Developments in Geotechnical Engineering, 80
Embankments on Organic Soils
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Developments in Geotechnical Engineering, 80
Embankments on Organic Soils
Edited by J. Hartlen
Swedish Geotechnical Institute, S-581 93 Linkoping, Sweden and W. Wolski
Department of Geotechnics, Warsaw Agricultural University, ul. Nowoursynowska 166,OZ-766 Warsaw, Poland
1996
ELSEVIER Amsterdam - Lausanne - New York - Oxford - Shannon -Tokyo
Further titles in this series: VOlUmeS 2, 3, 5-7, 9, 10, 12, 13, 15, 16A, 22 and 26 are out of print G. SANGLERAT - THE PENETROMETER AND SOIL EXPLORATION R. SILVESTER - COASTAL ENGINEERING. 1 AND 2 L.N. PERSEN - ROCK DYNAMICS AND GEOPHYSICAL EXPLORATION Introductionto Stress Waves in Rocks 11. H.K. GUPTA AND B.K. RASTOGI - DAMS AND EARTHQUAKES 14. B. VOIGHT (Editor) - ROCKSLIDES AND AVALANCHES. 1 and 2 17. A.P.S. SELVADURAI - ELASTIC ANALYSIS OF SOIL-FOUNDATION INTERACTION 18. J. FEDA - STRESS IN SUBSOIL AND METHODS OF FINAL SETTLEMENT CALCULATION 19. A. KEZDI - STABILIZED EARTH ROADS 20. E.W. BRAND AND R.P. BRENNER (Editors) - SOFT-CLAY ENGINEERING 21. A. MYSLIVE AND 2 . KYSELA -THE BEARING CAPACITY OF BUILDING FOUNDATIONS 23. P. BRUUN - STABILITY OF TIDAL INLETS -Theory and Engineering 24. Z. BAZANT - METHODS OF FOUNDATION EGlNEERlNG 25. A. KEZDI - SOIL PHYSICS - Selected Topics 27. D. STEPHENSON - ROCKFILL IN HYDRAULIC ENGINEERING 28. P.E. FRIVIK, N. JANBU, R. SAETERSDAL AND L.I. FINBORUD (Editors) -GROUND FREEZING 1980 29. P. PETER -CANAL AND RIVER LEVEES 30. J. FEDA - MECHANICS OF PARTICULATE MATERIALS -The Principles 31. Q. ZARUBA AND v . MENCL - LANDSLIDES AND THEIR CONTROL Second completely revised edition 32. I.W. FARMER (Editor) -STRATA MECHANICS 33. L. HOBST AND J. ZAJiC - ANCHORING IN ROCK AND SOIL Second completely revised edition 34. G. SANGLERAT, G. OLlVARl AND B. CAMBOU - PRACTICAL PROBLEMS IN SOIL MECHANICS AND FOUNDATION ENGINEERING, 1 and 2 35. L. RETHATI-GROUNDWATER IN CIVIL ENGINEERING 36. S.S. VYALOV - RHEOLOGICAL FUNDAMENTALS OF SOIL MECHANICS 37. P. BRUUN (Editor) - DESIGN AND CONSTRUCTION OF MOUNDS FOR BREAKWATER AND COASTAL PROTECTION 38. W.F. CHEN AND G.Y. BALADI -SOIL PLASTICITY -Theory and Implementation 39. E.T. HANRAHAN -THE GEOTECTONICS OF REAL MATERIALS: THE eg'k METHOD 40. J. ALDORF AND K. EXNER -MINE OPENINGS - Stability and Support 41. J.E. GILLOT - CLAY IN ENGINEERING GEOLOGY 42. AS. CAKMAK (Editor) -SOIL DYNAMICS AND LIQUEFACTION 43. A.S. CAKMAK (Editor) - SOIL-STRUCTURE INTERACTION 44. A.S. CAKMAK (Editor) -GROUND MOTION AND ENGINEERING SEISMOLOGY 45. AS. CAKMAK (Editor) - STRUCTURES, UNDERGROUND STRUCTURES, DAMS, AND STOCHASTIC METHODS 46. L. RETHATI - PROBABILISTIC SOLUTIONS IN GEOTECTONICS 47. B.M. DAS -THEORETICAL FOUNDATION ENGINEERING 48. W. DERSKI, R. IZBICKI, I. KlSlEL AND Z. MROZ - ROCK AND SOIL MECHANICS 49. T. ARIMAN, M. HAMADA, A.C. SINGHAL, M.A. HAROUN AND A.S. CAKMAK (Editors) - RECENT ADVANCES IN LIFELINE EARTHQUAKE ENGINEERING 50. B.M. DAS - EARTH ANCHORS 51. K. THIEL - ROCK MECHANICS IN HYDROENGINEERING 52. W.F. CHEN AND X.L. LIU -LIMIT ANALYSIS IN SOIL MECHANICS 53. W.F. CHEN AND E. MIZUNO - NONLINEAR ANALYSIS IN SOIL MECHANICS 54. F.H. CHEN -FOUNDATIONS ON EXPANSIVE SOILS 55. J. VERFEL - ROCK GROUTING AND DIAPHRAGM WALL CONSTRUCTION 56. B.N. WHITTAKER AND D.J. REDDISH - SUBSIDENCE - Occurrence, Prediction and Control 57. E. NONVEILLER - GROUTING, THEORY AND PRACTICE 58. v . KOLAR AND I. NEMEC -MODELLING OF SOIL-STRUCTURE INTERACTION 59A R.S. SINHA (Editor) - UNDERGROUND STRUCTURES - Design and Instrumentation 598 R.S. SINHA (Editor) - UNDERGROUND STRUCTURES - Design and Construction 60. R.L. HARLAN, K.E. KOLM AND E.D. GUTENTAG - WATER-WELL DESIGN AND CONSTRUCTION 61. I. KASDA - FINITE ELEMENT TECHNIQUES IN GROUNDWATER FLOW STUDIES 62. L. FIALOVSZKY (Editor) -SURVEYING INSTRUMENTS AND THEIR OPERATION PRINCIPLES
1. 4. 8.
63. 64. 65. 66. 67. 68. 69. 70. 71. 72 73 74. 75. 76. 77. 70. 79.
H. GIL - THE THEORY OF STRATA MECHANICS H.K. GUPTA - RESERVOIR-INDUCED EARTHQUAKES V.J. LUNARDlNl - HEAT TRANSFER WITH FREEZING AND THAWING T.S. NAGARAI - PRINCIPLES OF TESTING SOILS, ROCKS AND CONCRETE E. JUHASOVA - SEISMIC EFFECTS ON STRUCTURES J. FEDA - CREEP OF SOILS - and Related Phenomena E. DULACSKA - SOIL SETTLEMENT EFFECTS ON BUILDINGS D. MlLOVlc - STRESSES AND DISPLACEMENTS FOR SHALLOW FOUNDATIONS B.N. WHITTAKER, R.N. SINGH AND G. SUN - ROCK FRACTURE MECHANICS - Principles, Design and Applications M.A. MAHTAB AND P. GRASS0 - GEOMECHANICS PRINCIPLES IN THE DESIGN OF TUNNELS AND CAVERNS IN ROCK R.N. YONG, A.M.O. MOHAMED AND B.P. WARKENTIN - PRINCIPLES OF CONTAMINANT TRANSPORT IN SOILS H. BURGER (Editor) - OPTIONS FOR TUNNELING 1993 S. HANSBO - FOUNDATION ENGINEERING R. PUSCH - WASTE DISPOSAL IN ROCK R. PUSCH - ROCK MECHANICS ON A GEOLOGICAL BASE T. SAWARAGI - COASTAL ENGINEERING - WAVES, BEACHES, WAVE-STRUCTURE INTERACTIONS 0. STEPHANSSON, L. JlNG AND CHIN-FU TSANG (Editors) - COUPLED THERMO-HYDRO-MECHANICAL PROCESSES OF FRACTURED MEDIA
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ELSEVIER SCIENCE B.V. Sara Burgerhartstraat 25 P.O. Box 21 1,1000 AE Amsterdam, The Netherlands
ISBN: 0-444-88273-1
0 1996 Elsevier Science B.V. All rights reserved. No part of this publication may be reproduced, stored in a retrieval system or transmitted in any form or by any means, electronic, mechanical, photocopying, recording or otherwise, without the prior written permission of the publisher, Elsevier Science B.V., Copyright & Permissions Department, P.O. Box 521,1000 AM Amsterdam, The Netherlands. Special regulations for readers in the USA - This publication has been registered with the Copyright Clearance Center Inc. (CCC), 222 Rosewood Drive, Danvers, MA 01923. Information can be obtained from the CCC about conditions under which photocopies of parts of this publication may be made in the USA. All other copyright questions, including photocopying outside of the USA, should be referred to the copyright owner, Elsevier Science B.V., unless otherwise specified. No responsibility is assumed by the publisher for any injuty and/or damage to persons or property as a matter of products liability, negligence or otherwise, or from any use or operation of any methods, products, instructions or ideas contained in the material herein. This book is printed on acid-free paper Printed in The Netherlands
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1x
The Authors w. Wolski
Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland
Z. Lechowicz
Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland
A. Szymanski
Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland
T. Baranslu
Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland
J. Hartlen
Swedish Geotechnical Institute Linkoping, Sweden
U. Bergdahl
Swedish Geotechnical Institute Linkoping, Sweden
P. Carlsten
Swedish Geotechcal Institute Linkoping, Sweden
R. Larsson
Swedish Geotechcal Institute Linkoping, Sweden
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Preface Development of regions and countries is often associated with the necessity of civil engineering works on soft soils. This happens more and more frequently because of earlier utilization of areas with better foundation conditions. The most troublesome of soft soils are organic soils, mainly due to their high compressibility, much higher than in mineral soils and very low shear strength. The large diversity of organic soils with respect to their origin as well as properties make their classification, testing and engineering prediction of behaviour very difficult. That is why engineers try, in general, to avoid constructing on deep layers of organic soils; if forced by the necessities, they manage with light structures, e.g. embankments or low buildings. The authors of this book have been involved in a joint research project of test embankments on organic soils carried out in the North-Western part of Poland by the Swedish Geotechnical Institute and the Department of Geotechnics of the Warsaw Agricultural University. These studies prompted us to write the present book. The authors wish to gratefully acknowledge the help of the Polish Local Land Reclamation Office WZMUW in Pila in the performance of field works. The authors are very grateful to professor Sven Hansbo at the Chalmers University of Gothenburg, who has reviewed this book and who has given many valuable comments on the text. The authors also want to express their gratitude to the Swedish Road Administration, for them letting us use their experience and publications on the subject. The authors will also greatly acknowledge the work done by Mr Eugeniusz Koda, making the computation of examples in chapter 9, and Mr Ryszard Zycnowicz, making most of the drawing, both working at the Warsaw Agricultural University, and to Mr Jan Lindgren, the Swedish Geotechnical Institute, who has done the large job of reviewing and editing the whole book. Spring 1996 The authors
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xiii
Contents PREFACE .......................................................................................... NOTATION AND SYMBOLS ........................................................... INTRODUCTION (J. Hartlen & W. Wolski) ..........................................
xi Xxii 1
PART I: TESTING AND ANALYSIS
1. 1.1
1.2 1.3
ORGANIC SOILS (R. Larsson) ............................................................ 4 Geological origin ................................................................................... 4 General Biogenic matter Organogenous soil Chemical sediments Organic soils Engineering properties ....................................................................... 15 Soil classification ................................................................................ 16 1.3.1 Identification of soil type * Gyttja-bearing soils . Dy-bearing soils Peat Topsoils * Marl and shell soils 1.3.2 Classification according to composition * Organic soils * Calciferous soils * Sulphide-rich soils 1.3.3 Other classification systems for organic soils 1.3.4 Detailed classification of peat * Humification . Water content * Sedge fibres * Root threads
xiv
. Wood remnants Designation Organic content Tensile strength - Smell - Plasticity Acidity 1.3.5 Other classification systems for peat 1.3.6 Geotechnical classification of peat References .................................................... *
*
1.4
2. 2.1 2.2 2.3
2.4
2.5
2.6 2.7
2.8
.....................................
29
SITE INVESTIGATIONS (U. Bergdahl) ........................................... General ................................ ...................... ...................... Mapping, general survey ........... Soil layer sequence ......... .............................................. 2.3.1 Soil radar 2.3.2 Penetration testing 2.3.3 Dilatometer testing 2.3.4 Sampling Groundwater ...................................................................................... 2.4.1 Pore pressure measurement 2.4.2 Permeability measurements Strength and deformation characteristics ......................................... 2.5.1 General 2.5.2 Field vane test . . Monitoring equipment ....................................................................... Test embankments for design purposes .......................... 2.7.1 Introduction 2.7.2 Preparatory investigations 2.7.3 Location of the test embankment 2.7.4 Design of test embankments 2.7.5 Test embankment monitoring 2.7.6 Construction of the test embankment 2.7.7 Presentation of test results 2.7.8 Analvsis of test results and recommendations References .........................................................................................
31 31 32 33
53
60
66
82
xv
3. 3.1 3.2
3.3 3.4
3.5
3.6
3.7 3.8 4. 4.1
4.2
LABORATORY INVESTIGATIONS (2 Lechowicz, A. Szymanski & T. Baranski) ................................................... .............................. 85 General ............ ....................................... Routine tests ....................................................................................... 85 3.2.1 Soil density 3.2.2 Consistency limits 3.2.3 Organic content 3.2.4 Carbonate content 3.2.5 Ferrous sulphide Determination of stress history .......... .......................................... 96 3.3.1 Preconsolidation pressure 3.3.2 Coefficient of earth pressure at rest Determination of deformation and consolidation parameters by oedometer tests .................................................... .......... 101 3.4.1 Incremental loading oedometer tests Testing procedure . Deformation and consolidation parameters 3.4.2 Continuous loading oedometer tests Testing procedure Deformation and consolidation parameters Determination of deformation parameters by triaxial test ............. 112 3.5.1 Testing procedure 3.5.2 Young’s modulus and Poisson’s ratio 3.5.3 Bulk modulus and shear modulus 3.5.4 Yield envelope and creep characteristics Determination of shear strength .............................................. 3.6.1 Swedish fall-cone test 3.6.2 Laboratory vane shear test 3.6.3 Direct simple shear test 3.6.4 Triaxial test Determination of permeability......................................................... 127 References ...................................................................... ..... 131 STABILITY ANALYSIS (Z. Lechowicz) ......................................... General ....................................................................................... Shear strength used in stability analysis .......................................... 4.2.1 Problems in the evaluation of shear strength 4.2.2 Undrained shear strength
137 137 138
XVl
4.3
4.4 4.5
4.6
4.7
5. 5.1 5.2
5.3
4.2.3 Effective shear strength Methods of stability analysis ............................................................ 4.3.1 Types and scope of analysis 4.3.2 Simple procedures of stability assessment * Preliminary estimation of safe embankment height Steady seepage conditions * Sudden drawdown conditions 4.3.3 Swedish circle method 4.3.4 Simplified Bishop method 4.3.5 Janbu generalized method 4.3.6 Janbu simple routine procedure 4.3.7 Morgenstem-Price method 4.3.8 Non-circular slip surface Stability of single-stage embankment .............................................. Stability of stage-constructed embankments ................................... 4.5.1 Types and scope of analysis 4.5.2 Evaluation of the increase in undrained shear strength based on empirical relations 4.5.3 Evaluation of the increase in undrained shear strength based on laboratory testing Other approaches in stability analysis ............................................. 4.6.1 Three-dimensional analysis 4.6.2 Stability asessment by f h t e element analysis References .......................................................................................
ANALYSIS OF SUBSOIL DEFORMATIONS (A. Szymanski) .................................................................................. General ....................................................................................... Deformation and consolidation parameters .................................... 5.2.1 Selection of parameters 5.2.2 Parameters used in deformation analysis 5.2.3 Consolidation parameters Analysis of “final” deformation ........................................................ 5.3.1 Type and scope of analysis 5.3.2 Initial settlement and horizontal movement 5.3.3 Empirical prediction of ‘‘final” settlement 5.3.4 Prediction of settlement in one-dimensional consolidation
145
160 162
171
176
181 181 183
192
xvii
5.4
5.5 5.6 5.7 5.8
Consolidation analysis .......................... ...................................... 203 5.4.1 Type and scope of analysis 5.4.2 Empirical prediction of the consolidation course 5.4.3 Prediction of one-dimensional consolidation at small strains 5.4.4 Prediction of one-dimensional consolidation at large strains 5.4.5 Layered soils Consolidation analysis of subsoil with vertical drains .................... 223 Swelling analysis ..................................... ................................. 227 Development trends in deformation a olidation analysis .... 230 References .......................................... .................................... 233
PART 11: DESIGN AND CONSTRUCTION METHODS
6. 6.1 6.2 6.3
7. 7.1
7.2
METHODS OF CONSTRUCTION (J. Hartlen) .............................. General ....................................................................................... Choice of method .............................................................................. Review of basic concepts of embankment construction on organic soils ...................................................................................... 6.3.1 Load adjustment 6.3.2 Soil replacement 6.3.3 Soil improvement 6.3.4 Other techniques
240 240 241
LOAD ADJUSTMENT (P. Carlsten) ................................................ Profile lowering ................................................................................. 7.1.1 Introduction 7.1.2 Design considerations 7.1.3 Limitations Pressure berms .................................................................................. 7.2.1 Introduction 7.2.2 Design considerations and dimensioning 7.2.3 Limitations 7.2.4 Construction aspects 7.2.5 Calculation example
249 249
246
250
xviii 7.3
7.4
8. 8.1 8.2
8.3
8.4
9. 9.1 9.2
9.3
Lightweight fills ................................................................................ 259 7.3.1 Introduction 7.3.2 Different lightweight fill materials 7.3.3 Design considerations 7.3.4 Construction aspects 7.3.5 Limitations 7.3.6 Case history from lightweight fill in a transition to a bridge 7.3.7 Case history from use of lightweight fill in repairing an existing road References ....................................................................................... 272 REPLACEMENT (P.Carlsten) ........................................................ General ....................................................................................... Excavation and backfill ..................................................................... 8.2.1 Description of the method 8.2.2 Design considerations 8.2.3 Limitations 8.2.4 Construction aspects Progressive displacement .................................................................. 8.3.1 Description of the method 8.3.2 Design considerations 8.3.3 Limitations 8.3.4 Construction aspects 8.3.5 Case history References .......................................................................................
274 274 274
STAGED CONSTRUCTION (W. Wolski) ...................................... General ....................................................................................... Precompression technique ............................................................... 9.2.1 Introduction 9.2.2 Design considerations . Surcharging to eliminate primary consolidation settlement * Surcharging to compensate for secondary compression 9.2.3 Design parameters 9.2.4 Limitations 9.2.5 An example - the Dalarovagen road Vertical drains .................................................................................. 9.3.1 Introduction
293 293 294
281
292
304
XiX
9.4
9.5 9.6
9.7
9.8
9.3.2 Installation of vertical drains 9.3.3 Design considerations 9.3.4 Design parameters 9.3.5 Limitations. Antoniny case history Construction monitoring ............................................... 318 9.4.1 Introduction 9.4.2 Instrumentation 9.4.3 Interpretation of the monitoring results Construction aspects ............................. 322 Design example for staged embankment with the use of vertical drains ........................... ...... ........ 323 9.6.1 Introduction 9.6.2 Stress distribution under the embankment axis during the first stage 9.6.3 Prediction of the immediate and consolidation settlements at the first stage of embankment loading 9.6.4 Consolidation performance at the first stage of loading 9.6.5 Total settlement at the end of the first stage 9.6.6 Shear strength increase due to the first stage of loading 9.6.7 Stress distribution in the subsoil under the embankment centre line, due to the second stage of loading 9.6.8 Prediction of the total settlement S, under the second stage of loading 9.6.9 Consolidation performance at the second stage of loading 9.6.losettlement of the subsoil after a lapse of two years from the beginning of consolidation 9.6.11Final remarks Design example for the staged embankment with surcharging ...... 345 9.7.1 Introduction 9.7.2 Required height of the embankment at the second stage of loading 9.7.3 Prediction of the settlements under the second stage embankment (with surcharge) 9.7.4 Vertical stress distribution after removal of the surcharge 9.7.5 Prediction of the swelling behaviour after removal of the surcharge 9.7.6 Prediction of the secondary compression behaviour of the subsoil 9.7.7 Final remarks References .....................
xx 10. LIME AND LIMECEMENT COLUMNS (P. Carlsten) ................. 355 10.1 Description of the method ................................................................. 355 10.1.1 What happens inside a lime/cement column? 10.1.2 Installation of limekement columns 10.2 Requirements for field and laboratory investigations ........ 10.2.1 Field investigations 10.2.2 Laboratory investigations 10.2.3 Test columns ................................... 360 10.3 Design considerations....................... 10.3.1 Introduction 10.3.2 Demands on the admixtures 10.3.3 Choise of admixture for different kinds of soils 10.3.4 Stability calculations * Embankments on horizontal ground surfaces or ground surfaces with slight lateral inclination . Embankments along a slope 10.3.5 Settlement calculations - Magnitude of settlements Distribution of load - limekement column and unstabilised soil * Settlementhime relationship ............................ 373 10.4 Limitations ......................... Construction aspects ........... ................................................... 374 10.5 ............................ 375 10.6 Requirements for field mea 10.6.1 Determination of the shear strength of lime/cement columns 10.6.2 Inspection of settlements 10.7 Example: Dimensioning of lime columns for reduction of settlements and for stabilisation of a road embankment .................................... 377 on soft and organic clay ................... 10.7.1 Introduction 10.7.2 Dimensioning of lime column reinforcement Calculation of settlements in stabilised clay . Settlementhime relationship 10.7.3 Stability calculation * Stability during construction . Stability of the road when in use 10.7.4 Results from dimensioning . Transition to the culverts * Construction schedule . Inspection and follow-up . Comments
xxi Case history - Bridge foundation on soft clay stabilised with lime columns 10.8 References .......................................................................................
396
. 400 11. OTHER METHODS (P. Carlsten) ..................................... ............................ ..................... 400 11.1 Reinforcement ... 11.1.1 Tradition S 11.1.2 Geotextiles . Introduction and description of the method * Construction aspects . Applications .................................. ...................... 406 11.2 Pile foundation ........ 11.2.1 Description o ethod and constructi ................................... 11.3 References ...............
AUTHOR INDEX ..................................................................................... .410 SUBJECT INDEX .....................................................................................
4 17
xxii
Notations and symbols -
area ratio (constant for a specific cone)
-
drain thickness
- width of embankment slope area of column base
A
-true
AQC
- failure surface
b
- load factor -
width of slice
- width of embankment -
B
width of drain
-width
of loaded area
-reduced width of embankment C r
- effective cohesion intercept
%
- coefficient of consolidation at horizontal drainage
%~ Ca
-effective field coefficient of consolidation at horizontal drainage
Cv C
-
apparent cohesion intercept
-
coefficient of consolidation at vertical drainage
-
constant for the fall cone test depending on the apex angle of the cone and the condition of the soil (undisturbed or remoulded) K0-consolidated test
CKoU
-undrained
Co
- compression index
Cee CC
- modified compression index -
continuous consolidation oedometer test pore pressure gradient oedometer test
CG
-constant
Cr
- recompression index
Cre CRL
-modified recompression index -
constant rate of loading oedometer test
xxiii constant rate of strain oedometer test
CRS
-
Cs
- swelling index
Cse
-modified
C~
-coefficient of secondary compression
C~
-modified secondary compression index
Cs
-coefficient of secondary compression at recompression
swelling index
-diameter of drainage well -depth of cone penetration
d/L
-value defining the location of the failure surface in Janbu's simplified method
dA
-bottom area of vertical column
ds
-
diameter of disturbed zone
dw D
-
equivalent diameter of drain
- depth factor -
diameter of sample
-diameter
of soil cylinder with drain
direct simple shear test
DSS
-
Dv
- diameter of vane - void ratio void ratio
e0
-initial
cp
-void ratio at the end of primary consolidation
E
- deformation modulus - normal component of interslice force - Young's modulus
g
r
- drained modulus of elasticity tangent modulus
Ei E
-
secant modulus
Et
-
tangent modulus
E
-
undrained modulus of elasticity
ED
-
dilatometer modulus
ESL
-effective stress level
S
-initial
xxiv f
f fl
offset of the normal force from the center of rotation
-perpendicular -correction
factor in Janbu's simplified method
-coefficient in Steinbrenner's formula -coefficient for Hansbo's diagram -coefficient
in Steinbrenner's formula
f(x)
- function describing the way in which the ratio T/E varies in a slope
F
- factor of safety
Fo
-
initial factor of safety in Janbu's simplified method
F2D
- two-dimensional factor of safety
F 3D
- three-dimensional factor of safety
g
-acceleration due to gravity
G
-
h
-depth below ground surface of the point where u is estimated
he
- final height of embankment
ho h
- initial height of embankment
h+f
- thickness of embankment with surcharge
H
-
thickness of subsoil
-
degree of humification
shear modulus
- thickness of surcharge
- height of embankment -
height of sample
Hs
- safe height of embankment
Ht Hv
-factor -
height of vane
H0
-
initial height of sample
i
-
hydraulic gradient
I
-
influence factor
depending on decomposition
ID
- material index
Ih
-horizontal influence displacement factor
IL
- liquidity index
XXV
Ip
- plasticity index
I
-vertical
v
influence displacement factor
IL
-
k
- coefficient of permeability
incremental loading oedometer test
- coefficient of horizontal permeability
\
-coefficient of horizontal permeability in disturbed soil - coefficient of vertical permeability
K
-
bulk modulus of shear strength increase (depending on loading case and overconsolidation ratio)
-coefficient
Ko
-coefficient -
of earth pressure at rest
horisontal stress index
- length of failure surface at the base of a slice
1
-
length of drainage path
-
length of drain
- reduced length of embankment L/H
- draw-down ratio
Ls
-total
m
-
length of a slide
mass
- slope of ~fu/CY'v - OCR relation in log-log scales slope of the relation between log0;fu/Cr'v) and log(ESL) in the normally consolidated state (ESL _< 1)
mnQ
-
Ill o o
- slope of the relation between log0;fu/~'v) and log(ESL) in the overconsolidated state (ESL > 1)
mv M
- coefficient of volume change -
oedometer modulus defining the location of the critical slip circle in Taylor's stability chart
-value
-parameter in Cam Clay model M
r
ME
-
modulus number
-
resisting moment due to end area effects for each end plane
xxvi initial oedometer modulus
M0
-
n
-porosity
N
-total
normal force on the base of a slice
-
critical stability number
-
normally consolidated
NF
-
Cousin's stability number
NKT
-empirical
NT
- Taylor's stability number
0C
- overconsolidated
0CR
-
overconsolidation ratio
P
-
mean normal stress
P~
-equivalent isotropic pressure
Pa PSC
- atmospheric pressure
No NC
-
cone factor
Plane strain compression test
- deviatoric stress -
load applied to the subsoil
- embankment load
qQ
-measured
value of cone resistance
qf
-final embankment load
qj
-component of the specific discharge vector of pore water
%
-
%
-total
qw
-discharge
Q
-outflow of water from a drain
-
surcharge load cone resistance capacity of a drain
drain radius
r
-pore pressure ratio
R
-radius of dewatered soil cylinder
xxvii
-
R
radius (moment arm) associated with mobilized shear forces S m percentage of humification
-radius of cylindrical part of failure surface - radius of failure surface -normalized undrained shear strength in the normally consolidated state (OCR = ESL = 1) - s p a c i n g
-
of vertical drains
settlement
So
-settlement associated with primary consolidation
S~
-
S c(f+sr)
Sf Sh
final consolidation settlement
-final consolidation for embankment and surcharge load -
final settlement
- horizontal movement
Si
-
immediate settlement
-
initial settlement
Sm
- shear force mobilized at the base of a slice
Ss
S sw
St
- settlement associated with secondary compression - swelling value -
-
total settlement settlement at time t
t
- time
tf
-time to failure in a vane shear test -time at the end of the settlement process
tp
-time at the end of primary consolidation
ts
-time for swelling
tso
-designed life-time of the embankment -time at removal of the surcharge
tso
-time for 50 % primary consolidation
tgo
-time for 90 % primary consolidation
boo
-time for 100 % primary consolidation
tL
- vertical distance from the base of a slice to the line of thrust on the
xxviii left side of the slice
tR
-
T
-tangential component of interslice force
vertical distance from the base of a slice to the line of thrust on the right side of the slice
- applied peak torque during a vane test (TC)
-
triaxial compression test
(TE)
-
triaxial extension test
Th T,,
-time factor for consolidation at horizontal drainage -time factor for consolidation at vertical drainage -excess pore pressure - pore pressure pressure measured behind the conical tip during cone penetration
Uc
-pore
ub
-pore
pressure acting at the centre of a column base
-pore
pressure at an undrained bottom end in an oedometer test
uj
-
U
- degree of primary consolidation
Uh Up
-
UV
-degree of consolidation at vertical drainage
U(f+sr)
-degree of consolidation at surcharge removal
(V1)max
-highest
W. 9
-component
W. J
- displacement vector
WE
-
WN
-natural
WV W
-
WO
-total weight of a column
l,j
gradient of pore water pressure degree of consolidation at horizontal drainage of consolidation at surcharging required to produce settlements equal to primary consolidation for the embankment load
-degree
allowable rate of axial displacement in a drained triaxial test of the displacement gradient
liquid limit water content
plastic limit
- weight of a slice
xxix -horizontal distance from a slice to the center of rotation -vertical space coordinate -depth - d i s t a n c e
from the open end of a drain
- vertical coordinate
-
-
angle between the horizontal plane and the tangent to the centre of the base of a slice angle between the horizontal plane and the line of thrust on the right side of a slice
- angle of embankment slope -compressibility of the pore water -permeability change index - convective coordinate 8ij
- Kronecker's delta
Ae~,Aey, Ay~y - increments of strain components A~x,A~y,A % - increments of stress components Aq
- load increment
Ap
- pressure difference
E
-
strain
-
lateral strain
l~ij(e)
-
elastic strain increment
Eij(P)
- plastic strain increment
Eo
- strain at start of primary compression
E1
- major principal strain
e3
- minor principal strain
E1
-vertical creep rate
e~
XXX
Elf El00
-
axial strain at failure in a triaxial test
- strain at the end of primary consolidation
-
angle of internal friction
-effective angle of internal friction
7o %
- unit weight of embankment material -
unit weight of water representing the portion of the function f(x) which is used when solving for the actual factor of safety in the MorgensternPrice method
- c o n s t a n t
X,,p
-
dimensionless parameter
g
- correction factor
gl la2
-
factor including the effect of drain spacing
-
factor including the effect of soil disturbance
la3
- factor including the effect of well resistance
V
-
V'
-Poisson's ratio in drained conditions
Vu
- P o i s s o n ' s
P
-
radial coordinate
-
density of soil
Poisson's ratio ratio in undrained conditions
Pd
- dry density
P~
-density of solid particles
Pw
- density of water
%
-horizontal stress component
(Jv
-vertical stress component
Gvo
- i n
situ overburden pressure
- major principial stress (Y3
-
minor principal stress
xxxi -
0""h ij I~' N (y' p
effective stress
-effective horizontal stress -
effective stress tensor
-effective -
preconsolidation pressure
-initial
(y'
11
(Y'v l~ "VO
normal stress preconsolidation pressure
-effective pressure at%r pore pressure equalization -effective vertical stress -in
situ effective vertical stress
(Y'vf
- final effective vertical stress after removal of surcharge
(Y'vf+s
- final effective vertical stress due to permanent load and surcharge after pore pressure equalization -decrease of effective vertical stress due to surcharge removal
q~
"Cfd %
-
shear stress
-
available shear strength
-
drained shear strength
-
undrained shear strength
%c
-undrained
shear strength in triaxial compression
'l~fuD
-undrained
shear strength in direct simple shear
q~fuE
-
~fu(LAB) q~fv q7N
undrained shear strength in triaxial extension
- average undrained shear strength in laboratory tests, [0:f~C+q:f~D+q:f~E)/3l strength value obtained in field vane shear test
-shear -
shear stress
This Page Intentionally Left Blank
Introduction J. Hartldn, Swedish Geotechnical Institute W. Wolski, Department of Geotechnics, Warsaw Agricultural University
The purpose of this book is to introduce up to date knowledge of how to construct embankments on organic soil. Organic soil is a conception, which involves several types of soil from pure organic forms as peat and gyttja to transition forms towards the mineral soils, as clayey gyttja and organic clay. In the geotechnical practice, there is however not a generally accepted rule how to classify the organic soils into main groups. This book only covers questions relevant to organic soil. This means that much of the achievements reached during the last years about soft clay are not dealt with here as long as they are not relevant for design calculations associated with construction on organic soil. Embankments on organic soils are most often constructed for roads or for flood control dikes. There are dams for water retention as well as waste tailings dams founded on organic soils. In recent years temporary embankments have been utilized to preload the sot~ subsoil, including organic subsoil and in this way improving the bearing capacity before the structure is built. Such procedures are used not only for easily adjustable structures like e.g. coal yards but also for buildings. When an embankment is erected directly on soft organic soil layers, both stability and settlement problems will generally arise. The load increase and geotechnical properties of soft soil together with schedule of construction (available construction time) and acceptable future settlements are the important factors that govern the choice of construction method. Improvement of existing structures founded on organic soil create special problems. Formerly, the settlements of e.g. a road sometimes were of little importance and thus many roads were built as "floating" structures (on timber fascines) with a relatively low factor of safety. Organic soil occurs in many forms and with varying thickness. Due to this, the problems differ from site to site and the construction methods must be adopted to the conditions in each specific case. One also has to consider the experience of an individual country and the machinery available by the contractors.
The progress in understanding the behaviour of organic soils under load as well as in utilization of new techniques and materials makes it possible to undertake successfully projects which could not be performed a couple of years ago. This textbook presents the experience gained by the authors from practical consuiting work and from research projects. A research co-operation has been going on between the Swedish Geotechnical Institute and the Warsaw Agricultural University. This close cooperation has resulted in several reports and papers. The authors are all involved in the total text although the great work of making the manuscript to each chapter was divided. The textbook consists of two parts. The first part is dedicated to the testing, calculation and general behaviour of organic soils under load and the second part to design and construction methods. The first part of the textbook thus presents 9 Classification 9 Field tests 9 Laboratory testing 9 Stability analysis 9 Deformation analysis Based on these properties (basic information) different foundation methods are presented in the second part of the textbook. The methods are applicable to construction of 9 Road embankments 9 Dikes e.g. for flood control 9 Dams for water retention 9 Preloading of foundations for structures 9 Waste tailing dams on organic soils 9 Widening of existing embankments 9 Transition zones between settling embankments and stiff structures 9 Land reclamation with soft sea beds. The second part of this textbook is written in such a way that it shall be possible for a designer/contractor to pass over the basic chapters in part one. To improve the usefulness of the textbook, detailed examples are shown how to make evaluation and calculation.
PART I:
Behaviour of Organic So~lsTesting and Analysis
Chapter I
Organic Soils R. Larsson, Swedish Geotechnical Institute
1.1
GEOLOGICAL ORIGIN
General "Organic soils" or "soils with an organic content" has often been a concept with various meanings in geotechnical engineering and the rules for division into different groups have often been rather diffuse. Apart from purely organic forms of peat and gyttja, there are a large number of transitional forms towards the mineral soils. Examples are clayey gyttja, organic clay and flood- plain sediments. Soils rich in sulphides and so called "svartmocka", which mainly consists of sulphide rich silt, are often designated as organic soils. Sediments rich in calcium carbonates, such as marl and diatomaceous soil, are sometimes included in the concept and when the limits are extended very far, also shell soils are included. Topsoils which have some kind of organic content are also included among the organic soils.
Biogenic matter The organic matter in soils originates from living plants, animals and organisms, fonning biogenic matter in contrast to mineral matter. Animal life on the continents plays a relatively small role from a geological point of view. Marine animals, on the other hand, have a major role in the formation of coral reefs and shell soils, of which many are wave-washed sediments. Aquatic animals and organisms also form the basis for many other more or less biogenic sediments. A large number of formations originate directly or indirectly from plants. During transformation processes of the plants, organic products such as peat and coal are created, as well as various inorganic products.
Geological Origin
5
Organogenous soil The term "organogenous soil" is used to denote matter consisting of plant and animal remains which cannot decay into humus or gyttja. Especially in the tropical seas, this type of matter occurs in the form of coral reefs. More globally, the organogenous soil occurs in the form of shell soils. Diatomaceous soil consists of the small but resistant SiO 2 skeletons of diatoms and needles of sponge animals. Various forms of inorganic muds also occur in the deep seas.
Chemical sediments The chemical sediments, mainly calcareous soils, have been formed by a combination of mechanical accumulation and precipitation of water soluble calcium and iron compounds, with or without the aid of organisms. Marl occurs in areas with calcareous bedrock, usually in connection with peat deposits.
Organic soils Organic soils can easily be identified by their combustibility. They are formed during the decomposition of dead organic substances i.e. remnants of plants and animals. This process takes place in different ways, mainly through bacterial activity, and is intensified by a hot climate, suitable humidity and access to oxygen from the air. The processes are schematically as follows: Living plants and animals
Microscopical plants and animals
Dead organic substances
Dead organic substances
Complete oxidation mainly in the tropics
Incompleteoxidation and decomposition
Especially anaerob reduction in water
Humus CO 2, H20
etc.
#
In dry areas: Topsoils k
Relatively fast
Fig. 1.1.
In swampy areas and lakes: Peat and Dy
Gyttja
Y
Retatively slow
Schematic process during decomposition of biogenic matter. After Hallden (1961).
6
Organic Soils
Humus is a dark substance with a colloidal structure. The destruction process from dead organic substances to humus, which is the solid product of this process, is called humification. The process takes place with the aid of fungi, bacteria and other organisms. In dry areas, the further decomposition processes are so rapid that the thickness of the topsoils seldom exceeds a few tenths of a metre. In swampy areas, the processes are slower as lack of oxygen delays the oxidation. Access to air is prevented and the water in these areas may be almost totally devoid of free oxygen from dissolved air. In the upper layers, there is usually a certain amount of oxygen from precipitation, contact with the air, flowing water and fluctuating water levels. Some of this oxygen may penetrate to deeper layers. Also in the absence of free oxygen, the decay proceeds in the form of fermentation and putrefaction. This is evidenced by the evolution of gaseous products, such as methane and sulphuretted hydrogen. Other sulphides and compounds devoid of oxygen are also produced. These anaerobic processes are much slower than the decay process when there is access to free oxygen. The formation of peat areas mainly occurs in humid parts of the temperate climate zones. Mires will form and peat accumulate wherever the conditions are favourable, irrespective of altitude or latitude, but mainly in those parts of the world where the climate is relatively cold and wet and where suitable conditions are consequently most frequent, Table 1.1. The list is not complete as detailed information from several areas is unavailable - from all African countries, from several South and Central American countries, from parts of Southeast Asia and from parts of Australia/Tasmania.
