttensioned, the concrete portion of the element is put into compression and the steel portion of the element is put into tension. Putting the components of the element into this state prior to applying a service load puts the building materials in a stronger state resulting in a stiffer element that now has more capacity to resist tensile forces. The first successful use of unbonded post-tensioning systems dates back to 1920–1930 in Europe and it was not until the 1960’s that it became popular in North America. 2.2

Button-Head system

The term “Button-Head” refers to the shape of the wire’s end after it is anchored. The “Button-Head System” installed in this facility’s parking slabs consists of ¼″ diameter high strength steel (240,000 psi), cold drawn, stress relieved wires which are bundled together into tendons and wrapped with protective greased paper. Tendon sizes ranged

393

from 7 wires to 19 wires in each bundle and were typically spaced at 4 foot centres. The Button-Head System is an early type of post-tensioning technology and possesses very little protection against moisture and chloride attack. The post-tensioning system in the suspended slabs of this parking garage is an unbonded system meaning it is anchored to the concrete slab at the end anchor locations only. The post-tensioning in the concrete beams is a grouted or bonded system and thus not as susceptible to moisture deterioration as the unbonded slab system. Because the unbonded Button-Head post-tensioning system is susceptible to moisture and chloride attack, it was found that as the structure continued to age and experience this ongoing moisture and chloride ingress, the embedded post-tensioning strands were deteriorating and becoming de-stressed, resulting in a reduced load carrying capacity of the suspended parking decks.

3

HISTORY OF PAST EVALUATIONS

Read Jones Christoffersen’s involvement with this structure dates back to 1985. Over the years the structure and in particular the post-tensioning system was evaluated to identify signs of corrosion related deterioration, tension deficiencies in the post-tensioning and reductions in load carrying capacity.

of external post-tensioning had an extra two tendons (5 total) to accommodate the extra span and flexural moment due to the end conditions. At each external post-tensioning location, the tendon groups were threaded through the existing suspended slab and anchored at each end. The dead ends were typically located at the expansion joint where as the live ends (stressing end) were at the building exterior elevation to accommodate the stressing. The tendons had a draped profile with high points overtop of beam support lines and low points at mid-bay between beams. At high points, troughs were chipped in the slab surface to allow the new tendons to obtain the appropriate concrete cover. In order to achieve the appropriate drape, through slab openings were chipped in the suspended slabs to allow the tendons to be threaded down to the underside of the slab and back up at next beam line. The exposed portion of the tendons below the existing slab were then encapsulated in reinforced concrete beams or ribs to provide the required slab-strand integration and overall fire protection to the system. It should be noted that during the process, the existing post-tensioning system remained in place and was de-stressed simultaneously with the stressing of new external system to ensure the existing suspended slabs were not overstressed at any time during construction. The new tendons were stressed to a final effective stress of 173 ksi (1193 MPa). 5

4

4.1

EXTERNAL POST-TENSIONING REHABILITATION SOLUTION General system description

Once a quadrant was deemed to be in need of external reinforcement, calculations were performed based on the extent of mild reinforcing and transverse post-tensioning to determine the maximum spacing between the proposed external post-tensioning system (i.e. 8′–0″ centres). The new external posttensioning consists of 7-wire high strength (270,000 ksi) low relaxation steel strands in extruded plastic sheathing. The sheathing was high-density polyethylene with a minimum wall thickness of 0.06 inches. The tendons were placed at 8-foot centres and at each location there were 3 tendons which ran the full length of the quadrant. At end bays, each line

CONSTRUCTION COST AND SERVICE LIFE

The construction cost for this external post-tensioning solution was in the order of $28.00 per square foot of suspended slab (Canadian Dollars). We are of the opinion that the suspended slabs of this facility will remain in a serviceable condition, if properly maintained, for another 40 years. The maintenance being continued repair and renewal of the waterproofing as well as localized repair to corrosion related deterioration as it develops. 6

CLOSING REMARKS

Deteriorated parking structures can be repaired and/ or rehabilitated to extend their effective service life and maximize economic return through innovative thinking, we all can achieve practical results.

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Mountain pass slope failure retrofitted with half viaduct bridge structure E.J. Kruger South African National Road Agency Ltd., Cape Town, South Africa

A.A. Newmark & M. Smuts BKS Engineering & Management, Cape Town, South Africa

ABSTRACT: The Garden Route, a major tourist attractions and national road on the south coast of South Africa, experienced abnormally high rainfall in August 2006. A major slip failure occurred in one of the road cuttings, which severely damaged the roadway. The South African National Road Agency Limited (SANRAL) immediately took measures to restore this vital roadway link. Several preliminary design concepts were considered and a modified version of a tied pile anchor wall was finally decided upon. This was considered to provide the optimum solution, considering interaction with the adjacent road formation and the bedrock profile. The new viaduct, is supported on eight 1200 mm diameter reinforced concrete oscillator bored piles approximately 14 m long socketed into competent rock. The superstructure comprises a 7 m wide, 60 m long cantilevered deck that is counterbalanced by an integrated buried jockey slab. The structure is anchored with re-stressable rock anchors.

EXTENDED ABSTRACT The Garden Route, one of the main tourist attractions on the south coast of South Africa experienced abnormally high rainfall during August 2006. This route forms the main national economic arterial on the southern side of the country, serving heavy transport carriers and commuters. The steep mountainous topography in this region has resulted in very few alternative routes being available and those that can be used are of a substantially lower standard with several hours of additional travel time. The combination of the heavy downpours and the unfavourable angle of dip of the rock formation in one of the road cuttings, known as Kaaimans River Pass, resulted in a slip failure which severely damaged the roadway. The client immediately stepped in to restore this vital link and appointed consulting engineers who had experienced geotechnical and bridge engineering staff immediately available. Based on available knowledge from historical records and detailed site observations, the risk related to further damage became known and first one lane and then two of the original three lanes were reopened, partially restoring the link. Three design concepts were selected and further developed by the consultant’s bridge specialists, Abe Newmark and Martin Smuts and the client’s Edwin Kruger. Ron Tluczek, the consultant’s geotechnical specialist, provided critical geotechnical input. The first design concept comprised a tied pile anchor wall and reinforced concrete cantilever. The construction procedure for this concept involved the

excavation to a suitable work platform, to drive oscillator piles, by stabilizing the narrowed roadway with a temporary soil nail structure to ensure accommodation of two-way traffic during construction. It was envisaged that the horizontal stability of the piles would be secured by sloped rock anchors in this instance. The main advantage of this concept was the minimization of the risk as a result of the stabilizing effect of multiple anchored pile shafts in the rock formation. In addition, the substantial lateral force component minimized relevant movement between the roadway and reinforced concrete elements of the structure. The second design concept involved a more conventional retaining structure. This comprised of a tied mechanically stabilized embankment, constructed by excavating to a sound founding level and the construction of a stabilized earth type backfill on an

Figure 1.

395

Failure crack with subsidence clearly evident.

anchored-back reinforced concrete footing. It was considered possible to construct this option without the temporary soil nail structure, in view of the reduced work space requirements. The third concept comprised a half-width viaduct bridge structure. Similar to the first option, a soil nail structure (permanent in this instance) would be constructed adjacent to the road centre line to provide space for two-way traffic and a working platform to erect conventional piers. Longitudinal precast beams supported on the pier caps and abutments would carry the viaduct deck. The pier footings would be anchored back by means of rock anchors. The interaction of this structure with the adjacent embankment would require the provision of a longitudinal joint with resulting long-term maintenance requirements. The main geotechnical factor that influenced the choice of the final concept was that the steep angle of dip and highly weathered jointing could result in a possible block/wedge failure. This would significantly enhance the risk associated with founding at isolated locations with a pier type structure in a local contact zone. The second concept of a mechanically stabilized type of retaining wall was also rejected as unsuitable due to the possibility of future wedge or slip failures that could possibly occur under such a wall. After due consideration of the alternatives a modified version of the first concept was finally adopted. The final proposal as described below was considered to provide the optimum solution considering interaction with the adjacent road formation and the bedrock profile. The new half viaduct known as bridge B0015 as shown in Figure 2, which was completed in November 2007, is supported on eight 1200 mm diameter reinforced concrete oscillator bored piles spaced at 8 m centres. The piles have a total average length of 14 m and are socketed 4 m into competent rock. The superstructure comprises a 7 m wide, cantilevered reinforced concrete deck which spans between the piles and has an overall length of 60 m. A novel feature is the counterbalancing of the cantilever with an integrated buried jockey slab which eliminated a problematic longitudinal joint and also reduced deck torsion and bending effects on the piles due to eccentric traffic loading. This not only removed fill material, but also ensured that all road loading was transferred to bedrock via the structure and piles and not via any fill material. The structure is anchored laterally with re-stressable rock anchors to ensure the stability of the structure and rock formation. The anchor forces are distributed to the pile caps by means of the horizontal deck beam which is integrated with the cladding.

Figure 2.

Final design concept.

In order to minimize vertical movement between the roadway layer works and structure, the jockey slab is also tied down onto rock with vertical rock bolts. The construction proceeded largely as planned and unforeseen delays that were experienced during construction mainly related to the piling process which proceeded substantially slower than anticipated. On completion, the portion of the road that once seemed destined to slide into the Kaaimans River now has four permanent lanes. Post completion inspections concluded that the design objectives had been attained. The project provided the structural and geotechnical engineers with not only a challenging problem, but also an opportunity to practice their art by restoring a vital transportation link and so enhancing the lives of the people and towns to whom this road serves as an economic lifeline.

ACKNOWLEDGEMENTS The South African National Roads Agency has kindly granted permission to publish this paper. Vusela Construction is acknowledged as the main contractor and Franki Africa as specialist geotechnical subcontractor. Solid State Safety Consultants performed safety audits.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Rapid repair of RC bridge columns subjected to earthquakes A. Vosooghi & M. Saiid Saiidi University of Nevada at Reno (UNR), Reno, Nevada, USA

J. Gutierrez California Department of Transportation (Caltrans), Sacramento, California, USA

ABSTRACT: The middle bent of a large-scale two-span bridge model, which was damaged to the highest repairable level (visible bars, initial buckling in some longitudinal bars, and initial concrete core damage) in the previous tests, was repaired using unidirectional carbon fiber reinforced polymer (CFRP) jacketing and retested to evaluate the repair performance. The concrete was repaired by removing the loose concrete and using a fast set grout. Adjacent cracks were epoxy injected. The columns were then wrapped with CFRP sheets and were cured only for 54 hours (24 hours accelerated, and 30 hours regular curing) before testing. The specifications call for seven day curing but the objective here was to determine the effectiveness of the repair method for emergency repair. The tests showed that the lateral load capacity and the ductility capacity of the bent were fully restored, and the service level stiffness of the bent was restored to 87% of the undamaged bent stiffness.

1

INTRODUCTION

Past effort in the seismic design of concrete bridges has been on detailing of bridges to prevent collapse. During earthquakes, reinforced concrete bridge columns are designed to undergo cracking, spalling, and yielding of steel and provide significant rotational capacity at plastic hinges so that the integrity of the overall structure is maintained. With proper design and construction, this objective can be met. However, the serviceability of the bridge after the earthquake is in question. The level of damage to different columns of a bridge varies depending on the intensity of the ground shaking, type of earthquake, and the force/ deformation demand on individual members. Based on the inspection of the damaged columns engineers have to determine whether the bridge is sufficiently safe to be kept open to traffic. They should also recommend repair methods for the columns. Any delay in opening of the bridge to traffic can have severe consequences on the passage of emergency vehicles, detour lengths, and traffic congestion in the area. Rapid and effective repair methods are needed to enable quick opening of the bridge to minimize impact on the community. In this study, the middle bent (Fig. 1) of a largescale two-span bridge model, which was damaged to the highest repairable level in the previous tests, was repaired using CFRP wrapping. At this level of the damage, many spirals and longitudinal bars are visible, some of the longitudinal bars are beginning to buckle, and the edge of concrete core is damaged. No bars are ruptured.

Figure 1.

2

Middle bent (Bent-2) details.

REPAIR DESIGN

The repair system was designed with the objective of restoring confinement and shear strength of the columns by using unidirectional CFRP jacketing. Based on the California Department of Transportation (Caltrans) provisions, for regions inside and outside a plastic hinge region, it is necessary to provide a minimum confinement stress of 300 psi (2.07 MPa), and 150 psi (1.03MPa), respectively at a radial dilating strain of 0.004. Using Caltrans criteria for seismic shear design for ductile concrete members, the required thickness for the jacketing, tj, is determined as:

397

Relative Displ. (mm) 0

(2)

Where Vo is overstrength shear, Vc is the concrete shear capacity, Vs is the shear strength provided by the spirals, Ej is CFRP modulus of elasticity, D is column diameter, and φ is 0.85. Since spiral experienced a maximum strain for 74% of yield, and some of cracks were not repairable inside the core, Vc, and Vs were neglected. Vo was assumed to be associated with the maximum moment achieved in the pre-repair tests.

50

50.8

101.6

152.4

203.2

254

304.8 222

40

178

30

133 Experimental Envelope Elasto-Plastic Idealization

20

89

10

44

0

Base Shear (kN)

Vo φ − (Vc + Vs ) π 2 × 0.004 × E j × D

Base Shear (kips)

tj =

0 0

2

4

6

8

10

12

Relative Displ. (in)

Figure 2. Force-displacement relationship for pre-repair tests.

Relative Displ. (mm) 0

REPAIR PROCESS

50.8

101.6

152.4

203.2

50

Test-07 Test-05 Test-04

4

Straightening the columns Concrete chipping Pressurized epoxy injection of the cracks Fast set concrete patching Surface preparation for CFRP wrapping CFRP wrapping Curing.

Base Shear (kips)

− − − − − − −

Test-03

40

The entire repair process took approximately 30 hours spread over four days, and it involved the following steps:

Test-06

254

Test-08 Test-02

30

304.8

355.6 222

First CFRP rupture Bent Failure Test-09

178

Test-10 Test-11

Second CFRP rupture

133

Test-01

20

89

Experimental Envelope Elasto-Plastic Idealization

10

Base Shear (kN)

3

44

0

0 0

2

4

6 8 Relative Displ. (in)

10

12

14

Figure 3. Force-displacement relationship for post-repair tests.

5

REPAIRED BRIDGE PERFORMANCE

To allow the comparison of the responses with respect to lateral load capacity, service level stiffness, and the ductility capacity, the measured envelopes were idealized by elasto-plastic curves. The idealized elasto-plastic curves are shown in the Figures 2, and 3 for pre-repair and post-repair tests, respectively. The results show that the capacity of the bent was restored completely after repair. The ductility capacity of the bent was also restored. Although the achieved ductility capacity for post-repair tests is greater than the ductility of pre-repair tests, but it must be noted that the calculated ductility for the pre-repair tests is based on maximum displacement at the highest repairable level, and not failure. In addition, 87% of the elastic stiffness of the bent was restored by the repair. The maximum drift ratio in the pre-repair tests, and that of post-repair tests at failure were 10.4%, and 12.75%, respectively. It can be concluded that the ultimate drift capacity of the repaired bent was comparable to that of the original bent.

CONCLUSIONS

Based on the achieved data from testing the repaired model, the following conclusions are made: − The repair design method, providing minimum confinement pressure for 300 psi (2.07 MPa), and providing plastic shear strength, was effective and appropriate. − The repair process was practical and may be used for emergency repair of earthquake damage concrete columns. − The repair restored the strength, ductility capacity, and drift capacity of the model completely, and restored the service level stiffness up to 87% of the original stiffness. − Although the jacketing system was cured for only 54 hours (24 hours of elevated heat and 30 hours of ambient lab temperature), the jacket system had the modulus of elasticity equal to specified value after one week curing. The minimum ultimate strain was less than specifications, but it was larger than the maximum design strain. Note that the specified curing time for CFRP is seven days.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Rehabilitation problems regarding an over pass in Timisoara – Romania A. Bota “Politehnica” University, Civil Engineering Faculty, Timisoara, Romania

Alexandra Bota Design Office ALDOR SRL, Timisoara, Romania

ABSTRACT: The over pass Calea Sagului was realized in the year 1974, being part of the so called category of the “new bridges”. It is situated in the western part of Romania, at the exit from Timisoara, assuring the passing of the rail way and thus the connection with the south-west part of the Banat region, Romania, respectively with the capital of Serbia, Belgrade. Due to its emplacement on an artery for international traffic, the over pass bears a most intense and heterogenic traffic as structure and tonnage, an important ponderosity having the heavy traffic from trucks. The over pass is realized with 2 lanes for each direction, as 2 parallel structures separated through a longitudinal contraction joint. Each structure has 7–8 spans of each 21 m, freely supported beam realized from 4 girders of prestressed concrete. The span over the rail way is a composite steel-concrete box girder. Due to the constructive deficiencies in the contraction joint area and also due to the lack of maintenance works the over pass suffered a series of degradations. The paper presents the solution for providing the continuity of the prestressed concrete superstructure and also technological aspects specific for rehabilitation works for the over pass, which will start in approximately 1 month. 1 1.1

GENERAL ASPECTS Technical information

The structure, realized in 1974, assures the continuity over the rail way of the national road 59, which makes the connection between Timisoara and the capital of Serbia, Belgrade, respectively Resita, the residential town of the county Caras-Severin. At the same time it facilitates the access to the industrial and commercial area, which is developing in the southern part of the city.

Figure 1.

Cross section.

The over pass has 4 traffic lanes, arranged as pair on parallel independent structures. In cross section, for one traffic direction there are 2 traffic lanes of 3.50 m, one 1.80 m wide bicycle lane and a footway of 1.35 m (Fig. 1). The total length of the over pass is 651.60 m, out of which 356.75 m represent the length of the structure and the access ramps have a length of 2 x 147.35 m. Longitudinally, the over pass presents simply supported spans, which can be grouped in three areas: access to Timisoara, constituted of 8 spans of 21 m, central span with L = 41 m and access to Resita with 7 spans of 21 m (Fig. 2).

Figure 2.

399

Side view.

2

DEGRADATIONS OF THE STRUCTURE

The carriage way shows numerous degradations, extended on quit wide areas. Due to these, the carriage way looses its continuity the sealing role being thus diminished and the comfort for the users is not assured anymore. The coating on the foot ways and bicycle lanes is heavily degraded due to the lack of maintenance. The steel part of the pedestrian parapet, realized out of rectangular pipe, has numerous degraded areas and parts prostrated through impact. The contraction joints realized out of steel plates filled with bituminous cement, constitute a technical solution absolutely improper and unfavorable for a good behavior and preservation of the over pass structure. These degradations of the surface, together with the degradations of the contraction joint, facilitate the penetration of the water at the structural elements. Due to the fact that water penetrates the contraction joints, the rust which is formed at the steel bearings, leads to their blocking. A lot of damages, materialized through corroded reinforcement and exfoliated concrete, are localized on the bearing bench at the abutments but also at the piers. Degradations can also be seen at the covering concrete of the circular columns of the piers.

3

Figure 3.

Continuity of the concrete slab.

section in the articulation area being though sufficient for absorbing the efforts from the compression, which appear from the movement conditions of the bridge deck under the influence of the temperature variations (Fig. 3). The works at the substructure are divided in two categories, namely construction works and consolidation works. The construction works mean the modifying of the bearing system of the prestressed concrete girders and namely the replacing of the steel bearings with neopren bearings, works described earlier. The consolidation works aim to bring the substructure at the initial bearing capacity and to assure its good behaviour considering the perspective loads.

REHABILITATION SOLUTIONS 4

In order to stop the degradation process of the joint areas and to seal them, the continuity of the structure regarding the plate was designed. This means, on one hand, keeping the girders as simply supported elements without changing the distribution of the efforts and, on the other hand, the elimination of the joints between the prestressed concrete bridge decks only by realizing a continuous reinforced concrete slab. As a consequence the entire superstructure of the over pass, for each traffic direction, is divided in three sectors. Only 4 contraction joints are kept, in order to separate these: access to Timisoara with 8 spans out of prestressed concrete, the span over the rail way and access to Resita with 7 spans out of prestressed concrete. After placing all girders on the new bearing blocks (the procedure is not to be applied for the span over the rail way, where the existent bearing blocks are kept), the continuity of the slab over the bearing areas is realized; this area is realized as a double abutment hinge with adequate reinforcement and diminished

CONCLUSIONS

The solution chosen for the rehabilitation of the over pass Calea Sagului from Timisoara, has reached the following objectives: • the continuity of the slab allows the reduction of the contraction joints from 17 pieces to 4 pieces with a movement capacity of ±50 mm; • the comfort for the users is substantially improved, due to the fact that the minor dislevelments from the contraction joint area are reduced; • the sealing of the areas near the contraction joints is achieved and implicit proper bearing conditions are provided; • the bearing capacity of the substructure is assured in relation to the loadings from the actual traffic and the one in perspective; • the amelioration of the functionality of the over pass is achieved; • the aesthetic aspect of the entire structure is also improved.

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Seismic retrofit and rehabilitation

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

An approach of service life in repair of structures with concrete due to natural hazards N.G. Maldonado, N.F. Pizarro, R.J. Michelini & A.M. Guzmán National Technological University. Regional Mendoza Faculty. CEREDETEC, Mendoza, Argentina

ABSTRACT: In the last fifteen years, the problems generated by pathologies and vulnerability in reinforced concrete structures are increased due the aging of materials and the forces generated by natural phenomena. The new Argentine regulations of reinforced concrete define the service life in fifty years. They include durability as an action in the evaluation of environmental conditions. The problems of damage generated by a natural phenomenon imply new issues about the residual service life with important social and economic impacts. In this paper through laboratory tests different types of repairs in seismic-resistant structures such as: additional ties, reinforced steel or plastic mesh are evaluated. Structural safety, habitableness and durability with national standards are also evaluated. The results of testing repairs show that they are applied with measures of maintenance. In the evaluation of service life the results of all constructive process must be included together with cultural changes.