Peat originates from plants and denotes the various stages in the humification process where the plant structure can still be discerned. Dy denotes the stage where the plant structure is completely destroyed. Peat is a sedentary soil which has been formed in situ from the original material. Some types of dy are also formed in situ, where they constitute the highest degree of humification of the peat. Other types of dy have been transported by water and precipitated in a colloidal form in environments with low contents of calcium. Gyttja originates from remains of plants and animals rich in fats and proteins, in contrast to peat which is formed by remains of plants rich in carbohydrates. Dead microscopic aquatic animals are dissolved and decomposed with the aid of bacteria to a flocculent substance, in which mineral particles and less decomposed remains of plants and animals are embedded. Further decomposition occurs with the aid of organisms living in the substance, such as worms and larvae. Fermentation processes generating sulphuretted hydrogen and methane complete the formation of gyttja. Gyttja formed in nutritious water is greenish in colour. In less nutritious environ-
Geological Origin Table 1.1. Rank
Percentage of national area covered by peat in different countries in rank order (after Kivinen and Pakarinen 1980 and Taylor 1983). Country
Peatland area mill. ha > 0.3 m peat
Peat area order %
Finland Canada Republic of Ireland Sweden Indonesia
10.4 170 1.2 7.6 26
33.5 18.4 17.2 17.1 13.7
6 7 8 9 10
Northern Ireland Scotland Iceland Norway Wales
0.2 0.8 1.0 3.0 0.2
12.7 10.4 9.7 9.4 7.7
11 12 13 14 15
The Netherlands Malaysia USSR Germany (GDR) Poland
0.3 2.4 150 0.6 1.4
7.4 7.2 6.7 5.1 4.4
16 17 18 19 20
Germany (GFR) Cuba USA (including Alaska) England Austria
1.1 0.5 30 0.4
4.4 3.9 3.3 2.8 2.8
21 22 23 24 25
Denmark Switzerland Hungary New Zealand Belgium
0.1 0.1 0.1 0.2
2.8 1.3 1.1 0.6 0.6
26 27 28 29 30
Uruguay Japan Yugoslavia China Italy
0.1 0.2 0.1 3.5 0.1
0.5 0.5 0.4 0.4 0.4
31 32 33 34 35
Israel Czechoslovakia France Greece Romania
0.1
0.25 0.2 0.2 0.04 0.03
36 37 38 39
Argentina Spain Australia Bulgaria
0.016 0.012 0.002 0.001
8
Organic Soils
ments, the gyttja becomes brown from mixing with brown-black dy. Gyttja has a more or less elastic consistency which is sometimes almost jellylike. Dy has a stickier consistency. Gyttja formed in areas with calcareous soils often occurs as a transitional form between gyttja and marl, called calcareous gyttja. Depending on the content of mineral particles a number of different soils occur, such as clayey gyttja, organic clay etc. Due to the increasing biological activity during deglaciation, postglacial clays in general contain some organic matter. Another type of sediment with a highly variable organic content consists of the
flood-plain sediments. These have been deposited when streams, lakes and seas at high water overflowed their natural boundaries. The flood-plain sediments thereby contain mixtures of coarse and fine mineral particles and organic matter. A special type is flood-plain clay, which is considered to have become more common as cultivation has progressed. The run-off water from the fields has thereby transported away a large amount of fine grained material previously confined by the vegetation. Flood-plain clay mainly occurs in the low-lying parts of the cultivated areas. The peat areas and the thick deposits of more or less organic soils occur to a large extent in the northern parts of the world. In northern Europe, the deposition of postglacial clay started at a time when large parts of the land were submerged under the sea. At the same time as the inland ice retreated, the temperature rose and thereby also the biological activity in the areas now free from ice. This brought an increase in the amount of plant and animal remains in the sediments that were deposited. Lakes and seas slowly became shallower through the accumulation of a mixture of mineral and organic material at the bottom and because the ongoing land-heave bays were cut off from the sea, becoming lakes and marshes. Soils such as organic clays, clayey gyttjas and gyttjas were formed as the organic content increased. The postglacial clays, which are located in topographically lower areas, thus became overlaid to a large extent by soils with increasing organic contents. The thickness of the deposits varies from place to place. Deposits of postglacial clay 10 to 20 metres thick are common, but may in extreme cases exceed 100 metres. The overlying deposits of soils designated as organic mineral or organic soils may be up to 10 metres thick. In some areas, the lakes were overgrown, becoming swamps and fens where different types of peat were formed. Such areas were also created in higher regions by topographical conditions leading to high ground water levels and marshes. The formation of peat bogs has been described in detail by Hobbs (1986). The development of a mire from open water to a raised bog is shown schematically in Table 1.2. In the first stage, the mires develop in moving water in lakes, basins and valleys
Mire stages, morphology, flora and some associated properties of some British peak (Hobbs, 1986).
Geological Origin Table 1.2.
9
10
Organic Soils
under the control of the water level, nutrients being brought in by stream flow, runoff and percolating groundwater. The nutrient supply is generally rich. The landscape at the completion of this stage is marsh-like and is commonly referred to as "fen" and the peats as "fen peat". Such mires are generally underlain by very soft organic muds of gyttja and organic clay, which can cause severe engineering problems. In the second, transitional stage, the mire relies to an increasing extent on precipitation. It continues to receive nutrients via fluctuating ground water, although to a decreasing extent. In the third stage, the mire has grown beyond the maximum physical limits of the ground water and relies solely on direct precipitation for its water supply. The peat itself acts as a reservoir holding water above the level of the ground water. The nutrient conditions are deficient since rainfall or snow contains only minute quantifies of salts, dust and dissolved gases. These mires are called raised bogs and are acid in character. The third stage does not have to be preceded by the earlier stages, but mires can develop directly on the land surface provided that the climatic and topographic conditions are favourable. Such mires, which may be extensive, are known as blanket bogs. Another presentation of the normal development of a mire from a lake to a raised bog with emphasis on vegetation and type of organic soil is shown in Table 1.3. The corresponding stratigraphy is shown in Fig. 1.2. It should be observed that the presentation in this figure is made for convenience. The different stages do not develop simultaneously. Bog raising normally starts at the centre of the old lake and only after filling of the lake is complete. The development of a typical Swedish peat bog is shown in Fig. 1.3. The height to which a bog can be raised depends mainly on the topographic and climatic conditions. The relationship between raised height, diameter of the bog basin and mean annual precipitation for some raised bogs in southern Sweden is shown in Fig. 1.4. This relationship probably varies with the average temperature. Mires are highly complex systems and changes in their environments concerning water supply, temperature and supply of nutrients alter the course of their development. The different soil layers in a profile contain various amounts of calcium. Apart from the clay particles originating from lime-rich rock, also dissolved calcium was precipitated and already in the early stages various amounts of shell bearing organisms were deposited. The remains of these may be microscopic and evenly distributed in the soil, but also wave-washed layers of shell soil occur. The carbonate contents in different layers thus vary depending on the environment at deposition. In the
Geological Origin
Table 1.3. Vegetation and peat succession (Hobbs 1986).
~~
Notes
1 lherr
FIG 1. Lake filling-the hydroseral succession. Note that bog raising normally starts at the centre of the old lake after filling is complete but is shown in the lake margin here for convenience.
11
wll b e romr natural ovsrloppng at the vortous rloqer 7 Layers o f silt brought In dutmg tloodmq may be prestnl on t h e t e n stapes 3 Condittanr 1 1 1 1 b r partuularly f t r h In calcareous regions. rich m species
12
Fig. 1.2.
. . . . .
. . . . .
. . . .
_>
__>
.__>
->
I'
L
I-
! I
L~_
i
I'.< LU
0. UJ 14.
co
10
(D
_~
(3.
to
::3 (3
43
to
E ,,.* 9
9
.c t~ 43
(D G)
10
. . . .
......
-.->
__>
....-~
-->
o L
E
r-
.,==
o
43
u~
.c
Organic Soils
I I
I i II li li
I I I I
I I II II
i
~
to
43
rr
._~
10 a)
0 UO
(D L O.
10
.c
O~
:3 0
3=
--
10 L
--
r 10 L
~)
0
-.
.=.
10
Ck ::3
.c
__
___>
10
.c c 0 to o~ 0 43 10
.to nrv
. . . .
q)
to
E
~,
43
Lake filling - the hydroseral succession. Note that bog raising normally starts at the centre of the old lake after filling is complete, but is shown in the lake margin here for convenience. (Hobbs 1986).
13
Geological Origin
3
4 .s
Fig. 1.3.
Four stages during the development of a typical Swedish raised bog. 1) A lake with gyttja deposited at the bottom. 2) The lake has become a fen with sedge peat. 3) Sphagnum mosses have invaded and the fen has been transformed into a bog. The bog has grown in height, expanded over surrounding firm ground and finally become a vegetation of pine trees on the surface. Only a small pond remains of the former lake. 4) The bog has grown further and a typical bog for central Sweden has developed, surrounded by birch and alder vegetation. Pine trees grow on the outskirts of the bog and scattered dwarf pines grow in the central parts, (Magnusson et al 1963).
14
Organic Soils
1
I
I
I
I
I
I
I
I
I
I
I
I
I
(I) (D
E
r
J~.
I I
nn nn l
-r
--__.__
~O
9
Rainfall
s9
i
D
t
o
I
o
z~
I
550 - 600 700 - 800
1
1
1
1
Diameter
9 i
1
1000
500
Fig. 1.4.
450 - 500 9
I
i
( mm )
( me~
!
900 - 1000 1
1
1500
)
The relationship between degree of convexity of raised bogs in southern Sweden (plotted as height of cupola against diameter of bog basin) and the mean annual rainfall. The data are replotted from Granlund (1932) and many of the values to the left of the broken line have been omitted for clarity. The continuous lines define so-called limiting heights for a given rainfall range, (Tallis 1983).
dry crust and other zones affected by climatic conditions, the carbonate contents have later often been greatly reduced due to weathering and leaching. In lakes with a high content of calcium in the water, various forms of calcareous gyttja and marl were formed. Weathered calcium was transported long distances from the lime-rich bedrock and occurs in large areas. In mires formed under such conditions, a layer of calcareous marl is sometimes found between the peat and the gyttja layers. This layer has been deposited through photo- reduction of dissolved
Engineering Properties
15
carbon dioxide by submerged green plants in clear shallow water penetrated by sunlight. Such layers may be up to 1.5 metres thick and may constitute engineering hazards if the underlying very soft layers are not detected in field investigations. Other types of marl layers may occur in alterations of the normal sequence for mire formation. In clay sediments, there are also often layers containing remnants of decomposed organic material, which have been transformed during processes in a reducing environment to ferrous sulphide, among other substances. The ferrous sulphide, which in pure form is completely black, occurs as dark spots, patches, bands or completely colours the soil even at moderate contents. In Finland and northern Sweden, a special type of mass transport took place. Because of the land heave and the narrow fjord-like creeks, the rivers started to erode through the earlier deposited deltaic sediments. The sediments were redeposited further and further away as the coast line moved. The redeposition occurred in still water along the rivers and the coast. Thick layers of silt and clay mixed with dead algae and remains of animals were formed. The deposition took place in a reducing environment. The decomposition of the organic matter in this environment created a special *ype of soil called "svartmocka", which covers large parts of the coastal areas of the Gulf of Bothnia but is confined to this region. It consists of silt and clay with a relatively high content of amorphous ferrous sulphide and is black in colour. The content of gyttja varies, but is usually relatively low.
1.2
ENGINEERING PROPERTIES
The engineering properties of organic soils show a great variation depending on the type and amount of organic matter. The organic matter may occur in many forms from small amounts of amorphous or colloidal substance embedded in the pores of a mineral soil to fibrous peat with a structure resembling a coarse, loosely woven mat. The effect of the organic content on the engineering properties in relation to the properties of a pure mineral soil is in the former case mainly confined to a decreased permeability and a somewhat increased tendency to creep. In the latter case, the properties are quite different in most respects. As in mineral soils, the strength of organic soils is a function of the effective stresses acting in the soil, as well as its loading history. When normalized towards the stress history, the shear strengths of organic soils are usually higher than for mineral soils. Most organic soils, however, have no significant loading history as they are fairly recent deposits in waterlogged areas. Many of these profiles do not even have a dry crust and most of the soil layers have not been subjected to any load
16
Organic Soils
other than the weight of the overlying soil. The resulting effective stresses are relatively low because of the high ground water levels and the low densities of the organic and uncompressed soils. Consequently, most organic soils have very low strengths and are extremely compressible. They also exhibit large creep effects. Highly fibrous and undecomposed organic soil has a pronounced structural anisotropy. The fibres and the plant remains usually have a horizontal orientation. The fibres constitute a horizontal reinforcement and failure surfaces in such materials usually occur as vertical fractures or horizontal shear planes parallel to the fibres. The distribution of stress from loads on the ground surface with depth is relatively small because of the fibres. The permeability of the soil is relatively high and is often many times higher horizontally than vertically. Due to the high permeability and the fibres, stability is usually not a problem in the fibrous peat itself, provided that measures are taken to prevent punching or cracking under the loaded area and that the loading is not extremely rapid. Many fibrous peats, however, are underlain by other very soft soils and serious stability problems often occur. The compressibility of the peat is very high. Even for small extemal loads it is common for the settlements to amount to more than half of the original thickness of the peat layer. On top of this, there are considerable creep deformations with time, which for most engineering tasks cannot be accepted and have to be stopped. The effects of structural anisotropy decrease with decreasing content of fibres and increasing degree of humification. The permeability also rapidly decreases as the soil becomes more humified. The engineering properties of organic soils thus depend on the type of organic matter as well as the organic content. Structural anisotropy is important not only in peat but also in other organic soils such as gyttja, where significant effects may occur even at relatively low organic contents. The engineering problems in the more humified organic soils with low permeabilities resemble the problems encountered in soft mineral clays, but are often more accentuated because of the higher compressibility, the enhanced creep effects, the very low effective stresses and strengths and the sometimes very low permeabilities.
1.3
SOIL CLASSIFICATION
Different classification systems are used in different countries. The rules for classification presented here conform with the Swedish geotechnical classification system worked out by Karlsson and Hansbo (1981) in cooperation with the Laboratory Comittee of the Swedish Geotechnical Society.
Soil Classification
17
Understanding of the stratification and properties in a soil profile is made easier if the geological history and the environmental conditions at deposition of the sediments are known. When possible, the soils should therefore first be classified in this respect as, for example flood-plain sediments, wave-washed sediments, lowmoor peat (deposited in fens) or highmoor peat (formed in raised bogs). In order to classify the soils correctly in the laboratory, an initial requirement is a determination of the organic content, the content of carbonates and possibly the content of ferrous sulphide. The natural water content, the consistency limits and the density are also valuable aids to classification. For detailed classification of peat, a number of other determinations are required. As a rule, fresh samples of organic soils can be distinguished from pure mineral soils by their odour, which originates from decayed organic substances. The odour from a sample with a low water content can be accentuated by wetting and heating of the sample.
1.3.1
Identification of soil type
Gyttja-bearing soils: 9 Gyttja is normally greenish in colour, but may be brown or red. It bleaches on drying, usually to a grey colour. In the wet state, gyttja has an elastic, rubbery consistency. It has a brittle rupture. It shrinks strongly on drying to form hard lumps with low density. 9 Clayey gytt]a in the damp state has a green-grey colour. It differs from gyttja in the wet state in that it feels sticky, due to the clay content. 9 Gyttja-bearing clay in the damp state has a dull, slightly greenish, often dark colour, sometimes brown due to the presence of dy, sometimes black or with black patches due to ferrous sulphide. Gyttja-bearing clay is less elastic and less brittle than gyttja. On the soil surface, it often cracks in a characteristic cubic pattem. 9
Gyttja-bearing silt and sand are seldom encountered.
Alkali extracts of gyttja-bearing soils are light yellow or light green in colour. 9 Calciferous gyttja can be distinguished from marl by lowering a sample into a beaker containing dilute hydrochloric acid. If the sample consists of calciferous gyttja, it will retain its gyttja skeleton.
Dy-bearing soils: 9 Dy consists of a dense, black or dark brown soil which, besides dy matter, also contains peat or gyttja matter and mineral particles. Pure dy is seldom seen. On
18
Organic Soils
drying, dy retains its dark colour. In contrast to gyttja, dy is relatively inelastic and has a mushy consistency. Like gyttja, it shrinks strongly on drying to form hard, very light lumps. Sand or silt may often be mixed with dy, giving rise to intermediate forms, such as sandy or silty dy or dy with sand or silt layers. Dy-bearing clay is uncommon. Alkali extracts of dy have a dark colour.
Peat: In practice, the classification and division of peat is based on ocular inspection of the structure and consistency and on the squeezing test according to von Post (1924), (see Table 1.7). For many engineering purposes, only a coarse division is made into three types, (see Table 1.4). 9 Fibrous peat is low-humified and has a distinct plant structure. It is brown to brownish-yellow in colour. If a sample is squeezed in the hand, it gives brown to colourless, cloudy to clear water, but without any peat matter. The material remaining in the hand has a fibrous structure. (Degree of decomposition on the von Post scale; H l-H4.) 9 Pseudo-fibrous peat is moderately humified and has an indistinct to relatively distinct plant structure. It is usually brown. If a sample is squeezed in the hand, less than half of the peat mass passes between the fingers. The material remaining in the hand has a more or less mushy consistency, but with a distinct plant structure, (H5H7).
9 Amorphous peat is highly humified. The plant structure is very indistinct or invisible. It is brown to brown-black in colour. If a sample is squeezed in the hand, more than half of the peat mass passes between the fingers without any free water running out. When squeezing, only a few more solid components, such as root fibres, wood remnants, etc. can be felt. These constitute any material remaining in the hand, (H 8-H 10).
Topsoils: In the classification of topsoils, the humus content and the composition of the mineral components are stated, e.g. somewhat humus-bearing sand, humus-bearing clay (see Table 1.5). The colour of the topsoil may be darker or lighter depending on the humus content. For one and the same mineral soil, the colour will be darker with higher humus content. Even a somewhat humus-bearing sand has a fairly dark colour. On the other hand, a humus-bearing clay has about the same colour as a pure clay.
Soil Classification Table 1.4.
19
Classification of peat on the basis of decomposition on the von Post scale. After Karlsson and Hansbo (1981).
Designation
Group
Description
Fibrous peat
H1-H4
Low degree of decomposition. Fibrous structure. Easily recognizable plant structure, primarily of white mosses.
Pseudo-fibrous peat
H5-H7
Intermediate degree of decomposition. Recognizable plant structure.
Amorphous peat
H8-H10
High degree of decomposition.No visible plant structure. Mushy consistency.
Alkali extracts of topsoil humus (mull) are normally colourless or faintly brown. Marl and shell soils:
Marl and shell soils can normally be identified on the basis of colour, structure, mode of formation and place of formation. Diatomaceous soil is best identified with the aid of a microscope. Marl is almost entirely soluble in dilute hydrochloric acid. 1.3.2
C l a s s i f i c a t i o n a c c o r d i n g to c o m p o s i t i o n
Organic soils:
On the basis of composition, the organic soils are divided into three main types: gyttja, dy and peat. To these are added topsoils. Organic mineral soils and medium organic soils are classified on the basis of the content and nature of organic material, as well as the composition of the mineral material, Table 1.5.
20 Table 1.5.
Organic Soils Guiding values for the classification of soils on the basis of organic content. After Karlsson and Hansbo (1981).
Soil group
Organic content in weight % of dry material (< 2 mm)
Examples of designations
Low-organic soils
2-6
Gyttja-bearing clay Dy-bearing silt Humus-beating, clayey sand
Medium-organic soils
6-20
Clayey gyttja Silty dy Humus-rich sand
High-organic soils
>20
Gyttja Dy Peat Humus-rich topsoil
Calciferous
soils"
On the basis of the calcium carbonate content, the fine-grained soils can be classified according to Table 1.6. Table 1.6.
Guiding values for the classification of fine-grained soils on the basis of carbonate content. After Karlsson and Hansbo (1981).
Designation
Calcareous soils Clayey or silty marl Very marly or very calciferous clay or silt Marly or calciferous clay or silt
Carbonate content in % of material < 0.06 mm
>80 80-40 40-20 20- 5
For lime content below 5 %, the modifiers "somewhat marly" or "somewhat calciferous" can be used. S u l p h i d e - r i c h soils:
At present, there are no general guidelines for the classification of soil on the basis of sulphide content.
Soil Classification 1.3.3
21
Other classification systems for organic soils
A number of classification systems for organic soils are used in various countries and are based on similar grounds. Most of them, however, have not been specially designed for geotechnical purposes. Some classification systems used or suggested in context with soil mechanics in the USSR, Poland, Canada and USA are compared to the Swedish system in Fig. 1.5. O. 10.
peats
20" ,==,,
peats
30. 40.
F-
z
50"
z
60"
LU l--
0 (0
r
ch
70" 80. 90. 100.
highly organic 0
u~ >LLI
o_
__
peaty organic soils
high organic ( gyttja, dy, peat, humusrich topsoil}
organic soils
organic
Low organic
medium
organic
mineral with org. cont. Low organic mineral mineral
high ash
i, U3
C3
medium
O
Fig. 1.5
tow ash medium ash
IIIe]l[(lI
|
mineral
|
low
ash
Low
QI,
ash ~) Z
< (.9
medium ash
>,,
vl
o medium ash
mineral sediments
high
u:~
ash
mineralorganic
0co
~ ~
____m~,r.J~C
9
8
==.__ _ _ =
(9
Comparison of some classification systems for organic soils on a basis of ash contents, (Wolski 1988).
1. Konvalov (1980) (USSR) 2. Karlsson and Hansbo (1981) (Sweden) 3. Landva et al (1983) (Canada) 4. Andrejko et al (1983) (USA) 5. Classification used at the (Poland) Department of Geotechnics, WAU, based on the reports of Okruszko (1969, 1984) and Zawadzki (1970)
22
Organic Soils
1.3.4
Detailed classification of peat
According to Hobbs (1986), a detailed classification system for peat should include the following characteristics: -
T h e colour o f the p e a t in situ, which may change rapidly on exposure to air.
-
T h e degree o f decomposition (or humification).
Wetness, normally replaced by water content and, if possible, consistency limits in geotechnical engineering. -
-
M a i n constituents; fibres, wood remnants, amorphous and granular material.
-
M i n e r a l content a n d p o s s i b l e layers. The former is usually expressed in terms
of organic content, ash content or mineral content in geotechnical engineering. - Smell. Sulphides smell strongly but may be unevenly distributed. Methane requires a detector. - Chemistry. Measurement ofpH to detemame whetherthe peat should be described as alkaline or acid. -
Tensile strength. The resistance to tensile forces is an indicator of the structure
and the condition of the fibres. W h e t h e r a plastic limit test is possible or not. This is possible on fen and some transitional peats, but not on bog peat unless it is almost completely humified. - S p e c i a l characteristics including plant types if they can be identified. -
An example of such a classification as carried out in the field would be: Dark brown, moderately decomposed (Hs), wet, mainly fine fibrous PEAT with some amorphous granular matter and occasional rotten woody remnants, odourless, neutral, horizontal resistance slightly greater than vertical. Plastic limit test possible. Plant types not identified. A similar but even more detailed system was suggested by Landva et al (1983). A classification system covering some of these points was introduced by yon Post (1924). It has been much used in Europe and has lately also been introduced in Canada. The von Post system uses the following characteristics:
Humifieation (H) The degree ofhumification is graded on a scale from 1 to 10 and designated H 1 t o H10. The various degrees ofhumification are recognized as shown in Table 1.7.
23
Soil Classification Table 1.7.
Degrees of humification according to yon Post (1924), (Landva and Pheeny 1980). D e g r e e s of h u m i f i c a t i o n
Degree of humification
Decomposition
Plant structure
Content of amorphous material
H! H2 Ha
None Insignificant Very slight
Easily identified Easily identified Still identifiable
None None Slight
H4
Slight
Some
H.s
Moderate
H,
Moderately strong
H7
Strong
Not easily identified Recognisable. but vague Indistinct (more distinct after squeezing) Faintly recognizable
H~
Very strong
Very indistinct
High
H9
Nearly complete
Almost unrecognisable
H~,
Complete
Not discernible
Considerable Considerable
High
Material extruded on squeezing (passing between fingers) Clear. colourless water Yellowish water Brown. muddy water; no peat Dark brown, muddy water; no peat Muddy water and some peat About one third of peat squeezed out" water dark brown About one half of peat squeezed out; any water very dark brown
Nature of residue
Not pasty Somewhat pasty Strongly pasty
Fibres and roots more resistant to decomposition
About two thirds of pcat squeezed out; also some pasty water Nearly all the peat squeezed out as a fairly uniform paste All the peat passes between the lingers; no free water visible
Water content (B) In the field, the water content of the peat is estimated on a scale of 5 degrees. B 1 represents air dried peat, B 2 somewhat dried peat, B 3 peat with normal water content, B 4 v e r y w e t peat and B 5 largely free water with slime. In the case of actual water contents, Landva and Pheeney (1980) have later suggested the following range s B 2 less than 500 %; B 3 500 to 1000 %; B 4 1000 to 2000 % and B 5 greater than 2000 %.
Sedge fibres (F) The content of fibres and stems from sedge is given as F. Roots should not be included. F 3 denotes a peat entirely or mainly consisting of such fibres, F 2 a high but not predominant fibre content, F 1 a low fibre content and F 0 no macroscopically discernible fibres. Landva and Pheeney suggest that also fibres from mosses and shrub rootlets should be included, provided that they are properly specified as F(H) or F(S) for Hypnum and Sphagnum mosses and F(N) for shrub rootlets.
24
Organic Soils
Root threads (R) The content of root threads, R, was given as R 3 for almost pure root mat (felt), as R 2 for high and R 1 for low contents of root threads and R 0 ifthe content is nil. If the species of root threads can be determined, it should be given in brackets after the R symbol. Landva and Pheeney suggest that "sedge fibres" and "root threads" should be exchanged to "fine fibres" and "coarse fibres". The division between fine and coarse should then be at a diameter or width greater or smaller than 1 mm. No distinction should be made between fibres, stems and rootlets, but the plant origin should, if possible, be given in brackets after the F and R symbols.
Wood remnants (V) The content of wood remnants was given by the symbols V3, V2, V 1 and V 0 along the same lines as for symbols F and R. The species of wood and its consistency should, if possible, be given in brackets after the R symbol. Landva and Pheeney suggest a division of the V symbol into W for wood remnants and N for shrub remnants. Landva and Pheeney further suggest that the peat should be designated according to its plant origin:
Plant types Bryales (moss) Carex (sedge) Equisetum (horse tail) Eriophorum (cotton grass) Hypnum (moss) Lignidi (wood) Nanolignidi (shrubs) Phragmites Scheuchzeria (aquatic herbs) Sphagnum (moss)
= = = = = = = = = =
B C Eq Er H W N Ph Sch S
Designation With few exceptions, natural peats consist of a mixture of two or more plant types. The designation adopted is to list the plant types in the descending order of content, i.e. the first symbol represents the principal component. For example, a peat classified as ErCS consists mainly of Eriophorum remnants, while the content of Carex remnants would be lower and that of Sphagnum remnants relatively low. The designation is omitted when plant types cannot be identified.
Soil Classification
25
Hobbs (1986) suggests that the von Post classification system should be further extended through the addition of symbols for organic content, anisotropy, smell, plasticity and acidity as required:
Organic content (N)A It is not possible to estimate the organic content in the field unless the peat is obviously clayey, when the von Post humification test would not be realistic. Following ignition loss determinations, the organic content may be graded as follows: N 5 greater than 95 % organic matter; N 4 95 to 80 %; N 3 80 to 60 %; N 2 60 to 40 %; N~ 40 to 20 %. A) The suggested notation N is unfortunate as it may be confused with the notation for shrub remnants suggested by Landva and Pheeney (1980).
Tensile strength (TV and TH) The tensile strength in the vertical and horizontal directions may be judged by pulling specimens apart in these directions. The following scale may be used" T o zero strength; T 1 low, say less than 2 kN/m~; T 2 moderate, say 2 to 10 kN/m~; and T 3 high or greater than 10 kN/m ~.
Smell (A) The smell, which is an indication of fermentation under anaerobic conditions, may be scaled as follows: A 0 no smell; A 1 slight; A 2 moderate; A 3 strong. Note, methane, CH 4, the main indicator of anaerobic activity, has no smell. If specially detected, it should be reported.
Plasticity (P) Plastic limit test possible P1, not possible P0
Acidity (pH) Acid pilL; neutral PH0; alkaline pH H. An example of the use ofthe von Post classification in the field extended according to Hobbs is given below:
Soil description- Dark brown, oxidizing to black, moderately decomposed H 5, mainly fine fibrous PEAT with some coarse fibres and amorphous material. Low vertical tensile strength, moderate horizontally. No smell. Plastic limit test possible. Genera not identified.
26
Organic Soils
After determinations of ignition loss and pH, this becomes H5 B 2 F 3 R 1 V 0 /
N3 TV1 TH2 Ao P1 PHo
Lvon Post
Extension proposed by Hobbs
The classification in the example can be further elaborated if the modification of the von Post classification suggested by Landva and Pheeney (1980) is used. Peats are composed of the partly decomposed remains of plant communities containing varying morphology and texture. It is this structure that affects the retention or expulsion of water in the system, provides tensile strength and ultimately differentiates one type of peat from another. The von Post classification attempts to describe peat and the structure in quantitative terms and the extended system suggested by Hobbs is designed to provide a means of correlating the types of peat with their physical, chemical and structural properties.
1.3.5
Other classification systems for peat
The Russian handbook for peat (Lazarev and Kortjunov 1982) uses percentage of humification instead of the von Post scale. The following relation between the percentage ofhumification (R) and the von Post scale (H 1to H10) is given by Schneider (1967)" H (von Post)
R%
1
5
2 3 4 5 6 7 8 9 10
10 15 20 25-30 35 45 55 65
Radforth (1969) proposed a classification system for Canadian peat (muskeg) based on the structure of peat rather than its botanical origin. The system is described in Muskeg Engineering Handbook (MacFarlane 1969). According to Radforth, this approach makes it easier to classify structure and also leads to a better
Soil Classification
27
basis for estimating mechanical properties than a purely botanical classification system. Radforth divides peat into the main categories amorphous- granular, fine-fibrous (fibre diameter _< 1 mm) and coarse-fibrous according to the main character of the structure. Besides these main features, the concepts "woody" and "non-woody" are used to characterize the peat. A subdivision can be made into 17 categories, Table 1.8. Table 1.8. Classification of peat (The Canadian system after Radforth 1969). Predominant characteristic Amorphous-granular
Category 1
2 3 4 5
~~ 7
Fine-fibrous
8 9 10 11
Coarse-fibrous
12 13 14 15 16 17
Name Amorphous-granular peat Non-woody, fine-fibrous peat Amorphous-granular peat containing non-woody fine fibres Amorphous-granular peat containing woody fine fibres Peat, predominantlv amorphous-granular. containing non-woody fine fibres, held in a woody, fine-fibrous frame work Peat, predominantly amorphous-granular containing woody fine libres, held in a woody, coarse-fibrous framework Alternate layering of non-woody, linefibrous peat and amorpht~us-granular peat containing non-woody fine fibres Non-woody, fine-fibrous peat containing a mound of coarse fibres Woody, fine-fibrous peat held in a woody. coarse-fibrous framework Woody particles held in non-woody, linefibrous peat Woody and non-woody particles held in fine-fibrous peat Woody, coarse-fibrous peat Coarse fibres criss-crossing fine-fibrous peat Non-woody and woody fine-librous peat held in a coarsc-librous framework Woody mesh of fibres and particles enclosing amorphous-granular peat containing fine fibres Woody, coarse-fibrous peat containing scattered woody chunks Mesh of closely applied logs and roots enclosing woody coarse-fibrous peat with woody chunks
Organic Soils
28
The classification is facilitated by useful photographs of the various categories in the Muskeg Engineering Handbook. In the Radforth system, no mention is made of colour, wetness, degree ofhumification or organic content and these characteristics have to be supplemented in some way. Landva and Pheeney (1980) and Hobbs (1986) have found that the system is not generally applicable, even if it is useful for large areas in Canada. Radforth suggests that also vegetal cover and topsoil should be classified on the basis of structure rather than gefiesis. The vegetal cover is divided into nine different classes, Table 1.9. Since there is often more than one type of cover, the classification can be combined, e.g. ADE. The symbols should list the types of cover in descending order of amount, i.e. the first symbol represents the principal component, the second symbol represents the next largest component and so on. Types that cover less than 25 % of the area are not considered. Table 1.9.
Properties designating nine coverage classes (The Canadian system after Radforth 1969).
Coverage type (class)
Woodiness vs. nonwoodiness
Stature (approximate height)
A B
Woody Woody
15 fl or over 5-15 fl
C D
Non-woody Woody
2-5 fl 2-5 ft
Texture (where required)
Woody Up to 2 fl Non-woody Up to 2 ft G
Non-woody Up to 2 ft
H
Non-woody Up to 4 in. Non-woody Up to 4 in.
Leathery to crisp Soft or velvety
Growth habit
Tree form Young or dwarf tree or bush Tall, grasslike Tall shrub or very dwarfed tree Low shrub Mats, clumps or patches sometimes touching Singly or loose association Mostly continuous mats Often continuous mats, sometimes in hummocks
Soil Classification
1.3.6
29
Geotechnical classification of peat
In many cases, only a simplified version of the von Post scale for degree of humification is used in geotechnical engineering for the classification of peat, together with the normal geotechnical parameters used for all soils, such as water content, consistency limits, organic content, bulk density etc. This procedure has been recommended by Helenelund (1975) and Karlsson and Hansbo (1981), among others, and is normally used in Fenno - Scandinavia. In Canada, the Radforth system has been used to some extent and in other countries regional classification systems, often related to von Post in one way or another, have been used. In Scandinavia and Canada, as well as elsewhere, strong recommendations have been made to introduce the full von Post classification system also in geotechnical engineering, with or without suitable modifications and extensions, e.g. Landva and Pheeney (1980), Hobbs (1986), Landva et al (1983), Carlsten (1988).
1.4 REFERENCES Andrejko, M.J., Fiene, F. and Cohen, A.D. (1983) Comparison of ashing techtuques for determination of the inorganic contents in peats. Testing of peats and organic soils. ASTM Special Technical Publication 820, pp. 5-20.
Carlsten, E (1988) Geotechnical properties of peat and up-to-date methods for design and constrtiction on peat. State of the Art Report. International Conference on Peat. Tallinn.
Godwin, H. (1978) Fenland; Its Ancient Past and Uncertain Future. Cambridge University Press.
Granlund, E. (1932) De svenska h6gmossarnas geologi. Deras bildningsbetingelser, utvecklingshistoria och utbredning jfimte sambandet mellan h6gmossbildning och f6rsumpning. (The geology of Swedish raised bogs). Swedish Geological Survey. Ser. C. Nr. 373. Hallden, B.E. (1961) Allm~n Geologi. Kompendium Nr. 83. Tekniska h6gskolans studentkfir. Stockholm.
Helenelund, K. V. (1975) Geotechnical peat investigations. Proceedings, Baltic Conference on Soil Mechanics and Foundation Engineering, 1. Gdansk. Vol. 1. pp. 105-123.
Hobbs, N.B. (1986) Mire morphology and the properties and behaviour of some British and foreign peats. The Quartemary Journal of Engineering Geology. Vol. 19. No. 1.
30
Organic Soils
Karlsson, R. and Hansbo, S. (1981) (in collaboration with the Laboratory committee of the Swedish Geotechnical Society). Soil classification and identification. Swedish Council for Building Research. D8:81. Stockholm. Kivinen, E. and Pakarinen, P., (1980) Peatland areas and the proportion of virgin peatlands in different countries.In: 6th International Peat Congress, Duluth, Minnesota. Konovalov, P. A. (1980) Ustojstvo fundamentov na zatorfovannych gruntach. Moskwa. Stroizdof. Landva, A.O., Korpijaakko, E. O. and Pheeney, P. E. (1983) Geotechnical Classification of Peats and Organic Soils. Testing of Peats and Organic Soils. ASTM Special Technical Publication 820. pp. 37-51. Landva, A. O. and Pheeney, P. E. (1980) Peat fabric and structure. Canadian Geotechnical Journal, Vol. 17. pp. 416-435. Lazarev, A. V. and Kortjunov, S. S. (Editors) (1982) Spravochnik po torfy, Moskva Nedra, 1982. MacFarlane, I. C. (1969) Muskeg Engineering Handbook. University of Toronto Press. Magnusson, N. H., Lundqvist, G. and Regnell, G. (1963) Sveriges Geologi (Swedish Geology). Svenska Bokforlaget. Stockholm. Okruszko, H. (1969) Powstawanie mulrv i gleb mulowych. Roczniki gleboznawcze, Vol. 20, No. 1. Okruszko, H. (1984) Zaktualizowana klasyfikacja grtmtrw organicznych dla potrzeb budownictwa ziemnego. Mat. niepublikowane. yon Post, L. (1922) Upplysningar rOrande Sveriges Geologiska UndersOknings torvmarksrekognosering (Information on SGU peat inventory). Swedish Geological Survey. Ser. D. yon Post, L. (1924) Das genitische System der organogenen Bildungen Schwedens. Mrmoir. Nomenclat. et Classific. Sols 1924 pp. 287-304. Comit6 Intemat. Prdol. 4. Commiss. Nr 22. Radforth, N. W. (1969) Classification of muskeg. In: MacFarlane, I. C. (ed.) Muskeg engineering handbook. Canadian Building Series. University of Toronto Press.
31
Chapter2
Site Investigations U. Bergdahl, Swedish Geotechnical Institute
2.1
GENERAL
The results of a site investigation for an embankment project on organic soils must contain all the basic data for the design and construction of the embankment. Thus it should be possible to evaluate the following: 9 Local and total beating capacity of the soil. 9 Amount of settlement and time-settlement distribution. 9 Need for and design of reinforcement or improvement. 9 Design parameters for the embankment. 9 Choice of construction method and machines. 9 Costs for foundation and construction. 9 Influence on neighbouring area or structures. The field investigations must have such an extent that the whole area involved is covered regarding stability, settlements and the influence on neighbouring objects, Fig 2.1.
Investigation
area
_j
Fig. 2.1. Example of the area which it is necessary to investigate close to a creek.
Site Investigations
32
The site investigations are often performed in stages depending on the need for information in the different phases of planning, design, construction and maintenance of the embankment.
2.2
MAPPING,GENERAL SURVEY
At an early stage of a localization for a large embankment, it is important to have an overview of the soil* and groundwater conditions in the whole area involved. For this purpose, the aerial photo interpretation method may be recommended. Such an investigation is normally divided into three main parts (Viberg 1984): 9 inventory phase 9 aerial photo interpretation 9 field inspection
The inventory phase contains a compilation of existing knowledge from the areas of interest, e.g. maps and earlier soil investigations. From these sources, the aerial photo interpreter can obtain an overview of the geological and hydrological conditions of the area. The most common maps are topographical, geological, hydrological, economic and engineering geological maps. As most maps indicate the conditions close to the ground surface, previous borings in the area provide important additional information for understanding the soil profile at the site. Data from soil investigations can be obtained from different authorities, such as the national geological survey, road and railroad administrations, regional planning and building departments, geotechnical and geological consulting firms etc.
Aerialphoto interpretation is based on the fact that a number of indicators can be recognized in different areas of the site. These indicators may be topography, contours, vegetation, texture, colour or grey tone (black and white air photographs) and land use. Normally, the topography changes when the geological conditions change. Therefore, aerial photo interpretation should always be made in a threedimensional model using a stereoscope. During interpretation, areas of similar appearance as well as the limits between different areas are defined. The following areas can normally be identified: 9 rocky outcrops or rock covered by 0.5-1.0 m of soil. 9 tills or moraines 9 coarse sediments (sand, gravel) *) The word "soil" here means earth material
Mapping, General Survey
33
* fine sediments (silt, clay) . organic soils (peat, gyttja, mud). Areas of fine sediments can often be subdivided into areas of shallow and deep silt and clay layers. In areas with organic soils, the following indications can be recognized: 9 The ground surface is often fiat and nearly horizontal. The only exception is raised peat bogs, which are dome shaped. .
The vegetation is composed of plants requiring a great amount of water. Normal plants in the area often have poor growth.
9 The texture in peat areas is often tufty with closely spaced ditches. However, in areas of gyttja the soil often cracks, which makes the ditches unnecessary. 9 The colour or greyish tone is brown to black or dark to black in cultivated areas. However, dried gyttja is often light-grey. .
Organic soils are often found in the lowest part of an area, except for the peat bogs mentioned above. 9 Occasionally, peat bogs are partly excavated for fuel or soil improvement.
.
In populated regions, areas with organic soils are often cultivated also when they are occasionally flooded.
Good aerial photo interpretation requires overlapping aerial photos of high quality. Black and white films as well as colour films or infrared films can be used. The optimum scale of photographs for this purpose has been found to be 1:10.0001:15.000. In this scale, both an overview and a picture of ground detail can be obtained. After interpretation, it is necessary to make afield inspection on the site in order to check the composition and limits of the different areas indicated. For this purpose, lightweigt sounding and soil sampling equipment is used.
2.3
SOIL LAYER SEQUENCE
After the general survey of the area in question, it is important to obtain mformation on the soil layer sequence. Especially in organic soils, the characteristics may vary greatly from the ground surface to the denser bottom layers. To obtain a relevant profile of the soil layers, soil radar, some type of penetration testing and sampling can be used. Experience of soil radar is so far limited, except for investigation of peat layer thickness. Experience of penetration testing is good for estimation of the soil profile, but the possibilities for evaluation of the soil properties are
34
Site Investigations
very limited. Soil sampling is a necessity in organic soils, both for detailed soil identification and classification and also for determination of properties in laboratory investigations. 2.3.1
Soil
radar
Soil radar can be used to investigate the thickness of soil strata with quite different electrical resistivity or, more correctly, dielectric constant at shallow depths (< 10 m). By transmitting electromagnetic waves through the soil from the ground surface, different layers can be identified (Johansson 1987). Radar stands for radio detection and ranging. The georadar equipment consists of an antenna for frequencies between 80 and 1000 MHz, receiver and control unit. A low frequency antenna is used for depths greater than 5 m. The control unit contains an oscilloscope, amplifier, filter and recorder with plotter.
During the investigation, the antenna is moved over the ground surface with a speed of 1 to 10 kin/h, while the reflection of the electromagnetic waves is recorded on radargraphs, Fig. 2.2. Ground
surface
Road
~9 .~_~.2:'~: - :.:~ .- -- , ~ t ' , , . . . . 9 -::r." - ~ , "~ ~, ; ; ' .--- ---~';--...'---t-'~'_ ~ " - -=. . .9~ " ~ .=.,~-__--== = ~.~2-". -el. .. . . . .. .. .