1

INTRODUCTION

The service life of structures is subjected to performance and aging of materials. The evaluation of service life is based on experience and there are few regulations which define explicitness because the reinforced concrete is a relatively new material. The question is if there is advancement in pathological problems for reinforced concrete structures or if the performance of materials relatively new is evident. In regions with random natural phenomena such earthquakes, tsunamis, hurricanes, etc, special path loads are added. In non developed countries these effects make an impact on social and economical aspects (for examples: México Earthquake, 1985, Andaman Tsunami, 2004, Pisco and Ica Earthquake, Perú, 2007). The República Argentina has 2/3 of its territory under seismic risk, but the mid-west, Mendoza and San Juan provinces have the most seismic risk. The challenge of estimate the service life is necessary but it is not sufficient because it is difficult to prove that the service life can be reach without periodical inspections and adequate maintenance. Plus complex is when the phenomena had extraordinary registrations such the above earthquakes mentioned. The aim of codes is to establish the minimum technological requirements in order to guarantee a level of structural safety and the ability to foresee future service conditions. The new national code of reinforced concrete incorporates the environmental conditions as loads on the structure which must be identified to establish the criteria for its protection, as well as the requirement of a concrete design with a fifty-year service life. The environmental exposure conditions are classified as: general exposure conditions relate

to degradation of the structure by corrosion processes and specific exposure conditions related to freezing and thawing, as well as effects of chemicals or salts contained in soils and water contacting structures (CIRSOC 201 2005). The definition of service life in the code in force is: the period of time after construction during which safety, functionality or aptitude in service and acceptable aspect must be maintained, without significant operating expenses. The service life is a value which is established by owner in the preview design of construction. But to retrofit or rehabilitate the service life is not defined neither for innovative systems. Only in the Seismic-Resistant Code of Mendoza is explained the methodology to repair or rehabilitate of construction but only from the point of view of structures (SeismicResistant Code of Mendoza 1987). Although the repairing techniques for lateral bonded masonry have been investigated in our region since 1985 the new codes in force also include the evaluation of service life and the inclusion of new techniques expecting better performance of materials and structures. The experimental works for improving performance of the existing structure include the three basic classes of measures taken to retrofit a building: add elements, enhance performance of existing elements and improve connection between components. 2

EXPERIMENTAL STUDY

The original lateral bonded masonry is tested to horizontal static cyclic loads up to failure. In this damage state, the structure is repaired with different

403

techniques and it is tested at the same level of load or up to failure. The conclusions about repair technique used are obtained by comparison of both tests. The different techniques are verified through 1:1 laboratory tests applying horizontal cyclic load such as a Cantilever masonry beam. For closing cracks injection epoxy resin is used for reinforced concrete of columns. For the repairing of masonry walls we used different techniques: − for strengthening wall we used rhomboidal wire mesh with anchor ties (5 × 10–3 m) and wire (3 × 10–3 m), reinforced steel mesh # (4 × 10–3 m diameter at 0.40 m interval) and cement coating. − for strengthening wall we used high density polyethylene mesh # (0.34 × 10–3 m diameter wrapped fabric 2-row × 1-row or anti-hail mesh) adjusted with anchor tie (5 × 10–3 m diameter at 0.50 m interval in the half of wall) and wire (3 × 10–3 m) and cement coating. − for improved connections we used a reinforced concrete mid-beam anchored in tied columns. − for adding elements we used units of masonry without damage bonded with cement mortar for replacing damaged units.

3

ANALYSIS OF RESULTS

Section 8 of Seismic-Resistant Code of Mendoza (1987) presents four levels of security with its corresponding type of rehabilitation according to quality of materials, importance of building and structural safety. The reparation with mesh attains the sufficient level of security. The mid-beam repair and the replacement of units are not safety enough. The level of distortion of wall head does not surpass the limit established by code (1%) for all tests. For repairing with added elements or replacement of units the tests showed the highest value. The amount of steel was the limit for deformation capacity. Carbonation is considered the only cause of deterioration of concrete, but this attack can produce corrosion of reinforced steel of ties as well. The rate of carbonation is measured in normal cement concrete specimens but for pozzolan cement concrete is considered with a plus 10% because there are no tests in the region.

Concrete ties in corners and connections are more exposed and they will generate concentration of moisture, called thermal bridge, and the habitableness may be diminished. The risk of corrosion will be increased when reinforced steel mesh is used provided the cover is not sufficient to obtain the estimated service life. The type of design applied in social housing, with no coatings, thus this situation generates concentration of condensation of water and loss of hydrothermal conditions. The risk of corrosion will decrease if polyethylene mesh is used. The pattern of distribution of damages is better than the other reparations with steel mesh. The cost of repair with steel mesh is 13.5 dollars per m2 and the cost of repair with polyethylene mesh is 3 dollars per m2. If the replacement of units and the incorporation of ties are cheaper than other repairs, they are not recommending because safety is not sufficient. The reparation with polyethylene mesh is consistent with the state of the art of construction and the seismic resistant design. It is allowed to find a solution with local workmanship, with lower cost and acceptable structural and environment response. 4

CONCLUSIONS

− The reparation system must be evaluated in according the available materials and workmanship in the region. The choice of reparation system must be in account not only the cost also the durability of solution. − The reparation system of mesh diminishes and redistributes the splitting in the panel. − The reparation with polyethylene mesh allows a longer service life compatible with new regulations for social housing. − The reparation with steel mesh must include measures of maintenance to avoid corrosion of reinforcement. − The reparation with mid-beam attains a level of damage incompatible with estimated service life. − In the evaluation of service life of reparations the results of all constructive process must be included. − It is necessary to improve the study of technique applied to joint the meshes to the wall and the effect of repaired coating in the durability with the new used additions.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Seismic behavior of FRP-upgraded exterior RC beam-column joints Y.A. Al-Salloum, S.H. Alsayed, T.H. Almusallam & N.A. Siddiqui Department of Civil Engineering, King Saud University, Riyadh, Saudi Arabia

ABSTRACT: Shear failure of exterior beam-column joints is identified as the principal cause of collapse of many moment-resisting frame buildings during recent earthquakes. Effective and economical strengthening techniques to upgrade joint shear-resistance and ductility in existing structures are needed. In this paper, efficiency and effectiveness of Carbon Fiber Reinforced Polymers (CFRP) in upgrading the shear strength and ductility of seismically deficient exterior beam-column joints have been studied. For this purpose, two reinforced concrete exterior beam-column sub-assemblages were constructed with non-optimal design parameters (inadequate joint shear strength with no transverse reinforcement) representing pre-seismic code design construction practice of joints and encompassing the vast majority of existing beam-column connections. Out of these two, one specimen was used as baseline specimen (control specimen) and the other one was strengthened with CFRP sheets (strengthened specimen). These two sub-assemblages were subjected to cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged control specimen was then repaired using CFRP sheets. This repaired specimen was subjected to the similar cyclic lateral load history and its response history was obtained. Response histories of control, repaired and strengthened specimens were then compared. The results were compared through hysteretic loops, load-displacement envelopes, ductility and stiffness degradation. The comparison shows that CFRP sheets improve the shear resistance of the joint and increase its ductility.

1

INTRODUCTION

Inadequate shear reinforcement in the existing beamcolumn joints, especially exterior ones, is believed to be the prime cause of failure/collapse of moment resisting RC frame buildings. Hence, effective and economical strengthening techniques to upgrade joint shear-resistance in existing structures are needed. In the past, variety of techniques have been employed to upgrade shear capacity and ductility of RC (reinforced concrete) joints, with the most common being construction of RC or steel jackets. Plain or corrugated steel plates have also been tried. These techniques cause various difficulties in practical implementation at the joint, namely intensive labor, artful detailing, increased dimensions, corrosion protection and special attachments. To overcome the difficulties associated with these techniques recent research efforts have focused on the use of epoxy-bonded FRP sheets or strips with fibers oriented properly so as to carry tension forces due to shear. A detailed review of literature shows that systematic studies to determine the behavior of the repaired and/or strengthened members under cyclic loading are still limited. Moreover, the behavior of seismically excited FRP repaired beam-column joints is not well established at various stages of response e.g. before and after yielding of reinforcements, crushing of concrete, fiber fracture or debonding. The present paper is also an effort in the same direction. In this paper, efficiency and effectiveness of Carbon Fiber Reinforced Polymers (CFRP) in upgrading the shear

strength and ductility of seismically deficient exterior beam-column joints have been studied. For this purpose, two reinforced concrete exterior beam-column sub-assemblages were constructed with non-optimal design parameters (inadequate joint shear strength with no transverse reinforcement) representing preseismic code design construction practice of joints and encompassing the vast majority of existing beamcolumn connections. Out of these two, one specimen was used as baseline specimen (control specimen) and the other one was strengthened with CFRP sheets (strengthened specimen). These two sub-assemblages were subjected to cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged control specimen was then repaired using CFRP sheets. This repaired specimen was subjected to the similar cyclic lateral load history and its response history was obtained. Response histories of control, repaired and strengthened specimens were then compared. The results were compared through hysteretic loops, load-displacement envelopes, ductility and stiffness degradation. 2

EXPERIMENTAL PROGRAM

In order to carry out the experiments on exterior joint specimens, first a prototype member size was chosen and then a crude analysis was carried out to come up with the most reasonable scale for the test specimen that complies with the available testing facility and equipment. Half-scale beam-column joint was found

405

to be the most convenient. After deciding the size of the test specimens, two reinforced concrete baseline specimens were cast. Out of the two cast specimens, one specimen was used as control specimens (EC) and the other was strengthened with CFRP sheets (ES). CFRP sheets were epoxy bonded to joint, beams and part of the column regions (Fig. 1). These specimens were then subjected to cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged control specimen was then repaired using CFRP sheets. The repaired specimen (ER) was then subjected to the similar cyclic lateral load history and its response history was obtained. Response histories of control, repaired and strengthened specimens were then compared.

90 70

Lateral load (kN)

50 30 10 -10 -55 -45 -35 -25 -15 -5

15

-50 -70

25

35

45

55

Control (EC) Repaired (ER) Strengthened (ES)

-90

Lateral displacement (mm)

Figure 3.

3

5

-30

Envelope of hysteretic loops.

DISCUSSION OF TEST RESULTS

In this section, through various experimental results, the effectiveness of CFRP in improving the as-built joint shear strength and ductility has been studied. The results are presented and discussed under the

headings of general behavior, hysteretic loops, stiffness degradation, and load-displacement envelopes. Figures 2 and 3 show the envelope of hysteretic loops and stiffness degradation with lateral displacement for control, repaired and strengthened specimens respectively.

4

The results of the experimental program, presented in this paper, establish the effectiveness of CFRP sheets in upgrading deficient exterior beam-column joints. The results of CFRP repaired and strengthened specimens were compared with their corresponding control specimens and, in general, it was observed that CFRP sheets improve the shear resistance and ductility of the RC joints to a great extent. The failure of CFRP upgraded beam-column joints due to debonding (bulging) was examined and it was observed that at higher stage of loading there was significant yielding in beam reinforcing bars that allowed cracks to widen in the beam region which in turn ultimately tore the CFRP sheets.

Figure 1. Test setup for exterior joint specimens.

10.00 Control (EC) Stiffness (kN/mm)

8.00

Repaired (ER) Strengthened (ES)

6.00 4.00 2.00 0.00 0

Figure 2.

10

20 30 40 Lateral displacement (mm)

50

CONCLUSIONS

60

Stiffness degradation in the specimens.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

3D analysis of seismic response of RC beam-column exterior joints before and after retrofit R. Eligehausen, G. Genesio & J. Ožbolt Universität Stuttgart, Germany

S. Pampanin University of Canterbury, Christchurch, New Zealand

ABSTRACT: This paper presents an application of microplane based Finite Element (FE) approach for threedimensional (3D) modeling of reinforced concrete beam-column joints under cyclic loading. Experimental tests have shown that the structural behavior of poorly detailed joints is decisive for the structural response of older frame building. Due to inadequate reinforcement ratio in the joint, poor bond properties of longitudinal reinforcement (plain round bars) and deficiencies in the anchorage of reinforcement (bars with end-hooks) a brittle failure mechanism is expected. Tests on poorly detailed beam-column exterior joints, primarily designed for gravity loads only, were conducted at the University of Canterbury (UC) in order to evaluate their seismic response. The numerical analysis presented in the paper are performed with the FE Code MASA, developed at the Universität Stuttgart and capable of 3D nonlinear analysis of quasi-brittle materials, like concrete, based on a microplane material model with relaxed kinematic constraint. In this contribution, the experimental results are used for calibration of the model with particular focus on joint strength, failure mode mechanisms and strength and stiffness degradation. The influence of different parameters is investigated with the intention to quantify their influence on the performance of the joint. In the final part of the paper a retrofit solution, proposed and experimentally evaluated at the UC, is also numerically simulated and discussed.

1

INTRODUCTION

Reinforced concrete frame buildings designed before the introduction of modern seismic oriented codes in early 1970 s, offer an inadequate response to lateral loads typical of seismic events. In this work the attention is focused on the behavior of exterior beamcolumn joint, since it is recognized that they are the most vulnerable part of moment resisting RC frames, due to their lack of a reliable joint shear transfer mechanism. The experimental results of quasi static tests on poorly detailed beam-column joints performed at UC are used for development and refinement of the finite element code MASA, developed at the Universität Stuttgart and capable of three-dimensional (3D) nonlinear analysis of concrete-like materials and reinforced concrete structures. The program is based on the microplane model with relaxed kinematic constraint. 2

EXPERIMENTAL TESTS

The experimental tests considered for the validation of the FE model include three exterior beam-column joints with plain round and deformed bars, before

and after retrofit loaded under reversed cyclic lateral loading. The specimen TDD1 is characterized by deformed bars bent in the joint, as longitudinal reinforcement of the beam, and in the specimen TDP2 plain round bars with end-hooks were employed. The retrofitted specimen THR3 had the same detailing of TDP2. 3

FINITE ELEMENT MODEL

In the FE code used in the analysis, the microplane material model for concrete with 3D hexahedral elements and 1D three-linear constitutive law for reinforcement steel were used. The bond between longitudinal reinforcement and concrete was simulated using discrete bond elements. For transverse reinforcement, a rigid connection between steel and concrete was assumed. This assumption neglects the influence of the relative displacement between stirrups and concrete. The discrete bond model implemented in MASA consists of a 1D nonlinear springs with a bond-slip relationship, which depends on the state of stresses and strains in concrete and reinforcement, on type of loading and on geometry.

407

4

NUMERICAL ANALYSIS

4.1

Cyclic analysis

The hysteretic behavior of the test specimen TDD1 is shown in Fig. 1. The comparison between experimental results and numerical analysis shows an acceptable correspondence of the behavior until a top drift of 2.0%, with a difference of approximately 10% of the ultimate capacity of the joint. Consistently to the experimental behavior (Fig. 3a2), the cracking pattern obtained in the numeric simulation (Fig. 3a1) shows the formation of a flexural mechanism. In Fig. 2 the comparison between the experimental test and the numerical simulation of the test specimen TDP2 is shown. The simulation was able to reproduce with sufficient accuracy the main features of the cyclic behavior of the joint: (i) relevant strength degradation after 2.0% top drift and (ii) increasing pinching at higher drift levels. The resulting cracking patterns are presented in Fig. 3b1,b2.

Figure 3. Predicted and in experiments observed failure modes for: a1) FE analysis of TDD1 with formation of flexural hinge at the beam-column interface; a2) End of the experimental test TDD1; b1) Shear cracks in the joint panel in the FE analysis of TDP2; b2) End of the experimental test TDP2.

Late ralforce [kN]

Lateral displacement [mm] -80 25

-60

-40

-20

20

Specimen TDD1

0

20

40

60

80

15 10 5 0 -5 -10

Numerical Experimental

-15 -20

a)

-25 -4

-3

-2

-1

0

1

2

3

b)

4

Figure 4. a) Haunch retrofit solution; b) Flexural crack in the beam obtained from the FE analysis.

Top drift [%]

Figure 1. Hysteretic behavior of test specimen TDD1: comparison between experimental and numerical results.

4.2 Lateral displacement [mm] -80

-60

-40

-20

0

20

40

60

In order to avoid the shear failure of the connection in the joint panel a retrofit solution, based on a diagonal metallic haunch, was developed at the UC. Fig. 4a shows the implementation of the metallic haunch and the corresponding flexural hinge. The cracking pattern obtained with numerical simulation is shown in Fig. 4b.

80

25

Lateral force [kN]

20

Analysis of retrofitted specimen

Specimen TDP2

15 10 5 0 -5 -10

5

Numerical Experimental

-15 -20 -25 -4

-3

-2

-1

0

1

2

3

4

Top drift [%]

Figure 2. Hysteretic behavior of test specimen TDP2: comparison between experimental and numerical results.

CONCLUSIONS

The 3D numerical analysis of poor-detailed exterior beam-column joints under monotonic and cyclic lateral loading have been presented. The simulations were able to reproduce correctly the general behavior of the test specimens with deformed and plain round bars, before and after retrofit.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Seismic assessment of a RC school building retrofitted with innovative braces L. Di Sarno Department of Engineering, University of Sannio, Benevento, Italy

G. Manfredi Department of Structural Engineering, University of Naples, Naples, Italy

1

INTRODUCTION

Existing low- and medium-rise buildings in seismic regions worldwide consist of a large number of reinforced concrete (RC) framed structures designed for gravity loads only; hence such structures do not possess sufficient lateral stiffness, strength and ductility to sustain moderate-to-large magnitude earthquakes. The structural performance of multi-storey frames with inadequate seismic resistance may be enhanced by using either traditional or innovative strategies. The selection of the most suitable retrofitting strategies is generally a trade-off between the total cost of repair and the disruption of the occupancy of the building. This paper discusses the numerical results computed through extensive static and dynamic analyses, both linear and nonlinear, carried out on a sample RC framed school building located in the municipality of Avellino, near Naples, Italy. The building was designed originally for gravity loads only and retrofitted with the use of BRBs placed along the perimeter. The results of the inelastic static analyses are discusses in details hereafter.

2

Figure 1. Finite element modeling of the sample structure: as-built (top) and retrofitted (bottom) framed system.

CASE STUDY

The plan layout of the sample school building is geometrically irregular and consists of three main blocks: two T-shape block and a rectangular block. The structural system comprises spatial frames which were designed for gravity loads about 30 years ago. The site of construction of the sample school is characterized by moderate seismicity. A refined finite element (FE) model was used to perform the elastic and inelastic analyses of the sample structure, either as-built or retrofitted as shown in Figure 1.

3

SEISMIC RESPONSE ANALYSIS

The fundamental period of the retrofitted structure is significantly lower than the counterpart of the as-built system (0.397 sec versus 0.612 sec); the reduction is about 36%. Furthermore the estimated values of the modal participation mass of the retrofitted structure show that the presence of braces along the perimeter is beneficial to achieve a regular dynamic response of the structure.

409

Base seismic shear / Seismic weight (%)

Seismic base shear / Seismic weight (%)

45 40 35 30 25 20 15

Pushover +X M Pushover +XM Pushover -X M Pushover -XM

10 5 0 -1.0

-0.8

-0.5

-0.3

0.0

0.3

0.5

0.8

1.0

50 45 40 35 30 25 20 15

Pushover +Y M Pushover +YM Pushover -Y M Pushover -YM

10 5 0 -1.0

-0.8

-0.5

-0.3

0.0

0.3

0.5

0.8

1.0

Lateral displacement of control point / Height of control point (%)

Figure 3. Response curve of the retrofitted structure: X-direction (top) and Y-direction (bottom).

4

CONCLUSIONS

25.0

This analytical study has investigated the seismic response of an existing reinforced concrete (RC) framed school building designed for gravity loads only. The seismic response of the as-built spatial frames was assessed and the structural deficiencies highlighted. The building structure was then retrofitted with innovative buckling restrained braces (BRBs). Nonlinear static analyses demonstrated that the use of BRBs is very efficient to enhance the global over-strength (about 30%) and energy dissipation capacity of the RC frame under moderate-to-high magnitude earthquakes. These results were also confirmed by further extensive nonlinear time histories analyses carried out on the retrofitted structure.

20.0

15.0

10.0 Pushover +X M Pushover +XM Pushover -X M Pushover -XM

5.0

0.0 -1.50

-1.00

-0.50

0.00

0.50

1.00

1.50

Lateral displacement of control point / Height of control point (%) Base seismic shear / Seismic weight (%)

50

Lateral displacement of control point / Height of control point (%) Seismic base shear / Seismic weight (%)

The retrofitted building has a modal mass equal to 89.54% along X and 85.50% along Y, in the first and second mode, respectively. Conversely, the asbuilt framed system possesses a non-regular response along the X-direction. For both existing and retrofitted structures the first three modes account for the total modal masses. The computed response (pushover) curves for the as-built structure are provided in Figure 2 for the X- and Y-directions. Both positive and negative directions are considered as the beams have asymmetric longitudinal reinforcement bars. The points corresponding to the onset of displacement demands at damageability, life safety and collapse limit states are also included in the plots in Figure 2. The computed response curves show that the strength demand relative to the modal force distribution is higher than the uniform counterpart. The response curves of the retrofitted structure are provided in Figure 3 for X- and Y-direction. The computed results show a significant increase of the global over-strength of the structural system (about 30%) and a reduction of the displacement demand imposed on the structure at different limit states. The post-peak stiffness and strength degradation observed along the X-direction of the as-built structure is also eliminated thus enhancing the energy absorption and dissipate especially under high magnitude earthquake ground motions.

25.0

20.0

15.0

ACKNOWLEDGEMENTS

10.0 Pushover +Y M Pushover +YM Pushover -Y M Pushover -YM

5.0

0.0 -1.50

-1.00

-0.50

0.00

0.50

1.00

1.50

Lateral displacement of control point / Height of control point (%)

Figure 2. Response curve of the as-built structure: X-direction (top) and Y-direction (bottom).

This work was financially supported by the Italian Consortium of Laboratories RELUIS, funded by the Italian Federal Emergency Agency. Any opinions, findings and conclusions or recommendations expressed in this paper are those of the authors and do not necessarily reflect those of the Consortium RELUIS.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Seismic response of FRP-strengthened corner RC beam-column joints T.H. Almusallam, S.H. Alsayed, Y.A. Al-Salloum & N.A. Siddiqui Department of Civil Engineering, King Saud University, Riyadh, Saudi Arabia

ABSTRACT: In the present paper an innovative and practical technique for the seismic rehabilitation of poorly detailed beam-column corner joints using FRP composite sheets has been proposed. A full scale corner beam-column sub-assemblage was constructed with inadequate joint shear strength and no transverse reinforcement in the joint; representing pre-seismic code design construction practices of joints and encompassing the vast majority of existing joints. The corner joint specimen was tested under reversed cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged specimen was repaired using a suggested scheme and then subjected to the similar cyclic lateral load history. Response histories of the specimen before and after repair were then compared through hysteretic loops, load-displacement envelopes, ductility and stiffness degradation. The test results indicated that the suggested repair scheme is very effective in upgrading the shear capacity and ductility of the joint. The results also show that, with the proposed FRP scheme of repair, the repaired specimen achieves a substantially higher load carrying capacity, higher ductility and slower stiffness degradation.

1

INTRODUCTION

A concrete frame structure’s bearing capacity and ability to exhibit ductile behavior are highly dependent on the reinforcement detailing of the joint connections between its independent members, such as beams and columns. Accordingly, to obtain a sound structural behavior, the joints must be constructed to be at least as strong as the structural members connected to them and show ductile behavior in the ultimate limit state. In a moment resisting reinforced concrete framed buildings, these joints are of three types: interior, exterior and corner. Interior and exterior joints are found at bottom and intermediate levels, and corner joints at the roof level. The corner joints, if designed only for gravity loads and are based on pre-seismic codes, may suffer substantial damage during earthquakes particularly under opening cycles, i.e. those causing flexural tension on the inside of the joint. Several techniques of repair and strengthening of reinforced concrete joints, damaged by earthquakes, have been reported in earthquake prone countries such as Japan, Mexico, and Peru. Of the various repair techniques used, the most common involved were RC or steel jackets. Plain or corrugated steel plates have also been tried. These techniques cause various difficulties in practical implementation at the joint, namely intensive labor, artful detailing, increased dimensions, corrosion protection and special attachments. To overcome the difficulties associated with these techniques recent research efforts have focused on the use of epoxy-bonded Fiber Reinforced Polymer (FRP) sheets or strips with fibers oriented properly so as to carry tension forces due to shear.

In the present paper an innovative and practical technique for the seismic rehabilitation of poorly detailed beam-column corner joints using FRP composite sheets has been proposed. A full scale corner beam-column sub-assemblage was constructed with inadequate joint shear strength and no transverse reinforcement in the joint; representing pre-seismic code design construction practices of joints and encompassing the vast majority of existing joints. The corner joint specimen was tested under reversed cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged specimen was repaired using suggested scheme and then subjected to the similar cyclic lateral load history. Response histories of the specimen before and after repair were then compared through hysteretic loops, load-displacement envelopes, ductility and stiffness degradation. 2

EXPERIMENTAL PROGRAM

In finding out the size of corner joint specimen, first a prototype member size was chosen and then a crude analysis was carried out to come up with the most reasonable scale for the test specimen that complies with the available testing facility and equipment. A fullscale beam-column joint was found to be the most convenient. Having decided the size of the test specimen, a reinforced concrete joint specimen was cast. The specimen was then subjected to cyclic lateral load histories so as to provide the equivalent of severe earthquake damage. The damaged specimen was then repaired through injecting epoxy into the cracks and externally bonding the specimens with CFRP sheets.

411

90 70

Lateral load (kN)

50 30 10 -10 -30 -50 -70 -90 -50 -40 -30 -20 -10

0

10

20

30

40

50

Lateral displacement (mm)

Figure 1. Test setup for corner connection specimens. Figure 2.

The sheets were epoxy-bonded to joint region only and also effectively prevented against any possible debonding through mechanical anchorages as shown in Fig. 1.

Hysteretic plots for before repair specimen.

200 150

3

DISCUSSION OF TEST RESULTS

Connections lose their strength and stiffness due to forces imparted to them during earthquakes. The degradation in the strength and stiffness may be so critical that it may lead to failure of the overall structural system. The overall safety of the structural system, under these circumstances, would be expected to be controlled by the shear capacity of the joint. In the present section, through various experimental results, the effectiveness of CFRP in improving the joint shear strength and ductility has been studied. The results are presented and discussed under the head of hysteretic behavior, load-displacement envelopes, ductility and stiffness degradation. Figures 2 and 3 show the hysteretic curves obtained after the testing of before repair and after repair specimens respectively.

4

Lateral load (kN)

100

CONCLUSIONS

50 0 -50 -100 -150 -200 -50 -40 -30 -20 -10

0

10

20

30

40

50

Lateral displacement (mm)

Figure 3.

Hysteretic plots for repaired specimen.

observed that proposed method of repair can improve the load carrying capacity and ductility of joint by 88% and 97% respectively and can also decrease the rate of stiffness degradation significantly. The results thus show that use of CFRP sheets in upgrading shear deficient joints is very promising.

In the present study effectiveness of CFRP sheets in improving the load carrying capacity and ductility of shear deficient corner joint was studied. It was

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Experimental investigation on the seismic behaviour of SFRC columns under biaxial bending M. Palmieri & G.A. Plizzari Department DICATA, University of Brescia, Italy

S. Pampanin & J. Mackechnie Department of Civil Engineering, University of Canterbury, New Zealand

ABSTRACT: Research studies on Fibre Reinforced Concrete (FRC) elements have been mostly concentrated on the performance under static loading. Few information are available in literature on the behaviour, analysis and design under seismic loading . In the research work herein presented, four full-scale cantilever columns were tested at the structural laboratory of the University of Canterbury under reversing cyclic lateral loading. Hooked steel fibres were adopted (with a volume fraction, Vf=1%) in the plastic hinge zone to improve the seismic behaviour and reduce damage. Both unidirectional and bidirectional loading protocols were adopted to better simulate the actual response of a column subjected to a ground motion excitation. Test results indicate that fibres can significantly reduce the level of damage typically expected in the plastic hinge zone of a traditional RC element. By preventing the cover concrete to spall out at earlier stages and providing some additional anti-buckling restraint to the longitudinal steel bars, remarkable improvements are obtained in the hysteretic response of the element, with proper control on both stiffness and strength degradation. The observed limited damage would result into significant savings in terms of repairing costs and business interruption (downtime).

1

INTRODUCTION

When performing tests under reverse cyclic loading on structural member, it is worth reminding that the direction of the cyclic loading imposed during a seismic event is continuously changing. Thus, very seldom, if ever, a structural member is submitted to a state of stress corresponding to bending in one of the principal directions (typically referred to as unidirectional or uniaxial loading). This indicates the importance of evaluating the inelastic response of a structural member when subject to a more complex, but more realistic, bi-axial or bi-directional cyclic loading protocol. From several experimental investigations it was found that FRC has a high performance response in shear-prone elements, such as beam-column joint, squat walls or coupling beams, and can be used to significantly reduce the transversal reinforcement in flexural elements. As part of a more comprehensive collaborative research project between the University of Brescia (Italy) and the University of Canterbury (New Zealand) on the investigation of the seismic response of FRC structural elements and systems, the response of Steel Fibre Reinforced Concrete (SFRC) columns under seismic loading was experimentally investigated and compared with traditional RC columns. This experimental investigation on a standard structural element, as a cantilever column, aims to demonstrate the positive contributions provided by fibre

reinforcement under severe seismic loading conditions, both in terms of structural response (hysteretic behaviour, strength, stiffness and energy dissipation) and of limited level of damage. Special attention is given to the plastic hinge zone of the column units, in order to evaluate the structural enhancement due to the presence of fibres and the effects of different loading regimes. 2

EXPERIMENTAL INVESTIGATION AND RESULTS

Four full-scale columns, representing cantilever elements (from the foundation to the point of contra flexure), including two FRC elements (DS-1, DS-2) and two standard RC solutions (NZ-1, NZ-2), were subjected to quasi-static reverse cyclic lateral loads, using either uni-directional (DS-1, NZ-1) and bidirectional (DS-2, NZ-2) loading protocols. For the bi-directional testing regime, a four clovetype loop was used and preferred to a more traditional diagonal or s-type loop to better reproduce the real demand of displacement during a seismic event. From a comparison with lateral load-displacement results from uniaxial tests (Figure 1), it is evident as the biaxial bending actions are much more demanding than the uniaxial ones, confirming the importance of the use of an appropriate and more realistic loading protocol. An overall flexural strength reduction of approximately 35% was observed in the NZ-2 column when compared to the equivalent specimen NZ-1,

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Summary of Test Results

75

50

25

NZ-1

DS-1

NZ-2 45Þ

DS-2 45Þ

0 0

25

50

75

100

125

Displacement [mm]

Figure 1. Comparison of the envelopes of the loaddisplacements curves obtained from the four tests (displacements from biaxial tests are combined in x and y direction at 45ο).

Energy Evaluation 35

NZ - 1

DS - 1

NZ - 2

DS - 2

Energy Dissipated [ kJ ]

30 25

Figure 3. NZ-1, NZ-2, DS-1 & DS-2 (CW): damage of the column base at the end of each test.

20 15 10

Figure 3 shows the critical region of the four specimens at the end of the test; it can be notice the smaller damage of concrete when steel fibres are present at the column base.

5 0 0.10%

1.00%

10.00%

Drift Level

Figure 2. Comparison of Energy dissipated by the columns.

subjected to uni-directional loading. More importantly the response was not stable up to the maximum displacement, but exhibited a clear strength degradation at a low drift (1.5%), well below the design level of drift (2÷2.5%) adopted for a 500 years return period earthquake. The loss of loading bearing capacity continued reaching a 20% strength reduction at 3.26% drift. Incipient collapse was observed at a drift close to 4.5%, typically considered MCE level (2500 years). The structural response of the columns has been clearly positively improved by the presence of fibres in the concrete mix. In particular the FRC columns exhibited an enhanced capacity of maintaining their strength through the cycles without showing the damage occurred in the RC column unit (Figure 1). This enhanced behaviour was clearer when considering the more realistic case of bidirectional loading, which proved to be particularly demanding even for a well designed structural element. In terms of energy dissipated, the presence of fibres enhances the overall capacity of the column and guarantees a good structural performances up to a drift of 4.5% (Figure 2).

3

CONCLUDING REMARKS

The displacement-based biaxial testing regime showed to be significant for the general behaviour of the structural element, inducing asymmetry in the specimens and severe strength degradation and brittleness to the column. In fact, while the momentcurvature hysteresis loops for the uniaxial column units were “fat” and symmetric in both the column, the behaviour of both RC and FRC under biaxial bending was characterised by an evident asymmetry between the two principal directions of loading. This particular response shows that, when subjected to simultaneous biaxial bending, the column starts yielding in one direction at a load significant lower than the uniaxial carrying capacity due to the inelastic biaxial interaction. In other words, although a frame may be designed with a weak-beam-strong-column concept, column yielding may be possible during an earthquake due to the inelastic biaxial interaction. Finally, the biaxial bending speeded up the column degradation, increasing the strains in the longitudinal bars and letting the reinforcement buckling at lower level of drift. Moreover, it induced an asymmetry in the straining of the longitudinal bars.

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Retrofitting techniques and FRP systems

Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Feasibility study of a novel prestressing system for FRP-laminates Wine Figeys, Luc Schueremans & Dionys Van Gemert K.U.Leuven, Faculty of Engineering Sciences, Department of Civil Engineering, Laboratorium Reyntjens, Heverlee, Belgium

Kris Brosens Triconsult N.V., Lummen, Belgium

Ludo Van Schepdael Solico N.V., Oosterhout, The Netherlands

Johan Dereymaeker tDNE N.V., t. De Neef Engineering, Heist-op-den-Berg, Belgium

1

INTRODUCTION

2.2

The load capacity of a structure can be enhanced using the technique of externally bonded reinforcement. Today, mostly steel plates, CFRP sheets and laminates are applied. The high tensile strength of CFRP is seldom fully exhausted. By prestressing the high tensile capacity of the fibres is exploited more. Other advantages of this technique are the decrease in crack size and crack distances, and the higher stiffness of the element. Prestressing of the laminate introduces preceding stresses in the composite section: the concrete in the tensile zone is compressed and the internal tensile reinforcement is accordingly released. A novel system is presented in this paper and aims to overcome the inconveniences of existing prestressing systems: extensive preparation of the laminates; laborious preparation of the concrete element at the anchorages; long execution period, comprising consecutive steps with extended waiting times.

2 2.1

Principle of the new system

In the novel system, the laminate is curved and blocked with wedges; see Figure 1 [Figeys, 2008, Figeys 2005, Brosens, 2005]. The system consists of an active anchor and a passive anchor. The anchor blocks are fixed to the concrete with glue and anchor bolts. The laminate is inserted in the anchor block through a slit. In the anchor block, the laminate (dimensions: thickness up to 4 mm; width up to 120 mm; length indefinite) is bent away from the concrete. As a consequence, the distance between the laminate and the concrete is kept smaller than 3 to 5 mm. The laminates are blocked in the slot of the anchor blocks by means of wedges. 2.3

Advantages of the new system

The system can be applied in a short time. The number of actions is limited. Moreover, these actions are very simple. In this way, the advantages of the classical

NOVEL SYSTEM Requirements for a convenient prestressing system

A prestressing system that would be appropriate for application on the job site should fulfill the following minimum requirements, according to the experience of the authors: 1. reduced distance between concrete and laminate; 2. ease of anchorage; 3. reduced number of operations; The new system is designed and tested for a maximum transferable force of 200 kN.

Figure 1. Principle of the new prestressing system: active anchor.

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‘passive’ system of externally bonded reinforcement are preserved. Through bending of the laminate away from the concrete, the distance between the laminate and the concrete remains small, without the necessity of chipping a cavity. The wedges provide a full anchoring, so there is no need for additional anchorage by gluing the laminate.

3

EXPERIMENTAL STUDY

A feasibility study has been carried out in the Reyntjens Laboratoy of K.U.Leuven. A FEM analysis is performed to predict the stresses in the laminate and in the anchor block. The numerical results show that the laminate is not uniformly blocked. Tensile stresses locally reach high levels, which are higher than the tensile strength of the laminate. High in-plane stresses are situated in the transition zone which can cause tearing off in the laminate. Also in the anchor block, high stresses are locally present. The anchor block will irreversibly yield. The middle part of the anchor block opens almost 2 mm. 3.1

Testing the clamping system

The maximum transferable force of the clamping device is tested by means of tensile tests. In the anchor block, a CFRP-laminate is blocked by means of flat or rounded wedges. The test setups make use of different thicknesses of the anchor block (40 and 50 mm), different wedge shapes (flat and rounded) and different kinds of load transfer layers (steel plates, CFRP (Carbon Fibre Reinforced Polymer) and AFRP (Aramid Fibre Reinforced Polymer)). The experiments show that clamping of a prestressing force of 200 kN is feasible if: • a load transfer layer is applied; • the laminates are pushed into the anchor block before applying a prestressing force; • a stiffer anchor block is used to avoid large deformations. The general behaviour of the clamping system is well predicted by the preliminary FEM analysis. 3.2

Testing anchorage to the concrete

The anchor block will be fixed to the concrete by using a tab, bolts and/or glue. Tests are carried out on different setups. Two trends can be observed: • tab and glue versus only glue. Applying a tab increases the transferable force only with 10 kN. This contribution is limited compared to

the required labor for preparation (sawing a slot...) to be taken on site. Therefore, the tab will be omitted; • bolt(s) and glue versus only bolt(s) or only glue. – A glued connection transfers 112 kN. A maximum force of 136 kN (one bolt) and 132 kN (two bolts) is reached in case the connection is realized with only bolts. Failures take place by pulling out of the bolts. Applying bolts and adhesive can increase the transferable force significantly. – When only bolts are used, a large displacement of 3 to 4 mm is measured. Applying glue reduces this displacement to only 0.4 mm. The smaller displacement is necessary to limit prestressing losses. The maximum transferable (191 kN, test setup with adhesive and two bolts) force is close to the preset goal 200 kN. Additional tests are necessary to verify if 200 kN can always be reached.

4

CONCLUSIONS

Tensile tests on the system show that the objective of a clamping capacity of 200 kN is obtained. Tests show that the anchorage to the concrete is possible. A sufficient anchorage capacity is clearly within reach when gluing and bolting the anchor block. Further research is needed to optimize the design for on site applications.

ACKNOWLEDGEMENTS The authors express their thanks to the Flemish Institute for Promotion of Scientific and Technological Research in the Industry (IWT—Vlaams Instituut voor de Bevordering van Wetenschappelijk-Technologisch Onderzoek in de Industrie) for the financial support to this research, as well as for the doctoral grant, attributed to the first author.

REFERENCES Figeys, W., Brosens, K., Van Schepdael, L., Dereymaeker, J., Van Gemert, D., A novel system to prestress externally bonded reinforcement, Materials and structures, submitted Feb. 2008. Brosens, K., Dereymaeker, J., Figeys, W., Van Gemert, D. and Van Schepdael, L., Actief en passief verankeringsblok voor voorgespannen vezellaminaten—eindverslag (in Dutch, E: Active and passive anchor block for prestressed FRP strips), technical report, IWT KMOInnovatiestudie type 3, project 040360, 2005. Figeys, W., Brosens, K., Dereymaeker, J., Van Gemert, D. and Van Schepdael, L., UK Patent, 0513461.4, 2005.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

FRP tendon anchorage in post-tensioned concrete structures J.W. Schmidt Technical University of Denmark, Kgs. Lyngby, Denmark COWI A/S, Lyngby, Denmark

B. Täljsten Technical University of Denmark, Kgs. Lyngby, Denmark

A. Bennitz Luleå University of Technology, Luleå, Sweden

H. Pedersen COWI A/S, Lyngby, Denmark

ABSTRACT: Strengthening of building structures by the use of various external post-tensioning steel tendon systems, is known to be a very efficient method. However, FRP as material in external post-tensioning projects has been investigated during the last decade. The advantages for this material are the high effective Young´s modulus and the high stress capacity in the linear elastic range of the material. The use of external tendons increases the requirements on the anchorage systems. This is in particular important when using un-bonded tendon systems, where the anchorage and deviators are the only force transfer points. The demand for high capacity anchorage tendons is fulfilled for steel tendons, but no competitive mechanical anchor has yet been developed for FRP tendon. A new small, reliable and more user friendly anchor has to be developed, before FRP tendons can be utilized with all of its capacity. Thus, several attempts of developing a mechanical FRP anchor have been made worldwide with promising results. Some of these attempts are presented in this paper together with an insight into a present research collaboration project at the Technical University of Denmark, Luleå University of Technology, Sweden, and COWI A/S, Denmark.

1

INTRODUCTION

The implementation of external post-tensioning systems is a very efficient way to upgrade existing concrete structures. Traditional post tensioning steel systems have been used in this area for decades. Often the same anchorage systems as in ordinary bridge constructions are used. It seems obvious to look for an application where Carbon Fibre Reinforced Polymer (CFRP) rods are implemented due to their high corrosion resistance, high stress capacity and low weight. As for steel tendons the real benefit from the material´s high stress capacity can only be achieved when the CFRP rod is prestressed and installed in this state. By this the condition of the concrete structure is improved in the serviceability limit state (e.g. crack widths are decreased) and the load capacity of the structure is also increased to a desired level (due to the compression introduced into the concrete structure). Standard anchor systems for steel tendons have been in use for several decades. The modelling of such devices is simple, since steel is an isotropic material making it possible to determine the shear stresses obtained by introducing a normal stress perpendicular to the tendon. CFRP is, however, an anisotropic material with low compression strength in the transverse direction of the fibres. It is difficult, due to this fact, to establish a model, which determines the size of the shear stress which occurs when the rod is

subjected to a normal stress perpendicular to the rod. A sufficient shear resistance in the CFRP rod can therefore not be obtained in the same way as in traditional pre-stressing anchor systems using steel tendons. It is foreseen that development of a reliable standard anchor for CFRP rods shall to a large extent be based on experiments and tests. The tests shall comprise both the instant pre-stressing state as well as the long-term behaviour—decrease of prestressing force with time. Much research work has been, and still is, carried out in this field. However, a better understanding of the given CFRP anchorage systems is needed in order to develop newer systems which are commercially competitive in the civil engineering industry.The requirements for FRP anchorage system has been discussed by researchers Reda Taha (1997 and 2003) and Sayed-Ahmed (2002) and different standards ASTM (2006), ACI (2004) and associations such as IABSE-AIPC-IVBH(2003), and FIB (1993). 1.1

Mechanical anchors

The split wedge anchor system is the most widely used system in civil engineering application, and it is the most commonly used anchor type in steel pre-stressing. However, the use of wedge anchors implies compression of the FRP tendon, which causes problems due to the anisotropic properties. The strength properties in the

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transverse direction of the fibres are very poor, and mechanical friction anchors use large compression forces to anchorage the tendon. This introduces very high principal stresses in front of the anchor device, where both tension and compression forces are represented. FRP tendon anchorages are therefore, at present time, only reliable if they have a length exceeding 70 mm and a considerable presetting force is applied. It would be possible to use them in larger systems and compete with given steel pre- or post-stressing systems, if this length can be decreased with about 1/3, due to the fact that steel wedges require an anchorage length of 25–50 mm. The number of wedges and the material properties might be varied, in order to grab carefully around the FRP tendon. Enka (1986) used plastic wedges for pre-tensioning mechanical gripping anchors and reported these to work satisfactory. Kakihara et al. (1991) used a two-piece split-wedge anchor for pre-stressed concrete using AFRP, which gave an efficiency of 97%. The split-wedge anchor has the advantage of being easy to assemble and use on construction site and is therefore favoured in pre-stressing applications. Shear failure of the tendon due to excessive shear stresses in the anchor zone has been evaluated by Sayed-Ahmed et al. (1998), who also presented research of FRP tendon anchorage, which dealt with the optimizing of wedge anchor dimension through change of angle in the surface between wedge and barrel and the implementation of a copper sleeve between wedge and CFRP rod Sayed-Ahmed et al. (1998). The reason for the change in angle is to move the high compressive forces into the anchor-back end and through this prevent high principal stresses in front of the anchor, where both tension- and compressive forces are represented. The copper sleeve prevents crushing of the fibres due to direct contact between the wedges and CFRP rod. AlMayah (2007) and his team worked in general extensive on the development of the wedge anchor and gained great progress in developing a wedge anchor system. Al-Mayah et al. (2001–2007). Al-Mayah presented in 2008 a paper Al-Mayah et al. (2008) regarding the study of interfacial mechanics of CFRP–metal couples under different contact pressures using a clamping anchor Aluminium- and copper sleeves were used in direct contact with the rod. The tests showed that the shear stresses increased if smooth machined CFRP rods were used and also when soft sleeves were used together with a rod of high ultimate capacity.

2

PRESENT RESEARCH PROJECT

The research project is a co-operation between Technical University of Denmark (DTU), Luleå University of Technology (LTU) and COWI A/S and focuses on the

development of a new anchor type for FRP systems. The project is still ongoing but promising results have already been obtained. The project involves measuring of shear stresses in the rod through optical fibres, development of a new metal anchor, testing of different anchor lengths and verification through FEM-analysis and analytical theory. 2.1

Test setup

An 8 mm rod tendon is anchored by a clamp anchor in one end (dead end) and a wedge anchor in the other end. The presetting procedure is done though a special designed duct in which the CFRP tendon can be placed. The wedges in the anchor are placed against a support and the whole system is compressed with the decided pre-setting force. The test setup consists of the two mechanical anchors, the rod and the tension system with applied measurement equipment. The optical fibres are mounted in a groove at the rod surface and covered with epoxy for protection. The clamping anchor in the dead end has shown very good reliability in the performed tests, which is important for the test procedure of the new invented anchor. Pilot tests have been done on the clamping anchor in order to verify the reliability of the dead end before testing the wedge anchor. A clamping anchor was placed in both ends of the rods and tested in the test setup. 100% of the tests showed a fracture between the anchors which occurred at 130kN. 6 preliminary tests were made with the new developed wedge anchor, which showed a fracture at 110–112 kN at a presetting force of 100kN. These rods were however taken from another delivery batch, which could mean a change of capacity in the 8 mm rod. The wedges in the wedge anchor consist of aluminium with provided yield strength of 150MPa, and the barrel is produced in steel with provided yield strength of 235MPa. Further tests have to be done on the actual aluminium- and steel material for verification and implementation in the future FEM-model analysis and analytical approaches. The CFRP HR 2500 rod type HS from STO which is used for future tests is approved according to EN 2561. The measurements at present time have been made to ensure fracture between the anchors and reliability of the clamping- and wedge anchor. The barrel strain gauge measurements should result in values which we can use for the prediction of the hoop stresses and contact pressure which leads in the direction of predicting the friction between the barrel and wedges through analytical evaluations. The optical fibres are mounted in order to show the shear stresses in the rod. It is the goal, through these measurements, to verify the shear stresses at the rod surface. The deformation measurer (LVDT) measures the mutual movement of the individual anchor pieces and the rod.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Durability of CFRP strengthened concrete structures under accelerated or environmental ageing conditions K. Benzarti & M. Quiertant Université Paris-Est, Laboratoire Central des Ponts et Chaussées (LCPC), Paris, France

C. Aubagnac & S. Chataigner Laboratoire Régional des Ponts et Chaussées d’Autun, Autun, France

I. Nishizaki & Y. Kato Public Works Research Institute (PWRI), Tsukuba, Japan

ABSTRACT: The durability of concrete slabs strengthened by bonded composite materials has been investigated in the framework of an international cooperation between two French and Japanese research institutes. Time evolution of the concrete/composite adhesive bond strength was studied under both controlled and environmental ageing conditions, by using different mechanical characterization methods. The first results of this ongoing experimental campaign are presented.

1

− polished concrete + CFS, − polished concrete + CFRP plates.