--
.'::
- --
---
."
. . . . . .
-
9--..~e'Z-':.;...m,=: --* .x ~-,+w--
., . .9 ." " .
~. . . . . . .
,/'~-~,,
9
9 ,~: - ~ : -__ ~ : . . - ~ - - --~,---=, - ~ , ~ . .. . ~ . . -~ -~ 9"r ~ ~ . . . ..-~__ . . . ~: - . . L _,_ ~3 ._. . _ _ : .... 9 ...-
"- .~F..~---,,,. ~: ~ " .. r. -. -.- - - , - ~ . -
9" . . . ~ ~
~=~..~,.~-,~.4"-~.~,~..-,_--_-=-.:..'.."
....
. . . . . . . . . . .
Peat
"~
9
,-# "."
.,, o | * .... 9
;.~ .,..:.:...,r162 '
9 9 I.: t
'
99
" ' ' '
..
$'
Fig.
.. : , ~ , : 9 9-o "--''..~.., ' ....
~ "'"w'
2.2.
~ ,.~.
,~. ~
9 9. , .'." "'~"
"~"
Example
-. ..,
: ~,: :'. ~ ; . : "
, : ,,,,~." 9
Ti[[ ."
' :::
, * ; ~9 ' , ' - : ,
,.e :. " ." ~
~ ~ , -.
...' ." . . . . . '
.~.,~--,,.,;,, ,, " .,
of radargraph
:'. 9
9. .
from
peat
. .,,
:p~
~. .
~,--::. 9.,.: .4: -.,,.
.. ~
soil
9
, ;;.~=~y.~,~ ' :,~;-.,'.'.'-~"~"
:~'-~:~.-
9 9
' , ~9 - , . "9 . -,."..", , ,
~-,
k
:...# :.. -. ."...;. - . : ~ . ~ - - : : . . . .
. :..
9. . . . . . . .
. , . ,
.......
with
. . . . . . . . . .9. . . . . . . . . . . . .
9
" ..
.
'....r.;--.. ,.: : ..
. ,..
..
~ .
.
.
. , .{.'..,..
"':i
,....
underlying
.
. ,
". JI ~ .. , - t - ~ 9 . . I " -9 --. :. , " .
.-7 , . , ~ '.i l.e,..
-',&-.
t.
glacial
""
" .. :'." % " " :
.....
."
.--%
till.
For interpretation of the radargraphs, it is important to identify the different soil layers on the site by penetration testing and sampling, and to have thorough knowledge of the geology and groundwater conditions in the area. In the evaluation of the radargraphs, it is important to consider that the ground surface is shown as a horizontal line also when the terrain is hilly. Investigations (Bjelm et al 1982) indicate that it is possible to measure the thickness of organic soil layers, especially peat, and the depth to firm bottom layers or rock with the soil radar.
Soil Layer Sequence
35
Soil radar has been used for surveying peatlands. In Sweden, it has also been used to establish the thickness of existing roads over peat bogs. The method is quick and has shown good agreement with results obtained with sampling technique. It is essential to know the thickness of an existing road when widening and strengthening roads with low bearing capacity. Fig. 2.2 shows results from such measurements in Sweden. The radargrams show a clear boundary between the road material and the peat. The measurements by impulse radar were followed by soundings and sampling in the embankment. Very good agreement was obtained in this case between depths estimated from radar and actual depths. When widening existing roads, this method provides information on how the peat has reacted under a load, and it is thereby possible to gain experience from what could be considered as a very durable test embankment. In Finland and in Sweden, soil radar is also used to determine peat thickness and the topography of the mineral soils beneath the peat. In Finland, use is also made of a "radiowave moisture probe" to determine the dielectric constant, water content and dry matter content of peat. When comparing the measuring data with conventional methods in the laboratory, very good correlations have been obtained.
2.3.2
Penetration testing
Penetration testing or sounding is normally used to determine the thickness of different soil layers and the relative density or stiffness of the soil. A number of penetration testing methods are available. For the investigation of soft organic soils in general, static methods or light dynamic methods are preferred. A proposal for international Reference Test Procedures for four different methods has been produced by the ISSMFE Technical Committee on Penetration Testing of Soils and has been published by the Swedish Geotechnical Institute (Bergdahl 1988) among others. The four methods are: 9 Cone penetration test (CPT) 9 Standard penetration test (SPT) 9 Dynamic probing (DP) 9 Weight sounding test (WST)
The conepenetrometer was developed in the Netherlands at the beginning of the thirties. Originally, the cone penetrometer was mechanically operated, but today most penetrometers measure the resistance electrically. The penetrometer has a conical tip with a cross sectional area of 1000 mm 2, Fig. 2.3. Normally, a friction sleeve is mounted above the base of the cone with a shaft area of 15000 mm z. In addition, a pore pressure transducer may be mounted in the penetrometer tip with a
s filter located just above the cone base (a piezocone). Thus it is possible to measure simultaneously with penetration the cone resistance, local skin friction and generated pore pressure. Measurement of the generated pore pressures is of special importance in soft fine-grained soils, where high excess pore pressures develop during penetration. A new standard for the cone penetration test with special consideration to tests in very soft soils has been produced in Sweden (SGF 1993). Thrustmachine
I I
'1
I
I
-"----I Probe
Friction sleeve
I j
e--j
031
Fig. 2.3. Set-up for cone penetration using the piezocone principle.
The penetrometer is pushed into the soil with a constant rate of penetration of 20 mm/s using a thrust machine. The measured resistances are recorded continously. An example of resistance and pore pressure curves is shown in Fig. 2.7. The cone penetrometer is the most accurate sounding method and gives the best information on soil layer sequence and soil characteristics. When pore pressure measurement is added, it is possible to indicate also small variations in the soil. If the measured resistance is to be used to evaluate the soil characteristics, especially in soft soils, the measured cone and friction resistances must be corrected to total resistances according to Fig. 2.4. In estimating the soil layer sequence, the charts in Fig. 2.5 may be useful. However, the observations in organic soils are limited. According to Robertsson et al (1986), the corrected cone resistance qT, the friction ratio Rf and the pore pressure ratio Bq may be used to evaluate the type of soil penetrated.
37
Soil Layer Sequence
I
A N -Net area A T -Total area F c - M e a s u r e d cone resistance F T - Total cone resistance u
- Generated pore pressure
FT :F c +u(A T -A N)
Principle of correcting measured cone resistance to total resistance for the piezocone.
Fig. 2.4.
RU
(2.1)
~
qT U - Uo B
(2.2)
q _.. qT- (Yvo
where f qT = u = uo = ~vo =
unit skin friction resistance total cone resistance measured pore pressure static pore pressure total overburden pressure
The charts in Fig. 2.5 are to be used in parallel and may give contradictory indications of soil type. Another way of estimating soil type using similar parameters has been suggested in Sweden (Larsson 1992). In this method, the net cone resistance (qx- CYvo)and the total unit sleeve friction (fx) are normalized against the effective overburden pressure (~'vo) and the net cone resistance respectively. They are then plotted in the chart in Fig. 2.6 a. If the soil according to this chart is classified as "clay or organic soil", the second chart in Fig. 2.6 b is used to obtain a more detailed classification for this type of soil. In practice, it is very difficult to separate clays from organic soils on the basis of results from cone penetration tests alone.
Site Investigations
38
,,oo i oj/~.l, I
I
I
i
I
.,~ 8 z
10
0.1
I
I
11
Zone
1 -.-~....
3 _.....-1
'
0
<,
~z (/) i'Y
I
S'76
10
L
'
g
a
FRICTION RATIO, %
~- 100
~
i
9
z
~
/
I
11~
,
2
Ovo~
I
' ' '1
----u U
-
6
7
8 9
10
Uo
=7
12
8
~
5
Soil Behaviour Type sensitive fine grained organic material clay silty clay to clay clayey silt to silty clay sandy silt to clayey silt silty sand to sandy silt sand to silty sand sand gravelly sand to sand overconsolidated or cemented very stiff fine grained soil overconsolidated or cemented sand to clayey sand
o.
0
,
0./..
0.8
1.2
PORE PRESSURE RATIO, Bq
Fig. 2.5. Evaluation of soil type according to Robertson et al (1986).
39
Soil Layer Sequence
2001Very
dense
Sand
I (2.5)
150"1 Dense (2.0)
ID
-e ~oo IO i
Silt
I-O"
Medium dense
Very dense (2.1)
Clay and organic soil. Mediu___...mm dense Loose L
O.iO
0.05 fT / (qT - 6vo ) ,Heavily Overconsolidated or 4000-overcon_ very silty clay solidated
>~
.
3000-
Very stiff
2000"
Stiff
-
Normally Low plastic ;consolidated ~andlor highly clays or ~tsensitive clays slightly 't ov~-consotidat~ silty clays
I.-
. . . . .
Medium stiff . . . . .
-0.2
Fig. 2.6.
t t
__
1000-
0 I
"---.
.
.
.
,
. . . . . . . . . . . . . . .
.
......
0
.
0.2
i
. . . . .
1
--tI t
I -"-l-
- - = . .. .. .. .and . . . . . .- . . . . .Soft . . . . . . . . ~ _E -Gzttj.qs I - -. . . . . . 1' . . . . Very soft organtc clays,
~'*..,
. .
. . . . . .
0.4 0.6 Parameter Bq
I -,4-. . . . . . . . . i___ ~ ..... i ........ 0.8 1.0
1.2
a) Chart for evaluation of soil type according to Larsson (1991). b) Special classification chart for clay and organic soil (Larsson 1992).
40
Site Investigations
According to Landva et al (1986) the use of the static cone penetration test in fibrous peat is questionable owing to the fact that negative pore pressure is induced and that the peat in front of the cone is compressed before failure.
.10 0
CONE
RESISTANCE
RESSUREin
HPa
IN H P a 20
~,0
60
LOCAL FRICTION ft 0.0 0.2
IN H P a 0/,
FRICTION 0
R A T I O R f IN "/,. S tO
02
GL=S 09re.NAP
_ . , . u ,~-_
o. < z
- ~_--..=_~ - _ _.,_ _ _
.-= = == .. .=== . . . . . .
o I.cI ,., - I 0 n, uJ u. IL
---- "----
-10
-~__Z . . . .
Q: E
_z Q. 121
20
-20
-30
-30
Paezo cone nr : 1 0 / 1 - 3 5 S size of t i l t e r : h e i g t h 3 . 0 r a m . t h i c k n e s s 3.0ram l o c a t i o n of f i l t e r in t h e c y l i n d r i c a l e x t e n s i o n o f t h e cone m a t e r i a l of f i l t e r : s i n t e r e d s t a i n l e s s s t e e l " b e f - o r e ' t e r , t-T a f t e r test I capacity i Z. . . . . . . d,ng - ~ n e o~ ~ ! -o.o,o NPa I J o _ o _ - - E ~
i'~;~r~---~l~-;~i -~
DEL F T GEO TIcCliNICS WONINGEN TE HAASSLUIS CONE PENETRA T/ON TEST
Fig. 2.7.
1 -~.~6~ MPa lo.7 MP~__l
GO O2(RE)
date of t e s t
hme
: :
Remarks : fr,cr,on reducer : not apple,d abnormal interruptions : none observations : no s p e c i a l o b s e r v ~ t , o n s fall/excavation : old f i l l /.m t h , c k n e s s i n c l i n o m e t e r : no r e a d i n g s t a k e n c o n d i t i o n of p u s h r o d s / p e n e t r o m e t e r lip a f t e r t e s t : good waterievet in sounding h o l e : hole c o l l a p s e d near s u r f a c e backfilhnq - none
87-02-19 14-15 hrs
E x a m p l e o f t h e p r e s e n t a t i o n o f C P T t e s t results, a c c o r d i n g p r o p o s e d r e f e r e n c e test p r o c e d u r e , ( B e r g d a h l 1 9 8 9 ) ,
to t h e
The Standard Penetration Test (SPT) is the most common penetration testing method in the world. It is used to determine the bearing capacity of both piles and shallow foundations. As it is a heavy dynamic penetrometer, it is difficult to obtain minor variations in the characteristics of soft organic soils. However, disturbed samples are also obtained with this test procedure.
Soil Layer Sequence
41
The Standard Penetration Test is performed with a driving rig and a 63.5 kg hammer falling 760 mm onto the anvil fixed to the driving rods. The penetrometer itself contains a thick-walled sampler and the driving rods. The sampler has an inner diameter of 35 mm and a sample length of 457 mm, Fig. 2.8.
a]
Lifting head
b)
~ -~ ~
63.5kg hemmer----! I I .~,r
\\
V . m
o
,~ ~
,~ a
E C,4
L.
SPT sampler~~
Fig. 2.8. Arrangements (a) and cross section (b) of the Standard Penetrometer.
The SPT is normally performed in a casing. The sampler is driven into the soil by blows of the hammer. At each test level, there is first a seating drive of 0.15 m, followed by the actual test in the following 0.3 m of penetration. The number of blows required for 0.3 m of penetration is recorded as the penetration resistance, Fig. 2.9.
Site Investigations
42 7_
SPT +9.35 S< nd ;W
2 §
--z~
~735
P~ at
S~ind
i"~ "" z/i/i//2
Ground water level
68
"
S(Ind
i " ~ if/i//, "Ilia
50
!/i//ilL
~'/ii1111.
68
"lilllli,
Zlllll~
7i/.~
(50)cone
i
!
20
Number of blows outside range of scale
~lliillz
0
Fig. 2.9.
surface
79.09.05. Dote of ground water measurements
Sand
Gn vel
Elevation ground
GW§
~.o~ ....
79. Sand /
Bore hole number
Uncertain blowcount
n
/+0
60 bl/O.B0m
Example of results from a Standard Penetration Test.
The dynamic probing method includes a number of penetrometers with different weight of hammer, height of fall, cone and rod diameters. In the proposed reference test procedure, four types of dynamic probing methods are presented, Stefanoff et al (1988). As the driving energy for three of the penetrometers is rather high, they are normally not suitable for the indication of soil characteristics in soft organic soils. Only the Light Dynamic Probing method (DPL) can indicate small variations in the soil resistance. However, in deep and stiff organic soil layers, also the heavier versions of the dynamic probing method may be useful, especially for indication of depths to denser bottom layers. In this report, only the light dynamic probing method is presented. Information on the heavier method can be found in the report by Stefanoff et al (1988). The light dynamic penetrometer may be either manually or mechanically operated. It consists of a conical tip with an apex angle of 90 ~ and a cross sectional area of 10 cm z, Fig. 2.10. The rods should be o 22 mm hollow rods. The penetrometer is driven into the soil by blows from a 10 kg hammer falling freely from 0.5 m height. During the test, the number of blows per every 0.1 m of penetration (N10) is recorded. The results of such a penetration test are shown in Fig. 2.11.
43
Soil Layer Sequence
BLOWS PER 0.I m. N
OPL
10
20 ELEVATION.
12.3mNAP
E 1.0
6ROUNOWATER LEVEL:
I0.] mNAP
z 1.5 -'Jr).a. 2.0
PROJECT-NUHBER:
ZW "/
OAIE OF TESf
80-03-12
0.5
TYPEOFPENETROHETER: R DPL
O 2.5
Fig. 2.10. Scheme of cone for the light dynamic penetrometer.
:
NUtIBEROF TEST:
13
LOCATION.
XBOURG
Fig. 2.11. Example of presentation and rod from a DPL test, (Bergdahl 1989).
The weight sounding test was developed in Scandinavia in about 1915 as a tool for investigating the risk of landslides, mainly in soft soils. The weight penetrometer consists of a screw-shaped point, a number of weights (5, 10, 10, 25, 25 and 25 kg) a number of rods e 22 mm and a handle. Fig. 2.12. _~C~~L,~.~_~
I
=4s
-'~ To p
,.o?~ ~ - w ~ , g h t ,
,o'om
II
Pt
Weights 2,5 kg
,0 kg
E z 3 -xl-Q- S uJ c)
W S T 22
i ~ ~,~
,b(sp,,8o ~-i
4
(Rubber) 010 k~d
W S T 22
ht/GL2 m
0.80m J ~t
II [ i ~ R o d ~
22 rnrn
-~, ~ S c . r e w
point
V
Fig. 2.12. Details of the manually operated weight penetrometer, (Bergdahl 1989).
P! fb(.~oe00}
201,060
kN. ht/0.2
m
W E I G H T SOUNOING TEST. 22
mm
RODS
NUMBER OF HALFTURNS PER 0.2 m OF PENETRATION DRY CRUST OF ClAY PREBO~ING TO THIS LEVEL WITH 8Omm OIAM AUGER
DIAGRAM TO THE LEFT INOICATE LOADS APPLIEO IN kN
Fig. 2.13. Example of test results from a weight penetrometer test.
Site Investigations
44
The weight penetrometer is used as a static penetrometer in soft soils where the penetration resistance is less than 1 kN. If the penetrometer does not sink with this load, the penetrometer is rotated and the number of half turns for every 0.2 m of penetration is recorded. Nowadays, both petrol driven and hydraulic machines are used for the weight sounding test. In this equipment, the penetration resistance during the static phase is measured with a dynamometer. The penetrometer resistance from a weight sounding test is presented in the diagram shown in Fig. 2.13, with loads to the left and the number of half turns per 0.2 m of penetration to the right. The results of the weight sounding test are used to obtain a continuous soil resistance profile and indications of the layer sequence and the lateral extent of different soil layers. They are also used to determine the relative density of cohesionless soils and to estimate the relative strength of cohesive soils. Another type ofpenetrometer is the lime column penetrometer, which is used to check the homogeneity and strength increase of lime columns, c.f. Chapter 10. This penetrometer is a static mechanical penetrometer with a ~ 50 mm tip, which is designed to follow the centre hole in the column. About 450 mm from the tip the penetrometer is provided with 400 or 500 mm wide wings (depending on the column diameter) in order to cut the column into two parts, Fig. 2.14. 5001L.00
mm
.
10 o r 15 m m
Section A- A
L15 or 20 mm
36ram ~-
r~
50 mm
Fig. 2.14. Lime column penetrometer.
45
Soil Layer Sequence
The penetrometer is pushed into the column at a constant rate of penetration, 20 mm/s, and the total force is recorded. The shear strength of the column is taken as 1/ 10 to 1/11 of the unit penetration resistance. Fig. 2.15. The evaluation may be calibrated in undisturbed soil beside the lime columns.
Orga
-1-
da
o
Clay --
Clay
on/5, o, mn_l o lOO 200 Shear strength (kPa}
Fig. 2.15. Results of lime column penetrometer test in a column and, for comparison, also in the unstabilized clay.
The penetration test normally comprises 0.5-2.0 per cent of the total number of columns, cf Chapter 10. 2.3.3
Dilatometer
testing
Dilatometer testing is a relatively new in situ method of determining soil stratigraphy and soil properties. It was developed in Italy by Marchetti (1975) and has rapidly gained worldwide acceptance. The dilatometer consists of a spade- shaped instrument, which is pushed into the ground with a constant rate of penetration of 20 mm/s. The instrument, Fig. 2.16, is supplied with a flexible membrane on one side. At regular intervals, usually every 0.2 m, the penetration is stopped and the membrane is expanded and pushed out into the soil by the application of a regulated gas pressure on the inside of the membrane. The pressures required to overcome the
46
Site Investigations
earth pressure and make the membrane lift off from its base (P0) and the pressure required to expand the membrane 1.1 mm into the soil (P l) are recorded. The membrane is then deflated and the penetration is resumed. By using inflation tests at every 0.2 m depth, almost continuous curves of the variation ofthe pressures P0 and p~ are obtained, Fig. 2.17.
Fig. 2.16. The Marchetti dilatometer.
47
Soil Layer Sequence
Pressure
Pressure
(MPa}
Slgeffv
--UO 0
--PO
0.5
---PO
~Pl
0 0
1.0
0 ~
2
(MPa}
~Pl
0.5
1.O -
.
.
.
2.0
:1..5
1
.
I
~ i
, ,
i
9
2.5 1
-t - - - - ~ t
'
i i
,--,
~
k~/
t0
I
__)
,
t2
l,~
i4
16
16
18
18
20
20
a)
t
b)
Fig. 2.17. Results from dilatometer tests (a) soft slightly organic clay (b) sand with a clay layer.
The measured pressures are used together with the in situ stresses in terms of effective vertical stress (r'vo and pore pressure u 0 to calculate the parameters Material index
ID =
Horizontal stress index
KD = (P0 - u0)/Cr'vo
(2.4)
Dilatometer modulus
ED = 34"7(Pl - Po)
(2.5)
(Pl
-
P0)/(P0 - u0)
(2.3)
These parameters are then used for classification of the soil and estimation of soil properties. The internationally most widely used classification chart was presented by Marchetti and Crapps (1981) and was slightly modified by Schmertmann (1986), Fig. 2.18.
Site Investigations
48 2000.
EO . IOOO
iO(n
LINE A S C D
§
9
S..TY I ~ ' e " ~
lo| I 0) n
9
0.58S 0.621 0.6H? 0.696
S'ANDI
SILT
E.(~UATION OF THE LINES
1.737
2 013 2.289
2 ~64
CLAY
W o,;..:::3/ LId
,
<'g-----=
IOO
v
-IMUDIPEAT
'(y) - A p p r o x i m a t e
soil unit sho~n i n p a r e n c h e l e l
9 - If
tl 9 $0. 1s o v e l ' e J r
( I . 50)
0.1
Fig. 2.18.
0.2
Classification
0.5
, ,,,l MATERIAL
chart
I
,
INDEX
according
!
JD
2
weight
i n t/m 3
t h e n y i n t h e s e re$1ons by about O. 10 t/m ]
,
,
l
5
. . . .
to Schmertmann
(1986).
In Sweden, this chart has been modified mainly for clays and organic soils (Larsson 1989). The material index ID is then corrected for effects of overconsolidation according to IDCoo~)= ID - 0.075(K D - 2.5)
when depth < 2.0 m and K D > 2.5
(2.6)
ID(oorr) = ID - 0.035(K D - 2.5)
when depth > 2.0 m and K D > 2.5
(2.7)
The original material index is not corrected when its value is < 0.10 or when K D < 2.5. The soil is classified according to the chart in Fig. 2.19. Stiffer organic soils are classified as clays in the charts for the dilatometer test.
49
Soil Layer Sequence S~ILY WTTHTHIN SILTLAYEI'~E Y CLAY _
CLAY ]----~--__~] . . . . ~ ) - A N ~ ___L_~"~Y ~ t ~ _ s.... _L____ _-_--_--~
CORRECTED MATERIAL INDEX TO (CORR) 0.1 0.35 0.6 0.9 12 1.8
3.3
8
I
I
/
,.o
-
--
/
/ ,// ,I
/
,i"
~TIF/E~F/! / ' ! / I / _ ~ V E R Y L:X)S~ / L/_.~'/' i i
/ /,/;<x~-.~@ /
or
, r . , ~ " -
i
li
.................
.....
GYTTJA/PEAT
'
0,1-~/
/ ZONE 1 2 3 /, 5 6 ? 8 9
ORGANIC SOIL
!,
o!,
'
t CLASSI FICAT ION CLAY WITH THIN SILT LAYERS/SILTY CLAY/(ORGANIC CLAY ) CLAY/(0RGANIC CLAY) CLAY/( GYTTJA/PEAT ) SILTY CLAY/( GYTTJA / PEAT ) ORGANICCLAY/(CLAY WITH THIN SILT LAYERS/SILTY CLAY) ORGANIC CLAY / ( CLAY ) GYTTJA / PEAT/( CLAY ) GYTTJA/PEAT/(SILTY CLAY i ORGANIC CLAY/GYTTJA / PEAT
Z FOR CLAYEY AND ORGANIC SOILS. THE NOTATIONS VERY LOOSE. LOOSE MEDIUM 9 STIFF. STIFF AND VERY STIFF ARE GIVEN ON THE BASIS OF THE ESTIMATED UNDRAINED SHEAR STRENGTH
2.15/ 100
1.95 1.85 .85
2.1
1.75 ~.TS
10
/
.
/
/ 1.7
2
1.65
o:
c~ 1.0
,05
I
I I I I 0.1
025
0.6
1D
CORRECTED HATERIAL INDEX I0(CORR )
Fig. 2.19. Classification chart according to Larsson (1989).
Site Investigations
50
2.3.4
Sampling
In site investigation, the extraction of soil samples is a necessary complement to other investigation methods, both for the identification of soils and for laboratory investigation. There are a number of different tools for taking soil samples in different types of soils. However, the quality of the samples will vary between the sampiers and between the types of soil. In 1981, the ISSMFE Subcommittee on Soil Sampling presented an international manual on soil sampling of soft cohesive soils,(ISSMFE Subcommittee on Soil Sampling 1981). The degree of disturbance of the soil samples can be divided into three categories with the following general definitions. 9 Undisturbed samples = The soil retains the same fabric, type and proportion of constituents and physical and mechanical properties as in the field 9 Disturbed samples = The soil retains the type and proportion of constituents and water content, but the fabric may have changed. The physical and mechanical properties have changed. 9 Remoulded samples = The soil structure and its physical and mechanical properties have changed from the in situ conditions. The type and proportion of constituents and the water content remain unchanged. When choosing sampling quality and tool, it is necessary to consider the subsequent laboratory investigations. For soil identification only, disturbed or remoulded samples can be used. When the deformation and strength characteristics of the soil are to be investigated in the laboratory, it is necessary to obtain undisturbed samples. For undisturbed samples, the thin-walledpiston sampler with a fixed piston is normally the most suitable tool for non-fibrous soils. A number of piston samplers with different sample diameters are available, c.f. ISSMFE Subcommittee on Soil Sampling (1981). The piston sampler consists mainly of sampler head, piston, piston rod, piston extension rod or wire, sampling tube or liner and thrust equipment or machine. During sampling, the closed sampler is pushed into the soil to a level just above the sampling level. The piston is then fixed to the ground or thrust machine while the sample is cut out, Fig. 2.20. After a while, when part of the generated excess pore pressure has disappeared and the sampled soil adheres to the inside of the walls, the sampler can be extracted.
51
Soil Layer Sequence
II
Rigid ~ boring rl
on
Fig. 2.20. General set-up for fixed piston sampling with Swedish standard piston sampler, St I.
To be able to obtain high quality samples it is, among other things, very important to have a very sharp edge on the sampler and to have the piston carefully fixed to the ground or thrust machine. It is also important to have the sample cut out steadily and slowly, Kallstenius (1963). Samplers with small inner clearances and small area ratios should be used. One difficulty with the small diameter piston sampler is taking good samples in fibrous peat. Because of the force applied to cut off the fibres, the sample may be compressed at sampling. A peat sampler, o 100 mm, has been developed in Sweden. This sampler consists of a sharp wave-toothed edge mounted on a plastic tube with a driving head at the upper end, Fig. 2.21. Sampies are taken from the ground surface or at the bottom ofprebored holes.
Fig. 2.21. 100 mm diameter open peat sampler, Swedish Geotechnical Institute.
52
Site Investigations
After extraction of the sampler, the cutting edge and driving head are removed and the sample in the plastic tube is sealed. Laboratory tests show that samples of fibrous soils taken with this peat sampler have higher quality than samples taken with a small diameter piston sampler. Practical experience has also shown good correlation between laboratory test data from this kind of sample and measured field behaviour under embankments on fibrous peat. For soil identification purposes, it may be satisfactory to obtain disturbed or remoulded samples. Disturbed samples can be obtained with the split spoon sampier in the Standard Penetration Test. Disturbed samples can also be taken with screw augers, while remoulded samples are obtained with the peat drill. The screw auger consists of a steel rod on which a screw-shaped flange is welded. The length (0.25-1.0 m) and diameter (35-100 mm) may vary greatly. In sampiing, the auger is rotated into the soil down to the sampling level and then pulled out. During extraction of the sampler, soil from the shaft above the sampling level may stick to the outside of the sample. Thus the sample must be cut clean when extracted above ground. Fig. 2.22. One advantage of the screw auger sampler is that a long, continuous sample is obtained. Thus it is possible to recognize also thin layers of different materials in a soil layer sequence. In fibrous peat, difficulties may arise because the fibres are not cut off.
Fig. 2.22. Screw auger sampler after extraction from ground. The sample is cut clean for examination.
Soil Layer Sequence
53
In very soft organic soil, the peat drill may be used to extract samples for soil identification. The peat drill is an open, side intake sampler which can be closed with a shuttle. The sampler is closed during insertion in the soil down to the sampiing level. The sampler is turned while the shuttle first opens the sampler and then forces the soil into the sampler, Fig. 2.23. After one full turn, the sampler is extracted. The peat drill is often the most useful method in extremely soft organic soil, such as at the bottom of lakes or pools.
Fig. 2.23. Peat drill for soft organic soils with accessories, Borros, Sweden.
All samples should be sealed in airtight containers and carefully transported to the laboratory. Storage should be at ground temperature and the samples should be tested as soon as possible in order to prevent effects of chemical and biological changes with time.
2.4
GROUNDWATER
2.4.1
Pore pressure measurement
From the geotechnical point of view, groundwater or pore pressure measurements are of interest for the calculation of stability and settlements of embankments on organic soils. The permeability is also of importance for drainage and for the time-settlement relation for an embankment.
54
Site Investigations
Measurements are normally taken during the site investigation for a certain project, but can also be used for checking purposes during the construction of an embankment. There are principally three different ways of reading the groundwater pressure in a soil deposit. 9 Measurements of a free water level in rivers, lakes, pools or open boreholes. 9 Measurements in open standpipes or tubes within a pipe driven to a certain level. The standpipes can be provided with filter tips at the lower end. 9 Measurements with closed systems, piezometers, where the pipes are provided with filter tips. In order to obtain a complete picture of the pore pressure distribution in a soil profile, it is normally necessary to install more than one piezometer at each investigation point, especially when there is a mixture of permeable and impermeable layers in the ground and the deposit is in the vicinity of high ground. It is also important to note that the groundwater level or the pore water pressure in the soil may vary considerably with time. Therefore, the observation period must be long enough to cover the normal fluctuations (high or low) for the area involved in the project. Groundwater measurements in open boreholes are normally not accurate enough, as they only provide average values of the groundwater pressure within the penetrated depth and may also be affected by the amount of water flowing into the borehole during rain. In permeable soils, the pore pressure can be measured in open standpipes inserted to a certain level beneath the groundwater table. In organic soils, it is necessary to use open pipes where a small diameter (about 5 mm) tube is connected to the filter to limit the time lag, Fig. 2.24. With this type of open pipe, experience indicates that the time lag is less than 3 weeks, provided that the filter area is sufficiently large. The water level in the small plastic tube is measured with a coaxial electric cable and an indicator. When the water level is reached, the electrical circuit is closed. Due to the gas content in organic soils, so-called high air-entry filters, which prevent gas bubbles passing through the filter, ought not to be used. In permeable soils, it may be better to use a wider tube than the 5 mm diameter tube mentioned above in order to allow gas bubbles to pass freely through the pipe. For pore pressure measurements in organic soils and clays and when short time variations are to be measured, closedpiezometers should be used. There are a number of different types of piezometers: 9 pneumatic 9 hydraulic 9 electric with vibrating wire
55
Groundwater
// Plastic hose
Penetrometer
hollow rods
Sealing pipe
Filter
tip
Fig. 2.24. Open standpipe with filter tip and an inner small diameter plastic tube.
9 electric with electrical resistance strain gauges The working principle of these piezometers is shown in Fig. 2.25 - 2.28. In the hydraulic piezometer, the plastic hose is filled with a silicone oil, for example, and closely attached to a manometer. The pore pressure is calculated from the manometer reading, the levels of the filter and manometer respectively and the density of the oil. The effect of temperature changes is one problem with this equipment. Therefore the plastic hoses have to be carefully insulated. In the pneumaticpiezometer, the pore pressure acts on a membrane. Two hoses connected to a pressure gauge end at the membrane. During reading, the air or gas pressure is increased in the injection hose. When the gas pressure reaches the backpressure at the membrane, the air or gas can pass to the return hose and bubbles then appear in the water holder. In the electricpiezometerwith a vibrating wire, the membrane is connected to a wire in tension. The wire can be made to vibrate by electromagnets and the frequency of the vibration can be read with a frequency counter. High pressure on the
Site Investigations
56
---Manometer
T
Manometer
i--i !
' zlJc, y l l , ~ \
I "~-hold er
11,4~l,x"CYl,~',,,\'e.
Return hose
Plashc hoses
--
Fitter
~ ~ . _ Membrane Fitter Fig. 2.26. Pneumatic piezometer with water holder indicator.
Fig. 2.25. Hydraulic piezometer.
l
f
(~
t=~1 ~
\
Cable
1 Frequency counter
Vibrating wire Membrane Fitter
Fig. 2.27. Electric piezometer with vibrating wire.
•S__jgnal unit
-
~,~-.
Strain gauges Membrane
~
Fitter
Fig. 2.28. Electric piezometer with resistance strain gauges.
Groundwater
57
membrane decreases the frequency. The piezometer must be calibrated at actual ground temperature. In some of these piezometers, the membrane can be unloaded by back pressure. The zero point frequency can thereby be checked,which is important for long-term observations. In the electric piezometer with electrical resistance strain gauges, the strain gauges are fixed to the rear side of the membrane. The resistance of the strain gauges changes when the membrane is loaded. These changes in the electrical resistance can be measured using the principle of the Wheatstone bridge. In the B A T system, the filter tip and the sensor unit are separate, Fig. 2.29. The filter tip consists of a point of plastic or stainless steel, a filter of ceramic material or polyethylene and a nozzle with a self-sealing rubber membrane. The sensor unit consists of a pressure transducer with electrical resistance strain gauges connected to a hypodermic needle. During reading, the measuring unit is lowered through the extension pipe to the filter and the needle penetrates the rubber membrane. Thus the pore pressure can be transmitted from outside the filter to the pressure transducer. The time lag between insertion and a stable reading is about 15 minutes. If a continuous reading is to be obtained, the sensor is left in the tube and connected to a data acquisition system. There are two main advantages with this system: Cable
l
[Exte2on pjp_e_e
~
eedle
Rubber
...... -~ Nozzle !
~"
-.
*-
Permanently, installed
filter tip
Fig. 2.29. BAT piezometer system.
58
Site Investigations
9 The sensor can be recalibrated at every reading against the water level in the extension pipe. 9 The sensor can easily be exchanged, for example if the transducer is damaged, without pulling out the whole piezometer. When measuring pore pressures in settling soils under embankments on organic soils, the settlements have to be considered. It is often necessary to install the piezometers in separate casings which are driven to a depth of about 1-2 m above the filter tip. Otherwise, downward drag from the settling soil above may cause further penetration of the piezometer, which can create excessive pore pressure readings.
2.4.2
Permeability measurements
The drainage condition of the soil is of major importance in settlement analysis and for the design of vertical drains in embankments on compressible soils. Field permeability tests are a valuable complement to laboratory tests. The field test may be more reliable than the oedometer test, for example, especially when the horizontal permeability is of importance. The permeability in situ can be determined in a number of ways, c.f. Jamiolkowski et al (1985): 9 Pumping test (e.g. outflow and inflow tests in boreholes) 9 Tests using piezometers 9 Self-boring cells 9 Piezocone dissipation tests. 9 Back-analysis from field measurements For practical purposes, tests with piezometers are recommended in organic soils with comparatively low permeability. Pumping tests are normally not applicable because of the low permeability and piezocone tests give uncertain results. Backanalysis will be treated in Chapter 2.7. As most organic soils are normally consolidated or slightly overconsolidated, the inflow test is considered to be the most applicable test for embankment purposes, as it produces an increase in effective stresses. However, investigations by Tremblay and Eriksson (1987) have shown that, at least in clay, there is no significant difference between the k-values obtained from inflow and outflow tests. There are also practical problems connected with creating and maintaining a negative pressure in the system. Also the effect of gas bubbles is a source of error. Therefore, the outflow test with a constant and fairly low gradient is proposed for permeability investigations for embankments on organic soils. For this purpose, a piezometer filter is pushed to the desired depth and connected to constant head equipment, e.g. Mariotte's bottle, Fig. 2.30.
59
Groundwater Inner tube opened to ~"~'J~atmospheric pressure m
Scale for reading of the injected volume
i,
Partial vacuum
,.._.
h (constant head)
Ground surface /I/
~
iris----
ire---
iil~
Water table
Piezometer
Fig. 2.30. Piezometer tip provided with Mariotte's bottle for outflow tests with constant water head.
The permeability of the soil can be calculated from the following formula: q= F-k-h
(2.8)
where: q = measured flow k = soil permeability h = applied head F = size and shape factor 2.4p. L F
(2.9)
_._
ln[ 1.2L/D+~/ 1+(1.2 L/D) 2]
60
Site Investigations
2.5
STRENGTH AND DEFORMATION CHARACTERISTICS
2.5.1
General
There are no special tools available for the determination of the properties of organic soils. Therefore, the same equipments as for mineral soils, especially those developed for soft clays, have to be used. Also the methods of interpretation of field test results are limited in the literature, probably due to the reluctance to use areas with organic soils for building l~urposes. Landva et al (1986) have made a review of the applicability of different in situ tests for some organic soils. The main problems with organic soils are the heterogeneity and fabric of the soil, especially in peat soils and silty, sandy organic soils. Therefore, the number of field tests must be increased compared to normal investigations in clay soils. The in situ tests also must be completed with sampling and laboratory testing. Some results from in situ testing of organic soils are summarized below. As mentioned above, the cone penetrometer and especially the piezocone are useful tools for soil profiling. The results can also be used for estimation of the soil properties. However, testing of soft fine-grained soils involves measurements of very small tip resistances and sleeve frictions, where also the generated pore pressures significantly affect the measurements of the other parameters. Estimation of soil properties on the basis of cone penetration tests in this type of soil therefore demands very carefully calibrated piezocones, as well as correction of tip resistances and sleeve frictions for pore pressure effects. A number of methods have been proposed for evaluation of the undrained shear strength (~f~) of cohesive soils. The most commonmethod is to use the net cone resistance and evaluate the undrained shear strength from
"tr~ =
qT- (Yvo
(2.10)
NKT
where qT -- total corrected tip resistance ~vo = total overburden pressure NET = empirical cone factor The cone factor is sensitive to the plasticity of the soil. For clays, organic clays and gyttjas, it can be estimated from
Strength and Deformation Characteristics NI~x = 13.4 + 6.65.w L
61 (2.11)
where w L is the liquid limit of the soil (Larsson and Mulabdic" 1991). This formula is not applicable to peat. Alternatively, the undrained shear strength can be estimated from the generated pore pressures at penetration. In this case, it is preferable to use the pore pressure generated at mid-height of the conical face of the tip UVACE.In normally consolidated and only slightly overconsolidated clays, organic clays and gyttjas, the undrained shear strength can be estimated from
UFACE - U ~
q:f~ =
(2.12) 16
Similarly, this formula cannot be used for peat. The undrained shear strength obtained by both methods corresponds to the undrained shear strength in direct shear and also to what is obtained in corrected field vane tests. In highly humified non-fibrous peats, the undrained shear strength may possibly be estimated from cone penetration tests after local calibration against other types of tests. In more fibrous peats, the soil behaviour around the cone tip differs widely from normal undrained shear failures and the undrained shear strength cannot be estimated from this type of test (Landva 1986). The results from dilatometer tests can also be used to estimate the undrained shear strength. In this case, the undrained shear strength is estimated from
Pl
- O'ho
a:f~ =
(2.13)
where P l = pressure required to expand the membrane 1.1 mm Cyho= total in situ horizontal stress = u o + K ~ 9~'vo where the coefficient of earth pressure K o is also estimated from the results of the dilatometer test F
= empirical factor. F - 10.3 for clay and 9 for organic clay and gyttja, (Larsson 1989)
62
Site Investigations
As with the cone penetration test, it may be possible to obtain local correlations for the undrained shear strength in highly decomposed non-fibrous peats using calibration against other test methods. For more fibrous peats, the dilatometer test cannot be used to estimate undrained shear strength. The dilatometer test is often used to estimate deformation parameters in the ground. For clay and organic soils, its usefulness for this purpose is mainly limited to overconsolidated soils. According to Landva et al (1'986), fewpressuremeter tests in organic soils have been reported and this type of test is not recommended for peat soils. For stiff organic soils, pressuremeter tests could be useful. However, it may be necessary to change both the test procedure and the method of interpretation due to the special type of soil. According to Landva et al (1986) plate load tests are not applicable for the determination of strength and deformation characteristics in peat. However, the bearing capacity for concentrated loads, such as vehicle traffic on the peat mat, can be determined. Schwab (1976) performed both large scale and screw plate load tests, Fig. 2.31, in soft organic sulphide soils at three sites in northern Sweden along the coast of the Gulf of Bothnia.