INTRODUCTION

Composite materials are commonly used worldwide for the strengthening of civil structures. In Japan, FRP strengthening methods have been intensively used for the seismic retrofit of bridge columns after Kobe earthquake. In France, such methods are mainly applied to the rehabilitation or upgrading of damaged structures. Nevertheless, the durability of the adhesive bond at the concrete/composite interface is still a matter of investigation. In this paper, the first results of 2 ongoing studies conducted by the French and Japanese Public Works Research Institutes (LCPC and PWRI) are presented. On the French side, CFRP strengthened concrete slabs were exposed to accelerated ageing in saturated humidity. Time evolution of the adhesive bond strength was monitored by either pull-off or shear loading tests. On the Japanese side, similar specimens have been exposed to outdoor ageing for 14 years, and the residual bond strength was evaluated by pull-off or peeling tests. These data offer a unique opportunity to compare the effects of laboratory and environmental ageing on the concrete/FRP interface.

All strengthened slabs were exposed to saturated humidity at 40°C. After increasing ageing periods, characterizations of the adhesive bond were carried out by pull-off tests. Several trends were obtained: − the surface preparation of concrete played a significant role on the strength level as measured by pulloff tests: sand-blasted surfaces led to higher bond strength values than polished ones. Such effect was attributed to a higher cohesion of the superficial concrete after sand-blasting, − a slight decrease in the adhesive bond strength was observed after long ageing periods, especially for polished concrete slabs strengthened by CFRP plates. This result suggests that bonded CFRP plates are more sensitive to aggressive environments than carbon fabric reinforcements. − increased adhesive or mixed failures were observed after long ageing periods, which was consistent with a substantial adhesive bond degradation induced by humid ageing. 2.2

2 2.1

EFFECTS OF ACCELERATED AGEING

In a second set of experiments, CFS strengthened concrete blocs were exposed to saturated humidity at 40°C, and bond characterizations were carried-out by both pull-off and shear tests. It was found that:

Adhesive bond characterization by pull-off tests

In a first set of experiments, 4 types of model concrete/ composite interfaces were investigated: − sand-blasted concrete + CFS (carbon fiber sheet), − sand-blasted concrete + CFRP plates,

Bond characterization by shear loading test, and comparison with the pull-off method

− after 3 months ageing, there was no significant evolution of the maximum shear load or pull-off strength, − however, an evolution of the failure mode was observed on the shear tested specimens, from a

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cohesive to an adhesive failure. But the bond strength remained surprisingly unaffected, − such an evolution was not observed with the pulloff method: the failure mode remained cohesive in the concrete. This raised the question if the pulloff results are really representative of the adhesive bond behavior. 3 3.1

EFFECT OF ENVIRONMENTAL AGEING (b) adhesive failure.

Adhesive bond characterization by pull-off tests

Figure 2.

Concrete slabs were strengthened by 4 types of CFS strengthening systems from the market in 1992, and were exposed outdoor on the reference ageing site in Tsukuba. In 2006, after 14 years exposure, specimens were recovered and the adhesive strength was assessed by pull-off tests. It was shown that: − pull-off strengths of initial specimens were all above 3.0 MPa. The 14 years exposed specimens showed slight decreases in strength, with values ranging from 2.1 to 3.2 MPa. − all specimens (initial and exposed ones) showed substrate failures. 3.2

(a) mixed failure mode

Bond characterization by peel loading test, and comparison with the pull-off method

The 4 types of concrete-to composite interfaces were then characterized by peel loading tests: − exposed specimens showed values of the peel force per unit width between 0,10–0,29 N/mm, − different types of failure modes were observed, depending on the CFS system. On 2 specimens, concrete remained partially attached to the peeled sections of CFS, whereas adhesive failure were observed for the 2 others specimens (Fig. 2). These features differ from those of the pull-off test. This result raises again the question of the representativeness of the pull-off results.

Surfaces of fractured specimens after peel tests.

Unfortunately, it was not possible to compare the peel forces of exposed specimens to the initial values (peeling tests were not scheduled in the original study plan). So, it was decided to remake control specimens from the same raw materials originally used in 1992. Only one specimen (D) could have been remade with exactly the same constituents. Peel loading tests were then carried-out on this remade specimen D: the peel force per unit width was estimated at 0.51 N/mm and a pure substrate failure was observed. These features can be compared to the lower value of 0.16 N/mm and the mixed failure mode that were obtained for the 14 years exposed specimen D. Even if the remade specimen D is not fully representative of the initial one, this result suggests that the CFS/concrete interface of the exposed specimen might have been degraded over time. 4

CONCLUSIONS

French and Japanese researchers came to convergent conclusions on several aspects. Influence of the bond testing methods was discussed in the 2 cases: − alternative methods to the pull-off test, such as shear and peel loading tests, are suitable to assess adhesive properties and their evolution over time, − the representativeness of the pull-off test was questioned on both sides, as results were not consistent with those of alternative methods. It is believed that pull-off results are mainly influenced by properties of the near-surface concrete layer. Significant changes in the adhesive bond properties or failure mode were also observed on each side:

Figure 1. Surfaces of fractured specimens: control specimens tested by shear (a) or pull-off (b) tests, and specimens aged for 3 months and tested by shear (c) or pull-off (d) tests.

− after 3 months humid ageing (40°C, 95% R.H.) the failure mode of CFS strengthened concrete beams tested in shear evolved from a substrate failure to an adhesive failure. But the maximum shear load was not affected, − the peel strength of a 14 years outdoor exposed CFS-to-concrete interface was found to be lower than that measured on a remade specimen which simulated the initial state. This result may account for a degradation of the exposed interface.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Durability of GFRP composite made of epoxy/organoclay nanocomposite H.G. Zhu & Christopher K.Y. Leung Department of Civil Engineering, Hong Kong University of Science and technology, Hong Kong, China

J.K. Kim Department of Mechanical Engineering, Hong Kong University of Science and Technology, Hong Kong, China

ABSTRACT: With a suitable amount of organoclay introduced into epoxy resin, intercalated/exfoliated epoxy/ organoclay nanocomposite showing improved barrier property and thermal stability can be formed. GFRP composite with epoxy/organoclay nanocomposite matrix has been fabricated in this study. The main purpose of our work is to investigate the effect of epoxy/organoclay nanocomposite matrix on durability improvement of GFRP composites. GFRP composite laminates with neat epoxy matrix or 3wt% epoxy/organoclay nanocomposite matrix are conditioned in the following environments: (i) standard laboratory condition; (ii) heating for 1h at 40°C, 50°C, and 60°C; (iii) immersion in alkaline solution at 60°C. After pre-determined periods of conditioning, aged specimens are tested in uniaxial tension to obtain the tensile strength, tensile modulus and failure strain of GFRP composites. Results show that the tensile strength of aged GFRP composites is reduced to different degrees for different environmental exposure condition. However, the degradation rate is reduced when epoxy/organoclay nanocomposite is used as GFRP matrix. This can be attributed to the improved barrier property and thermal stability of the epoxy/organoclay nanocomposite. When GFRP composites are employed in the rehabilitation of concrete structures, the use of epoxy/organoclay nanocomposite matrix is likely to improve the durability of the rehabilitated structure as well.

1

INTRODUCTION

2

GFRP composites used in the rehabilitation of structures may degrade under exposure to elevated temperature, moisture, alkaline environment or thermal cycling. Concerns of durability hinder the widespread utilization of GFRP composite in the infrastructure with service life requirement of 80–100 years. A possible way to enhance the durability of GFRP composite is to use polymer (composites) matrix with improved barrier property and thermal stability. Polymer/organoclay nanocomposite is a new class of material exhibiting markedly improved mechanical and thermo-mechanical properties, enhanced thermal stability as well as improved barrier property. There is hence a good potential to develop polymer/organoclay nanocomposites to be the matrix of GFRP for durability improvement. In this study, the durability of GFRP composites with neat epoxy matrix and epoxy/organoclay nanocomposite matrix was studied. Specifically, the effects of elevated temperature and alkaline environment on the degradation of GFRP with and without organoclay in the matrix were compared.

2.1

EXPERIMENTAL PROGRAM Materials

The epoxy resin employed is a thermosetting epoxy resin (MRL A-115 system supplied by Reno, Taiwan) used for FRP matrix. The organoclay used for the fabrication of epoxy/organoclay nanocomposite is Nanomer I.30P obtained from Nanocor Inc. A plain weave glass woven fabric was used as fiber reinforcement. 2.2

Nanocomposite preparation

The in-situ polymerization method was employed to fabricate the epoxy/organoclay nanocomposite. The resin/organoclay suspension was stirred for 1 h at a shear rate of 3000 rpm at room temperature and was subsequently sonicated at 70 W and 42 kHz for 3 hr at 60ºC. This was followed by a degassing procedure for 2 hrs at 60ºC. Finally, the required amount of hardener was added into the resin/organoclay mixture to form epoxy/organoclay composite samples. The samples prepared through this process were characterized (by XRD and TEM) as intercalated/exfoliated nanocomposite.

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2.3

GFRP composite preparation

The vacuum-assisted hand lay-up method was adopted for fabricating GFRP laminar. The glass fabrics were wetted by epoxy resin (nanocomposite) and stacked together layer by layer until the desired thickness of GFRP laminar was reached. The uncured GFRP laminar was subsequently degassed for 1h at room temperature. All the fabricated GFRP laminates showed good quality (with little visible bubbles) and had a fiber content of about 35% in volume. 2.4

Specimens preparation and aging conditions

Two aging conditions were included in this study. The first one involved the conditioning of GFRP tensile specimens for 1h at 23ºC, 40ºC, 50ºC and 60ºC respectively, followed by testing at same temperature. The other involved the immersion of GFRP laminate in alkaline solution at 60ºC for 30 days. Tension specimens were obtained from both aged and un-aged GFRP laminates and tested at room temperature. For each case, GFRP composites made with both epoxy matrix and 3wt% organoclay/epoxy nanocomposite matrix were tested for comparison. 2.5

Figure 1. Tensile strength of thermal aged GFRP composite.

Tension test

Uniaxial tension testing was conducted on the MTS 810 machine according to ASTM D3039. Loading in the longitudinal direction of tensile specimen was applied at a pace rate of 0.25 mm/min. Longitudinal strain, loading and displacement were recorded concurrently.

Figure 2. Tensile strength of alkaline solution aged GFRP composite with epoxy matrix (blue column) nanocomposite matrix (red column).

3.2 3 3.1

RESULTS AND DISCUSSION Short term thermal aging at high temperature

Figure 1 shows the tensile strength of the two kinds of GFRP composite versus aging temperature. The GFRP with nanocomposite matrix had a slightly lower tensile strength at room temperature. This may be due to the higher viscosity of the nanocomposite resin, which increases the micro-void content in the formed specimens. With increasing aging temperature, the tensile strength of both kinds of GFRP composite decreased significantly due to the thermal softening of the matrix. However, when nanocomposite was used, the GFRP composite exhibited a smaller degradation rate of tensile strength. The difference between the tensile strength of the two kinds of GFRP composite became smaller with increasing temperature. At 60ºC, the tensile strength of GFRP composite with nanocomposite matrix was even slightly higher than that with neat epoxy matrix. This can be explained by the proper dispersion of organoclay platelets in the nanocomposite, which reduces the degree of thermal softening of the matrix.

Long term aging in alkaline solution at 60ºC

Figure 2 shows the tensile strength of the two kinds of GFRP composite versus aging period in alkaline solution (pH = 11.5). For virgin (un-aged) specimens, GFRP with neat epoxy matrix exhibited a slightly higher strength. After being aged for 30 days at 60ºC, the tensile strength of both kinds of GFRP composite were reduced significantly due to the ingress of alkaline solution. With the incorporation of organoclay into epoxy matrix, the GFRP composite exhibited a much slower degradation rate. The higher tensile strength of aged GFRP composite with nanocomposite matrix is due to the improved barrier property of the nanocomposite, which slows down the attack of glass fibers by the penetration of alkaline ions through the matrix. 4

CONCLUSIONS

The tensile strength of aged GFRP composite is reduced, but to different degrees under different aging conditions. GFRP with nanocomposite matrix exhibits a lower degradation rate of tensile strength than GFRP with neat epoxy matrix. The test results demonstrate the potential of using nanocomposite matrix to improve GFRP durability in practical applications.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Performance of Reactive Powder Concrete (RPC) with different curing conditions and its retrofitting effects on concrete member T.P. Chang, B.T. Chen, J.J. Wang & C.S. Wu Department of Construction Engineering, National Taiwan University of Science and Technology, Taipei, Taiwan, ROC

ABSTRACT: The material performance of Reactive Powder Concrete (RPC) with two different curing conditions, water-curing of 25ºC and steam-curing of 85ºC and 95% relative humidity, were studied experimentally. The reinforcing effects of the RPC with two wrapping thicknesses of 10 and 15 mm, respectively, on the surface of cylindrical concrete specimen were evaluated. Major experimental results show both the engineering properties and indices of durability of RPC with steam-curing at four different ages have substantially increased except for the supersonic pulse velocity and dynamic moduli of elasticity and shear. The ratio of increase of compressive strength of cylindrical specimens retrofitted with 10 and 15 mm of wrapping RPC are in the range of 9.5 to 38.0%.

1

INTRODUCTION

Reactive powder concrete (RPC), a cement-based composite material well known for the ultra-highstrength, high-durability and low-porosity, made its international debut in 1994 To produce a very high compressive strength of RPC, applications of pressure before and during setting and heat-treating after setting are normally required. Compressive strengths of 200 to 800 MPa, moduli of elasticity of 50 to 60 GPa and flexural strength of 6 to 13 MPa have be achieved with RPC. In this study, the material performance of RPC with two different curing conditions, water-curing of 25ºC and steam-curing of 85ºC and 95% relative humidity were evaluated by comparison on the changes of engineering properties and indices of durability. 2

EXPERIMENTAL PLAN

The experimental plan was composed two major parts. The first part was the evaluation of the variations of material properties of reactive powder concrete (RPC) resulting from two different curing conditions, watercuring of 25ºC and steam-curing of 85ºC and 95% relative humidity. The second part was the evaluation of the reinforcing effects of RPC when it was used as a retrofitting material to cast circumferentially on the surface of the cylindrical normal weight concrete specimen. The composition of normal weight concrete (NC) with water-to-cement ratio (W/C) of 0.6 Ordinary Type I Portland cement complying with ASTM C 150 was used. Natural crushed gravel (maximum size of 10 mm) and river sand (specific gravity of 2.65,

fineness modulus of 2.82) were used as the coarse and fine aggregates respectively. The NC was used to cast cylindrical concrete specimens with three different diameters, φ100 × 200 mm, φ80 × 200 mm and φ70 × 200 mm, respectively. After demoulding, the specimens were cured in saturated lime water for 28 days. Then, each of two latter types of smaller cylindrical specimens was placed into a φ100 × 200 mm steel mold and cast with fresh RPC to form a circumferential layer of thickness of 10 and 15 mm, respectively, as the retrofitting material. The composition of RPC with a water to cement ratio (W/C) of 0.193 was used. In practice, in order to reduce the cement hydration heat and increase the capability of sulfate resistance, either Type II or Type IV Portland cement can be used. In this study, Type II Portland cement of the fineness of 314 m2/kg using air permeability test was utilized. Imported silica fume was used which contains about 94.6% of SiO2. Its particle sizes were in the range of 0.67 to 5.02 µm with a specific surface area more than 20,000 m2/kg. Cylindrical φ50 × 100 mm paper mold was used to cast the RPC specimens for the test of compressive strength. Beam specimen of 40 × 40 × 160 mm in a steel mould was used to test the flexural strength. Cylindrical RPC specimens of φ100 × 200 mm were also cast for the tests of ultrasonic pulse velocity, electrical resistivity, and chloride permeability. After these fresh RPC specimens were cured in laboratory under ambient temperature for 24 hours, they were demoulded. In order to keep the strength and modulus of elasticity of RPC at hardened state be as close as possible to those of the base NC solid cylindrical specimens, no additional pressure was applied. The curing temperature was also kept at a relatively lower value of 85ºC.

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Two types of indices to evaluate the performance of RPC at ages of 3, 7, 14 and 28 days were conducted: (1) Engineering properties including the compressive strength, splitting tensile strength, flexural strength, dynamic modulus of elasticity and dynamic shear modulus; (2) Durability properties including electrical resistivity, ultrasonic pulse velocity, water absorption and chloride permeability.

for ultrasonic pulse velocity indicate that this property is not sensitive to the changes of other material with ages. The ratios of decrease for water absorption are in range of 49.0 to 66.7% imply this test method although is rather simple but seem to properly reflect the function of pozzolanic reaction activated by the steam curing. 3.3

3 3.1

RESULTS AND DISCUSSION Engineering properties of RPC

The results indicate that the compressive strengths of RPC specimens cured with the steam increase by ratios of 17.0 to 23.7% at four ages. Both the ratios of increase of 13.5 to 17.9% for the splitting tensile strengths and 19.3 to 25.8% for the flexural strength are also observed. Because the pozzolanic reaction resulting from the ingredients of silica fume, fly ash and blast furnace slag in the RPC mixture will be activated energetically by the high temperature and moisture of curing steam. Such pozzolanic reaction causes a denser microstructure of C-S-H cement hydrate and results in a faster development of strength gain. The strength development for the RPC specimen cured in the water under ambient temperature is apparently much slower. On the other hand, the increases of dynamic moduli of elasticity and shear, and Poisson’s ratio are insignificant, only in the range of –0.7 to 2.0%. These three measured values are not very sensitive to the changes of other material properties of RPC with the increase of ages. It is noted that, unlike the normal weight concrete specimens with the failure modes of either in crushed state or two separated pieces, these failed RPC specimens are still kept together by the steel fibers. Typical stress-strain curve for the RPC specimens 3.2

Durability properties of RPC

The durability properties of RPC under two different curing conditions indicate that the electrical resistivity of RPC specimens cured with the steam increases drastically by ratios of 333.8 to 5485.7% at four ages. Both the ratios of decrease of 0.04 to 3.4%

Retrofitting effects of RPC

Because the addition of steel fibers, RPC has a rather high tensile strength and impact resistance compared with the normal weight concrete. The failure mode will change from the brittle to the ductile. These two good material properties of RPC are the major advantages to consider it to be used as the retrofitting materials. On the other hand, unlike the conventional retrofitting sheet-type material such as Carbon Fiber Reinforced Polymer (CFRP), RPC needs to have certain amount of thickness and inherent high stiffness when it is used to retrofit the column-type structural members. The experimental ultimate compressive stresses of composite cylindrical specimens encircled with a retrofitting RPC layer of thicknesses of 10 mm and 15 mm, respectively, were calculated and shown in Table 6. The average compressive stresses were increased from 9.5% (10 mm thickness of layer) to 38.0% (15 mm thickness of layer)

4

CONCLUSION

Major experimental results show that the compressive, splitting tensile and flexural strengths of RPC with steam-curing increase substantially. However, the increases of dynamic moduli of elasticity and shear, and Poisson’s ratio are not very sensitive to the changes of other material properties of RPC with the increase of ages. Similar trend is also observed for the supersonic pulse velocity of RPC which is also insensitive to curing condition. Water absorption test is rather simple but seems to be able to provide distinguishable change of material properties resulting from the apparent pozzolanic effects due to steam cuiring. Increases of compressive strength of cylindrical specimens retrofitted with 10 and 15 mm of wrapping RPC are 9.5% and 38.0%, respectively.

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Residual strength of SBR-latex modified high strength concrete under repeated impact loading S.R. Shashikumara & Sadath Ali Khan Zai Faculty of Engineering-Civil, U.V.C.E., Jnanabharathi, Bangalore University, Bangalore, India

K. Muthumani Advanced Seismic Testing and Research Laboratory, S.E.R.C., Chennai, India

N. Munirudrappa Dayananda Sagar College of Engineering, Bangalore, India

ABSTRACT: In the field of concrete technology the conventional ingredients were slowly and safely replaced by new and more efficient materials, which have opened a wide scope for research on their specific uses. Once the concrete has subjected to any kind of external force, the deterioration or decay of concrete will start. The present approach is aimed to study the post-impact behavior of SBR-latex modified reinforced high strength concrete beams. Experimental study has carried out on six concrete beams with dimension of 150 mm × 230 mm × 3300 mm to obtain the load time variation under repeated impact loading, natural frequencies, and the load-deflection variation of test specimens under static loading. Experimental investigation has shown decrease in natural frequencies in all the beams, with increase in number of blows and Dynamic Response Index (DRI), which is an indirect measure of stiffness and strength of the test specimen, reduces with increase in number impact blows.

1

INTRODUCTION

Reinforced concrete structures, unlike steel structures, tend to fracture or fail in the relatively brittle fashion, as the deformation capacity of conventional concrete is limited. Among all polymer composites, styrene butadiene rubber latex (SBR) is performed well in augmenting the mechanical behaviors of the conventional concrete Zai S.A.K. et al (2007). Impact loading is a transient phenomenon, which is characterized by its short duration or sudden occurrence. Two—point static loading was practiced to determine the residual flexural strength of concrete. The reduction in strength due to repeated impact loading can be determined by studying the load— deflection behavior of concrete under two—point static loading test.

2

REVIEW OF LITRATURE

The present investigation deals with studies on combined effect of impact and static loading of SBRLatex polymer modified high strength reinforced concrete beams. Arun Kumar et al (2000) studied the mechanical properties of SBR-latex modified concrete both at the fresh and hardened state and observed that the toughness of SBR modified concrete was improved as compared to the control

concrete. Rajagopalan et al (1995) studied the stiffness degradation of reinforced concrete beams under lowenergy horizontal impact loading and suggested that the degradation of stiffness has been theoretically predicted in terms of natural frequency. Zai S.A.K et al (2007) carried out the experimental investigation on the impact behavior of SBR-latex modified steel fiber reinforced high strength concrete beams and concluded that the dynamic behavior of structural elements can be increased by using higher toughness and high absorbed capacity, which can be achieved by addition of fiber and latex to the concrete matrix. Further, Choubey et al (1995) carried out the investigation on the residual strength of reinforced concrete beams and concluded that with increase in impact loading, the rebound capacity of beams reduces and permanent deflection increases and also the dynamic response index reduces with increase in impact loading. The literature review conducted in the present work concentrated on the post-impact behavior of SBR-latex modified high strength concrete.

3

TEST PROGRAMME, INSTRUMENTATION AND RESULTS

Two High strength concrete beams (HSCB) and four SBR-latex modified high strength reinforced concrete beams (LMHSCB) are chosen with dimension of

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4

Table 1.

Details of natural frequencies. Natural frequencies (in Hz) HSCB B1

B2

B3

B4

B5

B6

0 25 50 100 150 200 250 300 350 400

31.74 31.42 30.00 27.52 26.91 – – – – –

35.29 28.44 27.62 27.15 26.66 26.54 25.91 25.75 – –

33.70 25.21 24.64 23.90 23.62 22.90 22.81 22.30 22.22 22.07

34.09 26.20 25.64 25.31 25.10 24.49 – – – –

35.29 27.65 26.90 26.66 26.43 24.39 – – – –

35.71 35.09 30.46 28.04 27.90 27.77 26.67 26.09 25.75 –

40 0blow

35 30

25blow 50blows

25 20

100blows 150blows

15 10

200blows 250blows

5 0

300blows 350blows

0.0

CONCLUSION Figure 3.

Cumulative Deflection (mm)

This work is aimed to understand the combined effect of both impact loading and the static loading and to assess

Pendulum type impact testing machine.

Displacement, mm

0.5 0.3 0.1 -0.1 -0.3 -0.5 3800

4000

4200 4400 4600 Time,msecs

4800

5000

Figure 2. Typical variation of displacement with time.