HYDRAULIC NNER
HOSE
PiPE
OUTER PIPE
I•BMEASURING BEAM _ _ H Y D R A U L I C
~
HOSE
SETTLEMENTGAUGES
PISTON It
_
,
ANCHOR FRAME EARTH ANCHOR
HYDRAULIC JACK
SCREW
0
5
lOcm
PLATE
Scale Ot , 510 , l ~ O c m
~
OUTERPIPE HYDRAULIC
JACK
Fig. 2.31. Screw plate device and test set-up, (Schwab 1976).
Strength and Deformation Characteristics
63
The results of these investigations indicate that, for this type of soil, the screw plate test could be a useful tool for determining the bearing capacity for foundations or embankments. The shear strength, ~f~, of the soil was evaluated as
~, =
- O'VO
(2.14)
Nc where % is the ultimate failure load, gvo is the total overburden pressure and N c is the beating capacity factor, which was found to be 9.
2.5.2
Field vane test
The field vane test is the most common method of determining the shear strength also of organic soils. However, the interpretation of the test results must be handled with caution, especially in fibrous peat and silty, sandy organic soils. An extensive investigation of vane testing in peat was performed by Landva (1980). There are different types of field vane equipment, Fig. 2.32.
[~
~J
I I/
Torque instrument 2/I
Penetrometer
rod
Casin9
77
-
Vane Cross section of the vane
Fig. 2.32.
Principle of field
Slip c--oupling Vane
vane equipment used in Sweden.
Site Investigations
64
In the SGI equipment developed by Cadlmg et al (1950), the vane is protected by a special sheath when driven into the soil. The rods connecting the vane to the torque instrument above the ground surface are enclosed in a casing. The second type of equipment in Fig. 2.32 was developed by Geotech in Sweden. In this equipment there is no casing or sheath protecting the rods and the vane. Therefore, the vane must be more robust in the Geotech type. In this case, the friction along the rods is measured by using a slip coupling close to the vane. Thereby, the torque necessary to turn the rods can be measured at each test level and subtracted from the total maximum torque at failure. Both recording and non-recording instruments are available. The vane normally has a height of twice its diameter. The undrained shear strength value, ~v, is then determined according to the following formula: 6 q:v = ~ 7
Mmax rt D 3
(2.15)
where Mmax is the maximum torque at failure for the vane and D is the diameter of the vane and half the height of the vane. It is assumed that the shear strength is reached at the same time along the cylindrical surface circumscribing the vane. The recommendation regarding the time limits for execution of the testing procedure is that rotation of the vane should start within 5 minutes after its installation at the test level. The rate of rotation should be such that failure is obtained between 1 and 3 minutes after the start of the rotation. In organic soils, the stress strain curves and failure strains differ very much depending on the organic content and it may be very difficult to estimate the proper rate of rotation beforehand. It may therefore be necessary to apply corrections for the actual time to failure in those cases where grosser misjudgements have been made. Based on investigations by Torstensson (1973) and Wolski et al (1988), such a correction for time to failure can be made according to Table 2.1. In this case, the measured value of Zv should be divided by the relevant correction factor in order to obtain the corresponding value of Zv at a reference time to failure of 3 minutes. To obtain the shear strength zf~ from the corrected strength values z~ the latter should be multiplied by another correction factor g, c.f. Chapter 4.2, (Larsson et al 1984).
65
Strength and Deformation Characteristics Table 2.1. Correction factor for the different times to failure.
Correction factor
Time to failure, sec
1.141 1.100 1.076 1.060 1 022 1 000 0.985 0.973 0 964 0 956 0 949 0 943 0.930
15 30 45 60 120 180 240 300 360 420 480 540 600 Thus q : ~ = p .1:v
(2.16)
where
1.2>p=(043
)~
> 0.5
(2.17)
WL
and w E - liquid limit of the soil For organic soils with a liquid limit > 200 %, the correction factor ~ is equal to 0.5. According to Landva (1980), the shear failure in peat was found to occur along a cylindrical body 7-10 mm larger than the diameter of the vane. This is one of the reasons why Landva does not recommend field vane tests in fibrous peat. However, for other types of organic soils the field vane test is considered to be a useful tool e.g. Schwab (1976), Wolski et al (1988) and Wolski et al (1989). Schwab pointed, out the necessity of making a greater reduction ofthe ~v values than that used at that time in Sweden for organic sulphide soils. However, the reduction rules have been changed since then. The selection of shear strength parameters for design is discussed in Chapter 4.2.
66
2.6
Site Investigations
MONITORING EQUIPMENT
Monitoring equipment is often used for embankments on organic soils both during normal construction and especially for test fills, Chapter 2.7. The reasons for this are mainly the large deformations and low bearing capacity of the soil, which make it necessary to use special foundation and improvement methods. An extensive description of different instruments for various purposes and constructions has been made by Hanna (1985). This book deals only with equipments that can be used for monitoring embankments on organic soils. Pore pressure measurements in the soil beneath an embankment can be made with the same equipment as described in Chapter 2.4.1. However, closed piezometers ought to be used. It is important to consider that large settlements in organic soils may cause increased pore pressures around the tip because of penetration of the piezometer, especially during filling, when the pore pressure may be critical for the stability of the embankment. To avoid this, the piezometer pipe can be provided with a casing to about 1 m above the filter. Also closed systems without pipes may be used or the piezometers may be installed with an inclination of, for example, 1:1. For the measurement of settlements, different kinds of benchmarks have been used, Fig. 2.33, both with steel plates on the ground surface and with screw tips that can be installed at certain depths. Benchmarks at greater depths ought to be provided with casings to protect the rods from the friction of the settling soil. In order to obtain continuous settlement distribution curves across the embankment and to prevent disturbance of the construction work, the hose settlement gauge can be used, Fig. 2.34. To perform these measurements, flexible tubes or hoses are placed in shallow ditches or under small sand fills on the ground surface across the embankment in the sections to be measured. If the tubes are extended outside the base of the embankment, also the heave of the ground surface outside the embankment can be measured. The measuring unit consists of two plastic hoses, one inside the other. The smaller tube contains air and an electric cable. The annular space between the two hoses is filled with a liquid (normally water). The lower ends of the hoses are connected to the measuring head containing the pressure transducer. This transducer measures the liquid pressure in relation to th e atmospheric pressure. The upper ends of the plastic hoses are connected to an open standpipe. Thus the difference in level between the measuring head, inserted in the flexible tube under the embankment, and the liquid level in the standpipe can be measured. The liquid level in the standpipe is in turn levelled in relation to a fixed point located outside the test area. An example of such measurements is shown. Fig. 2.35.
6?
Monitoring Equipment DEEP SETTLEMENT GAUGE {SCREW) -- [
[
SUPERFICIAL SETTLEMENT GAUGE (PLATE)
- Extension rod
Extension rod
J25 mm
25ram
lSOmm
50ram Protectinq tube
I Protect~nQ_ tube
I J
. .
.
I.
.
t 1t: / ....... ""
,.. i " .".. . /~. ' " " I ~3~176 j ~
Plate
.7
s~w
~,p
0 SOOmm
[
Fig. 2.33. Benchmarks for superficial and deep settlement measurements.
Read out unit
Atmospheric pressure Fluid level
--Cable
Atmospheric pressure
Fluid level regulator Pressure
Measuring head
Fig. 2.34. Principle of the measuring unit for the hose settlement gauge type SGI II.
Site Investigations
68 a)
Embankment
No.1
g_
! 4-
3.6
4-
3.9
L 2 ~. "
11 6 "..
4,
i
/..2
4-
III
J
. #
3.9
J-
3.6
4"
1/.
~
II
~
1.3
-0.2 0.0 u',
T ~r x
!
~
m
--4
~('~x _ . - - - - - - x
0.t. --4
._
rfrl
L_ !
"
,'--"z~
.I.
~.
~
0.8 m z
--4
..-
12 16
i
i
. .............................
. . . . . . . . . . . . . . . .
!
.. ~....-" L
z,5
I D (at e
i0907
{03 071 - - ~ ' - T ~ r * ' - q
---20 Ira)
11a4" ~ +,#~" i,,,~rF-T-nqar--T-~a~---1-~-,j-q-v~-T-T,]~r'r-q
io-ol - I:- to 4o, q-, -l-:-]----x IFig. 2.35. Example of settlement measurement obtained with the Swedis hose settlement gauges under a test embankment at Antoniny, Poland.
In order to measure the compression in different soil layers, the bellows hose settlement gauge or the magnetic screw settlement gauge may be used. This may be important in field tests where soil layers of different strength and deformation characteristics are found beneath the embankment. The bellows hose settlement gauge consists of a compressible, reinforced plastic tube 25 mm in diameter. The hose is cut into pieces corresponding to the distances between the observation points. The pieces are joined together again with special inner casings provided with a metal ring. This metal ring closes an electrical circuit when the sensor passes the ring, Fig. 2.36. The sensor is lowered in the pipe by a cable which at the same time is used as a measuring tape. In this way, the depth below the upper end of the pipe to the different metal rings can be measured from time to time.
69
Monitoring Equipment
"~-... Ir~dacafng casing with cap z m
~"--'--'-~
Measuring t a p e
One of ~ - . r a l r n e a ~ r i r ~ poin= =t pn~letermined c~=ntr~
Sounding rods fused during inr~dladonl
Lowes~ "point -
measuring ruled
35 m m i.d. ca~ng lu.~=d during in.~allaEon)
Steel 6p for driving and for hose at the bonom k=.vd
9fi~ng die
Fig. 2.36. Components of the bellows hose settlement gauge.
The installation process uses a system of inner rods and sometimes a casing. When the full depth is reached, the casing and inner rods are pulled out. One limitation of this method is that the compression of the plastic hose must be less than about 10 %. In a similar way, settlement measurements can be performed with the magnetic screw settlement gauge, which consists of a plastic pipe and a number of screw
70
Site Investigations
plates, Fig. 2.37. The screw plates are turned down to the desired depths around the plastic pipe. Magnetic tings are inserted in the plates close to the pipe. The level of the magnetic tings can be measured from time to time with a sensor on a measuring tape. Normally, the readings are taken in reference to the lowest plate where small or no settlements occur.
Settlement
indicator
_ -
r
,x.,u.t.~
Guiding plastic tube 9 ,x" t , , ~
.
,. . . . . .
Settling
screw plate m
~, ~
Reference plate Distance pipe
8 4
9 ,s
~
Screw tip
AAh
Fig. 2.37.
Magnetic screw
settlement gauge.
The main advantage of this type of equipment is that it can be used for larger compressions in the layers between the plates and that it is certain that the screw plates follow the settling soil layers. Fig.2.38 shows an example of magnetic screw settlement measurements where the two lower plates could not be measured atter 600 days due to buckling of the plastic pipe caused by the large settlements in the upper layers.
Monitoring Equipment
71
0.0 '~.- ~ = - ~ . - - . "- ~ ~ , _ . . . _ ~9.
0.2 0.4
.-. E E ~
~-.~.
O)
~ ~.,~,~
M-1
tube buckted
\ " ~
. ~
0.6
0.8
c
"~.
EMBANKMENT No 1.
H-2/__
1.0
l]J
4""
1.2
~'~
M-1
~- ~ ~
..
I 4"" Peat
"~,
"3.1
\.
co,c.,o,,
1.4
Sand 0
120
240
360
480
600
720
Time, (days)
840
Fig. 2.38. Results of settlement measurements with magnetic screw plates at different levels under the test embankment at Antoniny, Poland.
Considerable horizontal displacements also occur in the construction of embankments on organic soils due to shear deformations in the ground. The displacement of benchmarks at the ground surface can be measured with optical or electrooptical instruments or with a measuring tape using a fixed point outside the influenced area as reference. However, in order to measure the horizontal displacement below the ground surface, inclinometers have to be used. Various inclinometers are described by Hanna (1985) and here the inclinometer developed at the Swedish Geotechnical Institute is briefly described. The SGI inclinometer measures the horizontal displacements in a flexible plastic pipe which is installed in the ground at the point of interest. The pipe has an inner diameter of 42 mm without any tracks and is provided with a telescopic tip in order to avoid the influence of settling soil. The measuring unit consists of a cylinder with pairs of guide bosses, which are pressed against the plastic tube. Inside the cylinder, a pendulum is suspended in a plate-spring equipped with electrical resistance strain gauges. In this way, the deviation of the tube from the vertical axis can be measured accurately by means of a Wheatstone bridge.
72
Site Investigations
The inclination of the pipe, or change in inclination, is normally measured at every one or two metres of depth in two perpendicular directions, of which one is parallel to the direction of the expected main movements. To be able to measure in a certain direction from time to time, the inclinometer is inserted in the plastic tube with a series of torsionally rigid rods. The horizontal direction of the inclination measured is determined on a horizontal scale at the top of the pipe. Changes in the inclination in a certain direction correspond to the angular strain in the soil. The position of the pipe in relation to the tip can be calculated by integrating the inclination from the tip to the level considered. Examples of such inclinometer measurements are shown in Fig. 2.39.
ET
I-~
--
~
t'ttr i
2
PEAT
3
,,~
~'
2
2
3
3
II ilI:
~
~
" ~ ; " "7"
/r162
II
/S/f7 t'
/;12
I I ill i i-
\\ ~\ 11 ~
~~
. . . . . . . . . . . . . . . . . cm
cm
"
3
\ \ \\\\ \ I 1
Stoge
.
.
.
.
.
i; ~176
1;~o~o~o---o "o,~
,----_
i ~
i 8c'2 C7
cm
Fig. 2.39. H o r i z o n t a l d i s p l a c e m e n t s at different points under and outside the test embankment at Antoniny measured with the SGI type of inclinometer.
Test Embankmentsfor Design Purposes
2.7
TEST EMBANKMENTS FOR DESIGN PURPOSES
2.7.1
Introduction
73
Oiten, it is difficult to determine strength and deformation characteristics in organic soils by ordinary field and laboratory investigations. To establish the true stress and strain conditions, the construction of test embankments is often a useful method. This is especially the case for larger projects, e.g. where stage construction of embankments is planned. A number of test embankments are reported in the literature, e.g. Landva et al (1986), Chang (1981), Wolski et al (1988), Hudson et al (1989). Building embankments or dykes on organic soil is in most cases mainly a deformation problem. By some means, the main settlements must develop before the road embankment or the dyke is put into service. However, the bearing capacity of the soil may sometimes be a limitation. In 1989, a symposium was held in Malaysia on soil improvement techniques for road embankments, where different soil improvement methods were tested, (Hudson et al 1989). The aim of these test embankments was to clarify which techniques could be useful in future construction of main roads in Malaysia. This is one example ofthe use of test embankments. The test embankments were heavily instrumented during the construction period, which is very important for the evaluation of the test results. A number of monitoring equipments are presented in Chapter 2.6. A disadvantage of test embankments is that they often become expensive and time-consuming, which is the reason why they are not often used. Before the construction of a test embankment, it is important to clarify the purpose of the test and the limitations and requirements on the complete construction. In Malaysia, for instance, it was stated that the different test embankments should be built during the same period and in similar soil conditions. The performance requirements on the embankments during and after construction were also clearly stated. In the following, examples are given of different purposes of the construction of test embankments. The bearing capacity of the soil can be determined, Fig. 2.40. It is then very important to measure both horizontal and lateral deformations during construction of the test embankment to be able to estimate the yield level. In some soils, such as fibrous peat, the shear strength of the soil increases fairly rapidly when the soil consolidates under the additional load. Also the strength increase due to this consolidation can be measured. This means that the bearing capacity of the soil increases and, as a consequence, stage construction can be used to improve the soil. In con-
74
Site Investigations
nection with beating capacity tests, the effect of loading berms or flat slopes can be evaluated.
Fig. 2.40. The test embankment at Antoniny after failure.
Deformation characteristics of the soil can also be clarified from test embankments. In particular, the influence of horizontal displacements can be measured. Also the time-settlement curve for different layers and for the total soil deposit can be established. This may be important for planning the full scale project. The effect of surcharging, unloading and vertical drains on the time-settlement curves can be evaluated. If a certain height (level) of the top of the embankment or dyke is to be established and maintained, the construction method and necessary surcharge can be obtained from the test. The effect on both beating capacity and deformations of different improvement techniques can be evaluated, e.g. the effect of preloading, lime columns, vertical drains or geotextile reinforcement. For all these purposes, a test embankment is the most realistic test of the soil.
Test Embankmentsfor Design Purposes
2.7.2
75
Preparatory investigations
Before designing a test embankment, a preparatory investigation of the soil at the test site must be performed. All the parameters necessary for calculating the behaviour of the embankment with respect to the purpose of the test should be obtained. Index properties, shear strength and deformation properties of the soil, as well as the groundwater situation, are normally required. Methods for use in field and laboratory investigations are presented in Chapter 2.3-2.5 and Chapter 3 respectively. Specially designed laboratory tests can be used, for example when a special loading program is planned.
2.7.3
Location of the test embankment
When discussing the location of a test embankment, there are a number of aspects to be considered; both practical and geotechnical questions that may influence the results of the test and also the conclusions of the test. As a checklist for consideration, the following aspects can be noted: 9 The location should be in an area which is characteristic of the total area revolved in the project. The depth to firm bottom layer, the soil characteristics and the groundwater conditions ought to be representative. 9 The test embankment should not be placed in the vicinity of existing constructions where the soil is already influenced by loads or construction elements, such as piles. 9 It should be possible to drain the embankment or the ground to the same level as for the complete construction. 9 The test embankment should not normally be placed in the area of the plalmed construction as the soil characteristics will change under the test embankment. Otherwise, uneven settlements will occur at the site of the test embankment in the future. 9 Vegetation generally reinforces the surface layer of the soil. Therefore, it may be important to avoid areas of trees and dense vegetation where the influence of a root mat might have a disproportionate influence on the test results. 9 Finally, it may be wise to consider trafficability to the test site and the distance to suitable filling material and other facilities necessary for such a project.
2.7.4
Design of test embankments
In the design of a test embankment, the general concept for the main project, such as the height of the embankment, should be followed. If there is more than one concept for the realization ofthe project, it may be necessary to construct more than
Site Investigations
76
one test embankment, for example when testing more than one improvement technique. The design procedure can then follow the normal pattern of design for embankments on soft soils. Stability and settlement analysis must be performed in order to make reliable predictions for the behaviour of the embankment. This is important not only to check the test results, but also to be able to foresee measures that will be necessary during the construction of the test embankment. Also the choice of monitoring equipment is influenced by the expected behaviour of the test embankment. Some special questions are related to the use of test embankments: 9 The size of the test embankment must be large enough to be representative for the future embankment or dyke so that the boundary conditions do not influence the test results too much. With narrow dykes, for example, it is important to have a test embankment which is long enough, perhaps 3-4 times the width of the dyke. For highway embankment projects, where settlement behaviour is of major interest, the test embankment ought to have a width of at least 5-10 m plus the depth of the compressible layer that is of interest. 9 The working time for the different steps in construction of the test embankment should preferably be of the same order as the time available for the corresponding steps in construction of the full scale project. At least 80 % of the settlements should have occurred during the test period so that the time-settlement curves can be extrapolated. In order to obtain reformation on the secondary settlements, it may be wise to retain the test embankment during the construction period of the full scale project. In that way, additional measures can be taken during the construction of the final embankment to avoid the inconvenience of secondary settlements. 9 In the design and plalming of the test embankment, also the monitoring equipment has to be considered so that there is enough space and time available for installation and zero reading of the instruments.
2.7.5
Test embankment monitoring
The main idea of monitoring the test embankment is to measure those parameters or the course of behaviour of the embankment corresponding to the purpose of the test. In addition, extra monitoring equipment may be installed for research purposes. Many instruments are available for monitoring and some of these are described in Chapter 2.6. One example of an extensively instrumented test embankment is shown in Fig. 2.41. The type of equipment and the location of measuring points should be thoroughly considered in relation to the purpose of the test. In general, the accuracy of the
77
Test Embankmentsfor Design Purposes 35m y &O
~3.0 Z 3.6 ~ 3.__9W__4.2 __~___
Note6 River ~
{SN,
I - L ~ I
[~ P-6 -
116
/.~-
"S
S"
~S
'iS t -7
stage 3
P-4
S"
ME S M
"
M
M M
M MI~I P-3
"'S ~'S
~,,
Calcareous soil/Gyttja
f 4.0
.
3.6 X30%, 60
J
3.9 z
L
4.2
. . . . . . . . .
35.0 m 11.6
y
(~~
"S / *S
Sand
4
o
.
staqel
t
~
~
M !~,:!P--1 Peat S,I P-2 ,
, I:'-5
x # 3.9 # 3.6 Z 3 0
__ %, 42
/~-~':
""~.~
,K 4.2
Z 3.9 ~, '1
3.6 ~.0 -4'~ 3.0 I
-L.---
LEGEND -
,t, S-settlement gauge M-magnetic settlement gauge --- H -hose settlement gauge
i
~. .,:t, '
o I-inclinometer P- BAT piezometer
Fig. 2.41. Location and type of monitoring equipment for a test embankment at Antoniny, Poland.
78
Site Investigations
equipment should be about 10 times better than the accuracy absolutely required to evaluate the test results. The following general recommendations can be made for different monitoring parameters: 9 To evaluate the consolidation characteristics, settlement measurements ought to made close to the centre of the embankment where no shear deformations occur. The settlement distribution across the embankment can be measured with the hose settlement gauge. If there is more than one typical soil layer, it may be wise to install measuring points at the ground surface and at the boundaries between the different soil layers. Reference points must be installed a long distance from the test embankment or be made of steel rods to firm bottom layer. Measurements of settlements in the different soil layers in more profiles across the embankment, and particularly under the slopes, enables a more detailed evaluation of the deformation process. The horizontal displacement can be used to determine yield points of the bearing capacity and to estimate the volume of soil pressed outside the embankment. The latter information also enables an evaluation of initial and subsequent shear deformations. For this reason, inclinometer pipes are often placed in the middle of the slope where the shear deformations are greatest. Ifinclinometers are placed also outside the embankment, the horizontal compression of the soil can be evaluated. To enable an evaluation of the degree of consolidation, pore pressure measurements should be taken; preferably in the middle of typical soil layers and in the permeable bottom layers close to the centre of the embankment. Measurements should preferably also be made under the slope and outside the embankment area in order to obtain a total picture of the pore pressure distribution in the soil. If the embankment is expected to settle so much that the original ground surface is pressed down below the groundwater level, it is advisable also to measure the groundwater pressure close to the original ground surface in order to calculate the effective load.
The applied load is often calculated from the height of the embankment and from measurements of the density of the filling material. As there are rather large initial settlements in organic soils, it is advisable to use the thickness of the fill rather than the height of the embankment in this context. An even better method is to use total load cells on the ground surface. However, such cells must be large enough to indicate an average total pressure. If the changes in soil strength or consolidation parameters are to be investigated by in situ testing or sampling and following laboratory tests, it is advisable to install casings on steel plates on the ground surface before construction of the test embankment. The main reasons for this recommendation are the lack of space
Test Embankmentsfor Design Purposes
79
among the instruments and the limited possibilities for installing casings later in the fight place without disturbances. In the case of improved soil, casings may be installed on top of a lime column or in the centre of the spacing between vertical drains. During the construction of a test embankment, it is important that people responsible for the monitoring equipment are present at the site. They can then prevent damage to the equipment, check its operation and recalibrate it. Filling and compaction around the equipment must be performed with extra care. When planning the monitoring system, the interval between readings, as well as the mode of presentation of the test results, must be considered. Not all instruments can be connected to automatic data acquisition systems. However, it may be valuable for some transducers to be read automatically. In that way, the course of events between the manual readings can be followed. In general, readings should be taken more frequently close to changes of different kinds; improvement, filling in different stages, unloading etc. In an ongoing test, readings ought to be taken both before and after the change in question so that the effect of the change can be distinguished.
2.7.6
Construction of the test embankment
Construction of the test embankment should follow the assumed procedure for the actual full scale project. It is advisable to contact the contractor when planning the test embankment construction and emphasize this point. Furthermore, the importance of keeping the monitoring equipment operating satisfactorily must be stressed. Without proper measurements, the value of the embankment test is very limited. It is most important to follow the design of the test embankment, as deviations will result in undesired movements and pressures.
2.7.7
Presentation of test results
In connection with the installation of monitoring equipment and the construction of the embankment, the design drawings should be revised so that they describe the actual construction and installation both in plan and in sections. Dates of certain stages should be added. It should be considered that it is not possible subsequently to remember all the details needed for the explanation of the test results unless they are written down. All measurements of settlements, pore pressures and horizontal displacements should be related both to time and applied load. For long-term projects, it is some-
80
Site
Investigations
times necessary to use a logarithmic scale for the time. An example of such a presentation is given in Fig. 2.42. (]} E M B A N K M E N T
No
1.
I I:::.-..i : '< :i- -
i::.
~;.0: : ~ ~ .r !.-~..-..-...-.-. i ;.:..: - : .-. -.:-::!--.- ...-i - [o.-.--.-..-.:-..---.-~ I ~
,1,p-',
"
ol"t,
P.,.
P-3
CALC
.... ; 9. . . : :: .._ ..-.-.. -..-. : -:...... =-: :....:; - . . . . . - - . . : : : : I' "
E 3
~.
"~
~ ~2 1
"~
0
40
-~
;
f [o1a !
J
SOIL I GYTTJA -.:-.-..-.:.;:...::.1-.
:..
[
:-.-.-..--.:
SAND
II
,-- :-.
"'-i
P-8
III STAGE
STAGE ,--. . . . . . . . . . .
1983
~E.~T
J
o~
-.
"3-
:...
zoo
198Z.
~0 2;0 ~o
, .....
1985 "
~o
"
.,...-~-.
1986
7;0
~ 9 o o
DATA "
1100
TIME.DAYS
30
o
CL
20
P-I
100 9O O
Q_ -
8(2 70. 60
P-2
Fig. 2.42. The pore pressure dissipation in piezometers P-1 and P-2 under a test embankment at Antoniny related both to time and applied load,
Test Embankmentsfor Design Purposes
2.7.8
81
Analysis of test results and recommendations
The results from the test embankment should be continuously compiled and analysed during the construction and observation period so that necessary measures can be taken if unexpected events occur. At the end of the observation time or the available time, the results are finally compiled, presented and evaluated. Comparisons should also be made with the previous calculations both for stability and settlements. When evaluating the bearing capacity of the ground, both vertical and horizontal movements should be studied in an attempt to find a yield point where the deformations increase more than linearly with increasing load. It should be taken into account that failures in organic soils otten are progressive. Thus, considerable settlement and cracking normally occur before an actual slide takes place. In fibrous peat, there is rarely any sliding at all. The recommendations from the stability analysis of the test results for the full scale project may include the following: 9 increasing or decreasing the thickness of the fill totally or in different stages. * increasing or decreasing the time for consolidation before filling of the next stage .
presenting a control program for pore pressure dissipation or shear strength increase.
In the evaluation of the settlement characteristics of the ground, the degree of consolidation at different depths should be considered as observed from the piezometer measurements and the bellows hose or magnetic screw settlement gauges. In deep soil layers, primary consolidations in the middle layers continue for a long time while, at the same time, the upper and lower layers have reached the stage of secondary consolidation. The evaluation of the test results leads to a more reliable time-settlement process with or without improvement and with eventual surcharging and unloading, depending on the purpose and design of the test embankment. On this basis, new recommendations can be made for the full scale project, e.g.: 9 revised conclusion of the total settlements of the embankment or dyke. 9 increasing or decreasing the necessary height of fill or surcharge to obtain a certain level at a desired time. 9 decreasing or increasing the time for consolidation of the subsoil or decreasing the spacing of vertical drains. 9 shortening or omitting vertical drams. 9 presenting a control program for settlement measurements.
82
Site Investigations
2.8 REFERENCES Bergdahl, U. (editor) (1989) Report of the ISSMFE Technical Committee on Penetration Testing of Soils - TC 16 with Reference Test Procedures CPT - SPT DP - WST. Swedish Geotechnical Institute, Information No. 7, Link6ping. Bjelm, L., Follin, S., Svensson, C. (1982). The Soil Radar as soil investigation method. (m Swedish). Dept of Engineering Geology. Lund Institute of Technology, University of Lund, Sweden. LUTVDG/TVTG--3002/1982.
Campanella, R.G., Robertson, P.K. (1988). Current status of the piezocone test. Proc. First International Symposium on Penetration Testing, Orlando 20-24 March 1988. Balkema, Rotterdam.
Cadling, L., Odestad, S. (1950). The vane borer. Swedish Geotechnical Institute. Proceedings, No. 2. Stockholm. Chang, Y.C.E. (1981). Longterm consolidation beneath the test fills at V~isby, Sweden. Swedish Geotechnical Institute, Report No. 13, Link6ping.
Hanna, T.H. (1985). Field Instrumentation in Geotechnical Engineering. Trans Tech. Publications. Ser. on Rock and Soil Mechanics FRG. Vol. 10. .Holm, G., Bredenberg, H. and Broms, B.B (1981). Lime Columns as Foundation for Light Structures, Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm 1981, Vol. 3, pp 687-694.
Hudson, R.R., Toh, C.T., Chan, S.F. (1989). International symposium ontrial embankments on Malaysian marine clays, Kuala Lumpur, 6-8 November 1989, Proc. Vol. 1-3. ISSMFE Subcommittee on Soil Sampling (1981). International manual for the sampling of soft cohesive soils, Tokyo University Press. Tokyo 1981.
Jamiolkowski, M., Ladd, C.C., Germaine, J.T. and Lancelotta, R. (1985). New developments in field and laboratory testing of soils. Proc. l Ith International Conference on Soil Mechanics and Foundation Engineering, San Francisco 1985. Balkema, Rotterdam. Janbu, N., Senneset, K. (1973). Field Compressometer- Principles and Applications, Proc. 8th International Conference on Soil Mechanics and Foundation Engineering. Vol. 11 p 191- 198, Moscow. Johansson, H.G. (1987). The use of soil radar in different road projects - A state of the art report (in Swedish). Swedish Road Administration. Report 1987:59.
Kallstenius, T., (1963). Studies on clay samples taken with Standard Piston Sampier. Swedish Geotechnical Institute. Proceedings No. 21. Stockholm.
References
83
Landva, A.O. (1980). Vane testing in peat. Canadian Geotechnical Journal Vol. 17. No 1 Feb. 1980. NRC Canada. Landva, A. O. (1986). In situ testing of peat. ASCE Geotechnical Special Publication 6, pp. 191-205. Landva, A.O., Pheeney, P.E., La Rochelle, P. and Briaud J.-L. (1986). Structures on Peatland - Geotechnical Investigations. Proc. Advanced Peatland Engineering. Carleton University 25-26 August 1986. Canada. Larsson, R. (1989). Dilatometerf6rs6k. Swedish Geotechnical Institute, Information No. 10, Link6ping. Larsson, R. (1992). CPTU-sondering. Swedish Geotechnical Institute, Information No. 15, Link6ping. Larsson, R., Bergdahl, U., Eriksson, L. (1984). Evaluation of strength in cohesive soils with special reference to Swedish practice and experience. Swedish Geotechnical Institute. Information 3. LinkOping. Larsson, R. and Mulabdic, M. (1991). Piezocone tests in clay. Swedish Geotechnical Institute, Report No. 42, Linktiping. Marchetti, S. (1975). A New In Situ Test for the Measurement of Horizontal Soil Deformability. Proceedings of a Conference on In Situ Measurement of Soil Properties, June 1975, Raleigh, N.C., Vol. 2, ASCE New York. Marchetti, S. and Crapps, D. K. (1981). Flat Dilatometer Manual. GPE Inc. Gainesville, Florida, U.S.A. Robertson, P.K., Campanella, R.G., Gillespie, D. and Grieg, J. (1986). Use of Piezometer Cone Data. Proceedings of In Situ 86, a Specialty Conference on Use of In Situ Tests in Geotechnical Engineering. Blacksburg, Virginia. ASCE. New York. Schmertmann, J. H. (1986). Suggested method for performing the flat dilatometer test. Geotechnical Testing Journal, GTJ ODJ, Vol. 9, No. 2, June 1986. Schwab, E.F. (1976). Bearing capacity, strength and deformation behaviour of soft organic sulphide soils. Diss. Royal Institute of Technology. Stockholm. Stefanoff, G., Sanglerat, G., Bergdahl, U. and Melzer, K.-J. (1988). Dynamic Probing (DP): International reference test procedure. Proc. First International Symposium on Penetration Testing. Orlando 1988. Balkema, Rotterdam. Swedish Geotechnical Society (1993). Recommended Standard for Cone Penetration Tests. SGF Report 1:93E. Link6ping. Torstensson, B.A. (1973). Kohesionspfilar i 16s lera. Thesis. Chalmers University of Technology. Gothenburg.
84
Site Investigations
Tremblay, M., Eriksson, L. (1987). Use ofpiezometers for in situ measurement of permeability. Proc. 9th European Conference on Soil Mechanics and Foundation Engineering, Dublin 1987, Vol. 1. pp 99-102. Balkema, Rotterdam. Viberg, L. (1984). Aerial photo investigation (in Swedish). The building Handbook "Bygg, Geoteknik, G". Stockholm. Wolski, W., Szyrnanski, A., Mirecki, J., Lechowicz, Z., Larsson, R. Hartl6n, J., Garbulewski, K., Bergdahl, U. (1988). Two Stage-constructed Embankments on Organic Soils. Swedish Geotechnical Institute, Report No 32. Link6ping
Woiski, W., Szymanski, A., Lechowicz, Z., Larsson, R., Hartl6n, J. and Bergdahl, U. (1989). Full-scale failure test on a stage-constructed test fill on organic soil. Swedish Geotechnical Institute. Report No 36. Link6pi
85
Chapter3
Laboratory Investigations z. Lechowicz, A. Szymanski and T. Baranski, Department of Geotechnics, Warsaw Agricultural University
3.1
GENERAL
The common laboratory tests and equipments used for soft mineral soils can also be used for most organic soils. One condition for laboratory tests to be relevant to the field conditions for which the results are applied is that undisturbed and representative samples can be brought into the laboratory and mounted in the testing equipment. Particularly for peats, special samplers have to be used, as well as special trimming and mounting techniques. For fibrous and pseudo fibrous peats many of the common laboratory tests are not relevant. These soil types also often require larger specimens than usual to be tested and special apparatus, such as the compressiometer, to be used rather than ordinary oedometers. 3.2
ROUTINE TESTS
The initial ocular inspection and the very simple tests, with alkali extraction and lowering of samples into hydrochloric acid, required for preliminary soil classification, may be carried out in the field. In the laboratory, further quantitative analyses are carried out to enable a more precise classification and to give additional basic data. These will enable the engineer to obtain a better understanding of the soil behaviour and to use existing empirical knowledge of the properties of different types of soil. 3.2.1
Soil d e n s i t y
The bulk density 9 is determined by weighing an undisturbed specimen with known volume at its natural water content.
o =m/V where m = mass of soil V = volume of specimen
(3.1)
Laboratory Investigations
86
The dry density P d and the natural water content w N can be evaluated by drying and weighing the specimen. Drying is usually performed in ovens at 105 ~ but for highly organic soil it is often recommended that drying be performed at temperatures below 80 ~ to prevent loss of organic material. Pd = m / V
(3.2)
and
=mwlm
(3.3)
where rn = mass of solid material V = original volume of specimen mw= mass of water (m- m). The specific gravity p~ is usually determined by boiling the dry, pulverized sample with distilled water in a pyknometer or volume bottle and weighing the bottle after cooling and adding distilled water up to the volume mark. Especially for organic soils, the water is exchanged for kerosene and the boiling by heating is exchanged for treatment with vacuum.
p~ =
ms.P f
(3.4)
m~-m I + m e where m s = mass of solid material m I = mass of bottle, fluid and soil m 2 = mass of bottle filled only with fluidat the particular temperature Pf = specific gravity of fluid at the particular temperature. The specific gravity of peat soils can also be estimated from empirical relations between specific gravity and ash content (or organic content), (Fig. 3.1). The specific gravity of pure peat (cellulose) is in the order of 1.4 - 1.5 t/m 3 and the most frequently occuring minerals have specific gravities of about 2.7 t/m 3. Ac-
87
Routine Tests
30 Extreme
limits
according__
to Landva et at, (1983)
Hobbs (1986)
2.5 ....,.. O3
E
0_2, 2.0
>
cl
.m .m U
13. UO
1.5
Zawadzki (1970)
1.0 L
I
0
1
20
I
LO 60 Organic content (%)
1
80
I
100
Fig. 3.1. Specific gravity versus organic content.
cording to Hobbs (1986), the specific gravity of peat can for most practical purposes be taken as 3.8 o~ =
(3.5)
0.013 9Organic content + 1.4 By combining the natural water content, the bulk density, the dry density and the specific gravity, other properties such as the degree of saturation (or saturation ratio) Srthe void ratio e and porosity n, can be calculated. Sr = V~/Vp
(3.6)
Laboratory Investigations
88 p, (Wy + 1)
n=V
p
(3.7)
-1
e = Vp/V =
(3.8)
/V=I-
p, (WN+ 1)
where Vw= volume of water Vp = volume of pores V~ = volume of solids V = total volume ( V + Vp)
3.2.2.
C o n s i s t e n c y limits
The Atterberg limits are used to describe how the consistency of the remoulded soil changes with the water content. The liquid limit w L denotes the transition from a liquid to a plastic consistency, the plastic limit Wp denotes the transition from a plastic to a semi-solid consistency and the shrinkage limit w s is the limit where a further reduction of the water content results in air intrusion into the pores but no further reduction in volume of the soil. The liquid limit is determined either by percussion in the Casagrande apparatus or by fall-cone tests, (Fig. 3.2).
Stand arm vertically adjustable
l C
Plexiglass with a mm-graded scale
60g/60~faU-cone
[
I
.n J ;
Sample Nixing cup
Fig. 3.2. Fall-cone apparatus (Karlsson 1981).
Routine Tests
89
The fall cone test is the recommended method in many countries. In Sweden, where the method originates, the liquid limit is evaluated as the water content at which the penetration depth of a cone with an apex angle of 60 o and mass of 60 grams is exactly 10 mm. A one-point method has been worked out and, provided that the penetration depth is between 7 and 15 mm, the liquid limit can be calculated from
wL - M wy + N
(3.9)
where M and N are obtained from Table 3.1. Table 3.1.
Relation between cone penetration i (60g/60 ~ cone) and factors M and N in the formula w E = M w N + N (i=depth of cone penetration)
i mm
0
1
2
7. M N
1.21 -3.5
1.20 -3.4
8. M
4
5
6
7
1.19 1.18 -3.2 -3.0
1.17 -2.9
1.16 -2.7
1.15 -2.6
1.14 -2.5
1.14 -2.3
1.13 -2.2
1.12 -2.1
1.11 -1.9
1.11 1.10 -1.8 -1.7
1.10 -1.6
1.09 -1.4
1.08 -1.3
1.07 -1.2
1.07 -1.1
1.06 -1.0
N
1.05 -0.9
1.05 -0.8
1.04 1.04 -0.7 -0.6
1.03 1 . 0 3 1.02 -0.5 -0.4 -0.3
1.01 -0.3
1.01 -0.2
1.00 -0.1
10.M N
1.00 +0
1.00 +0.1
0.99 0.99 +0.2 +0.2
0.98 +0.3
0.98 +0.4
0.97 +0.5
0.97 +0.5
0.96 +0.6
0.96 +0.7
ll.M N
0.96 +0.7
0.95 +0.8
0.95 0.94 +0.9 +0.9
0.94 0.94 +1.0 +1.1
0.93 +1.1
0.93 + 1.2
0.93 +1.3
0.92 +1.3
12.M N
0.92 +1.4
0.92 +1.4
0.91 0.91 +1.5 +1.5
0.91 0.90 +1.6 +1.7
0.90 +1.7
0.90 +1.8
0.89 +1.8
0.89 +1.9
13.M N
0.89 +1.9
0.88 +2.0
0.88 0.88 +2.0 +2.1
0.88 +2.1
0.87 +2.2
0.87 +2.2
0.87 +2.2
0.87 +2.3
0.86 +2.3
14.M N
0.86 +2.4
0.86 +2.4
0.86 0.85 +2.5 +2.5
0.85 0.85 +2.5 +2.6
0.85 +2.6
0.84 +2.7
0.84 +2.7
0.84 +2.7
N 9. M
3
8
9
According to Hansbo (1957) this penetration depth of the specified cone is obtained when the undrained shear strength is 1.6 kPa. The liquid limit determined according to Swedish practice thus corresponds to a remoulded shear strength of 1.6 kPa. Cones with other geometries and weights are used in some other countries, where plots similar to that in the Casagrande test are used. The criteria of a liquid
Laboratory Investigations
90
limit corresponding to a remoulded shear strength of about 1.6 kPa, however, is roughly the same. The plastic limit Wp is defined as the lowest water content at which a soil specimen can be rolled out to a thread with a diameter of 3 mm without crumbling. Determination of the plastic limit is often difficult in organic soils and is not possible in fibrous peats. The plasticity index Ip is the difference between w L and Wp Ip
= w L - Wp
(3.10)
The liquidity index IL is a measure of the state of the soil in relation to its consistency limits. IL is defined by the equation I L = (w N - wp ) / Ip
100
I II
(3.11)
]
[
I
I
=
[
,,,-I,,~j \
80-
-
o~ ~-
~ (1;
~,~'~J........-.//,~o
~0-
~-.,~
... . .