5.0 10.0 Deflection in mm

15.0

400blows

Load—time variations.

Beam-B3

120

Beam-B1

100 80 60 40 20 0 0

Figure 1.

LMHSCB

No. of blows

DRl in kn

150 × 230 × 3300 mm. The characteristic strength of concrete used, is of 60MPa and 12 mm dia bar is used as main reinforcement along with 8 mm dia stirrups at 100mm center-to-center spacing. 53 MPa cement and 7% silica fume are used as binder. 15% of SBRlatex used in the investigation. Test beam are subjected to two different levels energy inputs. The impact responses and loads are monitored and stored during 1, 25, 50, 100, 150, 200 and upto 400 impact blows. The pendulum type impact testing machine used in the experiment is as shown in figure 1. To record the different responses during impact loading and static loading different instruments are used viz., linear variable differential transducer (LVDT), piezo electric accelerometers, load cell, dial gauge and data acquisition system. The results of the dynamic response of test beam consisting the responses like natural frequency and others.The natural frequencies are recorded during the free vibration test both at uncracked and cracked stage as shown in figure 2. It is observed that the peak displacement decreases as the number of impact blow increases as tabulated table 1. The static behavior in terms of load-deflection and DRI are as shown in figures 3 and 4.

Figure 4.

200 400 No. of blows

600

DRI—no. of impact blows.

the residual strength and it can be concluded with following experimental observations, With increase in impact loading, the rebound capacity of the test beam specimen reduces and permanent deflection increases. The natural frequency of the test beam specimens was reduced as the number of blows increases. The DRI reduces with increase in impact loading. which increases upto 5% for LMHSCB beams as compared to HSCB beams. The inclusion of SBR-latex to the concrete will improve both impact as well as static behavior as observed from test results.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Modeling the behavior of insulated FRP-strengthened reinforced concrete beams exposed to fire W.Y. Gao, K.X. Hu & Z.D. Lu Research Institute of Structural Engineering and Disaster Reduction, Tongji University, China

ABSTRACT: Externally bonding of Fiber-Reinforced Polymer (FRP) is becoming a popular technique for strengthening Reinforced Concrete (RC) structures all over the world. However, very little information is available on the behavior of FRP materials and FRP-strengthened reinforced concrete members in fire. This paper presents details and results of a nonlinear Finite Element (FE) model developed using commercial ANSYS program. The FE model include two portions: the initial portion is the calculation of the transient temperature distribution and the second portion is the structural response analysis due to the effect of thermal and mechanical load. The comparison of numerical results and full-scale proving tests shows that the proposed FE model is capable to describe the thermal and structural response of insulated FRP-strengthened RC beams in fire.

External bonding of fiber-reinforced polymer (FRP) plates or sheets (hereafter referred to as sheets only for brevity) has emerged as a popular technique for the strengthening of reinforced concrete (RC) members. However, very little information is available on the fire performance of FRP-strengthened RC members, and this is one of the primary impediments to using FRP in buildings. In this paper, three-dimensional finite element models are developed with ANSYS version 8.0 to predict the behavior of FRP-strengthened RC beams exposed to fire. The FE model consists of two calculation portions: temperature distribution analysis and structural response analysis. First, a nonlinear thermal analysis is carried out to determine the temperature distribution history within the element forming the beam under study. Second, the versions in the stiffness matrix due to changes in the materials properties of the structure are calculated accordingly, and a static analysis is performed in a number of time-steps until failure. The results from FE models are compared with the fire tests in terms of temperature-time and deformation-time curves of insulated strengthened beams and the model’s predictions are shown to agree with the test data satisfactorily. The temperature distribution T(x1, x2, x3) at time t is governed by the Fourier transient heat conduction second order partial differential equation: 3

∂ ⎛

3

i =1

i

j =1

∂T ⎞ ∂T =0 ⎟ + Q − ρc ∂t ⎠ j

∑ ∂x ⎜⎝ ∑ k ∂x

(1)

The transient heat flux at the boundary conditions can be regarded as a simple superposition of convection

and radiation heat fluxes, which can be represented by the following equation: ⎛ ∂T ∂T ∂T ⎞ k⎜ lx + ly + lz ⎟ = qc + qr x y ∂ ∂ ∂z ⎠ ⎝

(2)

qc + qr = α c (Te − Ts ) + ε ⋅ σ (Te 4 − Ts 4 )

(3)

For evaluating the effect of the flame convection and emissivity of the fire, the convection coefficient is assumed to be αc = 25 W/(m2-°C) for the fireexposed surfaces and αc = 9 W/(m2-°C) for the upper surface. The resultant emissivity of the strengthened beam surface and fire is assumed to be ε = 0.6. Then three-dimensional finite element models are developed for insulated strengthened beams using ANSYS. SOLID70, LINK33, and SOLID70 thermal element types represented concrete, reinforcing steel bars, and insulation layers, respectivity. In structural response analysis, SOLID65, 8-node solid elements with three degress of freedom per node are employed to simulate the concrete. Using such elements satisfies plastic derformation, cracking in three orthogonal directions, and crusing due to their quadratic interpolation functions. The steel reinforcement elements embedded in the concrete are represented by three-node truss elements LINK8. SHELL41 elements are used to model the FRP sheets. In the FE model, a ‘real strength model’ is established to consider the both factors, the FRP strength reduction and bond property deterioration, due to no work has been conducted on the accurate bond stress-slip behavior that

429

200

can be incorporated into a finite-element analysis. Nodes of the FRP shell elements are connected to those of adjacent concrete solid elements. For temperature distribution in FE model exposed to fire is non-homogeneous in reality, different mechanical and thermal material properties must be taken into account at each point. Numerical Gauss integration points are used to introduce these conditions into the stiffness matrix. After assembling the various elements and taking into account the boundary conditions, the equilibrium of the model is obtained by solving the following nonlinear equation in order to the displacement field: (4)

Tem perature (

)

160 140 120 100

Section C Section D

80 60 40 20 0

25

50

75

100

125

Exposure Time (min)

Figure 2. Comparison for temperature-time curves of Beam 2.

0

Mid-span Deformation (mm)

In nonlinear analysis, the total fire-resistance time applied to a FE model is divided into a series of time increments called time steps. At the completion of each incremental solution, the stiffness matrix [K ] of the model is a function of the mechanical properties and of the stress field, and it is updated at each internation step using the Newton-Raphson method. Prior to each solution, the Newton-Raphson approach assesses the out-of-balance forces {∆P}, which is the difference between the restoring forces (the loads corresponding to the element stresses) and the applied loads. Subsequently, the program carries out a linear solution, using the out-of-balance forces, and checks for convergence. In order to examine the accuracy of the proposed nonlinear finite element model, two simply support CFRP-strengthened RC beams with 4-point bending are carried out at the State Key Laboratory of Disaster Reduction of Tongji University according to ISO 834 standard fire and reported by Gao et al. The measured and predicted polymer matrix temperaturetime curves of Beams 1–2 are shown in Figs. 1–2; there is a reasonably good agreement between predicted and measured values. Fig. 3 shows the measured and predicted mid-span deformations of proving beams during fire exposure.

-40

-80

-120

-160

-200

Measured---Beam1 Measured---Beam2 Model Predition---Beam1 Model Predition---Beam2

0

25

50

75

100

125

150

Exposure Time (min)

Figure 3. Mid-span deformation-time curves of Beams 1–2. Measured Model Prediction

0 -4

Mid-span Defermation (mm)

[ K (σ )]{∆u} = {∆P}

Measured (1&4&5&8) Model Prediction (1&4&5&8) Measured(2&3&6&7) Model Prediction (2&3&6&7)

180

-8 -12 -16 -20 -24 -28 20

400 350

Temperature ( )

300 250

150 Section A Section B

100 50 25

50

75

100

100

140

180

220

260

300

Figure 4. Mid-span deformation-time curves: model predictions versus measured results.

200

0 0

60

Exposure Time (min)

Measured (1&5&6&10) Model Prediction (1&5&6&10) Measured (2&4&7&18) Model Prediction (2&4&7&18) Measured (3&8) Model Prediction (3&8)

125

150

Exposured Time (min)

Figure 1. Comparison for temperature-time curves of Beam 1.

It can be seen that the FE model predicts, reasonably well, the trend in the progression of the mid-span deformations with exposure time. Using the FE model presented in the preceding section, the predicted and measured curves of mid-span deformation versus fire duration time are shown in Fig. 4 with one of the four RC beams tested at the National Research Council of Canada (NRC) according to ASTM E119 standard fire and reported by Benichou et al. Good agreement is obtained between the predicted and test results.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Fracture optimization of RC beams strengthened with CFRP laminates with finite element and experimental methods A.A. Ramezanianpour & A. Gharachorlou Amirkabir University of Technology, Tehran, Iran

ABSTRACT: The use of epoxy-bonded FRP composite for structural repair is emerging as an efficient and cost-effective technique for restoring and upgrading the capacity of RC beams but research data on fracture optimization of this composite material is very limited. This paper presents the results of analytical FEM and experimental studies concerning the flexural strengthening of reinforced concrete beams by external bonding of CFRP laminates to the tension face of the beams. In this study using finite element program (Ansys) and several iterations lead to finding a model with optimized fracture mode. Furthermore, for checking the FEM results, 2 series of control and strengthened beams were fabricated and tested experimentally. The calculated load-deflection curves predict most of the available experimental data satisfactorily. Moreover, experimental fracture mechanism occurred by CFRP breaking and concrete crushing simultaneously which seems to be the best failure mode.

1

INRODUCTION

The continual deterioration of infrastructure has heightened awareness of the need for effective structure rehabilitation procedures. The use of externally bonded fiber-reinforced polymer (FRP) sheets and strips has recently been established as an effective tool for rehabilitating and strengthening reinforced concrete structures. Laboratory and finite element studies have demonstrated the effectiveness of externally bonded FRP plates in enhancing the flexural capacity of concrete beams. However, there are limited reports on fracture mode or optimization of the type of fracture. This paper presents the finite element method for optimizing the fracture mode of strengthened beams which is steel yielding at the first step and FRP rupture and concrete crushing at the next steps and checking the results by fabricating several real size beams.

2 2.1

BEAMS FABRICATION Finite element

After several iterations four rectangular, under-reinforced concrete beams modeled in this study. The specimens were divided into two series of control and strengthened beams, according to the specimen’s characteristics. ANSYS finite element program was chosen to perform the analysis. For defining concrete materials, concrete nonlinear material from ANSYS material library was chosen and filled by entering the elastic modulus, compressive strength and Poision’s ratio. The steel for the finite element models was grade 60 steel with 200 GPa elastic modulus and assumed to be an elastic-perfectly plastic material. FRP composites

are materials that consist of two constituents; one constituent is the reinforcement, which is embedded in the second constituent, a continuous polymer called the matrix. Therefore the FRP composites are especially orthotropic material. Beams with dimensions of 1000 × 150 × 150 mm, and with different compressive steel area of 100 and 157 mm2 (ρ = 0.005 and ρ = 0.008) were modeled corresponding to types I, II respectively. The finite element simulations were displacement controlled and the corresponding applied load due to the prescribed displacement was determined by monitoring the vertical reaction forces at the concrete nodes in contact with loading. 2.2

Experimental study

To check the main objective of this project, an intensive experimental work was conducted. The accomplished experimental work was divided into two main work packages deference in compressive steel area. Four specimens with dimensions of (1000 × 150 × 150) mm were prepared and cast from concrete mix having compressive strength of 30 MPa then reinforced with two F12 mm steel rebars as the tension steels and two F8 or F10 mm steel bars as compressive steels for series I and II, respectively. Furthermore, 8 mm bar at 10 cm space used for stirrups resisting against shear fracture. After 28 days curing period two RC beams retrofitted with CFRP plates while the other beams were used as control specimens. To determine the fracture mode, deflections and the ultimate failure loads of the beams strengthened with CFRP plates, four point loading system were used. All beams were loaded to failure and deflections were measured using test control hardware.

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3

FINITE ELEMENT AND EXPERIMENTAL RESULTS

3.1

Crack pattern at failure

ANSYS FEM and experimental results show that at the bottom of the beam at midspan, principal tensile stresses occur mostly in the x direction (longitudinally). When the principal stresses exceed the ultimate tensile strength of the concrete, cracking signs appear perpendicular to the principal stresses in the x direction. In general, flexural cracks occur early at midspan. When applied loads increase, vertical flexural cracks spread horizontally from the midspan to the support. Finally, compressive cracks appear at nearly the last applied load steps. The appearance of the cracks defines the flexural failure mode for the beams. 3.2

Figure 2. Stress in CFRP laminates in the strengthened beam of group I at crushing of the concrete.

Load deflection plots

Figure 1 illustrate the load-deflection curves of series I, strengthened and control beams respectively. The load-deflection curves in Figure 1 indicate that prior to concrete cracking CFRP strengthening does not influence the load deflection curve. After cracking, the strengthened specimen behaves stiffer and the load at yielding of reinforcement increased by 20 percent with respect to control specimen. After yielding, the strengthened specimen carried 69 and 71 percent higher load than the control specimens due to analytical and experimental results. As shown in Figure 1, the calculated load-deflection curves predict the available experimental data satisfactorily. 3.3

Failure optimization for FRP ruptures and concrete crushing simultaneously

that using proper compressive steel leads to achieving the maximum stress available in CFRP fibers. The ultimate stress of CFRP laminate reaches to 3850 and 3950 MPa for each group which shows that the best steel arrangement is two F8 bars leads to 3850 MPa stress in fibers which is nearest point to fiber maximum allowable stress (3900 MPa). Although in experimental result the second series achieve a little more strength before fiber rupture and that is because of the differences between material properties and experimental and analytical tests. From the experimental result it can be seen that both series of beams fail in the predicted failure mode, which is steel yielding at the first step and FRP rupture and concrete crushing at the next steps. However, the second series achieve 6 percent more strength than the same beams of series I due to more compressive steel.

Figure 2 illustrate the stress of CFRP laminate in failure position for series I. Finite element results show 4 MEASURED RESPONSE FOR GROUP I 90

Experimentall results Control Analytical results Control

80 70

Experimental results Strengthened

LOAD KN

60 50

Analytical results Strengthened

40 30 20 10 0 0

2

4

6

8

10

DISPLACEMENT mm

Figure 1.

12

14

CONCLUSIONS

This paper represent a general method and a finite element solution for best FRP and steel arrangement of RC beams strengthened in flexure leading to best kind of fracture. This kind of fracture starts by steel yielding and then FRP rupture and concrete crushing at the next steps. Thus this program can be used for arrangement of each beam using in experimental test for other purpose in particular flexural tests. The experimental results show good agreements with general behavior of the finite element models represented by cracking patterns, load deflection plots at midspan and the fracture mechanism occurred by CFRP breaking and concrete crushing simultaneously.

Load-deflection curves.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Comparing the behaviour of reinforced HSC beams with AFRP bars and confined HSC beams with AFRP sheets under bending R. Rahgozar University of Shahid Bahonar, Kerman, Iran

M. Ghalehnovi University of Sistan and Baluchestan, Zahedan, Iran

E. Adili Islamic Azad University of Zahedan, Zahedan, Iran

ABSTRACT: By increasing the use of FRP composites in civil engineering, they seem highly essential to be studied. The purpose of the study is comparison of the behavior of AFRP reinforced HSC beams (reinforced with AFRP bars) and steel reinforced HSC beams which confined with AFRP sheets under bending. eighteen beams have been modeled with ANSYS. Three beams are HSC which reinforced with AFRP bars. These beams are named AFn which AF is name of tensile bars of the beams and n is number of them. After modeling, the results have been compared with experimental results and then software has been calibrated. Then twelve steel reinforced HSC beams which confined with AFRP sheets (with different number of laminates) have been modeled. These beams are named SiCj which S is name of tensile bars(steel) and i is number of them, C shows cover layer of AFRP at the bottom of beams and j is number of layers. In addition three simple steel reinforced HSC beams have been modeled as the base of comparison. These beams are named STn which ST is the name of tensile bars(steel) and n is number of them. At the end behavior of aforementioned beams has been compared and corresponding graphs have been sketched. This table shows behavior of beams under bending.

INTRODUCTION Fiber-reinforced polymers (FRP) are using in the form of sheets or laminates to confinement and bars to reinforcement the concrete members. In both they have some advantages to steel jackets and steel bars. Steel is an isotropic material and its modulus of elasticity is high, thus the steel jackets stand the great part of axial forces which lead to buckling of steel. On the other hand, Poisson ratio of steel is greater then concrete, thus the two materials act separately. Corrosion and hard performance are the other problems of steel jackets. [1] Although using the FRP bars as the main reinforcement isn’t common yet, it seems they will play an important role as a main reinforcement. Fiber-reinforcement polymers (FRP) in the form of bars or sheets, usually made from one of the three basic types of fibers such as Aramid (AFRP), Carbon (CFRP), and glass (GFRP), represent one of the most promising new developments in the area of structural concrete. High strength, but lightweight fibers encapsulated in a polymer matrix possess noncorrosive, non-conducting, and nonmagnetic purpose structures. The non-corroding characteristics of FRP reinforcement could also significantly increase the service life of ordinary concrete structures. [2, 7]

In the case of flexure, the very high strength FRP bars, which exhibit elastic response up to failure, could perhaps be effectively used in combination with high strength concrete (HSC). However the majority of reported research works (Cosenza et al 1997.[5]; Toutanji and Saafi 2000 [11]) dealt only with normal strength concrete (f 'C ≤ 41 Μpa), while some other (Benmokrane et al. 1996 [12]; Masmoudi et al.1998 [6]; Grace et al. 1998 [13]) considered concrete with maximum compressive strength (f 'C ) of up to 70 Mpa. Only The’riault and Benmokrane (1998) [14] used concrete with (f 'C) as high as 100 Mpa. Some other researchers worked on the effect of confinement of RC beams (Dathinh and Starnes [4]). In this study behavior of HSC beams reinforced and confined with AFRP under bending have been compared. ANSYS 9 has been used for modeling the beams. CONCLUSIONS 1. Beams reinforced with AFRP bars (first group) have linear behavior up to failure. Their fracture is in brittle manner that can be a disadvantage but They have large deflection before failure which can be a caution.

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2. aximum deflection in HSC beams covered or reinforced with AFRP is higher than HSC beams reinforced with steel bars. Furthermore increase the number of bars. Furthermore increasing the number of tensile bars increases the Maximum deflection tensile bars increases the Maximum deflection of AFRP reinforced and covered beams (first and third groups) but decreases it in steel reinforced beams second group). 3. Failure force of AFRP reinforced and covered HSC beams are much higher than steel reinforced. Effect of tensile bars increasing on failure force in AFRP reinforced HSC beams is higher than AFRP covered and steel reinforced ones, furthermore it would be increased by increasing the number of tensile bars in first group and be decreased by increasing the number of bars in second and third groups.

4. Failure force in AFRP reinforced HSC beams is less than even one layer AFRP covered HSC beams. 5. Failure forces in third group are higher than first group and in all cases their maximum deflections are less than first group. Furthermore in third group effect of tensile bars increasing on failure force is less than the other groups. The mentioned effect become less and less when the number of AFRP layers increased because higher amounts of load are bearing by AFRP covers and number of tensile bars has less effect. 6. HSC beams with AFRP covers (third group) have higher ductility than uncovered beams (second group). Ductility factor (µ) increases by increasing the number of AFRP.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Corroded RC beam repaired with near-face mounted CFRP rods Amjad Kreit, Firas Al-Mahmoud, Arnaud Castel & Raoul Francois Université de Toulouse; UPS, INSA LMDC (Laboratoire Matériaux et Durabilité des Constructions), Toulouse, France

1

INTRODUCTION

The experimental work presented in this paper aims to study the possibility to repair one reinforced concrete beam (figure 1) corroded naturally since 1984 at the Laboratory of Materials and Construction Durability (L.M.D.C.) with a 6 mm diameter Near Surface Mounted CFRP rod. 2

MATERIALS

The average compressive strength and the elastic modulus obtained on cylinder specimens were respectively 45 MPa and 32 GPa at 28 days. The tensile strength, measured using the splitting test, was 4.7 MPa. Porosity was 15.2 %. The reinforcing cages were composed of naturally Fe40 type half-hard steels; ordinary ribbed reinforcing steel bars are used with yield stress equal to 500 MPa. One CFRP rod was used to repair the corroded beam: carbon-epoxy pultruded FRP with 6 mm diameter The modulus of elasticity and tensile strength of the CFRP rods were assessed by laboratory testing Ribbed Φ 6mm

3

EXPERIMENTAL PROGRAM

Two beams were tested in this experimental study, a control beam BT1 and a beam naturally corroded BCor. BCor was repaired by using one Near Surface Mounted 6 mm diameter CFRP Rod. Figure 3 shows the location of the concrete damage due to the expansion of the corrosion products. The corrosion damages are mainly characterised by a concrete cracking which propagates along the longitudinal steel reinforcing bar. Thus, the concrete located at each corner of the cross-section is often completely broken or at least significantly damaged. On the contrary, the concrete located far from the reinforcing bar in the middle of the beam cross-section is not significantly concerned by the steel bar corrosion. As a result, assuming that the concrete is still of a good quality, the NSM rod was implemented in the

1cm

2Φ 12mm

1cm 1cm 28cm

300cm

14 Stirrups Φ6mm

2Φ 6mm

Φ 6mm/220cm

and are equal to 146 GPa and 1875 MPa respectively. To modify the surface of the initially smooth rods in order to enhance the bond with the filling material, a surface sanding treatment was applied (figure 2). The smooth CFRP rods were coated with 0.2/0.3 mm sand by sprinkling it on to a thin layer of freshly applied epoxy resin (AL-Mahmoud et al. 2007). The filling material is an epoxy resin (Eponal380). After 7 days, the compressive and tensile strengths are 83 and 29.5 MPa respectively. The elastic modulus is 4.9 GPa.

Ribbed 12mm 15cm

Beam BCor

28cm

Figure 1. Lay-out of the reinforcement (all dimensions in cm) for type B beams.

Damaged Concrete

1cm

E=146 GPa σu=1875 MPa

Figure 2. Sanded CFRP rod used to repair the corroded beam.

CFRP Rod Φ 6mm 1.5cm 13.25cm 13.25cm 15cm

Resin

Figure 3. Location of the NSM 6 mm diameter rod in the tensile area of the corroded beam BCor.