./'~
v
.
.
"
."
..
o~
~.~.::..."-Y70o
w
20 -
o
~
- .
i..
9. . . . . . " "
~/
~o',
o
,,~.~'Y
\~
o
o ,
O0ooO~ .
~"/
20
o
o
I
1
60 Liquid
o
o o~
.
.
1
o
.
o
~
-
o~ o
o
oO
. o o o
o
100
o
o
o" 0
0
o
0o " o -
, .
o
o
O o
0
~ o o o
o O oO ~ o o
o-o
0
o
o
o oo _
Oo
o
oo
~
~176
oo
o
~
o Oo
o~
.
m i n e r a l soils
I
80 limit, w L
~
o
o o
::ipure
40
o
o
o o
o
9' . ..i
Fig. 3.3.
"
~y...::..Y/oo
~~...i~..~Jfoo
~0-
oO
o _"
~o~..::jSo
O0 -
o
oo;~J.-."......:..;,,//o
~ ~9
/
+""~'~ \ "/o~ ~ Y Y"Yo o - , , ~~ ". . : . . .. ."., S,~o o o
I
120
~ ~ o r g a n i c soils
I
140
(%)
Plasticity chart for Swedish soils. The line Ip = 0.75 (Wp-26), where Ip and w E are given in per cent, can be considered as the borderline between pure mineral soils and organic soils (Hansbo, 1957).
160
Routine Tests
91
The Casagrande plasticity chart is used for division of soils into different categories based on their consistency limits. The A-line in this chart basically separates pure mineral soils and soils with an organic content. The position of the A-line may vary somewhat for different regions. The consistency limits can be used for most fine grained and organic soils, but not for fibrous peat with a low degree of humification.
3.2.3
Organic content
The organic content is determined by dry or wet combustion of the organic matter. A much used method is the "loss on ignition method" where a sample is first dried and weighed (m 1) and then the organic material is burnt off (ignited) at a high temperature (400-900 ~ ). ARer cooling in a dessicator the sample is reweighed (m2) and the loss in mass at ignition is determined. This loss in mass is put in relation to the original mass and the relative value, called "loss on ignition", is assumed to correspond to the organic content. 9 Organic content
= (m 1 - m2)/m I
(3.12)
In highly organic soils, such as peat, it is common to use ash content instead of organic content as a measure of the mineral content in the soil. 9 Ash content = m 2 / m 1 (=1 - organic content)
(3.13)
There are conflicting opinions on the temperatures and associated times for ignition. At higher temperatures, not only the organic material is burnt off but also mineral components. On the other hand, high temperatures are required to ensure that all types of organic matter are combusted. The errors at loss on ignition increase with increasing mineral content, especially if the soil contains carbonates and sulphides. Unless the mineral content is high or the soil contains carbonates, the errors can usually be ignored. This is the case in peats and other highly organic materials. Taking into account the fact that large specimens can be ignited and that there is an uncertainty in converting the organic carbon detenmned in other types of analysis of organic content, the loss on ignition method can be considered to be the most suitable method for these soils. In soils with higher mineral contents and in calcareous soils, the colorimetric
92
Laboratory Investigations
method can be recommended for determination of the organic content. In the colorimetric method, a dry pulverized sample is mixed with a potassium dichromate solution in a retort. Concentrated sulphuric acid is then added, whereby the organic matter is wet-combusted. At oxidation of the organic carbon with dichromate, the colour of the oxidation fluid changes from orange to green. A simple but reliable measurement of the organic carbon content is obtained by measuring the intensity of the green colour with a colorimeter supplied with a filter for wavelengths close to 620 nm. The colorimeter is calibrated for the given test procedure with known amounts of organic carbon, Fig. 3.4. The main sources of error in this method are the relatively small amount of sample in each test and the fact that the conversion factor used to calculate organic content from organic carbon may vary somewhat. Usually, organic matter is considered to contain 58 % organic carbon. Other types of wet-combustion followed by titration or dry-combustion method with determination of evolved carbon dioxide may be used, but they are more complicated, more expensive and often require specially trained staff. They all suffer from the same general shortcoming as the colorimetric method and are in practice not more accurate. The colorimetric method can therefore be recommended for determination of organic content in soils with low and medium organic contents and for highly organic calciferous soils. The method is described in detail e.g. in SGI Report No. 27E, (Larsson et al 1987).
Fig. 3.4. Measurement in a colorimeter (Larsson et al., 1987).
Routine Tests
3.2.4.
93
Carbonate
content
The carbonate content is usually determined by wet combustion with a non- oxidizing acid, generally hydrochloric acid. All types of wet-combustion methods with measurement of the developed carbon dioxide may be used. The carbonate content may also be determined by comparing the losses on ignition at heating to 900 ~ for untreated samples and for samples where the carbonates have been removed by treatment with hydrochloric acid. An alternative method is to compare the losses on ignition with heating to 550 ~ and to 900 ~ respectively. In most of these tests, it is assumed that all carbonates are calcium carbonates. When the amount of dolomite or siderite is great, the error becomes important. Simple but reliable methods are available for quantitative determinations of calcium carbonate by wet-combustion. The simplest apparatus is the Passon apparatus (Fig. 3.5).
Outlet hole for acid
Water
Hydrof-H i-\ chtoric. /~r / acid /
~
Hose
clamp
~ Soil J sample
Reaction bottle
Measuring tube
Fig. 3.5. Passon apparatus (Thalme and Almen, 1975).
In the test, the hydrochloric acid is poured onto the sample by tilting the bottle. After the reaction, the amount of evolved gas is measured in the U-tube, where the pressure is first equalized by letting out water until it is equally high in both legs. 1 mol of CaCO 3 yields 1 mol of CO 2 and the carbonate content can be calculated from the evolved gas volume. The Chittick apparatus is a development of the Passon apparatus (Fig. 3.6). The tilting reservoir has been replaced by a drop funnel with pressure equalization and the measurement system has been improved.
Laboratory Investigations
94
Three-way valve
i
~=
Two-way valve
Burette
Drop funnel with pressure equalization
9
Leveling flask
E-retort Two-way valve Fig. 3.6. Chittick apparatus (Fredriksson and Kjellin, 1973).
In the Moum apparatus, Fig. 3.7, the reaction between acid and soil takes place in a closed system evacuated to vacuum. The amount of evolved CO 2 is then measured indirectly with a mercury manometer. Calibration of the apparatuses is performed with known amounts of pure calcium carbonate. Corrections have to be made for temperature and, except for the Moum apparatus, also for the atmospheric pressure. In the tests, there is also a possibility of appearance of manganese oxide or evolution of hydrogen sulphide. These errors can be eliminated by adding ferrous chloride and copper sulphate, respectively, to the hydrochloric acid. In these simple tests, also calcium carbonate and magnesium carbonate can be separated by reading the evolved gas volume after 30 seconds and after 15-45 minutes (depending upon whether stirring is performed or not). On the first occasion, the calcium carbonate is completely dissolved, but only about 4 % of the magnesium carbonate. On the second occasion, about 96 % of the magnesmm carbonate is dissolved. The separate contents can be calculated accordingly. The methods are described in detail in SGI Report No. 27E.
95
Routine Tests
Graduated burette
Vacuum
_.z
Vacuumtight valves
To vacuum pump
,;! ercurymanometer \,.
Fig. 3.7. Mourn apparatus (Moum, 1967). 3.2.5.
Ferrous sulphide
Determination of ferrous sulphide can be made by wet-oxidation, absorption of evolved sulphur and titration. In these determinations, it is important to use fresh samples which have been taken and treated in such a way that oxidation has been prevented. The specimens are placed in a reaction bottle where hydrochloric acid is slowly added. The solution is then boiled to dissolve the sample. Evolved hydrogen sulphide is transmitted by a nitrogen stream through a cooler to an absorption bottle with a cadmium chloride solution, where the sulphur is precipitated as cadmium sulphide. This continues for about 15 minutes after boiling has stopped (Fig. 3.8). After extraction, the amount of cadmium sulphide may be determined by filtering, drying and weighing the precipitate. The amount of cadmium sulphide can then be converted to content of ferrous sulphide. The cadmium sulphide can also be oxidized by adding a known amount of dissolved iodine to the absorption fluid. The surplus of free iodine is then titrated and the ferrous sulphide can be calculated. Simplifications of the method have been elaborated (SGI Report No 27E) but in spite of this there is no particularly simple method available for this type of determination.
Laboratory Investigations
96
Drop-funnel
1,~ ~
/ ~
~
Nitrogen N 2
Y ri
#
I
,
~, =__~. ,
Reaction vessel
Protective Absorption bottle bottle
j
~
fl[ Heater
Magnetic stirrer 21,
Fig. 3.8. Equipment for extraction in determination of ferrous sulphide (Larsson et al., 1987)
Further ocular inspections can be made with the aid of microscopy. Ordinary microscopes may be used e.g. to identify diatomaceous soils. For a closer study of organic soils and fine-grained organic mineral soils, electron microscopes may be used. Modem electron microscopes are also often equipped with X-ray diffraction facilities, whereby the various elements in the soil particles can be determined.
3.3
DETERMINATION OF STRESS HISTORY
3.3.1
Preconsolidation pressure
The preconsolidation pressure in soft and organic soils, at which large changes in compression characteristics occur, is the result of several processes such as consolidation due to loading, consolidation due to creep effects and thixotropic hardening or development of cementation bonds. These processes often bring that the soil be
97
Determination of Stress History
overconsolidated to some degree and that the subsoil can carry an additional load equal to Acy = ~'p-Cr'vo without any significant deformation. Determination of the preconsolidation pressure ~'p is performed from the stresscompression curve obtained in oedometer tests (Chapter 3.4). Methods for evaluation of Cy'phave been described by Casagrande (1936), S/illfors (1975) and Leroueil (1983). For determination of the preconsolidation pressure ~ p from stress-compression curves obtained in oedometer tests with incremental loading, the Casagrande method is used. This method is presented in Fig 3.9. The incremental tests should be performed with the normal loading procedure, where each increment is equal to the previous load and is applied for 24 hours. The vertical stress should be plotted on a logarithmic scale. R is recommended to plot the stress- compression curve also on arithmetic scales to verify the existence of a preconsolidation pressure and to enable a closer study of the shape of the curve before the parameters are evaluated from the semilogarithmic plot.
I I
A
tt~ I._
Detail
O t~ L.
Detail |
i
i
i
a
,ll
i
|
i
L
Effective verticat stress, 6"v log scare) Fig. 3.9. The Casagrande method for evaluating the preconsolidation pressure.
For oedometer tests performed with constant rate of strain, the preconsolidation pressure is evaluated according to S/illfors (1975). In order to evaluate Cy'p, the two straight parts of the stress-strain curve are extended and intersected (Fig. 3.10). An isosceles triangle is constructed between the lines and the compression curve. The intersection point between the base of the triangle and the upper line represents the preconsolidation pressure cy p. In this plot, the scales should be chosen so that the distance for 10 kPa on the stress axis corresponds to the distance for 1 % vertical
Laboratory Investigations
98
compression on the strain axis. Certain limitations apply for the maximum rate of strain to be used m the tests. v
I
6p
I
e
CO C"
Detail
t
Effective
I
vertical
i
stress, 6"'v
i
Fig. 3.10. The S~illfors method for evaluating the preconsolidation pressure (S~illfors, 1975).
It has been well established that the preconsolidation pressures evaluated by these two methods (the Casagrande method - incremental tests with standard loading procedure and the S~tllfors method - constant rate of strain tests) from tests on good quality samples are directly compatible with the preconsolidation pressure in situ. (Larsson and S~illfors 1985, Larsson 1986). If other tests, test procedures, evaluation methods or more or less disturbed samples are used, the evaluated preconsolidation pressures may deviate significantly from those obtained in the field. This problem was investigated by Leroueil et al. (1983) on the basis of comparatitive laboratory and field tests. The study presented by Leroueil et al. attempts to describe how the in situ preconsolidation pressure should be evaluated from the results of various laboratory consolidation tests. According to this procedure, correcting the laboratory preconsolidation pressure to field conditions requires a number of factors to be considered. Known factors influencing the evaluated preconsolidation pressure are the sample quality, the test method, the loading rate and the method of interpretation. The study shows that for tests and evaluation methods other than
99
Determination of Stress History
the two previously mentioned, empirical corrections have to be applied to the evaluated preconsolidation pressures. In organic soils, the determination of the preconsolidation pressure puts high demands on the quality of the samples and the tests because of the frequently very slight change in the oedometer curve at (y'p. The oedometer tests are often supplemented by the empirical relations presented by Skempton (1954) and Hansbo (1957). Skempton's formula gives a relation between preconsolidation pressure g'p, undrained shear strength l:f~ and plasticity index Iv 13"p= ~fu ( 0 . 1 1 +0.37 Iv )
(3.14)
Hansbo's formula gives the preconsolidation pressure cy p versus undrained shear strength values q:fv (uncorrected shear strength values obtained in field vane or fall cone tests) and liquid limit w L (obtained from fall cone tests) as follows r
I3 P -- q~fv
/(0-45wL )
(3.15)
The validity of these formulas is always uncertain and they should be used as supplements only. They are obviously not applicable in fibrous peat but especially Hansbo's formula has proved to be very useful in many organic clays and gyttjas, (Larsson 1990).
3.3.2
Coefficient of earth pressure at rest
The state of in situ stress anisotropy is usually expressed by the coefficient of earth pressure at rest K o. This parameter is important both for the reconsolidation of samples to in situ conditions, before starting the tests to obtain the deformation and strength characteristics, and for numerical calculations based on more sophisticated soil models (Wolski et al., 1985). Determination of the coefficient K ~ involves measurement of vertical and horizontal stress components in laboratory tests with no lateral strains in the samples. Test results obtained from oedometer or triaxial tests can be analyzed according to S~illfors (1975) by stress path plotting (Fig. 3.11). The steep part of the curve represents the increase in horizontal stress caused by the increase in vertical stress in the overconsolidated region before 13p is reached. The slope of the second part of the curve consequently represents the increase in the horizontal stress caused by an increase in the vertical stress in the normally consolidated region (A(Y'h/ACy'v = Ko"~
Laboratory Investigations
100
I
I
I
6v-ff~ 2
,
'2
'
2
Gv =6v--~ Ub
I
e-"
v~ i_
~
~
/
i_ t"
~ /
~ ~ <~t,e,.,
/ I I Effective horizontal
,--' \
-
Knc-.-_61~ -
% I 1 stress, ~1~
~
Fig.3.11. Determination of the coefficient of earth pressure at rest from laboratory test results (S~Ulfors, 1975).
The K o value in situ can be estimated with Schmidt's (1966) empirical formulae for K o and overconsolidation ratio OCR: K ~ = Kon~ OCR ~
(3.16)
The empirical coefficient cx is defined as the at-rest rebound parameter of the soil and is often assumed equal to sinq~'. A study made by Mayne and Kulhavy (1982) based on a great number of laboratory test results obtained by Adams (1965), and Edil and Dhowian (1981) gave typical values for this parameter between 0.09-0.18 for peat and between 0.3-0.5 for clay. Results obtained by the Department of Geotechnics WAU, for overconsolidated organic soils (peat and calcareous soil) in the Shllfors K o - oedometer and triaxial tests, indicate that the laboratory equipment used for soft clays can be used also to measure vertical and horizontal stress components in determination of K o for organic soils. Examples of K o test results for overconsolidated amorphous peat and calcareous soil are shown in Fig. 3.12.
101
Oedometer Tests 10
c~ 13_ .~
1
I
'
8
Z x/
1
I
Peat
x
x/x
x/
u3-
!~
u~ ~_
I x,, ~,z":,xlx
I
K nc = 0.45
~/x'''x
~//x
~,
r
Catcareous ..... ~
nc Ko
0.60
cO i l li"
0
l
"t,
5
1
10
~
1
15
. . . . .
1
20
,
,1
25
30
Effective horizontal stress, 5~ (kPa) Fig. 3.12 Oedometer test results for evaluating the coefficient K o. Peat and calcareous soil from the Antoniny site. (Lechowicz and Szymanski 1988)
3.4
DETERMINATION OF DEFORMATION AND CONSOLIDATION PARAMETERS BY OEDOMETER TESTS
3.4.1
Incremental loading oedometer tests
(a) Testing procedure: The main characteristics necessary in calculations of settlement and increase in effective stress are the relation between stress and strain and the development of strain with time. For the one-dimensional analysis of consolidation the basic parameters, such as the constrained (oedometer) modulus of deformation M (being the inverse of the coefficient of volume change m v ) or the compression index C , the coefficient of permeability k, and the coefficient of secondary compression C~ can be determined in oedometer tests. These tests can be performed either by incremental loading tests or by continuous loading tests. The normal test procedure of incremental loading, IL, with each load increment being equal to the previous load and a new load increment every 24 hours, was suggested by Terzaghi in 1925. For some highly compressible and low-permeable soils, such as gyttja, this procedure may be too fast to enable an evaluation of all compression characteristics.
Laboratory Investigations
102
Other loading procedures have been suggested to obtain more points and a better defined stress-strain curve. One such method was proposed for clay by Bjerrum (1972). This procedure uses smaller load increments until the preconsolidation pressure is reached. The increments up to the preconsolidation pressure are given a duration sufficient to obtain 100 % primary consolidation. When the preconsolidation pressure is reached the test is continued with doubled load steps with 24 hours duration as in the Terzaghi procedure (Fig. 3.13).
Verticat stress, 6 v (tog scare) !
LO .m I_ 4..,
cq
..--,,
u L.
>
smelt toad steps~ short duration
iI.
Fig. 3.13.
doubted steps standard duration
Compression curve illustrating the loading procedure suggested by Bjerrum (1972).
Any change in the loading procedure, however, entails a change in the stressstrain curve (Fig. 3.14).Thus the use of the Bjerrum procedure brings a curve which, if the Casagrande evaluation is used, gives too high a preconsolidation pressure, requiring correction, and a modulus that is lower than usual. If smaller load steps are used than in the standard procedure, it often becomes difficult to evaluate the end of primary consolidation and the rate of secondary consolidation. Taking these aspects into account, it is often better to use the normal loading procedure and more tests rather than more load steps m one test. The permeability of the soil is often estimated from the shape of the time- settlement curves for
103
Oedometer Tests
Vertical stress, 6"v (tog scale) I
v
~
E
121
\
~__~-r
I--
I/1
\
_=..,
0 u
-E >
Fig. 3.14.
\\ \ \\\ steps
A-doubted toad short duration (not futt consolidation ) B- Bjerrum method C- standard method D- smart toad steps tong duration
\ "\~\\
Compression curves obtained in oedometer tests IL performed with different loading procedures.
the individual load steps. This estimation is very crude. If the permeability is to be evaluated from this type of test, it is better to perform a number of falling - head permeability tests at the end of selected load steps and before the following load increments are applied (Tavenas et al., 1983 ). The tests are performed at such deformations that the relation between permeability and deformation can be evaluated for the range of deformations of interest. In oedometer tests, using fixed or floating rings, friction occurs between the sample and the ring. It is very important to minimize this friction, which is done by using very smooth rings and grease, or, in the case of fibrous peat, by using a compressiometer (Caflsten 1988). In the compressiometer (Fig. 3.15), the confining ring is replaced by a rubber membrane and lateral deformations are prevented by a number of thin rings spaced so that no vertical load can be transferred between the rings.
104
Laboratory Investigations P CI am .
.
.
~ .
Rubbermembrane
.....
Coarsefilter
.
. . . . .
Sample
~
Confining rings_
-
I:-
"
'
"
.-:-!
Fig. 3.15. The compressiometer apparatus.
(b) Deformation and consolidation parameters: The results from incrementally loaded oedometer tests are presented as the relation between void ratio e or strain ~ and effective vertical stress 6". The results are often first plotted on linear scales to enable a closer study of the shape of the curve. The oedometer modulus M, the permeability k, the coefficient of permeability change 13k, and coefficient of consolidation %, may then be evaluated in a similar way as in the CRS-tests (see Chapter 3.4.2). The results are then plotted with the effective vertical stress on a logarithmic scale to enable evaluation of the preconsolidation pressure G'p. From this plot the compression compression index Coot C e ,where Coe = Co/(l+e0),(Fig. 3.16 a), can also be evaluated. If the test is performed with a cycle of unloading and reloading, the swelling characteristics can be determined. These may be expressed as moduli or as swelling and recompression indices, The swelling and recompression indices are evaluated as shown in Fig. 3.16 b. The deformation process in organic soils often strongly deviates from the simple model used in Terzaghi's consolidation equation, which is the basis for the Casagrande and Taylor evaluations of primary consolidation and coefficient of consolidation.
105
Oedometer Tests a)
b) Vertio3[ pressure 6 v (log SCQ[e}
Vertical pressure 6v (log scale)
CO
(aJ
d
E] I..
I...
E~
U .m -i.., I_
A e or E 2
I._ (b >
> /
I_. 0
._d E~ L.
~" E l 0
E] .u_
O
z~e Cc- a log 6'v
o" E] I..
o~ -i..,
A
>
sE
E2
CcE : log 2
Ae r
Cr = A - l o g ~'"v
Fig. 3.16.
or
aEs
--'----
~Iog6"v
txEr CrE = ---------
~[ogB'v
Evaluation of compression index Cc, swelling index C s and rr index Cr"
Results of tests performed by several authors (Ozden and Wilson, 1970; Berry and Poskitt, 1972; Edil and Dhowian, 1979; Szymanski et al., 1983) indicate that in organic soils (a mixture of organic fibres, colloidal particles, water and gas bubbles) the deformation process involves elastic strains of gas, elasto- plastic primary strains of the soil and viscoplastic creep of the soil skeleton. These processes take place simultaneously, which significantly influences the stress-strain-time characteristics. Evaluating primary consolidation, coefficient of consolidation and permeability from the shape of the time-settlement curves may therefore be very misleading. In calculations of settlement, the Terzaghi equation is still normally used with calculation procedures that allow for large deformations and changing parameters and boundary conditions. In these calculations it is important that the degree of non-saturation be accounted for and that the changing permeability be accurately determined. The creep effects, which also can be accounted for are defined by the coefficient of secondary compression C a and can be determined from the time-settlement curves in the incremental load tests. Other rheological models and calculation methods for the process of on consolidation in organic soils have been suggested (Gibson and Lo 1961, Barden 1968, Berry and Poskitt 1972). Relevant parameters used in these models are evaluated from incremental oedometer tests.
Laboratory Investigations
106
The coefficient of secondary consolidation is evaluated for each load step. Fig. 3.17. This parameter is a function of stress history and deformation and the variation within the particular range of stresses and deformations to be applied in the field should be determined. r-
.,.--.
Cl
-,&
t 100
Time (tog scare) v
I,..
~ .,I,-, I..
primary consotidation described by c v
(P > I,-
0 0 ,,.I-, I,.,_
secondary compression .4~.~escribed by Co,,.
13 0
~
Fig. 3.17. Consolidation curve obtained in incremental loading oedometer tests.
The coefficient of secondary compression may be expressed as Ca= de/dlog t or as C e= de/dlog t, (C~ =C~e (1 +e0)). In long-term tests there is sometimes a downwards curvature of the log time-settlement curve m the secondary compression phase. This phenomenon is someteimes called tertiary compression. There is no field evidence of tertiary compression and it may therefore be considered a laboratory effect which need not be included in the test evaluation. The coefficient of consolidation cv is a function of modulus M and permeability k; cv = M. k / g" Pw and should preferably be calculated from directly determined permeabilities. Both M and k vary with stress and strain and consequently cv is a variable. Methods of estimating the coefficient of consolidation and indirectly also the permeability have been presented by Casagrande and Taylor. The results should, as mentioned above, be treated with great caution for organic soils.
107
Oedometer Tests
Continuous loading oedometer tests
3.4.2.
(a) Testing procedure: The incremental loading tests are time-consuming (one to two weeks). The need for time-saving tests and more accurate descriptions of the consolidation parameters and their variation has led to the development of oedometer tests with continuous loading CL, which are performed as: constant rate of strain CRS (Smith and Wahls, 1969), constant rate of loading CRL (Aboshi et al., 1970)~ constant gradient CG (Lowe et al., 1969) or continuous consolidation test CC (Tokheim and Janbu, 1980). These tests give continuous stress- strain relations and the continuous variation of the consolidation parameters. All of these tests give the consolidation parameters (modulus and permeability) and hence also the coefficient of consolidation, but the influence of secondary consolidation can not be separated or evaluated. As the loading procedures are different for the different types of tests, the strain rates become different and therefore also the stress-strain relations obtained in the various types of tests become different. (Fig. 3.18 and 3.19).
0
"-" 10
Effective vertical stress, 6 v (kPa} 100 200 300 !
i
i
400 ,
7.,~ Range for CRS test with tow rote of strain
E -p ~ 20. a (...) I,,.
Ub/6' v
30
Fig. 3.18.
Correlation between CC test with different relations UDI6'v and CRS test (Larsson and S~illfors 1985).
Laboratory Investigations
108 t
Effective vertical stress. 6 v
=10n-2 ~= 10n I~=10n+2 CO
d 0 L.
0 t.) .m .,4-., t_ Q)
II
CRS
- - - - - CGT CC Stress-strain relation for a constant rate of strain Fig. 3.19.
Schematic relation between a CRS test at low rate of strain, a CG test with a low gradient and a CC test with low relation UblG' v (Larsson and S~Ulfors 1985).
The only type of continuous test that to a large extent has been applied to organic soils is the constant rate of strain test. For this test it has been shown that the S~illfors evaluation of the preconsolidation pressure is applicable also for organic soils and that the stress-strain relations evaluated according to Larsson (1981) are directly compatible with the results from standard incremental tests, (Larsson and S~illfors 1985) Fig. 3.20. One condition for this type of test is that the pore pressure should not at any stage of the test exceed 15 % of the applied load. Therefore, also this test may require several days of testing time in highly compressible and low permeable organic soils, (Larsson 1990). The CRS-test is a standard test in Sweden and its use is spreading in other countries. No estimation of the secondary consolidation can be made from this type of test. Supplementary incremental load tests therefore have to be performed unless the empirical relations for C~ are considered sufficient.
Oedometer Tests
109
0
Effective vertical stress,
0
100 l
200
300 I
I
6"v (kPa) L.O0 I
Oo~\ 10.
Range for 24hour readings in standard incremental tests
20-
d ~
30-
0 u
1= > 40-
50-
8m Bisckebol 4 m
B~sckebol
. . . .
Vallda
Fig. 3.20.
Incremental loading 24 hours reading v . o
Correlation between standard incremental test results and results from CRS test (Larsson and S~illfors 1985).
(b) Deformation and consolidation parameters: The oedometer curves from CRS-tests are obtained as continuous relations between effective stress and strain, between modulus and effective stress and between permeability and strain. The average effective stress in the specimen is calculated, on the assumption of a parabolic distribution of pore pressure, according to the equation:
6"v = P / A - 2Ub/3 where P = applied vertical force A = cross-sectional area of specimen u b = pore pressure at undrained bottom
(3.17)
Laboratory Investigations
110
Taking into consideration the additional requirement for this test, i.e. that the pore pressure does not exceed 15 % of the applied vertical stress, the possible error resulting from Eqn. 3.17 is limited. The permeability is calculated from: g 9Pw 9H
de
2 ub
dt
kv =
(3.18)
and the coefficient of consolidation
kvM cv =
(3.19) g . [3w
where g Pw H de/dt M
= = = =
gravity (9.81 m/s 2) density of water height of specimen rate of vertical strain oedometer modulus.
From the continuous stress-strain curve obtained from a CRS-test (Fig 3.21) the preconsolidation pressure Cr'p is first evaluated according to S/fllfors (1975). The curve for stresses higher than cy p is then moved a distance c to the left in the diagram to compensate for rate effects and to become compatible with results from standard incremental tests. To describe the variation in modulus, the curve is divided into three parts. For stresses up to the preconsolidation pressure the modulus is constant, M o. For the part of the curve above the preconsolidation pressure where the stressstrain curve is a straight line (G'p - 6"L ) the modulus has another constant value, M L and for stresses above or"L the modulus increases linearly with the modulus number M'.
The permeability is evaluated by simplifying the log permeability - e curve to a straight line (Fig. 3.21 c). The initial permeability k i is evaluated at the intersection of the straight line and the horizontal line e = 0. The decrease in permeability with compression is expressed by the parameter 13k = -Alog(kv)/Ae. The coefficient of consolidation c v is calculated from Eqn. 3.19 and is thereby corrected by the correction of M (Fig. 3.2 ld).
Oedometer Tests
lll
PermeQbility, k v (m/s)
Effective vertica[ stress,~ (kPa)
I0-I0i~
1
201 p 80 120 '1 . . . . .
~'
0
160
I06, I04
,
gkv
N 10
.
A
CLJ
c
tJ c"
10,
I
~ 2o -~
0 L.
~E
m 15. 121 u
J'k =" A i ~ k v
I
30
L-
> 20-
40
25
2000, 1600
_
,' I/1 c,4
C
.o
%
E
/z
u
>
z 1200 ID
i 400
ML
Fig. 3.21.
. . . .
|
i
/]
0
0 u
A~v
0
/5~/ M'_aM
a csL
i
1
~v l
.
l
o Q;
0
CV=
I07
Mo kv gJ~w
168 169 0
,
Ii ' 80 ' 120
Effective vertical stress, 6v (kPa)
Interpretation of the CRS test results (Larsson and S~illfors 1985). Dashed lines indicate curves corrected for strain rate effects.
Laboratory Investigations
112
3.5
DETERMINATION OF DEFORMATION PARAMETERS BY TRIAXlAL TEST
3.5.1
Testing procedure
For calculations of two-or three-dimensional deformations, the parameters used in the applied soil models have to be determined. For calculations according to the theory of elasticity, Young's modulus E and Poisson's ratio v have to be determined. For elasto-plastic calculations, the yield criteria have to be established and for viscoelastic-plastic calculations, the creep parameters have to be determined. All these parameters are determined in triaxial tests designed to simulate in situ conditions and to measure the relevant parameters for the model to be used in the calculations. Depending on which case, drained or undrained, the calculations aim at, the tests should be carried out as drained or undrained tests. The difference between undrained and drained parameters should be distinguished. Undrained parameters are denoted here as E , v~ and drained parameters as E', v'. The symbols E, v are used in a general sense. Elastic parameters E, v, are evaluated from the results obtained in triaxial tests. The testing method is similar to that used to determine soil shear strength (Chapter 3.6).There are four important factors which have a significant influence on the accuracy of the evaluated E and v values 9 measurement of the vertical strain e~ 9 measurement of the horizontal strain e3 9 reconsolidation of the tested sample (to simulate "in situ" conditions) 9 testing procedure The first and second factors depend mostly on the type of the triaxial apparatus and the method of measurement. There are many methods for determination of the horizontal strain ~3" The methods can be divided into two main types: direct measurement of change in sample diameter reside a triaxial cell (Fig. 3.22; Felix, 1982; Paute, 1983) or measurements of change in distance between the cell wall or some other fixed point on the specimen perimeter (Fig. 3.23; Baranski and Wolski, 1985). Both types of method make it possible to determine volumetric and shear strains as well as dilatancy and creep characteristics in unsaturated soils and in soils saturated by pore water with gas bubbles, such as many organic soils. The compression characteristics of the soil can also be studied during the consolidation phases prior to tests performed to obtain the shear strength. This consolidation is usually performed in small steps with anisotropic stresses adjusted so that drained K conditions (i.e. no lateral strains) prevail throughout the consolidation process. These conditions are similar to those in an oedometer test.
Triaxial Test
ll3
~ ~1 ...... i!
I
7-
Gauges for lateral displacements
"l ~____ J
ik._
r'-
Fig.3.22.
Measurement of the lateral displacements of the sample inside the triaxial cell.
RESULT ON SCREEN
TRIAXIAL CELL
Atternate positions ~of ultrcL~nic head . . . . . ~ ~
-1" T
v//~'/)///~ /
V///////~
\\\
\\\ W--.c~*-~
iiiiiil
:--:---:'
/T~,ox,o~ k~o,, ceil
Fig. 3.23.
specimen
A
1///I 1~ 1
j: j. --
''o~k
on the screen
INTERPRETATION Disturbance area /~. . . . . . rlnl~lat snape
up spot ,.//of specimenEr
, VIII ~/7-~,//~
! I
A'
I!
~oo,ou~o,k
specimen
s ~0
The ultrasonic method for measuring the lateral displacements of the sample in the triaxial cell (Baranski and Wolski, 1988)
Laboratory Investigations
114
The values of the strength and deformation parameters obtained in a triaxial test depend to a large extent on the mode of loading during the test. In order to simulate the field conditions the tests can be performed as four main types: * active tests with increasing vertical stress (rv and constant horizontal stress 6a 9 active tests with constant ~v and decreasing crh . passive tests with decreasing 6v and constant 6h * passive tests with constant cyv and increasing ~h More complicated tests with simultaneous variation of both cy~ and ~h are also possible. The mode of loading is selected so that it closely corresponds to the loading conditions in the field in the particular zone of subsoil beneath the embankment (Fig. 3.24). a) loading
b) unloading zxq > 0
r
///,&,.7/,2A::~///,
zxq < 0
///,&-,7"//4,.'9"///
~
: ',F, Passwe zone
Active zone
7 / / . / / / / / / / / / / / / / / / / / /
/////
6v = const
Active zone
//////////////////////I/II/,,
///Z.,"7//,~//,
la6"v
a6"v < 0
Passive zone
Fig. 3.24. Stress conditions beneath embankment. (a) Loading, (b) Unloading. 3.5.2. Y o u n g ' s m o d u l u s a n d P o i s s o n ' s ratio The results from the triaxial tests are usually presented as the relation between deviatoric stress and vertical strain (Fig. 3.25). Young's modulus of elasticity E is expressed as: A E =
where (CYl-Cr3) = deviatoric stress E1 = vertical strain.
(3.20)
115
Triaxial Test
bl q = (61-63)
q =(61-6 3)
I
/!
/ / i ~(61-6 3)
//
.
.
.
.
.
aE I
J
li
il ~q
a(61-63) E t = ~
[
aE I
--
E. = a(61-631 -= - aE 1
Strain E 1
E1
c) q=(61 -63)
q =61-6 3
':/25 +`
tangent at
/ ~
'
"
n, AB
AF/' /
//
///
Ci~secant OB I _ ,cc~-G3 llaDI
i E~:'"~I{OO,
!
D
Fig. 3.25.
E1
S'" corrected origin
E1
v
Determination of Young's modulus from non-linear stress-strain relations (Head, 1986)
Depending on the method of calculation, three different types of elastic modulus can be evaluated: 9 initial tangent modulus E i - the slope of the tangent at the initial point (the origin) of the stress-strain curve, 9 secant modulus E - the slope of the line connecting two specified points in the curve, the lower of which could be the origin, 9 tangent modulus E t- the slope of the line drawn as a tangent to the stress-strain curve at selected stress level or strain. As has been mentioned above, precise strain measurements during the initial state of stresses are very important in modulus estimation. If the stress-strain characteristics in the initial phase of the test are incorrect, as is shown in Fig. 3.25, that part of the curve should be omitted.
Laboratory Investigations
116
The initial modulus E i has been found to vary with the confining pressure (r3, and this dependence may be expressed according to Janbu (1963) as" E i = m P a ( 0 " 3 / P a )n
(3.21)
where pa = the atmospheric pressure m and n = constants determined in the tests. A vertical compressive stress (Yl applied to an elastic material produces not only a vertical compressive strain el, but also a lateral expansion. The horizontal strains in two perpendicular directions are denoted by e 2 and e 3 and for an isotropic material e 2 = e 3. These transverse strains are related to the longitudinal strain by the equation:
g 2 = g 3 -- - VE 1
(3.22)
in which v is Poisson's ratio for the material. The minus sign indicates lateral expansion corresponding to longitudinal compression. Over a particular stress range, Poisson's ratio is calculated from measured stratus by the equation: AE 3
v -
~
(3.23) Ae~
Poisson's ratio v is numerically the slope of this curve and the value is calculated from a tangent or a secant using the same criteria as for determining Young's modulus (Fig. 3.26). For undrained fully saturated conditions, in which the volume remains constant, Poisson's ratio v has a value of 0.5.
3.5.3.
Bulk m o d u l u s and s h e a r m o d u l u s
Bulk modulus K and shear modulus G are often determined on the basis of E and V as
K=
(3.24) 3(1-2v)
and
117
Triaxial Test Stress 61
/
or
stress level
// ] //
//
I
I
46"1 Es
I
Lateral strain E3
Fig. 3.26.
I
-
a6"1 AE I
Axiat strain
E1
E3 E1
Typical graphical plots of axial stress and lateral strain against axial strain, showing determination of secant modulus and Poisson's ratio (Head, 1986)
E G =
(3.25)
2(l+v) Both moduli can also be determined from laboratory tests: the modulus K on the basis of isotropic compressive stress - volumetric strain characteristic and G from the deviatoric stress - deviatoric strain curve.
3.5.4
Yield envelope and creep characteristics
To predict deformations with an elasto-plastic soil model, the stress ranges for small strains (elastic) and large strains (plastic) should be distinguished. For this purpose, a yield envelope (locus of stress states in q-p' stress space which cause large plastic strain responses to loading) can be determined. Laboratory tests performed on organic soils (Lechowicz and Szymanski, 1988) show that the shape of the yield envelope is anisotropic, as for other natural soft soils. (Fig. 3.27). The yield envelope can be determined by a number of consolidated drained triaxial tests with varying stress paths.
Laboratory Investigations
118
150 -6 "-"
100
"/~\~ r I \
1,.0 ~'-
II
o0V Fig. 3.27.
, ,J/
50
~
modet
l
, 100
.Modified Cam-CLay
150
',tl
:
200
o
.
p'=11316~.26~1 (kPa)
Yield envelope obtained for calcareous soil from the Antoniny site (Lechowicz and Szymanski, 1988).
In organic soils and other soft soils, the deformations are of an elastic-viscoplastic character because of creep effects. To estimate parameters describing the visco-plastic behaviour it is necessary to use suitable laboratory tests. One of the tests employed to describe this behaviour is the triaxial creep test, which can be performed as drained or undrained with different modes of loading. These tests are performed in standard triaxial cells (Singh and Mitchell, 1968; Larsson, 1977). The test results obtained in creep tests depend on the applied stress conditions. Tests performed on organic soils by Lechowicz and Szymanski (1988) show a wide range of deformation rates depending on stress conditions and time after load application (Fig. 3.28). When the deviatoric stress is smaller than the stress that ultimately leads to creep failure the deformation rate regularly decreases, but when the deviator stress is higher than this value, the rate of deformation first decreases and then rapidly increases when failure is approached. The test results show that unless failure is approached, the relation between deformation rate and time for a given constant shear stress in log-log scales is linear. The relation between deformation rate and stress for a given time is described by an exponential function. This fact makes it necessary to use stresses in the creep tests which closely simulate the stress paths occuring in the subsoil.
Triaxial Test
119
a) I
I
I
i
Peat
60
60 0 13.. v
i
Calcareous soil
I
e21 200
o10 8~ 9
Lo 40i
-
e19
40
e7
LO
ii
o18 16217
e6 e5
1:31-
20-
14~15 o12 "13 o11
20
20 ~34 1o
_
1
0
I
I
20
!
I
-
/..0
0
p" 1/3 (6i +263) (kPa)
I
1
20
l
4O
-
I
60
p': I13 (6i+ 26"3} (kPa)
b) 1
t
i
I
9 ~ 101
9~
1
_
10-2
10-3
lCi3
/
~6 51
-,-.,_ l d 6
_
>
i0-I
Fig. 3.28.
!
16 ~ ~3
10-5
io .7 _
!
Calcareous soil
19
10-1
10-2
"~ 16/._ E 0 o
I
Peat
I 1
I 10
I 10 2
J , 10 3 104 105 Time (rain)
~6 e_
11~7 104
I
10
l
102
I
103 104 105 Time (rain)
Creep test results obtained for organic soils from the Antoniny site. Relation between strain rate and time. Higher numbers indicate higher shear stresses. (Lechowicz and Szymanski, 1989).
Laboratory Investigations
120
3.6
DETERMINATION OF SHEAR STRENGTH
3.6.1
Swedish fall-cone test
A simple method to determine undrained shear strength is the Swedish fall-cone test (Fig. 3.2). The fall-cone apparatus is equipped with four different cones, the mass and apex angle of which are 400g/30 ~ 100g/30 ~ 60g/60 ~ and 10g/60 ~ The 60g/60 ~ cone can also be used to determine the fall-cone liquid limit (Chapter 3.2.2). Based on the correlations with vane shear tests by Hansbo (1957), the undrained shear strength l:fomeasured by fall-cone depends on the depth of cone penetration i, the cone apex angle and the cone mass m as follows: ~fo = K m g/i 2
(3.26)
where K= constant, dependent on the apex angle of the cone and to some extent on the type of soil g = acceleration of gravity. Based on tests on remoulded clay, Hansbo (1957) suggested a constant K=0.25 for 60 ~ cones and 0.8 for 30 ~ cones. For undisturbed samples, the value of K for the 30 ~ cone was found to be 1.0. Ranges for this value between 0.8 and 1.2 have been suggested by Gameau and Lebihan (1977) and Wood (1981). A comparison of undrained shear strength values measured by the laboratory vane shear test and the fall-cone test presented by Wasti and Bezirci (1986) indicates that for 30 ~ cone the value of the constant K is equal to 1. Determination of undrained shear strength using Swedish fall-cone test can be performed together with the evaluation of other index properties. However, an application of this test for peat and especially fibrous peat is questionable. According to Swedish practice, the undrained shear strength values zfo obtained from fall-cone tests should be corrected in the same way as vane shear strength values.
3.6.2
Laboratory vane shear test
Another simple method used for determining the undrained shear strength of organic soils is the laboratory vane shear test. The vanes commonly used in laboratory tests have a height/diameter ratio of 2, as in most field vanes, or 1 (Hanrahan, 1954).The undrained shear strength value zf~ is typically calculated assuming full strength mobilization along a circular cylinder circumscribing the vane as
121
Determination of Shear Strength
....
L a b o r a t o r y v a n e s h e a r test 9F a l l - c o n e
10.
test
J
x ~9 05"IJ
.J
0 v
,
i
,
,
,
) , j
1
Fig. 3.29.
q~fv----2
10
!
,
,
u
vv
i
,
...----
100 r f c ' (kPa)
Comparison of shear strength values from laboratory vane and fallcone tests (Wasti and Bezirci, 1986).
Mma x / ( r t D
e (H + D/3))
(3.27)
where Mmax = applied peak torque Dv = vane diameter Hv = vane height The above equation includes the assumption that the soil has isotropic strength properties and that the distribution of shear stress around the sheared cylinder is uniform. The laboratory vane test is usually carried out on a sample remaining in the sampling tube. It requires a minimum of sample preparation, but the testing is performed in uncontrolled stress and drainage conditions. Because of this, the vane shear strength values can be considered as index values only. The results from laboratory vane shear tests should be corrected, as fall-cone tests and field vane tests. By performing laboratory vane shear tests m a triaxial cell, the vane shear test can be conducted in axi-symmetrically loaded samples under known stress conditions with controlled drainage conditions at the boundaries of the tested sample (Law, 1979). The laboratory vane apparatus can also be combined with the Rowe oedometer to examine the change in undrained shear strength during one-dimensional consolidation.
Laboratory Investigations
122
Also the results of laboratory vane shear tests are highly questionable in peat and especially in fibrous peat. The results of vane shear tests depend on the test procedure and it is important that the standard procedure be followed. Therefore the test should start directly after insertion of the vane and the torque should be applied at such a rate that failure occurs within 1 to 3 minutes after the start of the test. If the test is used in conjunction with triaxial apparatus or a Rowe oedometer, this procedure may not be feasible. In this case, correction factors have to be produced for the particular procedure or the results may be used to illustrate relative changes in strength only.
3.6.3
Direct simple shear test
The direct simple shear test has been extensively used for over 50 years to determine shear strength of soft soils. It has been criticized from time to time for nonuniformity of stresses and boundary conditions that hinder a closer evaluation in terms of stress-paths. The test is still very commonly used partly because of the relatively simple testing procedure, but mainly because of the well documented experience. The evaluated undrained shear strengths from this type of test are directly applicable to stability calculations, e.g. Ladd 1981. The results of direct simple shear tests are usually applied to the central part of a slip surface where it is more or less horizontal. As the results of direct simple shear tests are close to the average of active and passive triaxial tests the results are sometimes used for the entire slip surface. Furthermore, the results of direct simple shear tests on undisturbed samples reconsolidated to in situ conditions are normally very similar to the results of corrected field vane shear tests. In direct simple shear tests, a soil specimen is laterally confined by a rubber membrane and a series of thin and evenly spaced tings (SGI device) (Kjellman, 1951; Larsson, 1977) or by a cylindrical wire-reinforced rubber membrane (NGI device) (Bjerrum and Landva, 1966) in such a way that the diameter of the specimen is kept constant. The specimen is consolidated one-dimensionally for a vertical load. The sample is then sheared by moving the top cap horizontally at a constant rate or by incremental horizontal loads while the bottom pedestal is fixed (Fig. 3.30). The pedestal and the top cap may be supplied with short pins which protrude into the specimen to prevent sliding at these interfaces during shear. Shearing of the soil specimen can be conducted in undrained conditions with constant volume or in drained conditions. It is not possible to perform truly undrained tests with pore pressure measurements in the normal direct simple shear device. A comparison between truly undrained direct simple shear tests and conventional constant volume tests presented
123
Determination of Shear Strength
by Dyvik et al. (1987) shows that the test results obtained from these two methods are very similar.
a)
bl
Rejnmf~:C:de
6hc :t~6'vc
i n '
"'S
Filter
=
=
=
1'h
lttttltttI
sliding
Fig. 3.30.
<,c
Direct simple shear device. (a) A specimen maintained by wirereinforced rubber membrane. (b) Stress conditions and shear deformations.
Shear tests should be performed with different vertical consolidation stresses to cover the stress range anticipated in the field. The results of undrained direct simple shear tests on normally consolidated amorphous peat and calcareous soil from the Antoniny site (Fig. 3.31) show that the normalized shear stress at failure for normally consolidated organic soils is not a constant, but decreases with increasing vertical effective stress. The normalized shear stress at failure obtained for overconsolidated organic soils is also not a constant but increases with increasing overconsolidation ratio OCR.
0.5 _.>0.4 U
0.5
Peat 6"b =15 kPa
.~'vc: 20kPa
OLo .4.-" O
c_ 9
0 ~-
f
C"
Nb-,
0.3
N Ul 0. 2
0
c"
z m O.1
0 N
X
80kPa
/S~~'~~x 60kPa
//#//"
.Ne-~ 0.2
80 kPo
Z m 0.1
// r I
I
!
Shear strain, ~ 1%) Fig. 3.31.
/ Z c-2_0kP_o._ _
~
c
\60kPo
//
O E ~
~=20kPa
-e. ~ 0.z.
g ~ 0.3
/
~n ul
~P
Calcareoussoil
)
l
2
A
z,
l
l
10 12 Shear strain, ~ (%) 6
Results of undrained direct simple shear tests on normally consolidated organic soils from the Antoniny site,
8
Laboratory Investigations
124
Effective shear strength parameters obtained from drained direct simple shear tests carried out on organic soils from the Antoniny site are shown in Fig. 3.32. To estimate the corrected angle of internal friction (1)' the test results were corrected for volumetric changes according to Bishop (1954). The values of obtained (1)' from drained direct simple shear tests are somewhat lower than the corresponding values obtained in triaxial tests. Like all other tests, the results of direct simple shear tests are rate dependent. According to the standard procedure used in Sweden, the specimens are consolidated for 24 hours and the rate of shear in undrained test is about 0.6 % of the specimen height per hour. This rate can normally be used also for drained tests when the specimen height is 10 - 20 mm, but especially for highly compressible and low permeable soils, such as gyttja, it is necessary to check whether this rate is sufficiently low for the test to be drained. Tfd (kFb
Peat
i
e r = 27*
?fd
Calcareous
soil
e~- = 25.5"
(kPa) 30
30 20
l'fu = 5.2 kPa
10
9
20 Tfd = 0.35 6v
"rf~ : 6.6 kPa
10.
0 Fig. 3.32.
3.6.4
20
/,.0
60 -6"~, (kPa)
o
~
,
y
~
:0.3~.e~,
I_ _i._~.,....~"*" ..I
~(~;:20kPa
20
'
Lo
'
Co
6'v (kPa)
-
Results of drained direct simple shear tests on normally consolidated organic soils from the Antoniny site.
Triaxial test
In the conventional triaxial test, a solid circular specimen is loaded axi-symmetrically. The fluid pressure acting in the triaxial cell provides two of the principal stresses, while the third is provided by both the fluid pressure and the axial force imposed by the piston. Consolidation of the sample can be applied three-dimensionally at any given ratio of axial to lateral stress. Although in situ stress conditions are usually anisotropic, isotropic stress conditions are often used in routine triaxial tests for soils with the coefficient of earth pressure at rest higher than 0.8. Testing of the sample can be conducted at any ratio of principal stresses or at a constant mean stress (Bishop and Wesley, 1975). To model the loading conditions imposed by the construction of an embankment, active triaxial compression tests can be performed
125
Determination of Shear Strength
by increasing the axial stress or decreasing the lateral stress, and passive triaxial extension test can be performed in the opposite way (Fig. 3.33). To simulate the field drainage conditions, triaxial tests can be carried out in drained or undrained conditions (Head, 1986).
compression
compression
!
1461
I
U~
c5
c~
II
II
A~I
extension
Fig. 3.33.
9
extension
Various loading paths in triaxial compression and extension tests used to model the loading conditions beneath embankments.
Because organic soils are often extremely loose and sensitive to disturbance, special equipment should be used for mounting specimens in the triaxial cell. Due to difficulties in the preparation of organic soils with a high fibre content, the triaxial tests should be performed on specimens with a larger diameter than that used in standard tests (Landva et al., 1986). The specimen in a triaxial cell should be confined by a rubber membrane with very good resistance to water diffusion to avoid leakage problems. To reduce problems with leakage through the rubber membrane, the triaxial cell can be filled with liquid paraffin (Berre, 1981) or castor oil (Tavenas et al. 1983). Filter paper drains fitted to the side of the sample help to equalize pore pressures as well as provide short radial drainage paths. In the case of low permeable organic soils, they reduce the testing time both for consolidation and for the test phase. To reduce the effect of the side drains on the stress conditions, the filter paper drains should be placed in spirals around the specimen with inclination 1:1.3 for compression tests and 1:1.5 for extension tests (Berre, 1981). The results of triaxial tests on heterogeneous soils (mixed soils, coarse soils and fissured soils) are sensitive to the size of the specimen and in such materials large specimens should be tested.
Laboratory Investigations
126
The results of all triaxial tests are sensitive to the testing procedure and the imposed strain rate. The standard rate used in Scandinavian laboratories for undrained tests on specimens with a diameter of 50 mm and a height of 100 mm is a deformation of 0.6 % per hour. This rate has been found to give appropriate results for calculations of stability in soft clays. Investigations by Graham et al. 1983 have shown that in this rate region a tenfold decrease in strain rate results in a decrease in shear strength of 10 - 20 % for clays (Fig. 3.34). Similar results have been obtained for organic clays. Considering .that some organic soils such as gyttja and clayey gyttja often have coefficients of consolidation about one order of magnitude lower than soft clays, it would be prudent either to reduce the testing rate to 0.06 % per hour or to use a correction factor of 0.8 to 0.9 on results obtained with standard tests. Belfast clay___,._ ~
9
u
~,~ 0.5 -4--"
0.4
tn
i.. ETI r
0.3
_- - - o ~ ~ ~
~
.m
t_ r-
0.2
"a
~-
C
~__._.~a
0.1 -
L,. C
0
i
0.003
Fig.3.34.
0.01
I
0.03
1
0.1
9CAU o CAU 9CAU a JCAU 0.3
Triaxiat Triaxiai Triaxial Triaxiat J
compression-constant rate compression-relaxation compression-step changing extension-relaxation i I
1 3 Strain rate ( % / h )
10
Influence of strain rate on undrained shearing resistance from undrained triaxial compression and extension tests (Graham et al., 1983).
The results of undrained triaxial tests can be evaluated as undrained shear strengths from the stress-strain curves and as effective strength parameters at constant volume from the stress-paths in the tests (Fig. 3.35). Drained tests must be run slowly enough to allow water to drain out of the sample so that no appreciable pore pressure can build up. Filter paper side drains and drainage at both ends are used to speed up the water flow from the specimen. The highest permissible rate of axial displacement (VlmJ during drained tests in such condition can be calculated from the following expression:
Determination of Shear Strength
g.
L,O
127
~ Catcareous soil
Peat c'= 2kPa
"-" z'Oi
c'=3kPa
I
'
~ 2o 0
..-
Fig. 3.35.
0
20
o.s
40
60
I lkPo !
80
'-'
1
0
I,
,
|
,
20
,~
LO
,
,
60
,
,
~.
~_____
80
s':0.5(6'~-6'3 ) (kPa)
Stress paths in undrained triaxial compression tests performed on organic soils from the Antoniny site.
(V1 max ) = (H Elf )/(15 tl0 0 )
(3.28)
where H = height of sample e~f = expected axial strain at failure tl00 = time required for 100 % of primary consolidation. Data from the consolidation stage can be used to determine h00 and a suitable rate of strain for the drained tests can be calculated. For organic soils with low permeability, shearing to failure sometimes takes several weeks. Triaxial tests on fibrous peat are very difficult to interpret. The fibres act as a horizontal reinforcement, so failure is seldom obtained in a drained test. Only large compressions occur. In undrained tests, failure usually occurs when the pore pressure build-up is so large that tensile stresses occur and the sample cracks. This is very different to the behaviour of granular materials and clays and should not be interpreted in the same way.
3.7
D E T E R M I N A T I O N OF P E R M E A B I L I T Y
The permeability of a soil is a function of its void ratio. Many authors (Berry and Poskitt 1972, Larsson 1981, Tavenas et al. 1983) have found that for the working range of compressions the permeability can be simplified to a straight line in the log permeability - strain plot. The logarithm of the permeability can thus be written as: log k e - log k i - I]k" e
(3.29)
Laboratory Investigations
128 where ke = permeability at strain e k i = initial permeability ~k = permeability change index e = strain
The initial permeability is determined by field or laboratory tests. The permeability change index is determined by laboratory tests or may, for some types of soil, be estimated from empirical correlations with water content or void ratio. For some types of peat, both the initial permeability and the permeability change index can be estimated from empirical correlations with the natural water content (Carlsten 1988) Fig. 3.36. These values are to be used in preliminary calculations only and have to be verified by laboratory tests before they are used in design calculations. -7
I
I
I
I
w
I
V////~
I
I
8
I
Peat H5 - H 10
o ._~
-
- 50
1000
2000
Water content, w (%)
Fig. 3.36.
z, 0
1
1000 Water content, w
1
2000 (%}
Relation between permeability and water content for peat. (a) Initial permeability coefficient kvo. (b) Permeability change index 13k (Carlsten, 1988)
Depending on the type of design method to be used for the embankment and type of calculations to be performed, the horizontal permeability k h may have to be determined as well as the vertical permeability k v. The vertical permeability is normally determined by CRS-tests or by permeability tests at various deformation levels in the incremental oedometer tests. In special cases, where it may be assumed that creep effects have a negligible influence on the course of the primary consolidation and the soil is fully saturated, the vertical permeability may be roughly estimated from the time settlement curves in the incremental tests.
Determination of Permeability
129
Permeability tests in both directions can be performed on soil specimens in the Rowe cell (Fig. 3.37; Rowe and Barden, 1966) either with flow of water in the vertical direction (upwards or downwards) or with radial flow horizontally (inwards or outwards). The permeability tests are carried out after vertical consolidation for incremental loads at the end of each loading stage with flow and drainage conditions designed to simulate the field conditions.
,• Y
o)
II
!1
Back pressure system (1)
"~
......
(c tosed ) ~x'~X'~'~NxNXx~ ~.XXXXx~lj~NXXXx~~'~XXXXxN~ outflow~ p I) ~ (with volume change gauge) ~~--..~F/'~ rigid plate with centrat hole
|
~------~11 _---:----~
!'i:!i imi; i,' t'e
r upward flow (shown)... P2 > Pl for downward flow ........... Pl > P2
Back pressure system (2)
Back pressure system { 1 ) ~n flow ~)
(closed)
F~o~ou~plo~tio F~--;-~ ! /
~ r-lip t e~ , ' ~ ' ~
d:t::age hole plugged
K~,\\\\\\\\\\~
out flow (
sand drain
\
]
For flow radiaUy inwards (shown)..p1 > P2 For flow radially outwards ........ P2 > Pl
Fig. 3.37.
Back pressure system (2) (with volume change gauge)
Arrangement of Rowe cell for permeability test. (a) Vertical flow. (b) Radial flow (Rowe and Barden, 1966).
Laboratory Investigations
130
Measurement of permeability in the Rowe cell can be performed with four different types of testing procedures (Fig. 3.38): 9 with the flow of water vertically upwards, 9 with water flow vertically downwards, 9 with horizontal flow radially outwards, 9 with horizontal flow radially inwards. Back pressure
Diaphragm
Back pressure
b)
..............................;~//. Back pressure (2) ~ ~ (Inflow) ......
/, ~W"7)'7"~ ~YJ7);2J /, Back pressure --------d;b(... Outflow) (2) b.p.(1) > b.p.(2)
b.p.(2) > b.p.(1)
d)
/7 S,
"//////,
Back pressure (1)
S,, , @ (Inflow) Back pressure b.p.(2) > b.p.(1) (2)
Fig. 3.38.
| ~
.
[
~
Back /~ pressure 7
':i~,
(~ ,(Outflow) b. p.(1) > b.p.(2)
Back pressure (2)
Flow conditions for permeability test in Rowe cell. (a) and (b) Vertical flow. (c) and (d) Radial flow (Head, 1986).
Permeability tests are usually carried out under "equal strain" loading, i.e. with a rigid plate on top of the specimen to maintain a uniform sample thickness. The calculation of permeability from vertical and horizontal flow tests is performed for each of the tests at various deformations from measurements of water flow rate and hydraulic gradient across the sample. The vertical permeability k is calculated from:
Determination of Permeability
131
(3.3o)
k v - q/(A 9i) or
kv - q .
H/(A.
Ap)
(3.31)
where A = cross section area of sample i = hydraulic gradient q = flow rate of water H = height of sample Ap = in difference water pressure at the ends of the sample. The horizontal permeability is evaluated assuming axi-symmetrical radial flow of water. The coefficient of horizontal permeability k h is calculated from:
kh =
q ' P w "g
In (D/d)
(3.32)
2 7 z ' H " Ap where D = diameter of sample d = diameter of central drain Ap = applied pore pressure difference Pw = density of water g = 9.81 m/s 2 The relaton between the vertical and horizontal permeabilities is the permeability anisotropy of the soil. The anisotropy of the permeability is especially pronounced in peats with low degrees of humification and some varved and layered soils which have very high horizontal permeabilities k h as compared to the vertical k v. The anisotropy decreases with increasing degree ofhumification and, as in other homogenous soils, it becomes negligible for highly decomposed peats.
3.8 REFERENCES Aboshi, H., Yoshikumi, H. and Maruyama, S. (1970). Constant loading the consolidation test. Soils and Foundations, Vol. 10, pp.43-56.
132
Laboratory Investigations
Adams, J.I.(1965). The engineering behaviour of Canadian Muskeg. Proc. of the 6th International Conference on Soil Mechanics and Foundation Engineering, Montreal, Vol. 1, pp.3-7. Baranski, T. and Wolski, W. (1985). Ultrasonic testing for the strain characteristics of soil, ASTM Symposium on Consolidation of Soils, Fort Lauderdale, Florida, pp.516-525 Barden, L. (1968). Primary and secondary consolidation of clay and peat, Geotechnique, Vol. 18, pp.345-362 Berre, T., (1981). Triaxial testing at the Norwegian Geotechnical Institute. Norwegian Geotechnical Institute, Publication No. 134. Berry, EL. and Poskitt, T.J. (1972). The consolidation of peat. Geotechnique, Vol.22, pp.27-52. Bishop, A.W. (1954). Correspondence on shear characteristics of a saturated silt measured in triaxial compression. Geotechnique, Vol. 4, No. 1, pp. 43-45. Bishop, A.W. and Wesley, L.D. (1975). A hydraulic triaxial apparatus for controlled stress path testing, Geotechnique, Vol. 25, No. 4, pp. 657-670. Bjerrurn, L. (1972). Embankments on soft ground. Proc. of the ASCE Specialty Conference on Performance of Earth and Earth Supported Structures, Purdue University, Lafayette, Vol. 1, pp. 1-54. Bjerrurn, L. and Landva, A.O. (1966). Direct simple shear tests on a Norwegian quick clay. Geotechnique, Vol. 16, No. 1, pp. 1-20. Casagrande, A. (1936). The determination of the preconsolidation load and it's practical significance. Proc. 1st International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Mass, Vol. 3, pp. 60. Carlsten, P. (1988). The use ofpreloading when building roads an peat. Proc. of the 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn, pp. 135-143. Carlsten, P. (1988). Geotechnical properties of peat and up-to-date methods of design and construction on peat. State-of-the-art report. 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn. Also in Swedish Geotechnical Institute Varia No. 215. Link6ping. Dyvik, R., Berre, T., Lacasse, S. and Raadirn, B. (1987). Comparison of truly undrained and constant volume direct simple shear tests, Geotechnique, Vol.37, No.l, pp. 3-10. Edil, T.B. and Dhowian, A.W., (1979). Analysis of long therm compression of peats. Geotechnical Engineering, Vol. 10, No. 2, pp. 159-178.
References
13 3
Edil, T.B. and Dhowian, A.W. (1981). At-rest lateral pressure of peat soils, Journal of the Geotechnical Engineering Division, ASCE, Vol.107, No. GT2, pp. 201-217. Edil, T.B. and Mochtar, N.E. (1984). Prediction of peat settlement. Proc. of the Symposium on Sedimentation/Consolidation Models, San Francisco ASCE, pp. 411-424. Felix, B. (1982). Systeme de mesure des deformations radiales pour eprouvettes de sol. Laboratoire Central des Ponts et Chaussees, Paris. Fredriksson, D. and Kjellin, B., (1973). Meddelande till kunder vid jordartslaboratoilet. Swedish Geological Survey. Garneau, R. and Lebihan, J.P. (1977). Estimation of some properties Champlain clays with the Swedish fall cone. Canadian Geotechnical Journal,, Vol. 14, No. 4, pp. 571-581. Gibson, R.E. and Lo, K. (1961). A theory of consolidation for soils exhibiting secondary compression. Acta Polytechnica Scandinavica No. 41, pp. 1-15. Graham, J., Noonan, M.L. and Lew, K.V. (1983). Yield states and stress-strain relationships in a natural plastic clay. Canadian Geotechnical Journal, Vol. 26, pp. 502-516. Hansbo, S. (1957). A new approach to the determination of the shear strength of clay by the fall-cone test. Swedish Geotechnical Institute, Proceedings No. 14, pp.5-47. Hanrahan, E.T. (1954). An investigation into some physical properties of peat. Geotechnique, Vol. 4, pp. 108-123. Head, K.H. (1986). Soil laboratory testing, Vol. 3, Effective stress tests, Pentech Press. Hobbs, N.B. (1986). Mire morphology and the properties and behaviour of some British and foreign peats. Quaternary Journal of Engineering Geology, Vol. 19, pp. 7-80. Janbu, N. (1963). Soil compressibility as determined by oedometer and triaxial tests. Proc. 3rd European Conference on Soil Mechanics and Foundation Engineering, Vol. 1. Wiesbaden. Karlsson, R. (1981). Consistency Limits. Swedish Council for Building Research, Document D9:1981, Stockholm. Kjeliman, W. (1951). Testing the shear strength of clay in Sweden. Geotechnique, Vol. 2, No. 3, pp.225-232.
134
Laboratory Investigations
Ladd, C.C. (1981). Discussion on laboratory shear devices. Laboratory shear strength of soil, ASTM STP 740, pp. 643-652. Landva, A.O., Korpijaakko, E.O. and Pheeney, P.E. (1983). Geotechnical classification of peats and organic soils. Symposium on testing of peats and organic soils, ASTM, pp. 37-51. Landva, A.O., Pheeney, P.E. and Mersereau, D.E. (1983). Undisturbed sampling of peat. Symposium on testing of peats and organic soils, ASTM, pp. 141- 156. Landva, A.O., Korpijaakko, E.O. and Pheeney, P.E. (1986). Notes on the original von Post peat and peatland classification system. Proc. Advances in Peatlands Engineering, Ottawa, pp. 17-29. Larsson, R. (1977). Basic behaviour of Scandinavian soft clays. Swedish Geotechnical Institute, Link6ping, Report No. 4. Larsson, R. (1981). Drained behaviour of Swedish clays. Swedish Geotechnical Institute, Linkrping, Report No. 12. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Link6ping, Report No. 29. Larsson, R. (1990). Behaviour of organic clay and gyttja. Swedish Geotechnical Institute, Link6ping, Report No. 38. Larsson, R. and S~illfors, G. (1985). Automatic continous consolidation testing in Sweden. ASTM Symposium on consolidation of soils: Testing and evaluation, Fort Lauderdale. ASTM STP 892 Consolidation behaviour of soil. Larsson, R., Niison, G. and Rogbeck, J. (1987). Determination of organic content, carbonate content and sulphur content in soils. Swedish Geotechnical Institute, Link6ping, Report no. 27E. Law, K.T. (1979). Triaxial-vane tests on a soft marine clay. Canadian Geotechnical Joumal, Vol. 16, pp. 11-16. Lechowicz, Z. and Szymanski, A. (1988). Deformation analyses of organic subsoil in anisotropic stress conditions. Archiwum Hydrotechniki, Gdansk. Vol. XXV, pp. 125- 133. Lechowicz, Z. and Szymanski, A. (1988). Creep behaviour of organic soils, Annals of Warsaw Agricultural University, Land Reclamation, No.24, pp. 99 - 106. Leroueil, S., Collins, G. and Tavenas, E, (1983). Total and effective stress analyses of slopes in Champlain sea clays. Symposium on Slopes on Soft Clays. Swedish Geotechnical Institute, Link6ping, Report No. 17, pp. 293-321.
References
135
Lowe, J., Jonas, E. and Obrician, V., (1969). Controlled Gradient consolidation Test. Journal of the Soil Mechanics and Foundation Division, Vol. 95, No. SM 1, pp. 77-97. Mayne, P.W. and Kulhawy, EH., (1982). K0-OCR Relationships in soil. Journal of the Geotechnical Engineering Division, ASCE, No. GT6, pp. 851-871. Moum, J., (1967). Determination of inorganic and organic carbon in soil samples. Internal report, Norwegian Geotechnical Institute, Oslo. Ozden, Z.S. and Wilson, N.E. (1970). Shear strength characteristics and structure of organic soils. Proc. of the 13th Muskeg Research Conference, Fredericton, New Brunswick, National Research Council, Technical Memorandum No. 99, Ottawa. Paute, J.L. (1983). Comportement des sols supports de Chaussees a 1' appareil triaxial a'chargements repeter. Bulletin 123, Laboratoire Central de Ponts et Chaussees, Paris. Rowe, P.W. and Barden, L. (1966). A new consolidation cell. Geotechnique, Vol. 16, No. 2, pp. 162-170. S~ilifors, G. (1975). Preconsolidation pressure of soft, high-plastic clays. PhD Thesis, Chalmers University of Technology, Gothenburg, Sweden. Schmidt, B. (1966). Earth pressure at rest related to stress history. Canadian Geotechnical Journal, Vol. 3, pp. 239-343. Singh, A. and Mitchell, J.K. (1968). General stress-strain-time function for soils. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 94, No. SM 1, pp. 21-46. Skempton, A.W., (1954). Discussion of the Structure of Inorganic Soil. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 80, p. 478, New York. Smith, R.E. and Wahls, H.E. (1969). Consolidation under constant rate of strain. Journal of the Soil Mechanics and Foundation Division, ASCE, Vol. 95, No. SM 2, pp. 519-539. Szymanski, A., Fiirstenberg, A., Lechowicz, Z. and Wolski W. (1983). Consolidation of organic soils. Proc. of the 7th Danube European Conference on Soil Mechanics and Foundation Engineering, Kishinev, Vol.II, pp. 273-278. Tavenas, E, Leblond, P., Jean, P. and Leroueil, S. (1983). The permeability of natural soft clays. Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 629- 644.
Tavenas, F., Jean, P., Lebiond, P. and Leroueil, S. (1983). The permeability of natural soft clays. Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 645- 660.
136
Laboratory Investigations
Terzaghi, K. (1925). Principles of Soil Mechanics, Engineering News-Record 26. Thalme, O. and Almen, K-E. (1975). Jordartsanalys. Laboratorieanvisningar, Del 1. Kvart~rgeologiska institutionen, Stockholm University. Tokheim, O. and Janbu, N. (1980). A continous consolidation test. Norwegian Institute of Technology, Geotechnical Division,. Meddelelse No. 9. Wasti, Y. and Bezirci, M.H. (1986). Determination of the consistency limits of soils by the fall cone test. Canadian Geotechnical Journal, Vol. 23, pp. 241- 246. Woiski, W., Baranski, T., Garbulewski, K., Lechowicz,Z. and Szymanski, A. (1985). Testing of anisotropic consolidation in organic soils. Proc. l lth International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 2, pp. 699-702. Wood, D.M. (1981). Cone penetrometer and liquid limit. Geotechnique, Vol. 32, No. 2, pp. 152-157. Zawadzki, S. (1970). Relationship between the content of organic matter and physical properties ofhydrogenic soils. Polish Journal of Soil Science, Vol. III, No. 1.
137
Chapter4
Stability Analysis z. Lechowicz, Department of Geotechnics, Warsaw Agricultural University
4.1
GENERAL
For embankments and dikes on organic soils, an evaluation of the stability during construction is of major importance. Stability analyses may also have to be carried out for other particular loading situations, such as: (a) steady seepage conditions (b) sudden drawdown of the water level and (c) application of additional load. In estimation of stability of embankments on organic soils, two particular cases may be identified: single-stage loading - when the initial shear strength of the soil is sufficient to support safely the maximum embankment load, and multi-stage loading (stage-constructed embankments) - which requires evaluation of the shear strength increase with load and time for the design of a safe loading rate. In evaluation of the stability of embankments on organic soils, more attention should be paid to the appropriate selection of initial values of soil parameters and their changes due to consolidation than to the application of more complex methods of stability calculation. Therefore, this chapter gives an outline of the approach for estimation of stability used in current design practice with special focus on the selection of initial values of shear strength and prediction of the increase in shear strength, as these parameters are of major concern when loading organic soils. In geotechnical practice, the methods of slices, based on the limit equilibrium approach, are commonly used for estimation of stability. The presentation in this chapter is mainly concemed with these methods, but simple procedures of stability evaluation and trends in advanced techniques for stability assessment are also presented. Taking into account the way in which the available shear strength on a potential failure surface acts and how the resulting factor of safety is computed, two types of analysis are usually considered: (a) total stress analysis and (b) effective stress analysis. The total stress analysis assumes undrained conditions and consequently uses undrained shear strength parameters. The effective stress analysis can consider undrained or drained conditions using effective stresses and effective strength parameters.
13 8
Stability Analysis
4.2
SHEAR STRENGTH USED IN STABILITY ANALYSIS
4.2.1
Problems in the evaluation of shear strength
For design of embankments on organic subsoil, the information concerning in situ shear strength is often insufficient. Moreover, changes in shear strength resulting from future changes in the stress conditions are required. The shear behaviour of organic soils is strongly affected by the pore pressure, the effective stress state and the stress history. In engineering practice several different laboratory and field tests are used to estimate the shear strength. In situ methods, such as the vane shear test, dilatometer test and penetration tests, are often used with preference as they are costeffective and also avoid some of the problems associated with disturbance of soil samples during extraction from the subsoil. However, these methods can only be used to determine the shear strength at the prevailing stress conditions and the interpretation of the results often has to be calibrated against qualified laboratory tests. On the other hand, sampling and laboratory testing permit modelling of almost any stress condition and evaluation of the change in shear strength with changing stress conditions. Both in situ and laboratory testing are normally used simultaneously. The testing rate in in situ tests is usually so high in relation to the low permeability of soil and organic soils that the tests can be considered as completely undrained. Fibrous peat is an exception as the permeability may be so high that the tests become at least partly drained. In most laboratory tests, the drainage conditions can be controlled and therefore the undrained as well as the effective shear strength parameters can be estimated.
4.2.2
Undrained shear strength
It has been widely recognized that the shear strength value 1;f~ obtained from field vane shear tests cannot be used directly in the calculation of the stability of embankments. To evaluate undrained shear strength of organic soils from vane shear tests, the measured values have to be corrected according to: % = ~t % where Zfu = corrected undrained shear strength g = correction factor Zfv = shear strength value from vane shear test
(4.1)
139
Shear Strength used in Stability Analysis
Bjerrum's correction factor g is related to the plasticity index Ip and is often used, although it is mainly valid for normally consolidated low plastic clay deposits (Bjerrum, 1972) (Fig. 4.1). A revised version of this correction, which attempts to consider the influence of end effects in the full scale failures forming the empirical basis of the factor, gives corrected field vane shear strengths about 10 % lower than those previously recommended by Bjerrum (Azzouz et al., 1983).
:~. {.,.
1.2
I
I
I
I
1
- 1.1-
0
U
o 1 0 -. IE O
u
Bierrum (1972)
0.9-
I... I...
o
o
0.80.70.60.5
0
Azzouz
11983) I
20
et
at./ I
40
I
60
I
80 Plasticity index,
Fig. 4.1.
1
100 Ip
120 (%)
Field vane strength correction factor proposed by Bjerrum (1972) and revised version presented by Azzouz et al. (1983).
Based on Swedish experience in normally consolidated and slightly overconsolidated high plastic soft soils, the Swedish Geotechnical Institute (Larsson et al., 1984) has recommended the correction factor g calculated as a function of the liquid limit w L (Fig. 4.2). Correction factors higher than 1.2 should not be used without supporting evidence from complementary ilwestigations. The lowest correction factor is 0.5. All these correction factors have been derived from empirical experience and are averages of data with a considerable scatter. If there is any doubt about the validity of the correction factors in a particular soil of interest, the field vane test should be calibrated against results from triaxial and direct simple shear tests. Several investigations have been carried out to observe the mode of failure during field vane shear tests in organic soils. Field vane shear tests (Landva, 1980) performed by the 65xl 30 mm vane in sphagnum peat showed that the diameter of the
Stability Analysis
140
1.3
I
I
I
i
I
I
I
:~,
I
I
0.45
~
=
U
Ol. 1 C
"
.s "~ 1.OL. 0
"
0.9O.80.70.60.5
0
I
I 0.4
I
I 0.8
~
J 1.2 Liquid
Fig. 4.2.
l
J 1.6
2.0
limit, WL
C o r r e c t i o n f a c t o r f o r s h e a r s t r e n g t h v a l u e s o b t a i n e d by v a n e s h e a r t e s t or S w e d i s h f a l l - c o n e test ( L a r s s o n et al., 1984).
observed failure surface was 7-10 mm greater than that of the vane. Based on the measurements of the failure zone during vane shear tests, Golebiewska (1976, 1983) has proposed the correction factors for peats ~t = 0.5-0.55 and for gyttja and calcareous soil ~t = 0.6-0.8. In organic soil, the scatter in the results from field vane shear tests is often quite large. To obtain the undrained shear strength profile for design, the average of several corrected shear strength values should be calculated (Larsson et al., 1984; Baecher and Ladd, 1985). An example is shown in Fig. 4.3, where shear strength values obtained in field vane shear tests corrected according to the SGI recormnendations are presented. Fig. 4.3 also shows undrained shear strength obtained from triaxial compression tests (TC), direct simple shear tests (DSS) and triaxial extension tests (TE) as well as the average undrained shear strength obtained from laboratory tests. A comparison between the average undrained shear strength obtained from laboratory tests and the corrected shear strength from field vane shear tests indicates that for the organic soils at the Antoniny site, the correction factors estimated according to the SGI recommendations are in general agreement. Differences were obtained for calcareous soil where a reduction in addition to the general correction factor was necessary. The calibrated correction factor for calcareous soil estimated on the basis of the average laboratory undrained shear strength and field vane shear strength
Shear Strength used in Stabili.tF Analysis
0
0
~
I
I,I~L.
corrected
.c:: 4 r'~
field
vane
_.
5 - o (/) (/)
ov,ro,,
Lob.
I
//~J/clverag e f ~ / " ,measured I,,~'C field v a n e
~,~t,
2-a_
6-
I
,•/
0
0
Undrained shear strength, (kPa) 10 15 20 25
5
_
-E i3 -
141
II
"~x ~"~~>
Cr
l
"-"~-,.~>
.,,.p
,'Y/-
#1c if!
O (b 1..
8
7--~
O
_+one standard deviation I
1
I
I
Fig. 4.3. Undrained shear strength profile obtained from field vane shear tests at the Antoniny site.
values equals 0.6, while a correction factor obtained from the SGI procedure was 0.7. In order to estimate calibrated correction factors which correspond to different modes of shearing, the undrained shear strength obtained from different types of laboratory tests can be related to field vane shear strength values. In this way, more reliable correction factors for field vane shear strength values can be obtained. The cone penetration test with simultaneous measurement of pore pressure, i.e. the piezocone test, can be used to estimate the undrained shear strength, (see Chapter 2.5). Similar to the correction factor p to the vane shear test, the cone factor NKT can be calibrated locally for a specific type of soil in order to obtain more reliable values. An exalnple of a such calibration in a soil stratum with peat and calcareous soil is given in Fig. 4.4. In this particular case the cone factor was found to be 22 for the peat and 16 for the calcareous soil. The dilatometer test can also be used to obtain the undrained shear strength (see Chapter 2.5). Alternatively, the undrained shear strength ~fu can be estimated from
142
Analysis
Stability
Total point resistance (kPa)
qT 0
!
200 !
0(b CL
I
600
400 i
outside the ill[
.c.
10
I
I
,, 20
I
I
1
i,x /'t'fu (LAB}
~)3
x1~,
~'///////Z
--3 E
'
t/~"
~
o
it ],
N
v
40
i corrected field
vane test
,,~.
~k ~ e
trati~
test
0
-
-
-
under the fill i
qTfu
v
//~./calibrated corrected
rl M" ~
I.. 0
~=
30
::. :...:.~ /Ill~l/,
~
,
5-
Fig. 4.4.
w
J:
u
o
30
Estimated undrained shear strength (kPa)
1
I
1
I
I
I
I
i
I
Undrained shear strengths beneath and outside the test fill at the Antoniny site.
S (0.45 9KD)"2~
(4.2)
(Y'v where ~'v = vertical effective stress S = normalized undrained shear strength for the normally consolidated state ( O C R - 1) K D- horizontal stress index An example of a comparison between calibrated corrected field vane shear strength values and undrained shear strength from dilatometer test evaluated based on Eqn. 4.2 for a soil stratum with peat and calcareous soil is given in Fig. 4.5. In this case the Sparameter was found to be 0.5 for the peat and 0.45 for the calcareous soil (see Chapter 4.5.2). The relevant range of undrained shear strength of organic soils for design of small structures or as preliminary low-cost investigations can often be estimated from Swedish fall-cone tests or from laboratory vane shear tests. However, these
Shear Strength used in Stability Analysis
143
Horizontal Materiel index stress index ID KD 0;4 0.8 0 4 8 00 '
I
'
9
I
,
'
!
'
1
'
i
I
i
'
Ditatometer modulus ED, MPa 0 0.4 0.8 1.2 '
I
v
!
'
9
J
Undrained shear strength Tfu, kPo 0 5 10 15 v
!
I
v
|
_
Q.
E
,s I
11
i
!
|
i
v
i
i
) (
E3
1
1
bU
o (..) 9
Fig. 4.5.
I
n
I
I
I
J
9
i
,
I
9
I
,
i
,
,
!