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middle of the cross-section in the tension area, as shown in figure 3. The total length of the NSM rod is equal to 270 cm. Thus, the repair is stopped just before the supports. Finally, control beam BT1 is the reference reinforced concrete beam without corrosion or CFRP strengthening. The two beams were tested up to failure under a monotonic increasing load in three-points bending. The global behaviour is analysed through the study of the mid-span deflection. This mid-span deflection was measured by using a 100 mm capacity LVDT displacement sensor with an accuracy of 0.01 mm. The distribution of the steel cross-section reduction along both tensile bars of the corroded beam was measured after the mechanical test up to failure. To assess the steel corrosion, concrete was completely removed from reinforcement. Reinforcement mass loss was measured. 4

EXPERIMENTAL RESULTS

The maximum average cross-section reduction on both tension reinforcing bars was equal to 36.3% (in percentage of total initial cross-section 226 mm2). This maximum cross-section loss was located about 20 cm away from the mid-span, then almost where the bending moment is the maximum. As a result, about 40% reduction in the bearing capacity can be expected in comparison to the control beam BT1. Of course, this assumption could not be confirmed by the mechanical experiments as the beam BCor was repaired before to be tested up to failure. Figure 4 shows the deflection-bending moment curves obtained for the two beams. Both beams were already cracked due to their ageing condition under sustained loading since 1984. Tensile steel yielding bending moment was 25 kN.m and 31 kN.m respectively for beams BCor and BT1. The ultimate moment was 37 kN.m for BCor and 36 kN.m for the control beam BT1. The corroded beam was already cracked at the time of the test as it was kept under a high sustained loading level for 24 years. Point A is the tensile steel yielding moment. During this first stage, both the steel and the CFRP rod are working. On the contrary, during the second stage (point A to point B), the tensile steel is yielded, and then only the CFRP rod contributes to carry the increasing bending moment. That is why the slope of the linear curves is significantly reduced in comparison to the previous stage. The non linear behavior of the resin can also contribute to reduce the beam stiffness at this stage. Point C is the failure. During the third phase, between points B and C, the curve is not linear anymore. Significant damage of the concrete located where the NSM was implemented occurs. The non linear behavior of the resin. The global behavior of the control beam BT1 (without NSM rods) corresponds to a typical precracked reinforced concrete beam behavior.

Figure 4. Bending moment-deflection curves obtained for the two beams.

Even repaired with the NSM rod, the corroded beam yielding moment is still 20% lower than the one of the control beam due to the steel cross-section reduction resulting from the chloride induced corrosion. Indeed, the carbon cross-section added is not enough for the total compensation of the steel loss due to the corrosion. Contrary to the yielding moment, BCor ultimate moment is slightly higher than the one of the beam BT1. This result shows the efficiency of the NSM technique to repair corroded bended RC members. The ultimate capacity obtained is similar to the one of the control beam even for 40% reduction in steel cross-section due to the corrosion. For BCor, the ultimate deflection measured is significantly lower than the one of the control beam BT1. The NSM CFRP rod increases the bearing capacity but also reduces the steel and concrete strains as its behavior is elastic linear up to failure (no yielding phase). The ultimate concrete strain in compression can not be reached as the shear failure of the concrete layer where the CFRP rod is implemented occurs a lot before, at point C (figure 4). 5

CONCLUSIONS

NSM CFRP rod (6 mm diameter) allows increasing significantly the load bearing capacity of corroded beams. Indeed, the corroded beam with about 40 % steel cross-section reduction, shows the same ultimate capacity as the one of the control beam. A reduction in ductility (ultimate deflection) was observed on the repaired corroded beam in comparison to the control element. The NSM technique seems to be efficient to repair corroded beams. But, the possibility to place the NSM rods depends on the quality and the location of the concrete cover not damaged by the steel corrosion. REFERENCE Al-Mahmoud F. Castel A. François R. & Tourneur C. 2007. Strengthening of RC members with near-surface mounted CFRP composites. CONSEC’07 T2: 1735.

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Effect of damaged concrete cover on the structural performance of CFRP strengthened corroded concrete beams A.H. Al-Saidy, A.S. Al-Harthy & K.S. Al-Jabri Department of Civil and Architectural Engineering, Sultan Qaboos University, Sultanate of Oman

ABSTRACT: Corrosion of reinforcement is a serious problem and is the main cause of concrete structures deterioration costing millions of dollars even though the majority of such structures are at the early age of their expected service life. This paper presents the experimental results of damaged/repaired reinforced concrete beams. The experimental program consisted of reinforced concrete rectangular beam specimens exposed to accelerated corrosion. The corrosion rate was varied from 5% to 7.5% which represents loss in cross sectional area of the steel reinforcement in the tension side. Half of the damaged beams were repaired by bonding Carbon Fiber Reinforced Polymer (CFRP) sheets to the tension side to restore the strength loss due to corrosion. The other half of the beams were first cleaned from the contaminated concrete cover and a thorough cleaning of the rusted bars was done. A new layer of concrete was cast to replace the removed contaminated concrete. Then the CFRP sheets were attached to the new concrete layer. Corroded beams showed lower stiffness and strength than control (uncorroded) beams. Strength of damaged beams due to corrosion was restored to the undamaged state when strengthened with CFRP sheets for all strengthened beams. However, the beams with replaced concrete layer exhibited better performance in the load carrying capacity whenever bond was not the mode of failure.

Every year building owners and managers are faced with the costs of repairing concrete that spalls when the reinforcing steel corrodes, usually due to the presence of salt. Removal, patching and the application of waterproofing membranes are some of the treatments that, alone or in combination, have traditionally been used to rehabilitate corrosion-damaged concrete. Steel corrosion is a major cause of deterioration which disrupts the cover zone of reinforced concrete. As steel corrodes, there is a corresponding loss in cross-sectional area and in turn reduction in the flexural strength capacity. Repair or strengthening with fiber reinforced polymers (FRP) has gained some acceptance in recent years. It involves the external bonding of FRP sheets or plates to RC beams and slabs, or confinement of RC columns. Numerous studies have shown that repair and strengthening of corrosion damaged RC beams with FRP sheets or plates is efficient in restoring the strength of concrete members. This paper presents the experimental results of damaged/repaired reinforced concrete beams. The experimental program consisted of reinforced concrete rectangular beam specimens exposed to accelerated corrosion. The corrosion rate was varied from 5% to 7.5% which represents loss in cross sectional area of the steel reinforcement in the tension side. Half of the damaged beams were repaired by bonding Carbon Fiber Reinforced Polymer (CFRP) sheets to the tension side to restore the strength loss due to corrosion. The other half of the beams were first cleaned

from the contaminated concrete cover and a thorough cleaning of the rusted bars was done. A new layer of concrete was cast to replace the removed contaminated concrete. Then the CFRP sheets were attached to the new concrete layer. A total of 7 reinforced concrete beams were tested in this study as summarized in Table 1. Beam C(0%) was a control beam with no corrosion, while beams C (5%) and C (7.5%) were control beams with 5% and 7.5% corrosion (mass loss in reinforcement). Beams S (5%) and S (7.5%) were repaired by applying CFRP

Table 1.

Beams description.

Specimen Corrosion level designation (mass loss %) Remark C(0%) C(5%) C(7.5%)

0% 5% 7.5 %

S(5%)

5%

S(7.5%)

7.5 %

RS(5%)

5%

RS(7.5%)

7.5 %

437

Control beam Control 5% corrosion beam Control 7.5% corrosion beam 5% corrosion strengthened with CFRP sheet 7.5% corrosion strengthened with CFRP sheet 5% corrosion patch repaired and strengthened with CFRP sheet 7.5% corrosion patch repaired and strengthened with CFRP sheet

sheet at the bottom of the beam. On the other hand beams RS (5%) and RS (7.5%) were damaged with corrosion of 5% and 7.5%; then repaired by replacing damaged concrete with new layer of concrete and attaching CFRP sheets to the bottom of the beam as shown in Figure 1. The combined load-midspan-deflections for specimens with 5% corrosion are shown in Figure 2. Beam S(5%) strengthened by one layer of CFRP sheet without removing the damaged concrete due to corrosion cracks. This beam was able to sustain load higher than beam C(0%) (Beam with no corrosion) by 3% and higher by 9.4% than beam C(5%) (control beam with 5% corrosion). The yield strength was also higher in the strengthened beam (S5%) than both control beams (C0% & C5%). On the other hand beam RS(5%) performed better than beam S(5%). This beam was able to sustain load

P

150 100

CFRP sheet 150

950

500

950

150

Table 2.

Summary of test results.

Specimen

Cracking Failure Max. Deflec- Mode of load (kN) load (kN) tion (mm) failure

C (0%) C (5%) C (7.5%) S (5%) S (7.5%) RS (5%) RS (7.5%)

4 3 3 6 6 5 4

17 16 14 17.5 20 20 19.5

40 30 25 23 43 40 30

CC CC CC RT/CC DB/RT DB RT

CC = CONCRETE CRUSHING. RT= CFRP RUPTURE. DB= DEBONDING OF CFRP.

higher than beam C(0%) by 17.6% and higher by 25% than C(5%). The yield strength was also higher in the strengthened beam (RS5%) than both control beams (C0% & C5%). More effective load transfer between the CFRP sheet and the strengthened concrete was possible due to the new concrete cover zone. Similar results were also observed in specimens with 7.5% corrosion. All test results are summarized in Table 2 below.

2700 All dimensions in mm

CONCLUSIONS

Figure 1. Test Specimen.

This study presented test results on the structural performance of repaired/strengthened corrosion damaged reinforced concrete beams. Based on the test results it can be concluded that:

Figure 2. Load-deflection curves of beams with 5% corrosion.

• Strengthening of corrosion damaged beams using CFRP sheets is effective and all strengthened beams were able to reach ultimate loads higher than the ultimate of the damaged state. • The corrosion weakend the bond along the corrosion cracks at the interface of concrete and corroded steel rebars as indicated by the strain profile. • Replacing the damaged concrete in the cover zone with new layer of concrete prior to strengthening with FRP is more effective in the load transfer mechanism between the FRP and concrete.

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Role of U-shaped anchorages on performance of RC beams strengthened by CFRP plates A.R. Khan NED University of Engineering & Technology, Karachi, Pakistan

ABSTRACT: Use of Carbon Fiber Reinforced Polymer (CFRP) Plates, as externally bonded reinforcement, is a practically efficient and technically sound method of strengthening and upgrading structurally inadequate or otherwise damaged or deteriorating Reinforced Concrete (RC) members. The ultimate capacity of the strengthened beam is controlled by either compression crushing of concrete, rupture of the composite plate, local failure of concrete at the plate end due to stress concentrations and flexural shear crack-induced debonding of Concrete-CFRP interface. Another factor that could affect the ultimate strength and mode of failure is the shear span. Beams with U-shaped anchorages were tested in four-point bending to determine the ultimate load carrying capacities and modes of failure of the beams. Properly placed U-shaped anchorages at plate cut-off points and along the span were shown to be effective in improving the performance of CFRP strengthened RC beams in flexure and shear by optimizing deformability and strength characteristics.

1

INTRODUCTION

Reinforced concrete members strengthened in bending by bonding of FRP may present several failure modes: failure of material (reinforcing steel, concrete and composite material) or failure of the interface between concrete adhesive or adhesive-FRP. Nevertheless, experience gained from testing confirms that in most cases delamination prevails over the other possible rupture modes. Different delamination failure modes can be classified into two main types: due to high interfacial stresses near plate ends and due to flexural or flexural-shear crack (intermediate crack) away from the plate ends. It has been observed in the beams reinforced by externally bonded plates that the failure mechanism is governed primarily by the bond and anchorage efficiency rather than by the tensile strength of the shear strengthening plate material. U-shaped external anchorage have been used to enhance the shear resistance of external plate bonding but the reported results are not adequate yet to draw rational conclusions. This paper focuses on investigating the role of U-shaped anchorages on ultimate load carrying capacities, failure modes and performance evaluation of normal and strengthened RC beams in predominant flexure and shear loadings by varying shear span-to-depth ratio (a/d).

2

EXPERIMENTAL PROGRAM

Overall six beams, three control and three precracked and strengthened, rectangular in cross-section,

150 mm × 200 mm, 1600 mm long, with two 12 mm dia. bars as tension reinforcement, were tested under static loading to determine the ultimate load carrying capacity and failure modes of the beams. Beams were divided in three pairs. Each pair was tested as simply supported beams under four point bending with a span of 1400 mm and shear spans of 550 mm, 475 mm and 400 mm respectively.

3 3.1

RESULTS AND DISCUSSIONS Ultimate loads and failure modes

Beam CB1 failed at an ultimate load of 68 kN in conventional ductile flexure mode with yielding of the main reinforcing steel. Beam PB1S, strengthened by externally bonded CFRP strips with full depth anchors at the ends and at midspan, failed in desirable flexure mode by yielding of steel followed by crushing of concrete and carried an additional 33 kN prior to failure (32% increase in load carrying capacity). Beam CB2 failed at an ultimate load of 95 kN in mixed flexure and shear mode. Beam PB2S, strengthened by externally bonded CFRP strips with full depth anchors at the end and at midspan, carried an additional 33 kN prior to failure (35% increase in load carrying capacity). It failed in mixed flexure and shear mode but new flexural cracks were observed in pure flexural zone when the load exceeded 95 kN. Beam CB3 failed at an ultimate load of 92.3 kN in pure shear mode with a prominent diagonal tension crack starting from the support and moving towards the load. Beam PB3S, strengthened by externally

439

bonded CFRP strips with full depth anchors at the ends and at midspan, carried an additional 24.7 kN prior to failure (27% increase in load carrying capacity). It failed in mixed flexure and shear mode. New flexural cracks were observed in pure flexural zone when the load exceeded 92.3 kN but the failure was due to excessive shear cracking in the shear span that led to the failure of concrete at end anchorages. 3.2

PF = DF × SF

Theoretical and experimental ultimate loads

The theoretical value of both shear and flexural strength of the control beams were determined according to the classical RC theory. Flexural strengths of all beams were lower than their shear strengths and all of them were expected to fail in flexure. The mode of failure varied from flexure to shear with the variation in shear span with the beams with largest shear span (CB1 and PB1S) failing in flexure and one with smallest shear span (CB3 and PB3S) failing in shear. A critical evaluation of failure loads, failure modes and effectiveness of u-shaped anchorages at ends and along the span shows that all strengthened beams were able to develop their full potential flexure strength. 3.4

Performance evaluation

For the best understanding of structural performance, the strengthened structural system must be considered for both strength and deformability. In fact, both these two factors need to be optimized when selecting materials, design parameters, and detailing. Deformability factor (DF) and Strength factor (SF) can then be defined as follows: DF =

Deflection at ultimate limit state Deflection at ε c = 1000 µ m/m

Load at ultimate limit state Load at ε c = 1000 µ m/m

(2)

The overall structural performance of the strengthened composite beam can thus be evaluated by a global factor defined as Performance factor (PF), integrating strength and deformability, both weighted equally. This Performance Factor is defined as:

Load-deflection curves

Load-deflection curves for all the strengthened beams are stiff as compared to the respective control beams. All the precracked strengthened beams carried additional load as compared to the respective control beams with the increase in load carrying capacity varying from 27% to 35%. The U-shaped anchorages provided at ends and at midspan not only contributed towards increasing the ductility but they also increased the shear capacity of the section by adding an additional shear component due to end anchorages to the shear resistance provided by concrete and shear reinforcement. They also prevented the premature failure of the beams that would have taken place due to the debonding of the CFRP strips thereby improving the performance of the strengthened beams. 3.3

SF =

(3)

Comparison of performance factors for all strengthened and unstrengthened beams shows that well-designed external anchorage system can significantly improve performance characteristics, both in terms of strength and deformability.

4

CONCLUSIONS

The main conclusions drawn from this study are summarized as follows: 1. Ultimate load carrying capacities of the strengthened beams increased by as much as 35 % over respective control beam. 2. Observed mode of failure varied from flexure to flexure-shear and pure shear in the case of control beams depending on the shear span while it varied from flexure to flexure-shear in the case of precracked-strengthened beams. 3. U-shaped anchorages provided at ends and at midspan improved the structural performance of the RC beams strengthened with externally bonded CFRP strips through enhanced strength and greater ductility as can be seen in the case of all the strengthened beams. 4. Properly placed U-shaped anchorages allowed the concrete in the compression zone to reach its ultimate strain capacity leading to crushing of concrete in compression in the case of beam PB1S, while in other beams they increased the shear capacity of the beams thereby transforming the brittle mode of failure to ductile mode. 5. The U-shaped anchorage system applied to the strengthened beams enabled the bonded CFRP strips to remain attached to the beams through the entire loading phase up to failure thereby preserving the structural integrity of the strengthened beams up to failure.

(1)

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Shear-flexure interaction of RC circular columns strengthened with FRP sheets P. Colajanni & A. Recupero Dipartimento di Ingegneria Civile, Università di Messina, Italy

ABSTRACT: a new physical model for evaluation of N-M-V interaction resistance domain for concrete columns strengthened with FRP sheets is presented. The model provides a simple and efficient method for performance evaluation of strengthening action, and it can be used as a preliminary tool in retrofitting design of reinforced concrete circular columns. In order to point out the flexibility of the proposed method in evaluation of the resistance domain of r.c. elements strengthened with different FRP configurations, and to discuss the influence of N-M-V interaction, the resistance domain for a r. c. circular cross column, strengthened by several different layout arrangements of FRP reinforcements, is presented.

FRP composite systems appear to be a suitable way to increase the strength capacity and the ductility of structures that have been damaged or are structurally unsatisfactory. Use of composite materials is a clear option for external reinforcing because of reduced employment of building manpower, their light weight, high tensile strength and high durability. In the last two decades many papers, addressing this application, have been published. Behavioural aspects have been investigated thorough experimental works; analytical models have been proposed for prediction of normal force-bending moment or shear strength of columns strengthened with FRP sheets. However, the interaction effects between the internal forces have been systematically neglected. Their interaction effects can provide a premature collapse of the element, characterized by a failure mode different than those due to bending-normal force, or pure shear. Nevertheless, standard codes usually do not consider appropriately the N-M-V interaction effects and prescribe design formulas for N-M internal forces, and pure shear strengths. In the paper, a model for N-M-V interaction domain evaluation of r.c. elements having cross-sections of any shapes, already formulated for static loads and based on “stress field” plasticity theory, is extended to the case of elements strengthened with FRP reinforcements. When concrete elements are simultaneously loaded by axial force N, bending moment M, and shear force V, the actual stress-fields distribution in the cross-section is very complex, and an analytical model that may determine their exact distribution cannot be easily derived. Nevertheless, the ultimate strength of the structural element can be evaluated under the following simplifying assumptions: • classical contributions of strength due to dowel action, the aggregate interlock action and concrete traction resistance of the teeth are ignored;

• longitudinal and transverse reinforcements are subjected only to axial forces, including forces due to bending moments; their action is expressed by distributed stress-fields, provided that they are shortly spaced; the stress-fields are assumed uniform, according to the theory of plasticity of steel structural elements; • concrete in the external portion of cross-section is subjected only to normal compressive stress-fields, once again assumed uniform; • stress-fields of central portion has a variable θ degree inclination angle (yield surface) on the longitudinal direction, which may differ from 45 degrees due to actions transmitted along the shear fractures; • failure mode of the structural element occurs for concrete crushing or for yieldinf of reinforcements or both. By these assumptions, the analytical model of the structural element for rectangular, I and T shape cross-section, is a generalization of the truss model that replaces all the components (compressed chord, tension stringer, strut, and tie) by uniform stress-fields. The general criterion adopted consists in dividing the basic structural element cross section in several layers not defined a-priori, subjected to uniform distributions of stresses (normal σ, shear τ, or both), to obtain as a whole the equilibrium with the internal actions N, M, V. This approach, typical for design of in plane loaded plates, has been proved to be equivalent to the stress-fields approach. Any number of subdivisions of the cross section may be assumed, provided that the basic structural element is in equilibrium. In order to obtaining a more general model, able to interpret the behavior of elements with any cross section shape, the cross section is divided into suitable regions and a segment of the element is obtained by suitable section surfaces.

441

For clarity’s sake, here it has been specified for derivation of N-M-V interaction domain for concrete reinforced columns of circular shape cross section (fig.1), with an uniform continuous distribution of longitudinal reinforcements, retrofitted by external reinforcement with a FRP sheet. To derive the equilibrium equations between internal forces and stress fields, the cross section has been divided in three concrete regions (Sc1, Sc2, Sc3) and three region of longitudinal steel reinforcements (Ss1, Ss2, Ss3) A segment of a concrete element of circular cross section (fig.2a) is obtained by a section plane at the abscissa z orthogonal to the element axis (θ = 90º), and a composed section surface at the opposite end, obtained by a upper half plane with slope θ = 90º at the abscissa z+∆z+(D/2−y1)cosθ, and a lower half plane with slope θ = 90º and a upper half plane with slope θ, parallel to the concrete web stress field, at the abscissa z+∆z−(D/2−y2)cosθ. By using the static theorem of the plasticity theory, the so-called “lower bound solution” is obtained by the equilibrium equations, once both geometrical condition and mechanical inequalities for the concrete and steel stress fields are satisfied. In order to point out the flexibility of the method in evaluation of the N, M, V resistance domain for r.c. circular columns strengthened with different configurations, and to discuss the influence of the internal force interaction, the resistance domains of an actual bridge pier with circular cross-section, strengthened by several different layout arrangement of FRP reinforcement, is presented in the paper. The pier has a diameter D= 800 mm and steel reinforcements consisting in 12 φ 24, and a transversal reinforcement composed by stirrup φ 10/300 mm. Cylindrical concrete strength fc0k = 20 MPa and steel resistance fyk = 320 MPa was estimated by experimental tests. The strengthening FRP layout consists in circularwrap strips, having ffud = 3400 MPa, Ef = 228 GPa, tickness tf = 0.165 mm, with two different widths (wf = 50, 75 mm respectively) and the same spacing (sf = 300 mm). For a useful comparison the case of un-strengthened section is analysed (wf = 0 mm).

The proposed procedure can be applied to members of any shape, with transversal reinforcements having any slope, by a suitable choice of the section surfaces for the definition of the member segment for which the equilibrium equations are written. The N-M-V interaction domains stress the reciprocal influence of internal forces, and the strength dependence on the values of the transversal steel web reinforcements and FRP web strengthening. The results shown in the paper, allow us to conclude that the proposed model is able to assess, in an accurate way and with a unified approach, the combined shear, flexural normal force strength of FRP strengthened elements with rectangular, T, I and circular cross section.

q

1

M*

1

N* 3

V*

2

a) z

2

Aws

ws

Awf

wf

q

C1 M*

F1

N*

cw

K

V*

F3

b)

F2 C2

Figure 2. Concrete element sectioned parallelly to the web concrete stress field and to the stirrups direction.

wf=75mm

D

wf=50mm

wf=0mm

350

Dc

300

Rc Awf

Awf

V[kN]

250

100 50

2

Aws

ws

150

wf

1

wf

200

Aws

0

ws

0

200

400

600

800

1000

M[kNm]

Figure 1.

Layers of circular shaped cross section.

Figure 3. Resistance domains for n = 4 N/(π D2 fccd1)= 0.50.

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Strengthening of R/C existing columns with high performance fiber reinforced concrete jacket A. Meda University of Bergamo, Italy

G.A. Plizzari University of Brescia, Italy

Z. Rinaldi University of Rome “Tor Vergata”, Italy

G. Martinola Concretum, Zurich, Switzerland

ABSTRACT: The possibility of strengthening existing R/C columns with a new technique based on the application of a High Performance Fiber Reinforced Concrete (HPFRC) jacket has been investigated herein. Previous researches, carried out by the Authors, shown the effectiveness of the proposed solution applied to beam elements under flexure. In the present paper, the effect of High Performance Fiber Reinforced Concrete jacketing on columns subjected to both axial load and bending moment is analyzed by analytically drawing interaction envelopes. The results will be compared with the response of columns reinforced with a traditional R/C jacket.