,
i
l
.~._c
1,
.CI .1O
I~
l
i
=
1
l
,
,
"6
,
i
l
l
1
I
Undrained shear strength profile obtained from dilatometer test at Antoninv site.
tests are obviously not applicable for highly fibrous soils. According to Landva et al. (1986) the application of these methods is doubtful when the fibre content amounts to 40 %. From the Swedish fall-cone test, not only the undrained shear strength can be roughly estimated but also the remoulded shear strength and the liquid limit. Because of this, the fall-cone test is used as a routine test, mainly in the Scandinavian countries but to an increasing extent around the world. The appropriate mode of shear at different locations in the soil beneath and outside an embankment can be simulated by triaxial compression, direct simple shear and triaxial extension tests for evaluation of undrained shear strength. In order to simulate the initial in situ stress conditions, the specimens should be consolidated under conditions of no lateral strain, that is, under K 0- conditions. The results obtained from shear tests are often presented as normalized values (i.e. divided by the pre-shear vertical consolidation stress Or'v) (Fig. 4.6). Because the designer of stage-constructed embankments must estimate not only the initial undrained shear strength but also how it will change throughout the period of construction, the laboratory shear tests should provide the data to evaluate the normalized undrained shear strength versus the overconsolidation ratio. In order to estimate correctly the initial undrained shear strength at stresses below the preconsolidation pressure, it is important that the preconsolidation stresses are not exceed-
Stability Analysis
144 a) 2"01 Test Sym S m 1.5 TC 0.325 0.78 DSS o 0.2550.78
:~ TE 1.0 6
v
I
I
I
L..I
I
/~-
/
Intact'
l'~
s Iml
IDSSI 0.29 10.69510:2S /
'
'
'
~ ~ -
-[TE ] 0.235l0,82 1".0,20 A ~I ~ ' / ~ -
-
~
0.4
i " 7 -r ~
'rfu=qf for TC end TE Tfu=Tma x for DSS
.
1 2 3 456 810 Overconsolidation ratio OCR= 6p / 6'v !
Fig. 4.6.
s ]
-ITc 10.t.5 10. ! 0.33sl
0.2ooo.8~,,,~/
0.6
"~ 0 2 l ~ ::3 ,
1Desth/
- ~ DSS
o
-
o~
]
Condition ,IOCRITClDSSlTE
lhs~l
o I~
: D,estructumdl , / A I 9I 9 1 1.5 2.0 2.5 3.0 3.5 4.0 OverconsolidQtion rQtio OCR= 6';::)/~v i
Relation between normalized undrained shear strength versus overconsolidation ratio relationship for marine clay. (a) SHANSEP CKoU tests. (b) CKoU tests on intact and destructured samples (Jamiolkowski et al., 1985).
ed and that the soil is allowed to keep its structure. Once the preconsolidation pressure is exceeded, the soil becomes "destructured", which in soft soils entails large compressions. In some materials, this also entails a breakdown of a strength component termed bonding or cementing of the structure. In the case of embankments on soft soils, loading involves exceeding the preconsolidation pressure with consequent large deformations, and the destructured strength thereby constitutes the relevant strength. 4.2.3
Effective
shear strength
The effective shear strength used in effective stress stability analysis is expressed by the Mohr-Coulomb effective shear strength parameters c" and @'. To simulate the appropriate mode of shear behaviour below an embankment, the effective shear strength parameters can be investigated by triaxial and/or direct simple shear tests. The strength parameters c' and @" are normally determined in totally drained triaxial tests, but may also be estimated from the effective stress paths in undrained triaxial tests. Undrained triaxial tests give the effective shear strength parameters at constant volume, which underestimate the shear strength somewhat in overconsoli-
Methods of Stability Analysis
145
dated state and overestimate the shear strength that can normally be used in the normally consolidated state. The parameters for constant volume are used in calculations of stability in the undrained case in terms of effective stress analyses. Effective shear strength parameters c" and ~" can also be evaluated from drained direct simple shears. Parameters c" and ~" obtained from direct simple shear tests are normally somewhat lower than the values from triaxial compression tests.
4.3
METHODS OF STABILITY ANALYSIS
4.3.1
Types and scope of analysis
In the preliminary design stage, the stability of an embankment may be estimated with the aid of simple rule of thumb or stability charts. The estimation of a safe embankment height during the first stage of construction can thus be made using e.g. Terzaghi's equation or Taylor's stability chart. Having established the pore pressure conditions, it is also possible to estimate the embankment stability at steady seepage using Cousins' stability charts (Cousins 1978). Preliminary assessment of embankment stability during rapid drawdown of water level, e.g. for dykes can be carried out using Morgenstem's stability charts (Morgenstem 1963). For the design of embankments under complex subsoil and pore pressure conditions the two-dimensional methods of slices with assumed plane strain conditions are often used (Fredlund and Krahn, 1977). In the limit equilibrium analysis, the average factor of safety F may be defined as follows: F = ~f / ~
(4.3)
where 1:f = available shear strength along a shear surface I: = equilibrium shear stress along the same shear surface In the methods of slices, the potential sliding mass is subdivided into a number of imaginary vertical slices with width b. Each slice is acted upon by its own weight W and by the boundary interslice forces which have both a tangential component T and normal component E (Fig. 4.7). The forces acting on the basis of a slice with inclination cz and length 1 are shear resistance S~ and normal force N, Fig 4.7. The methods of slices most commonly used for stability calculations are: the Swedish circle method; the simplified Bishop method; the Janbu generalized and simple methods; and the Morgenstem-Price method. The Swedish circle method is
Stability Analysis
146
~---
I
I
~---
.....
-1-1
\ \
"x
.
Fig. 4.7. Forces involved in the methods of slices.
the simplest method since no interslice forces are considered. Because the Newtonian force principles at the interslices are not satisfied, this method may in extreme cases give errors in the factor of safety of as much as 60 % (Whitman and Bailey, 1967). Of the other methods listed above, in which interslice forces are considered, the simplest and probably the internationally most commonly used is the simplified Bishop method. The factors of safety calculated by the above mentioned methods (with the exception of the Swedish circle method) normally differ only marginally for low embankments on soft soils (Wright et al., 1973). The simplified Bishop method may also give errors in the order of 10 % as not all interslice forces are accounted for. The errors are on the "safe" side, i.e. the calculated safety factor is too low. 4.3.2
Simple procedures
of s t a b i l i t y a s s e s s m e n t
(a) Preliminary estimation of safe embankment height: A simple equation commonly used to estimate the safe height for an embankment on homogeneous subsoil in undrained conditions is: H~ = (NdXfu)/(F-~(~)
(Terzaghi, 1956)
(4.4)
Methods of Stability Analysis
147
where H = height with acceptable factor of safety F N~ = stability number l:f~ = undrained shear strength of subsoil F = factor of safety y~ = unit weight of embankment. For a circular failure surface the stability number is N~ = 5.52, on the assumption of no internal strength of the embankment soil. The safe height of an embankment on soft subsoil can also be set, ~ from stability charts. One such chart, applicable in total stress analysis of the stability of an embankment on homogeneous subsoil, was developed by Taylor (1956). In this chart, Taylor's stability number N T is related to the depth factor D for given values of the slope angle 13and the parameter M which defines the location of the considered slip circle (Fig. 4.8). The safe height of an embankment can be calculated as: H = ~ f . / ( F . NT'Tc) 0.19 0.18
(4.5)
13=530
-.
0.17 T
~>
~ ~
~'
\\
/...._...~~.-~ ~
-
0.16
"I-- 0.15
z
d
~
0.14
\
/
J3
E c-
0.13
..~
0.12
\ \\0
\
u')
0.11 l i l i l i l i l l l i l i l i i l
0.10
0.09
1
2
3
i/
4
Depth factor, D Fig. 4.8. Taylor's stability chart for ~u = 0 conditions (Taylor, 1956).
3
Stability Analysis
148
Nc.Tfu 5.52.8.5
H s = ~
I I
:J
~
-
".'":'H.s'=?.''."
....
:
=18kNl'm
7. s
,,
MH,
s
F
a)
"
, OH s
~'e 9
=
~
1.3.18
11=18.4'; D = 2 . 5 ;
'*
=2m
M=1.6;
N.,.:o.,+', H s =Tfu/(F N T ~'e)=
EI G y t t j a , ~
~ T f u
=9 kPa
9. S a n d . 9. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .
..:;
= 8.5/(1.3- 0.167.18) = 2.2 m b}
13=18.4'
D=5
M=4
N T = 0.179 H s = 9/(1.3.0.179.18) = 2.1m
Fig. 4.9.
Example of estimation of the safe height for first stage of a dyke based on Taylor's chart.
The use of Taylor's chart is demonstrated by estimation of the safe height of a dike considered in Chapter 9 (Fig. 4.9). In order to find the approximate values of the parameters D and M in the first stage, the embankment height was calculated from Eqn. (4.4).
(b) Steady seepage conditions: Several types of stability charts designed to account for known pore pressure conditions and using effective stress analysis have been introduced. On the basis of the stability charts, an estimation of the factor of safety for known soil parameters, pore pressure conditions and geometry of embankment slope can be made, or a suitable slope inclination having a prescribed safety factor can be selected. To assess embankment stability under steady seepage conditions, Cousins' stability charts can be used (Cousins, 1978). Based on Cousins' stability charts, the safety factor F can be obtained from the following equation: F = N F [c'/(T 9H)]
(4.6)
Methods of Stability Analysis
149
where Nv = stability number; c" = effective cohesion intercept 3' = unit weight of soil (average) H = height of embankment. Cousins' stability charts are given in terms of stability number N F as a function of slope angle 13, depth factor D, dimensionless parameter ~ and pore pressure ratio r~. Dimensionless parameter ~,~ and pore pressure ratio r~ are defined as follows" 9H ) ]
=
(4.7)
and r. =
h)
(4.8)
where ~" = effective angle of internal friction u = estimated pore pressure (average) h = depth of the point below the soil surface for which u was estimated Fig. 4.10. shows stability charts considering critical toe circles and circles with specified depth factor D = 1; 1.25 and 1.5 and different values of pore pressure ratio r = 0, 0.25 and 0.5. (Cousins 1978). For the pore pressure ratio r > 0.5 the estimation of stability number N r can be made by using a new dimensionless parameter defined as follows: ~,'~ = (1 - r ) ~,~
(4.9)
In this way, the stability number N F can be estimated from stability chart for r = 0 by using parameter ~,'~ instead of~,~r The use of parameter g ' r together with the chart for r~ = 0 is recommended only for a slope angle less than 17.5 ~ An example of estimation of the factor of safety for a dyke designed in Chapter 9 under steady seepage conditions by using Cousins' charts is shown in Fig. 4.11. The pore pressure ratio r is calculated using the pore pressure at the lowest point of the slip surface and the average of the unit weights in the embakment material and the peat layer.
Stability Analysis
150
500
I
I
I
I
I
I
I
I
/
ru =0
u_ 300 z
,.5
0 .13
E :3
c
100
-
,,\
~
\Oi.4
1
50
oHIH
,
~
20
""
~
6
10_" " ' ? ~:cc.- - -...._._. ===__.=_~. 0
o,~ r-.,o,:,,,~,~or~~ ~.3oo I
"'~~
......
~.~_~1001|
~\
ru='o.~',' t '~176 3o0 =50
.,oo ,o
~~'~.~--~~ ~,o~ ,
, lo
, ~ - ~ - - - . ~ 2~ 3o
, ,..o
Slope angle, J3 (Degrees)
, o
, ',o
,
, 2~
, 30
,
, ,..o
SLope ongte, J3 (Degrees)
Fig. 4.10. Stability number N F considering specified depth factor D = 1; 1.25 and 1.5 when r u = O; 0.25 and 0.5 (Cousins, 1978).
Methods of Stability Analysis H=3m;D=2.0
151
~=12.5'.
ru>0.5
~
c'/(~' H)=0.042 Xc~) = 13.7
ru =0.6.
~.'cr162
I
f
.
.
NF =38
"
.
.
.
.
.
H=3m H
/ Gyttj(a 9
9S a n d .
.
I F = NF (c'/Ct~" Hi) = 38" 0.042 = 1.59 .
.
9. . . .
. ;...-
.
.
9
..
.
. . . .
i
.
9
.
. . - . .
,
,
.
.
9
.
.
".'.
E]
.
:
.
."
''
"
.
. ..
J-
9
. .
:...
Fig. 4.11. Example of estimation of the safety factor for a dyke under steady seepage conditions based on Cousins' charts.
(c) Sudden drawdown conditions: In cases with dykes or embankments where the fill material has a low permeability, it may be important to investigate the stability under sudden drawdown conditions. For preliminary design purposes, stability charts presented by Morgenstem (1963) can be used. These charts were elaborated for the special loading condition of sudden drawdown from full submergence with the assumption that no dissipation of pore pressure occurs. The charts are given in Fig. 4.12. The values of the safety factor have been plotted against drawdown ratio L/H for ~ '-values ranging from 200 to 40 ~ with values of c'/(y ~ H) equal to 0.0125, 0.025 and 0.05, and slope inclinations from 1:2 to 1:4.
4.3.3
Swedish circle method
The Swedish circle method is the simplest method of slices and is based on the assumption that the interslice forces can be neglected. This assumption causes an error in the estimated safety factor but, due to simplicity in calculation, the method is often used for approximate calculations. The safety factor F can be calculated using the following equations:
Stability Analysis
152
3.5| --
u_
~
/I-
'~
,
,
~
1:2
3~
.
.
.
.
O owOown
~
c'/( ~'e H) = 0.0125
.._
c./(~,eH}=O.025
"-
~'(lr~ H'= 0.05
o
prior to drawdown
.
,
- --
2.s ~. \ \ Level
.... i "'
y
-'-
2
,?
,
1.5:
-
.
"-~-
~ - ~ ~
-
-
0.5O-
J
0
I
0.2
I
I
I
0.4
I
i,.
0.6
Drawdown 7
T
-
LL
~6-
T
I
I
I
I
1
I
# ....
71-
I
LL
1"3 ,
l
|
!
I
I
I
0.8
I
1.0
ratio. L I H !
I'
'1
1"4
>~ _
"-"O
5-
"6
i.. O ..i-.
u 4 El
LL
_\
LL
3
00
i
~
'
L
0.2
I
I
oz,
i
I
06
Drawdown
I
i
0.8
1.
ratio. L/H
1.o
00
" "~ ,,, ,~
I
0.2
.
I
i
0,4
~'~.
I
"
-9- - - - ~ . . _ . . . _ .
I
I
1
,1
0.6 0.8 1.0 Drawdown ratio, LIH
Fig. 4.12. Drawdown stability chart for specified slope inclinations 1:2, 1:3 and 1:4. (1) ~'=20~ (2) ~'=30~ and (3) ~'=40 ~ (Morgenstern, 1963).
Methods of Stability Analysis
153
9 for the total stress analysis E,f~ 1
(4.10)
F
Z W sm~ 9
for the effective stress analysis E(W coscz - u 1) t a n r
+ E c' 1 (4.11)
F _~
E W simz where zf~ = undrained shear strength for cohesive soils u = presumed pore pressure at failure t~" = effective angle of internal friction c" = effective cohesion intercept. For the total stress analysis of the stability of embankments on soft soils, the normal procedure is that the material in the granular fill or granular layers in the subsoil is assumed to be drained and the soft soils undrained (Fig. 4.13).
~ ~176
~
e
.G,I
i|174c : Q " ......." ~W
"
"
c o s oC
Fig. 4.13. Stresses along the failure surface for granular and cohesive soils, which should be considered in total stress analysis.
In such a case, the available shear strength along a failure surface consists not only of undrained shear strength in the soft soils, but also of effective shear strength in the granular soils. When large deformations are expected, the friction angle as-
Stability Analysis
154
sumed for the granular fill material should be related to the critical state, i.e. shear strength at constant volume. In rough estimations of stability for low embankments, such as when using Terzaghi's equation (4.4), the shear strength in the fill material is neglected, as it is in Eqn. (4.10).
4.3.4
Simplified Bishop method
In the simplified Bishop method (Bishop, 1955) only the normal component E of the interslice forces is taken into account, and the variation in the tangential component T is ignored (Fig. 4.7). The resulting equation, from moment equilibrium about the centre of rotation, for the safety factor F is: 9 for the total stress analysis
F=
(4.12) X W sinc~
9 for the effective stress analysis Z[(W- u b) tan~" + c'bl/m~ F=
(4.13) Z W sinc~
where m~ = coscz (1 + tanc~ tank'/F) Since F appears on the left as well as the fight hand side in the equation, calculations start with an assumed value of F and a new value of F is calculated. With this new value, the calculations are repeated and such iterations proceed until there is little difference in successive values of F. An example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site is shown in Fig. 4.14 and in Table 4.1. In this example, the organic subsoil was divided into three zones with different magnitudes of undrained shear strength. The soils in these zones were divided into layers with characteristic undrained shear strengths. In the calculations, the effective shear strength in the fill material was also taken into account. The calculations were made for a third stage of an embankment when the soil below was partially consolidated for the previous stages. The undrained shear strength has therefore become unevenly distributed across the embankment and the centre of the critical circle is not necessarily located above the midpoint of the slope.
Methods of Stability Analysis
155
t
I
R=20m
Zone
,
r
C
S6nd.
~ '
~ "
"
Gyttja
Peat
A
16
12
B
13
10
C
6
8
" "..'.
" " ::
I
Gyttja
(kPa)
fu
~ ""
. "..
" . i
"'.
. .-'.
Fig. 4.14. Example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site based, on the simplified Bishop methods.
Table 4.1.
Example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site.
Slice W No.
r
122 370 202 448 415 181 316 249 84 56
I
Wtan~" m a
Assumed F = 1.25 WtanO'/mot m a WtanO'/mot or
(o) 1 2 3 4 5 6 7 8 9 10
Wsin(x Xfu
48.6 36.9 26.4 17.5 5.74 -2.87 -11.5 -23.6 -33.3 -40.9
~ 91.5 222.2 89.8 134.7 41.5 -9.1 -63.0 -99.7 -46.1 -36.7 E 325.1
(1) F = 393.5/325.1 = 1.21
(kPa) (m)
(k~
14.8 10 10 10 10 8 8 8 6
85.4 -
5.1 5.4 1.8 4.2 4.1 2.0 4.1 4.3 2.4 4.5
or
xf. ~) 1.08
(2) F = 392.7/325.1 = 1.21
79.1 80 18 42 41 20 32.8 34.4 19.2 27 E 393.5
Xfu~ 1.09
78.3 80 18 42 41 20 32.8 34.4 19.2 27 E 392.7
Stabili .ty Analysis
156
4.3.5
Janbu generalized method
In the Janbu generalized method (Janbu, 1973), which can be used for slip surfaces of arbitrary shape, both components of interslice forces are taken into account by making an assumption that the points at which the interslice forces act can be defined from a "line of thrust" (Fig. 4.15). The position ofthe line of thrust is defined by vertical distances from the base of the slice to this line i.e. t c on the left side and t R on the right side of the slice, and angle c~t between the line of thrust on the right side of the slice and the horizontal.
w
E
-.
,,
line
of
thrust
b/2 tan o( l/tl_/ / / / / /
/
/
Fig. 4.15. Forces acting on each slice in the Janbu generalized method (Janbu, 1973).
The normal force N on the base of the slice is derived from the SUlnmation of vertical forces: 9 for the total stress analysis N = [W - (T R - T L ) - -CfuI sin~/F]/cosc~ (for cohesive soils)
(4.14)
Methods of Stability Analysis
157
9 for the effective stress analysis N = [W - (T R - T L ) - c ' l smt:x/F + u 1 tan~' sinct/F]/moc
(4.15)
The safety factor equation based on force equilibrium derived by summing forces in the horizontal direction is" 9 for the total stress analysis
X%b F=
(4.16) Z N sint~
9 for the effective stress analysis Z [(N - u 1) tan~" + c" 1] cosct F=
(4.17) E N sin~
Forces, directions and their location are evaluated by iterative procedure.
4.3.6
Janbu simple routine procedure
The Janbu simple routine procedure (Janbu et al., 1956) assumes that in Eqn. (4.14) and (4.15), where the normal force N is defined, the interslice tangential forces can be ignored. A correction factor fo accounts for the effect of the interslice tangential forces and is used to correct the initial safety factor F o. The correction factor is related to cohesion, the angle of internal friction and the shape of the failure surface cl~ (Fig. 4.16). An initial safety factor is calculated on the basis of Eqn. (4.16) for total stress analysis or Eqn. (4.17) for effective stress analysis using a reduced form of Eqn. (4.14) or (4.15). The corrected safety factor is obtained as: F = f o Fo
(4.18)
15 8
Stability Analysis 1.2
i
1
i
1
i
I
i
L
O ~--
_
.s 1.1 _ s u r f a c e
r = 0
U 0
LL
-
_
-
5......-- c=0 1.0
0
I 0.1
~ Ratio
I 0.2
I
I 0.3
O.Z,
d/L
Fig. 4.16. Correction factor f0 as function of curvature ratio d/L and type of soil (Janbu
4.3.7
et al., 1 9 5 6 ) .
Morgenstern-Price
method
The Morgenstem-Price method (Morgenstern and Price, 1965) which is the general approach of limit equilibrium analysis assumes that there is an arbitrary mathelnatical function to describe the direction of the resultant of the interslice tangential component T and nomlal component E of the following form: T/E = )v f(x)
(4.19)
where f(x) = a function that describes the rammer in which T/E varies across the slope X = a constant representing the portion of the function f(x) used when solving for the safety factor The equations for the safety factor are obtained using the summation of forces tangential and normal to the base of a slice and the SUlmnation of moments about the centre of the base of each slice. The equation for the safety factor at force equilibrium is the same as Eqn. (4.16) in the Janbu generalized method for total stress analysis and as Eqn. (4.17) for effective stress analysis. The application of the Morgenstem-Price method unavoidably involves the use of an appropriate numerical programme which, as well as more details concerning the Morgenstem-Price method, is out of scope of this book.
Methods of Stability Analysis 4.3.8
159
N o n - c i r c u l a r slip s u r f a c e
In soils containing weak layers of limited thickness or for embankments with pressure berms, the safety factor for the case of non-circular failure surfaces has to be considered. Both the Janbu method and the Morgenstem-Price method can be used for failure surfaces of arbitrary shapes. Fredlund et al., (1981) presented another method of slices commonly used for non-circular composite slip surfaces. The composite slip surface starts and ends with a circular portion and has a central linear portion (Fig. 4.17).
i R/
\\
Fig. 4.17. Forces acting in the method of slices with composite slip surface.
In such a case, the equation of the two-dimensional safety factor for moment equilibrium in the methods of slices presented above becomes: 9 for the total stress analysis Z%
1R
F-
(4.20) ZWx-ZNf
9 for the effective stress analysis ~ [ ( y - u l) tan~" + c'l] R V=
(4.21)
ZWx-ZNf
Stability Analysis
160
where R = radius or the moment arm associated with mobilized shear forces Sm x = horizontal distance from the slice to the centre of rotation f = perpendicular offset of the normal force from the centre of rotation In the case when the critical failure surface has a long horizontal portion, the wedge method may be used. The shape of the wedge failure surface can be estimated as in Fig. 4.18.
9
or
9
9
9
9
9
9
9
9
I
9
S a n d
~
I .
9
~
o
"
o
.
.
.
9
.
~
.
.
.
.
.
~
9
o
.
"9
~
9
9
~
9
"-
" "
~
I
oq>
9
-
"
'
"
.
."
.
"
.
I
.
"
9
"
"
~
,
,
9
9
"
. . . .
"i
Fig. 4.18. Failure surface in wedge method of stability analysis.
4.4
STABILITY
OF SINGLE-STAGE
EMBANKMENT
When the initial shear strength of the soil is sufficient to ensure the required stability for the maximum embankment load, the embankment can be raised to the total height in one stage. In most cases with organic soils, only low embankments can be constructed in a single stage because of the low initial undrained shear strengths. The permeability of organic soils becomes relatively low regardless of whether the initial value is high or not as it decreases significantly at load application and subsequent compression. If the permeability is low with reference to the rate of loading, the stability analysis of an embankment during construction should be carried out as an undrained analysis. At loading of overconsolidated soils, some partial drainage may occur, whereby the effective stresses increase. When the vertical stress reaches the preconsolidation pressure, no further increase in effective stress will take place from a practical point of view. Additional loading brings about increases in pore pressure equal to the increase in total vertical stress and in those zones of the subsoil where the principal stresses rotate, the increase in pore pressure may be even higher.
Stability of Single-Stage Embankment
161
A drained analysis assumes that all excess pore pressure has dissipated. An effective stress analysis using measured or predicted pore pressures often does not account for the further generation of pore pressure at loading up to failure (Fig. 4.19). Therefore, modified total stress analysis so-called Undrained Strength Analysis (Ladd 1991) using undrained shear strengths should be used. This entails estimation of the in situ undrained shear strength 1:fu. Methods commonly used for determination of undrained shear strength, summarized in Chapter 4.2, are presented in Chapters 2 and 3. The estimation of undrained shear strength on the basis of empirical relations is discussed in Chapter 4.5.2.
~ d
: c'+ (G~p*AG) tan r
. . . . . _._. ~ . . ~ . '
.Tff = c', 6"vptan r
-
/
b., d ul
Effective Stress Anatysis
/
/
. .~ . . .
~,~..._~
~),
'
/ /Drained
f-L. rL/3
Undrained Strength Anatysis
..__
Effective norma[ stress, '6' Fig. 4.19. Assumed shear strength at failure r in Drained Analysis, Effective Stress Analysis and Undrained Strength Analysis (Ladd 1985b).
For preliminary studies, rapid stability estimations for embankments during construction can be made using Terzaghi's equation or Taylor's stability chart. For final design, an appropriate method of slices should be applied using the undrained shear strength m layers with organic soils or other soft low-permeable soils. In undrained strength analyses, the required safet7 factor during the construction period is typically of the order of 1.3-1.5, although values as low as 1.2 are sometimes allowed when the site conditions and undrained shear strength are very well established. Much higher values, in the order of 2 or more, are employed if significant shear deformations must be avoided. In design of an embankment, also other loading conditions with different acting forces and drainage conditions than the end-of-construction conditions may have to
Stability Analysis
162
be checked. For constructions such as flood control levees, pond dykes and waterretention dams, additional stability analyses for steady state seepage conditions and conditions of sudden drawdown of the water level should be carried out. For preliminary studies, stability charts presented in Chapter 4.3.2 can be used. For final design, effective strength analyses with computed pore pressures should be made. In analysing stability of road embankments, additional loads arising from road construction and traffic should be considered. The influence of additional loads on embankment stability is significant for low embankments on very soft soils.
4.5
STABILITY OF STAGE-CONSTRUCTED EMBANKMENTS
4.5.1
Types and scope of analysis
When the stability analysis for the end-of-construction stage indicates that the initial shear strength of the soil yields an inadequate safety factor, construction may be divided into two or more stages. The estimation of the maximum load for the first stage is then carried out as for a single-stage embankment, but the stability analysis for the following stages of construction requires prediction of shear strength increase developed with time due to partial consolidation. For most cases of stage-constructed embankments, the most critical stability condition occurs during actual construction. All stability calculations for stage-constructed embankments during construction should be carried out as undrained analyses. However, as for single-stage embankments, the stability for conditions after construction should also be ensured. Different levels of sophistication can be employed in estimating and using the increase in undrained shear strength in design of stage-constructed embankments. Firstly, they depend on the method of stability analysis and secondly on the method used to estimate the initial shear strength and the increase in shear strength during consolidation (Ladd, 1991). The methods for estimation of the increase in shear strength can be divided into two groups of sophistication and expense : 9 empirical relations, and 9 laboratory testing to obtain complete strength-stress data.
Stability of Stage-Constructed Embankments
4.5.2
163
Evaluation of the increase in undrained shear strength based on empirical relations
Past experience, summarized by Larsson (1980) and Ladd (1985b), shows that for soft normally consolidated mineral soils the following empirical relation can be used to predict shear strength increase: 1:f. = K "o'"v
(4.22)
where zf, = undrained shear strength K = coefficient of shear strength increase (depending on the loading case) (r"V = vertical effective stress. For overconsolidated soils, the coefficient of shear strength increase (Ladd, 1985b) can be expressed as: K = S (OCR) m
(4.23)
where S = zf,/cr' - normalized undrained shear strength for the normally consolidated state (OCR= 1) OCR = C'p/6'v - overconsolidation ratio cr p = preconsolidation pressure m = slope of l:f~/6"v - OCR relationship on log-log scale. Examples of values of coefficients S and m for soft mineral soils presented by Jamiolkowski et al. (1985) are shown in Fig. 4.6. The coefficient of shear strength increase for normally consolidated soft mineral soils varies with the plasticity, but can be assumed as constant for a given soil (Larsson, 1980; Ladd, 1985a) (Fig. 4.20): K = constant = S
(4.24)
For normally consolidated mineral soils, the undrained shear strength can then be calculated as: 1:f~ = S - ~ p
(4.25)
Stability Analysis
164 .4 . . . . &&
~- 0.3
_
lX
IX
~
I,.
Ix
TC &
IX
-
&
&
IX
O
oiX
IC
'-- 0.2
&
0
0
0
v
DSS
-~.... ~
_ _ _
_ - -
~
- - -
~--o
v
~ g~v~--" "~ V
.N _
,~ Triaxial compression (TC). o Direct simple shear (DSS). v Triaxial extension (TE).
0.1
E L
O Z
0C
I
10
.-
l
20
I
30
1
1,0
,1
50
I
60
Tfu =qf 'f'fu ='trh(max) Tfu = qf 1
70
I
80
90
Plasticity index, Ip (%) Fig. 4.20. Normalized undrained shear strength versus plasticity index (Ladd, 1985a).
Similar relations have been found for organic soils (Larsson 1990), and the normalized shear strength then varies with the organic content, (Fig. 4.21). Experience from organic soils (Bergdahl et al., 1987) indicates that the normalized undrained shear strength changes not only in the overconsolidated states but also in the normally consolidated state. For a better description of the change in undrained shear strength with stress, especially in the normally consolidated state, the concept of the effective stress level ESL (Lechowicz, 1986) was introduced: ESL = (O"p)o/(Y'v
(4.26)
where (6'p)o = initial preconsolidation pressure. The increase in undrained shear strength with effective stress level ESL can be estimated according to Eqn. (4.22) for which the coefficient of shear strength increase is obtained from the following relations"
165
Stability of Stage-Constructed Embankments
3 0. 1
~,,-.
/
e/ I
c O./.,
o
/
I,,. ,.m,-,,
~."
K.,,
u~ 0.3
~f s~
70
/
&,-
"0 C
/ ,or
l
l
9Triaxia[ compression o Triaxial extension o Direct simple shear
Mineral soil
"o 0.1
o._Eganic - Mineral soil
N
Z
/
O
.c_ a 0.2-
L.
0
l
0
JE
E o 0.0
9
/
....
/__
-T
7-
mineral-organic soil
organic soil
l-
20
8 Organic
3b
content,
(%)
~o>0.5 ;9
t..,'+-
.E
,-- O.Z, C I...
~O
4
0.3
/
I I
0.2. OJ
-o 0.1
Ba _,,1o
//
~ ~ " ~
0
"~ " -
m
m
0
o
O/
/
/
l
/
//
l
/
/
115~, /
/
/
/
/
0
Empirical relations for inorganic clays (Larsson 1980)
N
o~
O
E o00 z
0 ,..,..~.
t /
JE cu~ O K_ "10 c
0
I
0
2Go
Liquid limit, (%)
9Triaxia[ compression o Triaxia[ extension o Direct simple shear
3oo
Fig. 4.21. Normalized undrained shear strength in organic soils (a) as a function of organic content (b) as a function of liquid limit. (Larsson 1990).
Stability Analysis
166 K = S (ESL) m~162
ESL> 1
K = S (ESL) ~r
ESL_
(4.27)
or
where S = normalized undrained shear strength in the normally consolidated state at ESL=I moo= slope of the relation between log(a:fu/or'v) and log (ESL) in the overconsolidated state (ESL> 1) mno = slope of the relation between log (1:f./c~"v ) and log (ESL) in the normally consolidated state (ESL
Stability of Stage-Consmucted Embankmenm ,
,
,
,
,
,
,
,
167
,
0.50
,.2
% o.~o
_,~ o.8
~? 0.30
0.6
~
~176~
0
-~
0 . 5 I 0 I~ J0 "I 1 I 5 , L I
%
%
9
t:o.i'l
I
I
I:
0.50-
j
I
,
-
,
,
,
,
,
mnc = 0.20 .
.
.
,
~
~
-
.
0.40
.
.
o ~,
.
,
2,
.
,
;
,
o.
~,
,
Effective
Fig. 4.22.
Values
~ stress
of K s as functions
level
o
0,15
/
o..,o
, ESL :
, 0.2
,
, o.~
,
: o~
,
o:8,
(61~)o 16' v
of ESL for different
S, moc
and
mnc
~.o
Stability Analysis
168 7 5-
4-
2
Test
I
2 - cgVT
.c "o
1-
Peat
.~_ 0.8- DSS o
,3 O.6 c ~
Calc. soil
-
I
I I
I
I
!
I
I
I
I
ESL..
m~
mnc
I
0.30
0.43 0.83 0.23
3
0.47 0.80 0.16 9 0.43 0.75 0.14
5
9 0.38
0.11
6
9 0.36
0.10
3 _~
0.4
1
--5
02 t
Z
I
0.4g 0.84
3a
Peat CKoUTC' C.,aic6it 4
I
S
IQ Peat Cats~C.i 1 2o
s
I
ESL > I
Sym
Soil
~c~ 3u~
I
I
~'-
0.1
I
0.3
0.2
1
I
i
0.4 0.50.6
I
I
I
I
I
0.8 1
2
1
3
4
5
J
I
6 7 8
10
Effective stress level ESL = (6;]o/6"v Fig. 4.23. Normalized undrained shear strength versus effective stress level from in situ and laboratory tests (Bergdahl et al., 1987). Undrained
Vertical effective stress. (kPa) 0
0
\
'
=-
'
40 '
'
/
~
.
0
60 ' I
'
, .,~vo I ~ I/ / 6"v before ' I , ' 1(6P )o I / stage 2
I
-Q_
20 '
~
.
shear strength,
Tfu (kPa) 10 2O
/
-..-.-I
~~,. ~'t
31
~ ~ o
.~_
56-~
7~
,r i
I
- -
',
I I -
-/- 4 --
-
-
a
..
-?--4
k]
Fig. 4.24. Calculated effective stresses and predicted shear strengths at different stages of a test fill. Assumed values of coefficients; Peat: S=0.5, moc=0.85, m,c=0.2; Calcareous soil: (1) S=0.45, moc=0.85, mnc =0.2; (2)S=0.45, moc=0.8, mnc=0.2.
Stability of Stage-Constructed Embankments
169
"I'fu = 0.45 9if;
O
3 II
1_o :3
2
0.2 1"fu = 0.45.6"v (ESL)
1
'Tfu = 0.45.6'v (OCR) 0.8 I
o o12 0:5
;, I
,
i
~
i
o:s
3
2,
o13
ESL= (6"~))o16'{,
i
Fig. 4.25. Predicted undrained shear strength versus effective vertical stress (Wolski et al., 1988).
Finally, it should be mentioned that a more global description of currem stress state includes both vertical and horizontal effective stresses (Wroth and Houlsby, 1985; Becker et al., 1987; Lechowicz 1994). Therefore, it would be better to define the increase in shear strength as a yield surface changing in both shape and orientation in stress space (Runesson, 1978; Larsson and Sfillfors, 1981). In this way, a more accurate description of the stress state and the undrained shear strength would be obtained. To estimate undrained shear strengths to be used in the stability analysis, the subsoil is divided into several zones. The division of the subsoil into zones should model the layering of different types of soil, the distribution of effective stresses and the overconsolidation ratio. Such a system of zones may be further elaborated by allowing for the effects of principal stress rotation on the distribution of undrained shear strength.
4.5.3
Evaluation of the increase in undrained shear strength based on laboratory testing
The undrained strength-deformation behaviour of soils below and outside the loaded area during stage-construction can be simulated by using three types of lab-
Stability Analysis
170
oratory tests. Undrained shear strength for different areas beneath an embankment can be estimated from: 9 triaxial or plane-strain compression shear tests for the area directly beneath the embankement; 9 simple shear tests for the area where the shear surface is nearly horizontal usually below the slope of the embankment; 9 triaxial or plane-strain extension shear tests for the area outside the toe of the embankement. The tests require detailed knowledge of the initial stress history of the soil and when they are to be used for prediction of future increases in shear strength, an accurate prediction of the future stress history is also essential so that real soil conditions can be simulated. Based on the knowledge of existing and expected stresses in soil, shear strength tests should be carried out on soil samples with prevailing stresses and with expected future stress levels. In this way, the normalized stress-strength relations for current and future stress levels are obtained and undrained shear strength parameters S and m may be evaluated. A simple approach for estimating the shear strength increase m soft subsoils beneath embankments was proposed by Aas (1976) and Larsson et al. (1984). The failure surface is divided into three different segments. The increase in shear strength for each of them is evaluated in a different way as shown in Fig. 4.26. For the steep part of the slip surface beneath the embankment, the shear strength increase is estimated from plane-strain compression or triaxial compression tests and for the next flat part beneath the embankment out to the mid-point of the slope the increase is estimated from DS S tests.
J
T "
A'rfu =0 ~
"
!
9
~176176 ". ~ ~
~ o
~ ~
~ " ,
i ~
!
6v: (Gv)o*"6v
~6p
.6;= 6v C%1o
BV
Fig. 4.26. Simplified estimation of increase in shear strength due to consolidation (Larsson et al., 1984).
Other Approaches in Stability Analysis
171
Most geotechnical laboratories do not possess plane strain devices, but triaxial tests are routinely used. The simplest approach to the estimation of initial strength and the prediction of shear strength increase for evaluation of stability is to use the data from CKoU direct simple shear tests for the whole failure surface.
4.6
OTHER APPROACHES IN STABILITY ANALYSIS
4.6.1
Three-dimensional analysis
The methods of stability analysis presented above are based on the assumption of two-dimensional (plane strain) geometry. In most practical problems, the length of the embankment is limited and, within this length, the geometry, load and shear strength are also often not uniform. For a more economic and accurate design, the stability of embankments with complex geometries and variable soil and pore pressure conditions may be estimated with three-dimensional analyses (Azzouz et al., 1981). In such cases, instead of the infinite cylinders assumed in two-dimensional circular arc analysis, the stability analysis may consider failure surfaces consisting of cylinders with ends of different shapes (Baligh and Azzouz, 1975) (Fig. 4.27).
al
b)
c)
Fig. 4.27. Cylindrical failure surface with different ends" (a) plane end surface (b) conical end surface (c) ellipsoidal end surface (Baligh and Azzouz, 1975).
The simplest way to account for the end effects is to assume a circular slip surface with plane ends (Gens et al., 1988). To calculate the three-dimensional safety factor, the resisting moments from each end plane are then added to those from the cylindrical surface (Fig. 4.28). In this case, the three-dimensional factor of safety F3D can be expressed in terms of the two-dimensional factor F2D (Gens et al., 1988) as:
Stability Analysis
172
F3D = F2D (1 +
(4.28)
0) Ls
where Ro
L= 0 ME AQC
= = = = =
radius of the cylindrical part of the failure surface overall length of the slide 2 M E/(AQC 9Ro2 ) resisting moment of area of each end plane failure surface.
b)
o)
'/ /;J
0
d
H
Fig. 4.28. Geometry of three-dimensional cylindrical slide with plane ends. (a) oblique view (b) cross-section (Gens et al., 1988).
The assumption of plane ends generally overestimates the end effect because the most critical case involves curved ends. Gens et al. (1988) have analyzed the end effects using a large variation of the shape of the end surfaces. Minimum safety factors were obtained using a family of power curves to generate the end shapes. According to Gens et al. (1988) the errors involved in neglecting end effects by treating slope stability problems two-dimensionally can be as high as 30 %. When the considered length is small, the three-dimensional safety factor is always significantly greater than the two-dimensional factor. An extension of the simplified Bishop method of slices to three-dimensional analysis was presented by Hungr (1987). Thus the existing computer programs for twodimensional analysis can be easily modified for three-dimensional analysis. In this approach, the sliding body is divided into a series of vertical columns of a rectangular cross-section. The base shape of the individual column results from the compound shape of the sliding body consisting of cylindrical and semi-ellipsoidal parts (Fig. 4.29).
Other Approaches in Stability Analysis
173
z'
z[
c)
d) Axis of rotation 9- - - . I - T - , . ; I i:l =
I I
I
I ; I I ', I
--r-----T
J I I
Axis of rotation
9- - r -
~ , I ,/ I ."
a I I, ' ] I ! I./Crest ~-
H-
Crest
Ic
E V) ,...
o
~X
~"
/
1
Toe
E: o Q.
x
y
Fig. 4.29. Failure surface for three-dimensional simplified Bishop method.(a) one half of the sliding body (b) forces acting on a single c o l u m n (c) vertical cross-section of the sliding body in the x-z plane (d) and in the y-z plane (Hungr, 1987).