INTRODUCTION

Strengthening of existing R/C structures is becoming a major issue in civil engineering. As an example, this may arise from the necessity of retrofitting existing building to meet safety requirements in seismic areas where older constructions were not designed for earthquake actions. Other typical applications of strengthening techniques can be found where the bearing capacity has to be increased because of the higher vertical loads. 2

ULTIMATE BEHAVIOR OF EXISTING R/C COLUMNS

Columns in buildings are generally subjected to both horizontal and vertical forces and, as a consequence, they are loaded by axial forces (N) and bending moments (M). The ultimate behavior of these elements will be studied by considering M–N envelopes that represent the combination of internal actions leading the section to failure. An existing column having a square section of 300 × 300 mm, reinforced with 8 ∅ 16 mm bars (Fig. 1) is considered. The actual compressive strength of concrete is equal to 15 MPa, while the yielding strength of the steel rebars is equal to 400 MPa. This is a typical situation in existing constructions built in the ’60 and ’70. The material safety factors, introduced by

the codes, are considered equal to 1 in order to simulate the actual behavior of the existing element. Figure 4 shows the M-N envelope of the analyzed section. An innovative High Performance Fiber Reinforced Concrete is adopted for the strengthening of the analyzed column. In particular, a jacket of HPFRC having a thickness of 30 mm is considered (Fig. 2). The composite material, already adopted for the strengthening of R/C beams, is characterized by a hardening behavior in tension and by a high strength in compression. The results from uni-axial tests on the adopted HPFRC, highlight a value of tensile strength fct of about 11 MPa. Compression tests on cubic elements provided a maximum compression stress higher than 170 MPa. According to the experimental outcomes, a perfect bond between the existing section and the new

3φ16 300 mm

1

300 mm Figure 1.

443

st. φ8/30''

3φ16

Geometry of the reference column.

3

500 M [kNm] 400 HPFRC

300 200 un-reinf

100 0

-1000 0 -100

N[kN] 1000 2000 3000 4000 5000 6000 7000 8000

-200 -300 -400 -500

Figure 4. M–N envelope for the reference section reinforced with the HPFRC jacket. 3φ16 3φ16

300 mm

60 mm

jacket is assumed. In order to ensure a perfect bond between the old concrete and the new high performance material, a sandblast of the surface of the existing element was carried out, providing a roughness of about 1–2 mm. It is worth noting that no resin or primer interposition is needed. The tensile stresses in HPFRC are considered constant along the tensile zone; the classical stress-block assumption in compression is done (Fig. 3). Figure 4 shows the M–N envelope for the strengthened section, compared with the un-reinforced one. The effectiveness of the proposed technique is evaluated by comparing it with the results obtained with of a classic R/C jacketing. The adoption of a R/C jacket, probably the most used technique for the strengthening of column, has some inconveniences. In particular the jacket thickness is governed by the steel covers (both external and internal). This often leads to thickness higher than 60–100 mm, with a consequent increase of section geometry, and mainly increase of mass, that can give some problems for the seismic behavior. As an example, the reference section of Figure 1 is strengthened with a 60 mm thick concrete jacket, reinforced with 8∅16 bars (Fig. 5). The concrete used for the jacket can be considered of class C30/37, while the steel is characterized by a yielding stress of 500 MPa. The M–N envelope obtained for the reinforced section is drawn in Figure 6. Due to the presence of steel rebars in the jacket, the maximum tensile force is about 3.5 times higher, respect to the reference section.

300 mm

3φ16

Figure 5. Geometry of the section reinforced with the R/C jacket. 500 M [kNm] 400 300

HPFRC R/C

200 100 N[kN] un-reinf 0 -1500 -500 500 1500 2500 3500 4500 5500 6500 7500 8500 -100

CONCLUDING REMARKS

-200

The use of a High Performance Fiber Reinforced Concrete jacket for strengthening existing R/C

-300 -400 -500

Figure 6. M–N envelope for the reference section reinforced with the R/C jacket.

300 mm

30 mm

3φ16

3φ16

300 mm Geometry of the reinforced section.

300mm

300mm

3 φ 16

Figure 3. patterns.

30mm

εc εst εtHPFRC

concrete HPFRC εcHPFRC fc fc εsc fy 0.8x

Figure 2.

fy

fct

Reinforced section: strain and stress design

columns has been analyzed. It was shown that a 30 mm thick jacket allows a significant increase of the bearing capacity both under flexure and axial force. The proposed technique was compared with two other techniques such as the traditional R/C jacket (that requires higher thickness of the jacket) and the externally bonded FRP that is effective for the flexural strengthening only. The HPFRC jacketing results of great efficacy particularly for the axial force strengthening. This solution requires jacket with a very small thickness. Due to the very good surface quality that can be obtained with the HPFRC material (almost a self compacting material) the jacket can substitute the plaster layer with no significant change in the column size. Beside the strength increase, the HPFRC allows a significant enhancement of the durability of the structural element.

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Retrofitting of a precast industrial building M. di Prisco, M. Lamperti & S. Lapolla Department of Structural Engineering, Politecnico di Milano, Milano, Italy

ABSTRACT: Four years ago, a precast industrial building collapsed due to the failure of a pre-stressed beam. A wide experimental investigation was carried out on site to judge the remaining beam conditions; a full-size TT beam, produced according to the same design specifications, was tested up to the maximum load that caused the disaster, retrofitted and tested up to the element failure to clarify the collapse motivations. Another TT beam was extracted from the roof of the industrial building, suitably retrofitted and also tested up to collapse. Even the thin-webbed Ω roof elements supported by the beams were investigated due to an excessive deflection produced by the only self-weight. Three of them were also extracted from the roof and tested up to collapse. They were retrofitted by using CFRP strips and suitable ties. The paper is focused on the discussion of the problem causes and their solution by using experimental and numerical evidences. In north Italy, during a relatively moderate snow fall (about 40 kg/m2) a prefabricated roof of a new industrial building collapsed on February, Friday 20th 2004 after only few months from its erection.

The collapse of a TT prestressed beam characterized by a span of 23.2 m was due to the failure of the support regions. The sixteen cross-wise pre-stressed Ω roof elements, 10.15 m long, resting on the TT beams, fell down dragging the eight roof lights to the ground. The failure occurred in the night; after few hours other two beams and the connected roof elements fell down. At the end, about 1300 m2 of precast roof were collapsed and the observable scenario was really impressive (Fig.1): no victims were involved, but the fire brigade sequestrated the site. The geometry of the structural mesh (12.25 m × 23.2 m; Fig.2) consists in two TT beams supported on four 10 m high columns resting on pocket foundations. The cross wise Ω roof elements distribution considers the presence of four skylight elements in the TT element span. The external wings of the TT element are 10 cm thick, while the horizontal top slab is only 8 cm

thick: this can be regarded as the main cause of the disaster. A beam model based on a plane section approach is able to predict the non linear mechanical behaviour induced by the transversal deformability of the TT element, with and without retrofitting: a good fitting of the relative displacement between the two bottom chords guaranteed it. The failure occurred in situ can be justified by taking into account creep

roof light

A

A TT beam element

section A-A

GG

q

GG

a Figure 1.

b

Situation after structure breakdown.

Figure 2.

445

Roofing system.

GG

effects on the redistribution of transversal bending moment, which progressively induced an increase of torque acting in the web and in particular on the bottom chords which were already seriously damaged by prestressing diffusion phenomena due to the lack of ducts. No specific reinforcement was introduced to absorb this action! It was retrofitted by means of steel plates, 0.8 mm thick, embracing the webs in the support region for a length of 1.5 m fixed to the webs by means of 12 Φ14 (8.8) steel bolts and an epoxy resin. Two steel struts were also located at the first and the second third of the whole span. After one week a maximum load corresponding to 1.82 times the total load acting at SLS was imposed: at failure two longitudinal cracks propagated from one end at the connection between the wing and the web along the wing intrados. A second test was carried out in situ on a beam directly extracted from the same industrial building that failed, choosing a region close to the collapsed one. It was retrofitted by adopting the same technique previously described and the load applied reached 1.70 times the total load acting on the beam at SLS. The failure was caused by the collapse of the wings as in the previous case. The roof elements are characterized by a 10 m long span and a thin walled open cross section. The top horizontal slab is only 6 cm thick and is reinforced at the intrados with a wire fabric and 5 bars concentrated at the ends (1Φ6/20 + 5Φ8). The longitudinal reinforcement consists in 2+2 0.6” strands and 1+1 Φ10 4 m long bars, placed in the middle. They are simply supported at the ends and the acting loads are the dead weight (GΩ) and the snow loads. The precast covering structure uses TT beams which are longitudinally over-reinforced, but are transversally deformable. The two webs are interconnected by a very thin slab, only 8 cm thick, where the reinforcement location cannot be regarded as univocally determined. Even with a gross cover of 20 mm the transversal bending moment acting at the Serviceability Limit State greatly exceeds the cracking value: the bending moment caused by the only dead load of the whole structure (M = 8.09 kNm) is about the double of the one that causes first cracking. A drastic simplification that reduces the shear acting in the two webs to simply support reactions (Fig.3) allows the designer to neglect the redistribution that induces a torque in the webs and in the wings due to transversal deformability of the cross section and computes the transversal bending moment by means of pure equilibrium. Transversal deformability problems interested also the Ω roof elements placed on TT beams. A significant crack pattern mixed to a not negligible deflection suggested a check of the safety factors. First of all in

Mu

Mcr A AB

Mcr = 5.8 kNm AB

Mu = 33.6 kNm

Mu

M ULS

Mu Mcr

M SLS

Mcr C

B BC

Mcr = 3.6 kNm BC Mu = 17.3 kNm

D CD

Mcr = 5.8 kNm CD

Mu = 33.6 kNm

SLS =16.4 kNm M MAX ULS =23.7 kNm M MAX

Figure 3. Transversal equilibrium bending moments (characteristic values).

situ tests were carried out at SLS. The tests suggested to extract from the roof the already two tested elements: the first cracked without any retrofitting and the second cracked, retrofitted by means of FRP strips placed at its ends. Also a third element, less cracked than the others, was extracted and tested. The interest was aimed to understand the influence of the crack pattern exhibited by the elements on their ultimate behaviour. Fifty cycles between 0 and 30 kN were first applied to the elements to reproduce the serviceability conditions in rare combination for a certain expected life (20 years). After this, the tests were displacement controlled up to failure. The bending response of the three elements showed in situ or in the laboratory an elastic behaviour very close to the predicted one. The cyclic loads after imposed in the laboratory (span length = 9.6m) changed drastically the stiffness especially for the most damaged element, even if it was retrofitted with FRP strips. Only the less damaged element reached a peak in the bending moment larger than the design ultimate moment. This experimental evidence highlights the key role of transversal cracking on the ultimate behaviour of such structural elements. Finally to all the structures, a steel tie was added to the ends, after having sawn the end failed. Transversal bending affected the ultimate behaviour of Ω elements causing the collapse of the inclined webs due to shear and transversal bending for a load smaller than that indicated by the Codes in relation to Ultimate Limit State. CFRP strips could scantly improve the mechanical behaviour of Ω elements, while steel ties located at the element ends allow the designer to satisfy the ultimate checks.

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Strength and ductility of unreinforced concrete columns confined with CFRP composites under uniaxial loading P. Sadeghian Islamic Azad University of Qazvin, Qazvin, Iran

A.R. Rahai Amirkabir University of Technology, Tehran, Iran

M.R. Ehsani University of Arizona, Tucson, AZ, USA

In recent years, the use of fiber reinforced polymers (FRP) as an externally wrapping have achieved considerable popularity for the strengthening and repair of concrete structures. The FRP composites have been used successfully for rehabilitation and strengthening of deficient reinforced concrete elements. The potential market for such applications is huge since the estimated annual cost of repairing bridges in the United States alone is 9.4 billion dollars. [1] One popular technique of FRP strengthening is the wrapping of reinforced concrete columns to increase their axial strength, shear strength, and seismic resistance. In this application, the FRP sheets are generally wrapped around the columns with fibers oriented mainly in the circumferential (hoop) direction. The fibers confine the concrete and increase the axial strength by creating a triaxial stress condition. The FRP wraps also increase the shear resistance of columns and prevent premature failures when columns are subjected to lateral loadings typical of those observed during earthquakes. [2] Reinforced concrete columns need to be laterally confined in order to ensure large deformation under applied loads before failure and to provide an adequate bearing capacity. In the case of a seismic event, energy dissipation allowed by a well-confined concrete core can often save lives. On the contrary, a poorly confined concrete column behaves in a brittle manner leading to sudden and catastrophic failures. [3] When the FRP-wrapped concrete is subjected to an axial compression loading, the concrete core expands laterally. This expansion is resisted by the FRP wrap, and therefore the concrete core is changed to a three dimensional compressive stress state. In this state, performance of the concrete core is significantly influenced by the confinement pressure [4]. Several parameters influence the confinement effectiveness of the FRP wrap, which include concrete strength, wrap thickness or number of FRP layers, and wrap angle orientation [5]. Many investigations have been conducted into the behavior of FRP-wrapped concrete.

For a column subjected to a uniaxial compressive load, it is well established that fibers should be aligned along the hoop direction to confine the dilation of the concrete core. In practice, however, almost all the columns are subjected to an eccentric axial load, which can be resolved into a uniaxial compressive load and a bending moment. Because of this, almost all the columns should be treated as beamcolumns. For beam-column, hoop direction is the optimal fiber orientation only for uniaxial compressive load; for a bending moment, fibers in the axial direction are more favorable. Therefore, fiber orientation is an important variable in the structural design of FRP-wrapped concrete columns. [6] In the present paper, focuses are on applications of carbon FRP (CFRP) wraps with various fiber orientations and wrap thickness to increase the axial strength and ductility of unreinforced circular concrete columns. In this study a total of 23 CFRP-wrapped and 7 unconfined control concrete cylinders with a diameter of 150 mm and a height of 300 mm were prepared and tested under uniaxial compression loading. The main experimental parameters were included wrap thickness and fiber orientation. Four different wrap thicknesses 0.9, 1.8, 2.7, and 3.6 mm (1, 2, 3, and 4 layers) and four fiber orientation of 0º, 90º, +45º, and −45º with respect to the hoop direction were investigated. Target compression strength of plain concrete at 28 days was 30 MPa. However, actual compression strength of plain specimens at the test day were measured 35 MPa up to 45 MPa. A unidirectional carbon fiber sheet was used to prepare the CFRP wrap. Mechanical properties of CFRP wraps have been recorded through tension tests on CFRP coupons by Sadeghian et al. [12] The concrete was produced with a similar mixture design before casting in standard cylindrical formworks. After curing in humid room for 28 days, the surface of specimens were cleaned and prepared for wrapping. In order to prevent anchorage rupture in

447

CFRP-wrap, a lap splice equal to 75 mm was used in fiber direction. The carbon fibers were configured in predefined orientations as shown in Table 1, and then were impregnated with the epoxy resin. Epoxy resin should be cured in laboratory temperature for a minimum of seven days. In order to perform uniaxial compression testing on specimens, a hydraulic testing machine was used in the Amirkabir University of Technology (Strength of Material Laboratory). Bottom jaw of the machine is adjustable in the vertical direction and it is attached to an actuator, while the top jaw is fixed. The specimens were tested using a 3,000 KN capacity compression machine and the data were monitored using an automatic data acquisition system. The tests were continued up to failure under a monotonically increasing concentric load in a displacement control mode. The force and displacement data were obtained by data collecting system of the machine during the test and were stored for future reduction and analysis. The performance of the cylinders under axial load was consistent. At the early stages of loading of the confined specimens, the noise related to the micro cracking of concrete core was evident, indicating the start of stress transfer from the dilated concrete to the CFRP wrap. Prior to the failure, cracking noises were frequently heard. The failure was gradual, ending with a sudden and explosive noise. The failure of the wrap initiated away from the overlap region at mid-height of the specimen and progressed to the top and bottom of the specimen. The sudden and explosive nature of the failure indicates the release of extraordinary amount of energy as a result of the uniform confining stress provided by the wrap. Inspection of the broken samples showed good contact between the wrap and the concrete indicating that no debonding took place at any stage throughout the loading process. By wrapping the concrete with an external continuous CFRP wrap, the fibers in the hoop direction resist the transverse expansion of the concrete providing a confining pressure. At low levels of longitudinal stress; however, the transverse strains are so low that the FRP jacket induces little confinement, if any. At higher longitudinal stress levels, the dramatic increase in transverse tensile strains activates the CFRP wrap and the confining pressure becomes more significant. The general confining pressure induces a triaxial state of stress in the concrete. It is well understood that concrete under triaxial compressive stress exhibits superior behavior, in both strength and ductility, as compared to concrete in uniaxial compression. In a circular concrete column, the confining pressure is constant around the circumference, provided

small variations due to factors such as the inhomogeneity of concrete are ignored. When the CFRP wrap ruptures, this confining pressure reaches its maximum value given by fr =

2tf j D

,

(1)

where fr is confining pressure, t is wrap thickness, fj is ultimate tensile strength of wrap in hoop direction, and D is diameter of concrete core. In this paper two global forms for ultimate stress and strain of CFRPconfined concrete are used and constant factors were calculated by a regression analysis. The proposed models for circular section and transverse wraps are presented in the following equations. 0.7

⎛ f ⎞ f cc′ = 1 + 5.18 ⎜ r ⎟ , f co′ ⎝ f co′ ⎠ ⎛ f ⎞ ε cc = ε co + 0.039 ⎜ r ⎟ ⎝ f co′ ⎠

(2)

0.92

(3)

where f cc′ is confined concrete strength, f co′ is plain concrete strength, εcc is confined concrete strain at f cc′ , εco plain concrete strain at f co′ . Figure 9 shows the comparison between experimental data and proposed model for normalized ultimate stress. As the conclusion, the enhancement of strength and ductility of CFRP confined concrete is significant. The observed stress-strain response of CFRP confined concrete can be divided in three distinct zones. The first zone is approximately linear and the second zone is nonlinear as a transition zone. The third zone depends on the wrap behavior. If the wrap has a linear behavior, slope of the third zone will be constant. If the wrap has a nonlinear behavior, the third zone will has a decreasing slope. The specimens wrapped with hoop orientation have a bilinear behavior with a nonlinear transition zone. When the angle of fibers is changed to ±45°, the behavior is accompanied by a larger flat region as a plastic deformation prior to failure. This behavior can be very useful in cyclic loading and hysteretic damping against seismic loading. The combination of hoop and angle orientation is not useful and the angle layers can't produce plastic deformations. At the end, an analytical model for ultimate stress and strain of confined concrete has been proposed.

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Experimental investigation on repair of RC pavements with SFRC G. Boscato & S. Russo University Iuav of Venice, Italy

ABSTRACT: This research shows the first results of experimental tests on the cracking performance of RC slabs reinforce with SFRC, employed as industrial pavings. The aim of this study is to evaluate the influence of SFRC repairs of different thicknesses on the mechanical performance of RC slabs, especially with respect to the crack pattern and level of cracking load. To better understand the influence of SFRC, in terms of performance and variation of cracking load after repairing, a comparison with a reinforced concrete slab without fiber reinforcement was made. The study shows also the mechanical characterization of SFRC through conventional testing, to evaluate compressive strength, fracture energy, tensile strength and toughness. Concerning the application of SFRC on the concrete slab surface, the bond was improved by removing a small amount of superficial material . Finally, the experimental results on cracks distribution, displacements and level of cracking load are shown.

TEST SETUP

Table 1.

The ordinary RC slabs are 300x300 cm wide and 15 cm thick (Figure 1). The concrete belongs to 25 MPa strength class; the traditional reinforcement consists of two layers of welded wire of 8 mm diameter and 200x200 mm mesh. Three different layers of SFRC reinforcement were taken into account, respectively 1, 2 and 4 cm thick (Table 1); the percentage of fibers versus the concrete volume is Vf = 5%. The load was applied in the centre by means of a jack, with interposed distributing panel and a Teflon layer to better ensure adherence to the slab surface (Figure 2). Figure 2 shows the test setup scheme with the general sizes of slab, the position of transducers that

Slab names

Thickness of SFRC layer (mm)

RC SFRC1 SFRC2a, b, c SFRC4

/ 10 20 40

Figure 1.

Figure 2.

Setup of experimental test.

449

Characteristics of slabs.

Scheme of test setup.

obtain the vertical displacements (Vertical Displacements and position respect to the edge VD0, VD750 and VD1500) and the supports for the elastic soil. Under the slab-pavement, several point supports were positioned in order to simulate a Winkler elastic soil (Figure 2), with only 16 supports to simulate a particularly soft subfloor or a generic condition of damage in the pavement.

EXPERIMENTAL RESULTS Figure 4.

Load-vertical displacement comparison.

Mechanical characteristics of SFRC material The tests aimed to the mechanical characterization of concrete-matrix fiber reinforced material were performed in according to Italian Standard CNR-DT 204/2006, UNI 1139, 2003. Table 2 reports the mean values of mechanical parameters of SFRC. Table 2.

Material characterization of SFRC.

density

γ

2153 kg/m3

compression stress

σc

87.5 MPa

tensile stress

σt

5.3 MPa

fracture energy

Figure 3.

Cracking RC pavements.

32500 N/m

Cracking and mechanical behaviour of pavements The structural response of the slab element is enlightened by the crack pattern as represented in drawing of the cracks’ distribution of Figure 3. Figure 4 displays the relationship between load and vertical displacement of each of the tested slabs.

FINAL CONSIDERATIONS The following considerations can be made on the basis of the first experimental results: The crack pattern of slabs reinforced with SFRC enlightens the excellent adherence between the materials; in fact, no cracks due to tensile stress on the upper face were detected. For little thickness of SFRC material is evident a structural behaviour of composite two-dimensional structural element similar to RC slab but, anyhow, with an important increment of strength both with the first crack and collapse load. The symmetrical behaviour of the reinforced slabs with respect to the two median axes is evident, and on the whole it shows a quite uniform distribution of the load through the slab thickness and the different layers. The cracks on the intrados of the slabs indicate the different behaviour due to the various thicknesses of the SFRC reinforcement. The slabs with a thicker SFRC layer show a much more diffused crack pattern, this mainly due to a better distribution of the point load between the two collaborating materials. Such a structural response predicts the good service state behaviour of the element. With respect to the RC slab taken as reference, an increment of the first crack load is attained of 10%, 57% and 71%, corresponding respectively to 1 cm, 2 cm and 4 cm of SFRC reinforcement thickness.

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Experimental research of performance and strength of damage – Concrete beams strengthening with CFRP using two different epoxies A.R. Rahai, M.R. Saberi & R. Mahmoudzadeh Department of Civil Engineering, Amirkabir University of Technology, Tehran, Iran

ABSTRACT: Application of FRP material is an effective method for strengthening of deficient elements and to improve their performance. In this paper several beam specimens were strengthened with CFRP plate using two different resins. First some of the specimens are loaded and then they strengthened with the composite layers. In the next stage, the specimens are reloaded And their load-deformation curve are prepared. The performance of these specimens is compared with non-damaged specimens which are strengthened primarily and made under loading conditions. The results of two series of specimens are compared which show the appropriate effectiveness of CFRP strengthening.

1 1.1

EXPERIMENTAL INVESTIGATION Specimen design

Ten beams were tested. Two of these beams were tested in its virgin condition to serve as reference, while the remaining eight beams were tested after being strengthened using carbon fiber. All of the RC beams were designed to have the same nominal dimensions: 2000 mm long, 200 mm wide and 300 mm deep, with a span of 1800 mm. As shown in Figure 1, the flexural reinforcement consisted of 2T12 deformed bars. The shear reinforcement consisted of 1T8 distanced of 50 mm. Repair of specimens was achieved by the external strip of unidirectional carbon fiber sheets.

Figure 1.

were wrapped round the beam at 90 angle. Electrothermal blanket was used in some cases to maintain the curing temperature above 25°C Wu et al (2006). 1.3

1.2

Specimen preparation

All eight RC beams were cast with a normal density concrete mix and the casting followed normal practices After the RC beams had been cured for about 4 weeks the Repair of specimens was achieved by the external strip of unidirectional carbon fiber sheets. Debonding in a FRP bonded concrete system is a complex phenomenon for solving this problem it used at the bottom of the concrete core, CFRP strip with high strength were axially bonded to carry tensile load and assure the stiffness of the member. Outside the CFRP strip, CFRP strip is hoop directionally wrapped to provide confinement to the whole concrete core and also to prevent the premature debonding of CFRP strip (Fig. 2). Certain layers of carbon fiber sheets were bonded a long the axial direction (at 0 angle) at the bottom surface of the concrete core by two different resins and then carbon fiber sheets

Detail of RC beam bonded with CFRP.

Material properties

The concrete was prepared in the laboratory. The cube compressive strength of concrete was 40 MPa. The flexural reinforcement consisted of 2T12 deformed bars of yield strength 485 MPa. The shear reinforcement consisted of 1T8 distanced of 50 mm deformed bars of yield strength 430 MPa. The tensile properties of the CFRP materials are listed in Table 1, which were either supplied by the manufacturer or determined according to ASTM D3039/D3039M95a (1995) using 25 mm wide flat test coupons and calculated based on the nominal thickness of the corresponding fiber or fabric sheet and the actual widths of the test coupons. Beams B3 and B4 were strengthened with CFRP strip and epoxy resin these beams were strengthened by 2000 mm long and 200 mm wide CFRP strip and those strips that wrapped to provide confinement to the whole concrete core and also to prevent the premature debonding of interior CFRP strip has 300 mm wide (Fig. 4).

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Beams B5 and B6 were strengthened using CFRP and polyester resin. In these beams isolated CFRP strips were similar to beams B3 and B4. The beam models B7 to B10 were loaded before strengthening with the maximum crack width was reached to 0.4 mm (Fig. 5), next the damaged beams were strengthened with carbon strips then the beam specimens B7 and B8 were strengthened with CFRP strips and epoxy resin similar to B3 and B4. The beam models B9 and B10 were strengthened similar to B5 and B6. For specimens B7 to B10 moment causing flexural cracking at section due to externally applied loads was 42.3 KN.mm and its load was 70.5 KN according to ACI 318-005 (2005). 1.4

Instrumentation and test procedure

Tests were a four-point bending. The beams were simply supported over a span L of 1800 mm and tested in flexure under two symmetrical point loads, thus giving a L/h ratio of 6 and an a/h ratio of 2, where a is the bending span Islam et al (2005). The load was applied using a servo controlled hydraulic actuator with a maximum capacity of 1000 KN. According to ASTMC78-00 (2001) load the specimen was continuously and with out shock and applied at a constant rate to the breaking point. Apply the load at a rate that constantly increased the extreme fiber stress between 0.86 and 1.21 MPa/min. The All beams were instrumented by electrical resistance strain Gauges for measuring strains in the main tensile at mid-span and strains in the strengthening materials at different critical locations Mid-span deflection of the beams. The load was applied monotonically up to failure. One linear variable displacement transducer (LVDT) was placed under the mid-span and one LVDT was similarly placed under the beam at the load position. Fig. 5 gives the positions of LVDTs. Two strain gauges of 60 mm in gauge length placed under and above the mid-span and four strain gauges placed under and above the quarter-span and symmetrically located on the two sides of the centre line of the beam Yao & Teng (2007).

2

TEST RESULT AND DISCUSSION

All beams demonstrated a nearly linear response up to about 125% of the yield load. However, as expected, strengthening by external bonding of different CFRP

Table 1. Test results.

Beam

First crack Load (KN)

Yield Load (KN)

Increase in yielding load after strengthening (%)

B1, B2 B3, B4 B5, B6 B7, B8 B9, B10

33 53 48 35 35

61 76 74 70 68

– 24.6 21.3 14.8 11.5

systems resulted in an increase in stiffness, the highest increase being exhibited by beams B3 and B4 that were containing CFRP and epoxy (ML-506) resin. The loads corresponding to the first appearance of flexural are presented in Table 1. The first crack appeared at about the mid-span of under beams next propagated to top of beams. All the beams failed in bending. The parent beams B1, B2 failed by crushing and bending of the concrete. B3, B4 which were strengthened by CFRP and epoxy resin failed at a load 25% higher than the parent beam. Beams B5, B6 also show a 21.3% increase in yielding strength. Specimens B7, B8 failed at a load lower than B3, B4. Their yielding loads were almost half of yielding load of B3, B4 because of they were damaged. B9, B10 show an 11.5% increase in yielding strength. Similar to B7, B8 their yielding load were almost half of yielding load of B5, B6. The failure of all of strengthening beams was initiated due to failure of the bottom strip.

3

CONCLUDING REMARKS

This paper deals with bending strength of the reinforced concrete damaged beams by using an externally strip CFRP system in the beam bottom. Test result of ten beams that retrofitted with CFRP using two different resins in states of before and after damage are presented and discussed. This paper clearly shows that there is no difference between using CFRP with epoxy resin and using CFRP with polyester resin in concrete beams; however, there is a high difference between retrofitting of before and after damage and it is better than our retrofitting the beams before damage.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Strengthening of shear wall with high performance RC jacket A. Marini University of Brescia, Italy

A. Meda University of Bergamo, Italy

ABSTRACT: A new technique for the strengthening of existing R/C shear walls based on the application of thin high performance jackets is presented in this paper. The strengthening jacket is made of high performance concrete, having a compression resistance higher than 150 MPa, and reinforced by means of an high strength steel mesh. The experimental study is carried out on a 1:3 scale R/C wall, proportioned to resist the vertical loads only, and reinforced by means of a 15 mm thick high performance jacket. Cyclic loads of increasing magnitude are applied to the experimental shear wall up to collapse. The experimental results show the efficiency of the proposed strengthening solution. The strengthening jacked allow doubling the structure resistance, and the structure ductility is increased up to nine times the ductility of the unreinforced structure. The remarkable increase in the structure ductility is the result of the large deformability of the jacket and the absence of strain localization.

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INTRODUCTION

A new technique for transforming existing R/C walls into seismic resistant shear walls is presented in this paper. The proposed solution is based on the use of high performance jacket made of high strength steel mesh, having a tensile resistance higher than 1200 MPa, embedded in a thin layer of high performance fiber reinforced concrete, having a compressive strength higher than 150 MPa. By using these high performance materials, the jacket thickness can be significantly reduced (30-40 mm). In this paper, the proposed technique is validated by means of an experimental test on a 1:3 scaled shear wall. The experimental specimen was designed by reference to an existing three storey R/C building, which was proportioned to resist the vertical loads only. The high performance jacket was designed to entirely resist the seismic actions. The experimental results showed the efficiency of the technique in increasing the structure bearing capacity and ductility.

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As for the jacketing, a high strength steel mesh was used. The mesh is made of 2 mm diameter bent wires, assembled with a spacing of 20 mm (Fig. 1). The measured maximum strength was always higher than 1200 MPa. For the reinforcing jacket a high strength fiber concrete with a very compact matrix and with a maximum aggregate size of 2.7 mm was adopted. The resulting high performance fiber concrete exhibits a hardening behavior under tensile forces, and shows a compressive strength, measured on 100 mm cubes, higher than 150 MPa. In order verify the effectiveness of the proposed strengthening technique an experimental test was carried out. As a reference, the typical R/C wall of an existing three storey building was considered. The R/C experimental wall was built in the reduced 1:3 scale. The specimen, reproducing the typical stair block element of an existing building, was designed to resist the vertical loads only. The geometry of the scaled

WALL SPECIMEN AND TEST SET-UP

When performing a scaled test, a correct choice of the materials is necessary for the scaled model to be effectively representative of the full scale structure. For the construction of the R/C wall, the concrete mix design was defined by reference to a scaled aggregate grading curve, by adopting a maximum aggregate size of 15 mm.

Figure 1. High strength steel mesh made of bent wires. Single wires are weaved with each others to mechanically prevent their threading from the mesh.

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50 100

-20

Load [kN]

75 20 50 -50 -150

-100

25 -50

Top displacement [mm] 50

100

150

-25 -50 -75 -100

Figure 3. Applied horizontal load versus top displacement curve.

Figure 2. model.

Front view and cross section of the experimental

specimen is shown in Figure 2. The wall specimen has a height equal to 3.2 m (reproducing a 9.6 m real R/C wall) and a 100 mm × 800 mm cross section. The reinforcement is made of 5 mm diameter longitudinal rebars, having a spacing of 70 mm, and 4 mm diameter stirrups having a spacing of 100 mm. The R/C wall foundation block is anchored to the testing bench. Upon completion of the R/C wall casting and curing, the reinforcing jacket was applied. In order to ensure perfect bond to the high performance jacket, the R/C wall surface was previously sandblasted, thus obtaining a surface roughness of approximately 1–2 mm. Because of technological issues, a 15 mm jacket thickness was selected. This value is the minimum thickness which allows to obtain both a homogeneous casting and a sufficient mesh cover. The selected thickness results in a 45 mm jacket in a real scale application. 3

EXPERIMENTAL TEST

The experimental test was carried out by applying cyclic loads of increasing amplitude up to the structure collapse. Figure 3 shows the horizontal load versus top displacement curve. It can be noticed, that the behavior is almost linear up to 50 kN with a limited

dissipated energy. By increasing the cycle amplitude the dissipated energy increases and the behavior becomes remarkably non linear. The structure yielding is recorded at a top displacement equal to 12 mm (δy); whereas the collapse is reached with a top displacement of 107 mm (δu) and a maximum load equal to 81 kN. The structure collapse is induced by the crushing of the strengthening jacket at the wall base (Fig. 9). The structural ductility (δu/δy) was estimated as equal to 9. 4

CONCLUDING REMARKS

The experimental study allows drawing the following concluding remarks: • the use of a very thin high performance concrete jacket reinforced with high strength steel meshes allows to double the structure ultimate resistance and to largely increase the structure ductility; • the structure failure is characterized by the wall base concrete crushing and by a crack pattern uniformly extending over a critical zone. The depth of the critical zone is approximately equal to 1.5 times the shear wall base. Unlike traditionally reinforced shear walls, no damage localization in a single critical section is observed; • the structure behavior up to collapse was governed by the bending moment, with no appreciable influence of the shear effects; • the proposed technique can be easily used in structural applications, provided that its construction requires neither special works or special man labor. As for the materials, the high strength steel meshes can be fold by the producer prior to its transfer to the construction site and the high performance concrete can be easily pumped.

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Concrete Repair, Rehabilitation and Retrofitting II – Alexander et al (eds) © 2009 Taylor & Francis Group, London, ISBN 978-0-415-46850-3

Author index

Abbasnia, R. 353 Abecasis, D. 327 Adili, E. 433 Ahmadi, J. 353 Ahn, T. 121 Ahn, T.H. 125 Alexander, M.G. 67, 147, 151, 157, 163, 253, 271 Alexandra Bota, 399 Al-Harthy, A.S. 437 Ali Khan Zai, S. 427 Al-Jabri, K.S. 437 Al-Mahmoud, F. 435 Almeraya-Calderón, F. 165 Almusallam, T.H. 405, 411 Alonso, C. 199, 289 Alonso-Guzmán, E. 165 Al-Saidy, A.H. 437 Al-Salloum, Y.A. 405, 411 Alsayed, S.H. 405, 411 Alshebani, M.M. 195 Altmann, F. 117 Ambroise, J. 325 Anderberg, Y. 249 Andrä, H.-P. 49 Andrade, C. 31, 75, 155 Angst, U. 149 Annerel, E. 245, 381 Anton, H. 207, 269 Arici, M. 197 Arifovic, F. 383 Ariza-Aguilar, L. 165 Arntsen, B. 299 Aubagnac, C. 421 Audenaert, K. 107, 109, 159 Aydin, E. 91 Azhari, S. 195 Bachmaier, S. 219 Baias, M. 335 Balagija, A. 241 Balancan-Zapata, M. 165 Ballim, Y. 67 Baltazar, M. 165 Banic, D.I. 243 Banic, Z. 167, 243 Bänziger, H. 305 Barhum, R. 111

Barisic, E. 113 Barros, J.A.O. 387 Baukov, A. 217 Beck, M. 187 Bennitz, A. 419 Benzarti, K. 421 Beushausen, H.D. 67, 157, 163, 209, 369 Bhutta, M.A.R. 329 Bissonnette, B. 275, 345, 347, 351 Bjegovi , D. 89, 133, 169, 269, 317 Blümich, B. 335 Bode, K.A. 97 Boel, V. 107 Bohner, E. 161, 175 Boscato, G. 449 Bota, A. 399 Bouhicha, M. 389 Bouichou, M. 311 Bovassi, M. 321 Breitenbücher, R. 365 Brissaud, D. 311 Brosens, K. 417 Bruedern, A.-E. 327 Bruns, M. 225, 301 Burkert, A. 187 Büttner, T. 333 Cairns, J. 377 Capener, J.C.-M. 319 Carevi , M. 169 Castel, A. 435 Castellote, M. 31, 155 Castro-Borges, P. 165 Chang, T.P. 425 Chataigner, S. 421 Chatenoud, G. 357 Chen, B.T. 425 Chini, M. 299 Christodoulou, C. 297 Colajanni, P. 441 Concu, G. 213 Çopurog˘lu, O. 235 Courard, L. 347, 355, 367 Craeye, B. 101 Curbach, M. 391 Czarnecki, L. 343

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da Silva, V.A. 281, 303 Dalfré, G. 387 Dauberschmidt, Ch. 141 Davison, N. 307 De Belie, N. 131, 291 De Bock, J. 131 De Domenico, V.H. 293 De Muynck, W. 131, 291 De Nicolo, B. 213 de Oliveira, R.A. 193 de Paiva, S.C. 193 de Rooij, M.R. 123 De Schutter, G. 101, 107, 109, 159 de Viedma, P.G. 31 Dehn, F. 135 Dereymaeker, J. 417 di Prisco, M. 445 Di Sarno, L. 409 Dias, J.P. 387 Dimmig-Osburg, A. 97 Dobmann, G. 255 Dobnikar, V. 359 Domsa, J. 105 Drakuli , M. 169 Dreveny, I. 385 Ehsani, M.R. 447 Ekolu, S.O. 239, 267 Eligehausen, R. 407 Emmons, P.H. 275, 315 Evi , E. 207 Fan, K.Q. 261 Felicetti, R. 293 Féron, C. 357 Figeys, W. 417 Fischboeck, E.K. 191 Fischer, C. 183 Flohr, A. 97 Flohrer, C. 295 Fowler, D.W. 349 Francois, R. 435 Friese, M. 215 Gagné, R. 351 Gao, W.Y. 429 Gaona-Tiburcio, C. 165

Garbacz, A. 367 Gard, W.F. 123 Gehlen, C.D. 179, 183, 225 Genesio, G. 407 Ghalehnovi, M. 433 Gharachorlou, A. 431 Glass, G.K. 307 Goebbels, J. 187 Goodwin, F.R. 315 Goyns, A.M. 281 Gräfe, B. 215 Granata, M.F. 197 Grosse, C.U. 219 Gulikers, J.J.W. 153 Gutierrez, J. 397 Gutsch, A.-W. 251 Guzmán, A.M. 403 Hänisch, L. 85 Hao, H. 259, 261 Hariche, L. 389 Harmuth, H. 191 Harnisch, J. 177 Hauser, S. 295 Heiyantuduwa, R. 147 Hela, R. 93 Helene, P.R.L. 127 Herschelmann, F. 251 Hillemeier, B. 85, 339 Ho, A.C. 95 Hoppe, G.E. 303 Hosoda, A. 121 Hu, K.X. 429 Hubertova, M. 93 Humphries, W.S. 279 Ikeno, S. 121 Imamoto, K. 329 Imperatore, S. 203 Ioani, A. 105 Janz, M. 287 Jelcˇi , M. 169 Johansson, A. 287 Jonkers, H.M. 119 Kaipio, J.P. 229, 231 Kameni , N. 227 Kapteina, G. 141 Karhunen, K. 229, 231 Kato, Y. 421 Khan, A.R. 103, 439 Khanzadi, M. 353 Kim, J.K. 423 Kind, T. 221 Kirilenko, A. 217 Kishi, T. 121, 125 Klemm, A.J. 129 Klemm, P. 129

Kobayashi, K. 121 Koenders, E.A.B. 143, 235 Komatsu, S. 121 Krause, M. 215 Kreit, A. 435 Krolo, J. 99 Kruger, E.J. 279, 303, 395 Krüger, M. 219 Kuperman, S.C. 127 Kurz, J. 255 Lackovi , V. 99 Lamperti, M. 445 Lapolla, S. 445 Larive, C. 357 Lee, C.-C. 323 Lehikoinen, A. 229, 231 Lenaers, J.-F. 355 Leskovar, I. 359 Leung, C.K.Y. 423 Li, M. 363 Li, V.C. 363 Lieboldt, M. 111 Lindvall, A. 145 Liu, H. 123 Liu, Y.-W. 323 López-Vázquez, E. 165 Lu, Z.D. 429 Mackechnie, J. 413 Mackie, K.P. 257 Mahmoudzadeh, R. 451 Maier, M. 49 Maldonado, N.G. 309, 403 Maliehe, R.S. 283 Maltese, C. 321 Malumbela, G. 253 Manfredi, G. 409 Marchand, J. 345 Marie-Victoire, E. 311 Marini, A. 453 Martínez, I. 31 Martínez-Madrid, M. 165 Martínez-Molina, W. 165 Martinola, G. 443 Masuku, C. 369 Materazzi, A.L. 247 Matthews, S.L. 277 Matthys, S. 39 Matusinovi , T. 227 Maury, A. 131 Mavar, K. 241 Mayer, K. 215 Mayer, T.F. 139 Mechtcherine, V. 111, 117, 327 Meda, A. 443, 453 Meinel, D. 187 Menzel, K. 179 Michelini, R.J. 309, 403

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Mielentz, F. 215 Milmann, B. 215 Minelli, F. 377 Mircea, C. 105 Mistretta, F. 213 Mmusi, M.O. 163 Moczulski, G. 367 Moghaddam, H. 375 Mohebbi, S. 375 Monteiro, E.C.B. 193 Monteiro, P.J.M. 229, 231 Moodi, F. 201 Moravcik, M. 385 Morency, M. 345, 347 Moreno, E.I. 165 Morin, R. 351 Mork, J.H. 299 Morlidge, J.R. 277 Möser, B. 87 Moyo, P. 67, 151, 209, 253, 369 Muehlbauer, C. 379 Muigai, R.N. 151 Müller, H.S. 161, 175 Munirudrappa, N. 427 Muthumani, K. 427 Netinger, I. 207, 269 Newmark, A.A. 285, 395 Newtson, C. 331 Niedermeier, R. 379 Nieves-Mendoza, D. 165 Nishizaki, I. 421 Noll, R. 225 Nuta, A. 345 Obladen, B. 143 Ohama, Y. 329 Orgass, M. 135 Orlowsky, J. 335 Orosz, K. 263 Oslakovi , I.S. 89, 133 Osselmann, P. 339 Osterminski, K. 173, 181, 185 Otieno, M.B. 157 Ottelé, M. 143, 235 Ou, J.P. 261 Ožbolt, J. 183, 407 Palic, S.S. 241 Palmieri, M. 413 Pampanin, S. 407, 413 Pani, L. 213 Pann, K.S. 323 Pastorcic, S. 167 Pedersen, H. 419 Péra, J. 325 Perez, F. 347 Pérez-López, T. 165

Pérez-Quiroz, J. 165 Peroš, B. 169 Pfeifer, C. 87 Pistol, K. 135 Pistolesi, C. 321 Pizarro, N.F. 309, 403 Plizzari, G.A. 377, 413, 443 Pontes, R.B. 193 Qian, S. 123 Quiertant, M. 421 Rahai, A.R. 447, 451 Rahgozar, R. 433 Rak, M. 99 Ramezanianpour, A.A. 431 Raue, E. 373 Raupach, M. 61, 177, 225, 255, 301, 333, 335, 337 Recupero, A. 441 Reichling, K. 255 Reuter, U. 117 Rinaldi, Z. 203, 443 Roberts, A.C. 307 Robertson, I.N. 331 Ronné, P.D. 283, 285 Roskovi , R. 89 Rubio-Avalos, J.C. 165 Russo, S. 449 Saberi, M.R. 451 Sadeghian, P. 447 Saiidi, M.S. 397 Samadi, M. 375 Sanchez, L.F.M. 127 Sánchez, M. 289 Sanjeevan, P. 129 Santhanam, M. 265 Sarvinis, P. 393 Schaurich, D. 223 Schießl, P. 139, 173, 181, 185 Schladitz, F. 391 Schlangen, E. 119, 363 Schmidt, J.W. 419 Schröter, H. 373 Schubert, K. 339

Schueremans, L. 417 Schulze, G. 305 Scrivener, K.L. 11 Seferovic, E. 113 Seppänen, A. 229, 231 Serdar, M. 133, 317 Sesar, P. 167 Shashikumara, S.R. 427 Shi, C. 159 Sibanda, B. 209 Siddiqui, N.A. 405, 411 Siebert, B. 365 Silfwerbrand, J. 287 Šipuši , J. 227 Skazlic, M. 317 Smith, H.G. 271 Smuts, M. 395 Soddemann, N. 175 Sodeikat, Ch. 141 Sosa-Baz, M. 165 Srinath, B. 265 Stander, H. 361 Stark, J. 87 Stavinoha, R. 93 Stoppel, M. 221, 255 Streicher, P.E. 303 Šušterši , J. 359 Szilagyi, H. 105 Taerwe, L. 39, 245, 381 Taffe, A. 221, 223, 225 Täljsten, B. 263, 383, 419 Thibault, M. 351 Tian, W. 181 Timmler, H.-G. 373 Tkalcic, D. 243 Tornello, M.E. 309 Torres-Acosta, A. 165 Trägårdh, J. 287 Troconis-Rincón, O. 165 Tschötschel, M. 295 Turatsinze, A. 95 Valdez-Salas, B. 165 Valek, L. 133 van Breugel, K. 235, 363

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Van Cotthem, A. 101 van de Kuilen, J.W.G. 123 Van Gemert, D. 417 van Grieken, A.N. 237 Van Humbeeck, H. 101 Van Schepdael, L. 417 van Zijl, G.P.A.G. 115, 361 Vasseur, L. 39 Vaysburd, A.M. 275, 315, 345 Venanzi, I. 247 Vennesland, . 149, 299 Verstraete, W. 131 Viljoen, E.J. 285 Vinajera-Reyna, C. 165 Vogel, M. 161 Vogelsang, J. 305 Volkwein, A. 181 von Greve-Dierfeld, S. 179 Vosooghi, A. 397 Vu, D.C. 95 Walraven, J.C. 3 Wang, J.J. 425 Wang, Y. 261 Warkus, J. 177 Wens, R. 85 Weritz, F. 223 Whitney, D.P. 349 Wiggenhauser, H. 19, 215, 221, 255 Wilsch, G. 223, 225 Wolff, L. 337 Wu, C.S. 425 Ye, G. 363 Yuan, Q. 159 Zajc, A. 359 Zakorshmenny, A. 211 Zappia, M. 247 Zhou, J. 363 Zhu, H.G. 423 Zhu, X.Q. 259, 261 Zilch, K. 379 Zollinger, D. 349