The resulting equation from moment equilibrium for the safety factor F3D is: 9 for the total stress analysis Zzf~ A F3D =
*
X,W~ sin%
(4.29)
for the effective stress analysis Z[(W o - u b A cOSyz) tanr + c" A cOSyz]/M= (4.30)
F3D =
x w o sin%
17 4
Stability Analysis
where W ~ = total weight of the column A
=Ax Ay (1 - sin20~x sin2C~y)~
coSC~y) - true area of the column base
cOS~z = [ 1/(tan2C~y + tan2% + 1)]~ Mu
=cOSyz [1 + (sin~y tan@')/(F cOSyz)]
ub
= pore pressure acting in the centre of the column base
An example of the use of the simplified Bishop method modified for three-dimensional analysis to estimate the safety factor of test fill during the failure test is shown in Fig. 4.30. The fill constructed in the failure test had approximately the shape of a truncated pyramid. Half of the sliding body of the fill was divided into cylindrical part 1o wide and a semi-ellipsoidal part 1 wide. Both parts were divided into a series of colunms arranged in rows of uniform width. The organic soils were divided into zones with different magnitudes of undrained shear strength created by previous load stages. The soils in these zones were also divided into layers with specified undrained shear strength. In stability calculations, the shear strength in the fill material was also taken into account. The three-dimensional calculations of stability for the final stage in the failure test indicate that the factor of safety was about 1. The two-dimensional calculations carried out for the cross- section of the cylindrical part yielded a safety factor of about 0.75. F3D =1.04 Zone of subsidence \ ",,
~ ~ r
F
o~k ,~,"1,I, I./{.
'",N
IJ
CyLindricaL
I \part
Fig. 4.30.
Y
Z~B-B
iA
.....
I F
!i/,~ 7-" Crack ari~. due to failure 1X
F20 =0.TS
Zone of heave
I/c-I1! c.I
~4 g: ~
BIX
zl A-A
Axis of l/rotation
_J._\Semi-ellipsoidal - - part
u
Z C-C
9 9
.
9 9
~
~
Y
~ v
Example of estimation of the safety factor for test fill in the failure test based on the three-dimensional simplified Bishop method.
175
Other Approaches in Stability Analysis
4.6.2
Stability assessment by finite element analysis
Stability evaluations based on the limit equilibrium methods have an artificial separation from the stress distribution-consolidation analysis. The latter is used to predict subsequent changes in shear strength at certain stages during construction, after which the equilibrium analysis is used to calculate the safety factor at that stage. Numerical analyses such as the finite element method potentially offer a possibility to evaluate embankment stability as a coupled continuous effect of stressstrain-strength-time analysis. Information on the state of stresses obtained from the finite element analysis is then used to evaluate the stability conditions. Because the normal effective stresses 6"N and shear stresses ~N in each element are calculated, a local safety factor may be calculated as the ratio of the local shear strength to the local shear stress (Chen and Chameau, 1982): C"-F(Y"N tan(~"
FN =
(4.31)
The mean safe~y factor for the assumed failure surface may be obtained as the ratio of the total shear strength to the total shear stress as follows: +
tan ') a h
F=
(4.32) Z"[ N dA
where Z = summation over the whole failure surface dA = bottom area of a vertical column. There are other ways, based on the finite element analysis, in which the stability conditions may be estimated (Chowdhury, 1978). One of them is the evaluation of stability conditions on the basis of propagation of plastic zones (Lo and Lee, 1973; Lacasse et al., 1977; Almeida and Ramalho-Ortigao, 1982; Teunissen et al., 1986). Stability conditions have also been evaluated on the basis of the development of effective stress paths related to yield and failure surfaces (Tavenas and Leroueil, 1977; Tavenas et al., 1978; Folkes and Crooks, 1985; Crooks 1987; Jardme and Hight, 1987).
Stability Analysis
176
4.7
REFERENCES
Aas, G. (1976). Totalsp~inningsanalyser, prinsipp, gmnnlag. Norske Sivilingeniorers Forening. Kurs i Jordartsegenskaper- bestemmelse og anvendelse. Gol 2022/5. Aas, G., Lacasse, S., Lunne, T. and H6eg, K. (1986). Use of in situ tests for foundation design on clay. Proc. of In Situ '86, a Speciality Conference on Use of In Situ Tests in Geotechnical Engineering, Blacksburg, Virginia. ASCE Geotechnical Special Publication 6, New York, pp. 1-30. Almeida, M.S.S. and Ramalho-Ortigao, J.A. (1982). Performance and finite element analyses of a trial embankment on soft clay. Proc. International Symposium on Numerical Models in Geomechanics, Ziirich, pp. 548-558, A.A. Balkema. Azzouz, A.S., Baligh, M.M. and Ladd, C.C. (1981). Three-dimensional stability analyses of four embankment failures. Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 3, pp. 343-346. Azzouz, A.S., Baligh, M.M. and Ladd, C.C. (1983). Corrected field vane strength for embankment design. Journal of the Geotechnical Engineering Division. ASCE, Vol. 109, No. GT5, pp. 730-734. Baecher, G.B. and Ladd, C.C. (1985). Reliability analysis of the stability of embankments of soft clays. Proc. for MIT Special Summer Course 1.60S, Recent Developments in Measurement and Modeling of Clay Behavior for Foundation Design, Lecture 1. Baligh, M.M. and Azzouz, A.S. (1975). End effects on stability of cohesive slopes. Journal of the Geotechnical Engineering Division, ASCE, Vol. 101, No. GT11, pp. 1105-1117.
Becker, D.E., Crooks, J.H A. and Been, K. (1987). Interpretation ofthe field vane test in terms of in situ and yield stresses. International Symposium on Laboratory and Field Vane Shear Strength Testing, Tampa, Florida, ASTM STP 1014.
Bergdahl, J., Hartlen, J., Larsson, R., Lechowicz, Z., Szymanski, A. and Wolski W. (1987). Shear strength increase in organic soils due to embankment loading. 8th Krajowa Konferencja Mechaniki Gruntow i Fundamentowania, Wroclaw, Vol. 1, pp. 21-32. Bishop, A.W. (1955). The use of the slip circle in the stability analysis of slopes. Geotechnique, Vol. 5, No. 1, pp. 7-17. Bjerrum, L. (1972). Embankments on soft ground. Proc. ASCE Specialty Conference on Performance of Earth and Earth Supported Structures, Purdue University, Lafayette, Indiana, Vol. 2, pp. 1-54.
References
177
Chen, R.H. and Chameau, J.L. (1982). Three-dimensional slope stability analysis. Proc. 4th International Conference on Numerical Methods in Geomechanics, Edmonton, Canada, Vol. 2, pp. 671-677. Chowdhury, R.N. (1978). Slope analysis. Developments in Geoteclmical Engineering, 22, Elsevier, Amsterdam-Oxford-New York. Cousins, B. F. (1978). Stability charts for simple earth slopes. Journal of the Geotechnical Engineering Division, ASCE, Vol. 104, No. GT2, pp. 267-279. Crooks, J.H. (1987). Some observations on the stability of structures founded on soft clays. Proc. International Symposium on Prediction and Performance in Geotechnical Engineering, Calgary, pp. 27-38. A.A. Balkema. Folkes, D.J. and Crooks, J.H.A. (1985). Effective stress paths and yielding in soft clays below embankments. Canadian Geotechnical Journal, Vol. 22, No. 3,pp. 357- 374. Fredlund, D.G. and Krahn, J. (1977). Comparison of slope stability method of analysis. Canadian Geotechnical Journal, Vol. 14, No. 4, pp. 429-439. Fredlund, D.G., Krahn, J. Pufahl, D.E. (1981). The relationship between limit equilibrium slope stability methods.Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 3, pp. 409-416. Gens, A., Hutchinson, J.N. and Cavounidis, S. (1988). Three-dimensional analysis of slides in cohesive soils. Geotechnique, vol. 38, No. 1, pp 1-23. Golebiewska, A. (1976). An application of vane shear testing in organic soils. Ph. D. Thesis, Warsaw Agricultural University (in Polish) Golebiewska, A. (1983). Vane testing in peat. Proc. 7th Danube European Conference on Soil Mechanics and Foundation Engineering., Kishinev, Vol. 1, pp. 4953. Hungl, O. (1987). An extension of Bishop's simplified method of slope stability analysis to three dimensions. Geoteclmique, Vol. 37, No. 1, pp. 113- 117. Jamiolkowski, M., Ladd, C.C., Germaine, J.T. and Lancellotta, R. (1985). New developments in field and laboratory testing of soils. Theme lecture. Proc. 1l th International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 1, pp. 57-153. Janbu, N., Bjerrum, L. and Kjaernsli, B. (1956). Veiledning vid losning av fundamenteringsoppgaver. Norwegian Geotectmical Institute, Publication 16, Oslo. Janbu, N. (1973). Slope stability computations. Embankment-Dam Engineering Casagrande volume. John Wiley and Sons, New York.
17 8
Stability Analysis
Jardine, R.J. and Hight, D.W. (1987). The behaviour and analysis of embankments on soft clay. Special Publication on Embankments on soft soils, Bulletin of the Public Works Research Center, Athens, pp. 33-158. Lacasse, S.M., Ladd, C.C. and Barsvary, A.K. (1977). Undrained behaviour of embankments on New Liskeard varved clay. Canadian Geotechnical Journal, Vol. 14, No. 3, pp. 367-388.
Ladd, C.C. (1985a). Overview of clay behavior. Proc. for MIT Special Summer Course 1.6OS, Recent Developments in Measurement and Modeling of Clay Behaviour for Foundation Design, Lecture 2. Ladd, C.C. (1985b). Stability evaluation for staged construction, Proc. for MIT Special Summer Course 1.6OS, Recent Developments in Measurement and Modeling of Clay Behaviour for Foundation Design, Lecture 15.
Ladd, C.C. (1991). Stability evaluation during staged construction. The 22nd. Karl Terzaghi Lecture. Journal of Geoteclmical Engineering, ASCE, Vol. 117, No. 4, pp. 540-615.
Landva, A.O. (1980). Vane testing in peat. Canadian Geotechnical Journal, Vol. 17, No. l, pp. 1-19.
Landva, A.O. (1986). In-situ testing of peat. Proc. of In Situ '86, a Speciality Conference on Use of In Situ Tests in Geoteclmical Engineering, Blacksburg, Virginia, ASCE Geotechnical Special Publication 6, pp. 191-205.
Landva,,A.O., Pheeney, EE., La Rochelle, P. and Briaud, J.L. (1986). Structures on peatland - geotechnical investigations, Proc. Advances in Peatlands Engineering, Ottawa, 31-52. Larsson, R. (1980). Undrained shear strength in stability calculation of embankments and foundations on soft clays. Canadian Geotechnical Journal, Vol. 17, No. 4, pp. 591-602. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Report No.29, Link6ping. Larsson, R.(1990). Behaviour of organic clay and gyttja. Swedish Geotechnical Institute, Report No.38, Link6ping. Larsson, R. and S~illfors, G. (1981). Hypothetical yield envelope at stress rotation. Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 1, pp. 693-696. Larsson, R., Bergdahl, U. and Eriksson, L. (1984). Evaluation of shear strength in cohesive soils with special reference to Swedish practice and experience. Swedish Geotechnical Institute, Information No.3, 32 pp. Link6ping. Also in shorter version in ASTM Geotechnical Testing Journal, Vol. 10, No. 3, 1987.
References
179
Lechowicz, Z. (1986). Evaluation of the increase in shear strength in organic subsoil loaded by embankanent. Speciality Conference on Land Reclamation, Land Reclamation Facult~ Warsaw Agricultural University, 3, 77-84 (in Polish). Lechowicz, Z. (1994). All evaluation of the increase in shear strength of organic soils. Proc. International Workshop on Advances in Understanding and Modelling the Mechanical Behaviour of Peat, Delft, pp. 167-179, A.A. Balkema. Lechowicz, Z., Szymanski, A. and Wolski, W. (1984). Consolidation-strength analysis for soft soils. Proc. Sedimentation/Consolidation Models, ASCE, San Francisco, 107-120. Lo, K.Y. and Lee, C.F. (1973). Analysis of progressive failure in clay slopes. Proc. 8th International Conference on Soil Mechanics and Foundation Engineering, Moscow, Vol. 1.1, pp. 251-258. Lunne, T. and Kleven, A. (1981). Role of CPT in North Sea Foundation Engineering. Proc. Cone penetration testing and experience, St Louis, MO. Also in Norwegian Geotechnical Institute, Publication No. 139, Oslo. pp. 1- 14. Mayne, P. W. and Mitchell, J.K. (1988). Profiling of overconsolidation ratio in clays by field vane. Canadian Geotechnical Journal, Vol. 25, No.l, pp. 150- 157. Morgenstern, N.R. (1963). Stability charts for earth slopes during rapid draw down. Geoteclmique, Vol. 13, No. 2, pp. 121-131. Morgenstern, N.R. and Price, V.E. (1965). Analysis of the stability of general slip surfaces. Geotechnique, Vol. 15, No. 1, pp. 79-93. Runesson, K. (1978). On nonlinear consolidation of soft soils. Chalmers University of Tectmology, Department of Structural Mechanics, Publ. 78:1, Gothenburg. Spencel; E. (1967). A nlethod of analysis of the stability of elnbanklnents assuming parallel inter-slice forces. Geoteclmique, Vol. 17, No. 1, pp. 11-26. Tavenas, E and Leroueil, S. (1977). Effects of stresses and time on yielding of Clays. Proc. 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, Vol. 1, pp. 319-326. Tavenas, E, Blanchet, R., Garneau, R. and Leroueil S. (1978). The stability of stage-constructed embankments on soft clays. Canadian Geoteclmical Journal, Vol. 15, No. 2, pp. 283-305. Tayloh D.W. (1956). Fundamentals of soil mechanics, Jolm Wiley and Sons, New York. Terzaghi, K. (1956). Theoretical soil mechanics, Jolm Wiley and Sons, New York. Teunissen, J.A.M., Bauduin, M.H. and Calle, E.O.E (1986). Analysis of failure of an embankment on soft soil: A case study. Proc. 2nd International Symposium on Numerical Models in Geomechanics, Ghent, pp. 617-626.
180
Stability Analysis
Whitman, R.V. and Bailey, W.A. (1967). Use of computers for slope stability analysis. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 93, No. SM4, pp. 475-498. Wolski, W., Szymanski, A., Mirecki, J., Lechowicz, Z., Larsson, R., Hartlen, J., Garbulewski, K. and Bergdahl, U. (1988). Two stage-constructed embankments on organic soils. Swedish Geotechnical Institute, Report No. 32, Link6ping.
Wolski, W., Szymanski, A., Lechowicz, Z., Larsson, R., Hartlen, J. and Bergdahl,U. (1989). Full-scale failure test on stage-constructed test fill on organic soil. Swedish Geoteclmical Institute, Report No.36, Link6ping. Wright, S.G., Kulhawy, F.H. and Duncan, J.M. (1973). Accuracy of equilibrium slope stability analysis. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 99, No. SM 1O, pp. 783-791. Wroth, C.P. and Houlsby, G.T. (1985). Soil mechanics-property characterization and analysis procedures: Theme Lecture. Proc. 1lth International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 1, pp. 1-55.
181
Chapter 5
Analysis of Subsoil Deformations A. Szymanski, Department of Geotechnics, Warsaw Agricultural University
5.1
GENERAL
The design of embankments on organic soils involves estimation of the magnitude and rate of the subsoil deformations. In order to model the existing subsoil as well as the boundary conditions, two main cases are usually considered in these calculations (Fig. 5.1): case 1 - one-layer subsoil, and case 2 - multi-layer subsoil.
o)
J Xau=O Compressible layer &u=0 / ;" . . :" .".".".".".. <.':i
Compressible foyer d--V-u- 0 dx
Y/////~////////,
b)
'
\~u-0
Compressible
layer
Compressible
layer
/
A u =0
-...-.........."
'
\~u.0
Compressible
layer
Compressible
layer
du -0 a--x" -
Y///7///J/////,
Fig. 5.1. Boundary conditions in deformation analysis. (a) One-layer subsoil. (b) Multi-layer subsoil.
182
Analysis of subsoil deformations
A study of deformations in organic soils under embankments demonstrates large vertical and horizontal displacements in the loaded zone of the subsoil. Observations of the deformation development show that the greater part of the horizontal displacements appear during loading and shortly afterwards (Fig. 5.2). These results indicate that, except for the early phases of construction, the horizontal movements do not play any significant role in the consolidation process. Therefore, the calculation of settlement and excess pore pressure dissipation with time can be performed with one-dimensional analysis taking into account initial elastic deformations. Time, (days)
0
0
A
E d C
E
a)
24
48
I
72
!
Peat
0.6
96
!
Calcareous soil :Sand . . . .
l
'5 1.2-
1.8 C
0.6 b}
S
E
gE o.~ 0 IC 0
EL
,,..,_,,
.N_ 0.2 I,.,. 0 r-
C
~eo0 0
~4 0
Y Fig. 5.2.
~ Q,I2
0
I--
ii
f
construction schedule 480 Time (days)
III
960
Development of subsoil deformation under an embankment at the Antoniny site. (a) Settlement course under the centre of the embankment. (b) Maximum horizontal movements under the embankment slope.
Due to the significant variability of the soil parameters during a consolidation process and the large vertical displacements in the subsoil, the consolidation prediction should be based on a method which can account for the non-linear soil characteristics, the large deformations in the subsoil and preferably also the creep effects.
Deformation and Consolidation Parameters
183
In particular cases, the consolidation analysis can be supplemented by a twodimensional deformation analysis based on a realistic soil model. This may be helpful in deformation analyses for relatively narrow embankments on deep soft soils, where large horizontal displacements may appear during a large part of the deformation process.
5.2
DEFORMATION AND CONSOLIDATION PARAMETERS
5.2.1
Selection of parameters
Organic soils originate in wet conditions as a result of plant decomposition (peat) or plankton dissolution (gyttja). The variety of material from which organic soils originate influences their behaviour under load. It is rather difficult to describe this behaviour in terms of constitutive relations in theory of elasticity and plasticity. The significant secondary compression (visco-plastic creep in the soil skeleton) in organic soils calls for the application of non-linear consolidation approaches. A full descripton of the consolidation process in organic soils leads to very complex differential equations, which can only be solved by numerical methods. Solutions of these equations, without clearly defined material characteristics of the soils, are not very reliable. Therefore, for most practical cases the deformation and consolidation processes are predicted on the basis of one- dimensional strain analyses supplemented with estimation of initial "elastic" vertical and horizontal deformations because of the more easily definable parameters. The main characteristics which must be known in the calculations are the relationship between stress-strain and time expressed by parameters such as compression index Co or oedometer modulus M, coefficient of permeability k and coefficient of secondary compression C a. These parameters are obtained in oedometer tests, (Chapter 3.4). The initial deformations are estimated using a modulus of elasticity Eu determined in triaxial tests or from empirical relations (Chapters 3.5 and 5.3.2). The evaluation of compression and consolidation parameters for organic soils in oedometer tests is very often controversial due to sampling disturbance and difficulties in the interpretation of the obtained results (Hanrahan and Rogers, 1981; Landva and La Rochelle, 1982). Therefore, for a preliminary estimation of these parameters, several empirical formulae and diagrams based on relations obtained in the laboratory can be used (Chapters 5.2.2 and 5.2.3).
184
Analysis of subsoil deformations
However, it is necessary to remember that the empirical formulae are elaborated on the basis of a number of tests for particular types of soil. Therefore, their application to other soils, different to those investigated, could be misleading. Particularly in the case of peat and peaty soils, the empirical formulae should be applied with great caution. In special cases, when critical embankments on deep organic subsoil are beeing designed, an evaluation of the deformation process in plane strain conditions may be required. It is then necessary to employ triaxial tests in order to determine soil properties conventionally used in the theory of elasticity, i.e. two of the four elastic constants: Young's modulus E, Poisson's ratio v, shear modulus G and bulk modulus K. In engineering practice, the first pair (E, v) are normally used, whereas the second pair (G, K) are more fundamental mathematically because they separate pure shear from pure compression. Some typical examples of soil parameters required for this type of deformation and consolidation analysis are sketched in Fig. 5.3. It should be pointed out that the range of soil properties employed in calculations for the design of embankments depends on the requirements on the structures, the subsoil conditions and the method used in design calculations, as well as the availability of laboratory equipment and numerical programs.
h
,
"
~q
"..
one-dimensiona[ analysis -initial settlement
~[ane strain analysis
Eu
- consolidation
M or Cc
f
f
kv C~ p'
v
- consolidation kv, kh, C~x "r/
i
r
Fig. 5.3.
/ I I / / / / 1 1 1 / / / / / / 7 r 1 6 2 1 6 2 1 6 2 1 6 2
I I 1 / / / 1 / r 1 6 2w l r
,r/1/,'1~
Deformation and consolidation parameters required in design tions.
calcula-
Deformation and Consolidation Parameters
5.2.2.
185
Parameters used in deformation analysis
The calculation of settlements under embankments requires evaluation of soil parameters from the compression curves obtained in oedometer tests (Chapter 3.4). The results of IL (incremental loading) oedometer tests are usually presented as the relationship between void ratio e or strain ~ and effective vertical stress ~'v. The vertical effective stress may be plotted on a linear scale to determine the coefficient of volume change m v and oedometer modulus M or on a logarithmic scale to determine the compression index Co. It must be noted that the preconsolidation pressure ~'p is evaluated differently in the two plots. In continuously loaded oedometer tests, continuous curves are obtained for the relation of strain versus effective vertical stress. From this curve, a continuous relationship for the variation of the oedometer modulus with effective stress can be evaluated (see Chapter 3.4). The curve is divided into three parts with different values of the oedometer modulus M for each stress interval: 9 in the part CY'vo-cy p where M=M 0 9 in the part ~ p - ~'L where M=M L 9 in the part or'v > or'L where M=M L + M'(cr' v -G'L) A preliminary estimation of compression parameters can often be made with empirical formulae based on relations obtained in the laboratory with due regard to the applicability of the relation to the type of soil under consideration. The oedometer modulus M 0 may be estimated from empirical relations such as M0=250 ~f~ or M0=50 Cr'p for clay, (Larsson, 1986), and the compression index Co from relations such as Co =0.0115 w N for soft mineral soils (Azzouz et al., 1976) or the relationship between Co and void ratio e for peat ground presented by Kogure et a1.(1986), (Fig. 5.4). For some kinds of peat, the coefficient of volume change m v and oedometer modulus M may be estimated from empirical diagrams presenting the relation m v (Fig. 5.5), the O'p- (YL values (Fig. 5.6), the modulus M E (Fig. 5.7) and the modulus number M' (Fig. 5.8) to natural water content w N and degree of humification H (Flaate, 1968; Carlsten, 1988).
Analysis of subsoil deformations
186 20
10
5 0
o -
x"
0
_
33
oo
r
os
.o @ 13.
E
o
0.5 9Surface
peat stratum
o Lower peat
stratum
9Medium atternation strata 0.1
I
1
~
z
1
~ ,
5
Void ratio,
L,I
1
10
30
e
Fig. 5.4. Relation between C c and void ratio e (Kogure et al., 1986).
FI
!
-
---
,
I-
~ ~ -J ~ , / ' _ , , ~ 1 : 1
0.010 -,
-
q,kPo
~
_.
0008
13..
o
o.oo6
x
0.005
~-~.
"-
E
o.oo~
~" ~ ~ ~
" ~~
0.003
,CI~
.
.,<>_ ;
~ / ~
.~~:... ~'-
J,--~
~ . -, ~ _ " i ~ 8 0
~
- -_.. .
~
I-
,. " ~
,,..~.>..-
i , _
., 120
.~.4. ~16~176
.;-'. -_.,<.'..;.~_~8o I~ ~ "
1i_
. . . , 7 0
/
~
~'P
" -'-J'-
~
11.
.~-
i
-
..,
~,~
I
_
"
"J
"~.~
..
:;,,- " - " L,..""-
-~--
I H5_7=~ Ht= 2
0.002
0
200
400
600
800
i
1000
1200
1400
._~_w ( % Ht
Fig. 5.5.
C o e f f i c i e n t of v o l u m e c h a n g e m v as a f u n c t i o n of a p p l i e d s u r f a c e load q, n a t u r a l w a t e r c o n t e n t w N a n d d e c o m p o s i t i o n f a c t o r H t (Flaate, 1968).
Deformation and Consolidation Parameters 40
i
i
i
i
i
i
i
i
187 Fig. 5.6.
i
H5 - H10 H1 H / , O 13.. ........
Relation between (a" L a'p) and natural water content w N (Carlsten, 1988). H - Degree of d e c o m p o s i t i o n according to van Post (1924).
"@ 20 I . d
Lo _ _
0
1000
2000
Water content, w n (%)
Fig. 5.7.
200
Relation between modulus M L and natural water content w N (Carlsten, 1988). H = Degree of decomposition according to van Post (1924).
1
[
1
i
l
!
w
I
1
I
I
I
!
J
Peat H,-Hlol
,..-.,. El 12.
I
.__1
100 D "lD O
0 10.0
1
i
i
7.5
1
1
0
!
I
1
I
1
I
I
1000 2000 W a t e r c o n t e n t , w n (%)
_
J
E 5.0-
rC)
Fig. 5.8.
o') D
xJ 2.50
0
0
I
Peat
H5-H10
Peat
H1-H&
I
10b0
j
,
,
Water content, w n (%)
J
2000
Relation between modulus number M' and natural water content w N (Carlsten, 1988). H = Degree of decomposition according to van Post (1924).
Analysis of subsoil deformations
188 5.2.3
Consolidation
parameters
For prediction of the deformation process and the effective stress paths in the subsoil under an embankment, it is necessary to evaluate also the consolidation characteristics, which are usually obtained from oedometer test results. The consolidation curves obtained in incremental oedometer tests are usually divided into three parts (Fig. 5.9): 9 the initial deformation (~ < e.0) which is immediate, 9 the primary consolidation (eo < e _< El OO) describing the compression process during excess pore pressure dissipation, 9 the secondary compression (e > el00) which develops due to creep deformations after dissipation of excess pore pressures.
EO ~-
Time, (tog scare) ~ Previous toad " Initiat deformation s
c-"
Primary consotidation
L_
-6 s ~
I.....
\
Secondary
compression
>
Fig 5.9. Consolidation curve of soft soil.
The results of deformation tests (Ozden and Wilson, 1970, Berry and Poskitt, 1972, Edil and Dhowian, 1979, Hobbs, 1986) indicate that in organic soils, as in clays, all these phases of deformation are observed simultaneously at varying gradients of pore pressure. In consequence, it is difficult to give simple empirical or theoretical stress-strain characteristics for particular stages of deformations. Terzaghi's theory supplemented by secondary compression is often used for calculation of the consolidation process because of the easily definable parameters. The coefficient of consolidation cv can be estimated in incremental loading tests by one of two commonly used methods (Casagrande's or Taylor's). In continuous loading tests or incremental tests with direct measurements of permeability, c v is determined from the product of modulus and permeability, divided by water density.
189
Deformation and Consolidation Parameters
The coefficient of secondary consolidation is evaluated from the slope of the consolidation curve after the end of "primary" consolidation (e > el00)" The coefficient of secondary consolidation is expressed as either: 9 C~=de/dlogt from consolidation curves plotted as e versus log t or 9 Co~=da/dlogt from consolidation curves plotted as ~ versus log t.
The relation between these coefficients of secondary compression is C~ = C ~ (l+e0). The comprehensive analysis of tests results performed by Mesri (1973) and Mesri and Godlewski (1977) showed that the coefficient of secondary compression Co~ , similar to the compression index Co is mainly dependent on water content w~ (Fig. 5.10). This relation was further confirmed by tests on Swedish clays (Larsson 1981, 1986). These results gave the average relation C~/Co for organic soils in the range between 0.035 and 0.1, the lower values referring to amorphous peat and other nonfibrous organic soils, and the higher values referring to fibrous peat. 0.5
I
I I
I
I
I
I II|
I
I
k ( ~ Clay, organic ctay Land gyttjo
0
O.1
'l|
(Larsson 1990)
\ ~\
\
8 o o.o~
0.001
Fig.5.10.
10
,,, i , i I I i jjJl 1000 2000 100 Natural water content, wn (%)
Modified secondary compression index C=~ versus water content (Mesri and Godlewski, 1977; Wolski, 1989 and Larsson 1990).
WN
Analysis of subsoil deformations
190
Investigations on Swedish clays, organic clays and clayey gyttja, (Larsson 1990), have shown that for this type of soil the coefficient of secondary consolidation can be estimated from: Co~= 1.35e- 1
,%
(5.1)
In this formula, the active void ratio e is calculated using the specific densities 2.7 t/m 3 for mineral soil and 0.7 t/m 3 for organic matter, (Fig. 5.11).
Co~E (max): 1"35e-! y / -
T
9
9
u C~ 0
x 0
3
2-
(.9
I
0 0 Fig. 5.11.
9 / v/ . " -
.
Assumption: ?s = 2.7minerat matter .,os = 07 organic matter ,,..._
I
89
3
Void ratio, e
4
5
Relation between coefficient of secondary consolidation and estimated active void ratio (Larsson 1990).
For peats, the coefficient of secondary consolidation was found to be related more to the degree of humification than to the water content, Fig. 5.12. The prediction of a consolidation course when the deformations are large requires a description of the variation of the void ratio e and the permeability k during the consolidation process. These characteristics are usually obtained in oedometer tests performed with constant rate of strain, but can also be determined in incremental tests with direct measurements of permeability at various deformations (Fig. 5.13).
191
Deformation and Consolidation Parameters
J
9
C~ O
9
3
o
J~J
J~f
J
I
J
J
J
J
I
9
Legend
9
9 w N
< 500 %
9 wN I
/
o
O
....
0 Fig. 5.12.
500-1000%
o w N 1000-1500%
f
3
}.
~
T
DEGREE
'
s
~
WN>1500
~
" l " ~ ""--''T . . . .
6
7
8
OF HUMIFICATION
H
Coefficient of secondary consolidation versus degree of humification in peat. (Larsson 1990). H = Degree of humification according to yon Post (1924). Permeability, 6 0.1
9
I
!
_
1
pllll
I
k (x 10 m/s)
!
1
95
l
~ ltrl'~
X
0
X
O" ..8..,
x
0
o L-
0
o
>
10
X X X
0
X X
o 1
100
ll'll
O
0
I
1
X
0
l
l
0
0
|
i
Xx
0
4
!
o e= f(k) I
I
1 I II
10
l
. I
l
I
I l[lJ
x e= f (6") 1_.
100
t
1
1
IllL
1000
Effective stress, 6" (kPa) Fig. 5.13.
Consolidation characteristics of peat from Antoniny (Yong et al., 1988).
Empirical relations for initial permeability and permeability change index for peat can be found in Chapter 3.7.
192
Analysis of subsoil deformations
5.3
ANALYSIS OF "FINAL" DEFORMATION
5.3.1
Type and scope of analysis
For embankments on a thin layer of soft subsoil, an evaluation of the final settlement is sufficient to design the embankment. In this case, the settlement is estimated by one-dimensional analysis or empirical formulae. When embankments are to be constructed on deep organic subsoil, a full description of the behavoiur of the soft subsoil is required to evaluate the performance of the embankment. In order to evaluate the vertical and horizontal displacements as well as the effective stresses in the subsoil, calculation methods accounting for two-dimensional deformations and changes in geometries during the consolidation process should be used (see Chapter 5.7). If the simplified solution of the consolidation theory is used, the one- dimensional deformation prediction should be supplemented by an estimation of the initial plastic movement occurring in the initial stage of loading because of shear and lateral deformations in the subsoil. The total settlements should be estimated as a sum of the initial movement Si and long-term displacement Sf created by primary consolidation S~ and secondary compression S~: S = S i q- Sf
(5.2)
or S = S i + Se+ Ss
(5.3)
The initial movement and the secondary deformation of the subsoil under an embankment play significant roles in the consolidation process in organic soils. The deformations can also be estimated by use of finite elements and theory of elasto-plasticity. The applied soil models may then for example be hyperbolic - strain hardening (elastic Duncan - Chang model, Duncan 1980) or Critical State models (Cam Clay model, Wroth and Houlsby 1980) for soils with isotropic properties or anisotropic models, (e.g. Runesson 1978), for soils with anisotropic properties. The selection of model also depends on the effective stress level and the expected soil displacements under the embankment, (range of elastic and plastic deformations). Elastic models cannot be used when the plastic deformations in the subsoil are large. In such cases elasto-plastic models should be applied.
Analysis of "Final" Deformation
193
Initial settlement and horizontal movement
5.3.2
If the applied construction load has a limited lateral extension, shear and lateral deformations occur in the subsoil, which results in initial vertical and horizontal displacements. Such deformations are mainly associated with undrained conditions. Therefore, the computation of initial settlement S i and horizontal movement Sh is often based on theory of elasticity using Poisson's ratio v = 0.5 and an undrained modulus of elasticity E u. The initial deformations calculated by the theory of elasticity become:
S~- Iv q B/E.
(5.4)
and Sh =I h q H/E u
(5.5)
where q = stress applied to the subsoil B = width of the loaded area I
= influence factor which depends on the geometry of the problem ( I and Ih refer to vertical and horizontal direction respectively)
E -
undrained modulus of elasticity,
H = thickness of the compressible layer. Application of this theory to the prediction of initial movements requires the estimation of modulus E from laboratory tests or from empirical correlations with the undrained shear strength. Larsson (1986) points out that the undrained modulus E values vary from 80 ~fu for organic soils to 2000 ~f~ for low-plastic clays. The results of investigations carried out by Foott and Ladd (1981) indicate that the undrained modulus for normally consolidated soils can be calculated from the empirical formula: E u = 1:f~215 lnF/Ip
(5.6)
where ~f~ = undrained shear strength from vane shear tests or direct simple shear tests, F = calculated factor of safety against shear failure, Iv -
plasticity index.
Analysis of subsoil deformations
194
When the soil conditions vary with depth, a harmonic mean value of E u is often used.
Calculation of the initial deformation in soft subsoil under an embankment with Eqn. 5.4 and 5.5 requires evaluation of the displacement influence factors Iv and Ih from the theory of elasticity. Diagrams for various conditions have been presented by Steinbrenner (1934), Janbu et a1.(1964) and Poulos (1972), Figs. 5.14, 5.15 and 5.16.
a
3.0 ..,, ......... .......... C,_,,~o,,~,, ~ L-length 2.0
H ~
,-
0.9
fo.
~ 21
Iv=)J1 )u 0
5
~ :o.s
1.o
o.%.~ 0
Fig. 5.14.
01
dL
1 l lllll]
1
J
I
lllllll
10 H/B
l
l
I IJJ|II
100
l
[
IJIIU
~o
~oo
Hf/B
~ooo
1000
Influence factor Iv for evaluating the vertical displacements in the centre of a loaded area (Janbu et al., 1964).
00
x/B
1
2
00
=
~,
Influence factor, I h 0.1
0.2
O.Z,
-0.1
N
0.6 0.8
0.2
.5
1.0~''_q._.
[ B , 9
,,-1
fil"~l ~liJ J I i l J
Sh:Ih
0.3
2
0.2
'E
qH/Eu }
I x_ H
/////////////i////////////////,4
Fig. 5.15.
~
Case Oistrib.
of
Eu
with depth
'
1
'
2
3
~] 1.5Eu
0.5Eu
Influence factor I h f o r evaluating the horizontal displacements under a uniform strip load (Poulos, 1972). (a) Horizontal surface displacements. (b) Horizontal displacements beneath edge.
195
Analysis of "Final" Deformation
]' 2
/ r--
I T-li~176176 I
!, I
!
.~ b "1" ]
I
61~
~ =1
I i
i,a
! !
10 .I 0
~l
i
I
I
0.2
0.4
0.6
0.8
fl,f2 Fig. 5.16. The coefficients fl and 1'2 for Steinbrenner's formula.
Using Janbu's diagram to estimate the influence factor Iv with good accuracy is difficult for small values of the H/B ratio. In such a case, therefore, the factor Iv can be calculated directly from the analytical solution of the elastic theory according to Steinbrenner (1934). The settlement under a comer of a uniform strip load is expressed as:
(5.7)
(q B/E u) ((1-v2)fl + (1- v - 2v 2) f2)
Sic =
where"
fl = ~
1 fL
-In
B
(1+~1)~C1+Ch-2 + In((L/B)+'~I)'~h (L/B) (I+~CI+Ch-I)
1
(5.8)
L/B + ~CI+Ch-I
and 1/B f2 ~
(5.9)
(H/B) arctan 2rt
(H/B)~C
1-I-Ch- 1
where C 1 = 1 d-(L/B)
2
(5.1o)
Analysis of subsoil deformations
196 and Ch =
1 + (H/B) 2
(5.11)
where L = length of the loaded area, B = width of the loaded area, H = thickness of compressible layer. The coefficients fl and fz can also be determined from the diagram presented in Fig. 5.16. For the special case ofv - 0.5 equation 5.7 is reduced to Si~ = (q" B . 0.75 9fl) / E,. To use the equation (5.7) for a point within a loaded area, the area should be divided into four rectangles with comers at the considered point. To obtain the settlement at the point, the comer settlements of the four rectangles should be calculated and added.
5.3.3
Empirical prediction of "final" settlement
The final settlement of the subsoil can be estimated with empirical formulae, based on results of compressibility tests or collected field observations. Such empirical methods are presented by Ostromecki (1956), Drozd-Zajac (1968), Flaate (1968), Niesche (1977) and Carlsten (1988). The Ostromecki formula gives settlements of peat with thickness less than 4.5 m and loaded to 10-50 kPa. The equation is as follows: S = 1.08. C t o
(5.12)
where S = settlement (m) C = coefficient (a function of H/t 0 and ~'d) t o = q/7.55 normalized depth of dewatering (m) H = peat thickness (m) q = stress applied to the subsoil (kPa) 7a = dry unit weight of peat A nomogram for evaluation of the coefficient C is presented in Fig. 5.17.
Analysis of "Final" Deformation
197
0.9
]
r
08 '
,
. . ~ ~i. ~ , ~ > ' [ |
/,"
'
j
0.7
....
o.6
~
,.,,=
~9 o.~
~
f
0.3
.
/
/--/"
-/
i1/"
..,,I
/,//,
I-" ~
.i
~ ,
~
~
l
.
,
~_~
,
'
t
!
~I i . , /
~
~~''"
o.1 oo
~
0
~
0.2 0./, 0.6 0.8
_.-~ ~ ~
,
~
u ~ ~ ~ .
~
~;;t'~
---1.2; ,,_. _ . - ~ r - t ~ 1.66
_~~.----.-.---1
"
0.9~
~ ~ ~ ~ 0.2
I
~ -
i
-'--2
.,,,,,,~ ~ r " '.~
~
"
,
,
_~"",
--~.~
, ,
,
, " ------.-----------" -
. . . . . . . ~ .
,
3
/*
Coefficient C for the Ostromecki (1956) formula.
Settlement observations of peats of different degrees of decomposition enabled Niesche (1977) to elaborate an empirical approach to evaluate the final deformation of peat at loading. On the basis of field observations, Niesche gives a nomogram for the evaluation of deformation when the load q and degree of decomposition are known. This nomogram is presented in Fig. 5.18. An estimation of final settlement in organic clay and gyttja can be made with the Drozd-Zajac method (1968). The authors give the following empirical equation: (e 0 -0.36) 1"1
S = 3.6H
lg(3.12q ~e0) l+e 0
where S = settlement (ram) H = thickness of compressible layer (m) e 0 = initial void ratio q = stress increase (bar)
,
,---r-
H/t,
Fig. 5.17.
,
(5.13)
Analysis of subsoil deformations
198 70
60
--
-
50
,....
E "6 E
z.O
30
3
0
10
20
30
40
50
60
70
80
90
100
Degree of decomposition, (%)
Fig. 5.18. Niesche's (1977) diagram for settlement calculation.
The settlement investigations in peats with different water contents performed by Flaate gave an empirical nomogram for evaluating the coefficient of volume change m v (Fig. 5.5). In this approach, the calculation of the total settlement is made with: Sf = m v q H
(5.14)
The final settlement in peat when the water content is in the range 700 - 1500 % can also be estimated on the basis of the empirical diagram (Fig. 5.19) presented by Carlsten (1988) to estimate the relative vertical deformation ~. In this case the final settlement is calculated as" Sf = e H
(5.15)
The diagram anticipates that the peat is normally consolidated. In the case of slightly overconsolidated peat, the following correction to evaluate the calculation load q" for estimation of settlement was suggested: q'=q-(g'p -or'0) where q = applied load ~ p= preconsolidation pressure ~'0 = in situ vertical stress before load application
(5.16)
Analysis of "Final" Deformation
199
100 Apptied l
toad (kPa)
~9. ~ - - . ~ - , , - ~
~- 80 60
~-- ~ ' - ~
W
E
_~
O .m
"6 5O
Applied
E L.
~. ~
toad (kPa)
100 /
O
I
~ 30 ~
.-~ ~.-
I,
20
10
123
0
9 560
.
.
.
.
1000
Water content Fig. 5 . 1 9 .
.
.
.
.
15100 .
.
.
.
(%)
C a r l s t e n ' s (1988) d i a g r a m f o r c a l c u l a t i o n t i o n ~ in p e a t .
2000
of vertical deforma-
Empirical methods are based on field settlement measurements of particular constructions. They can be used to predict settlements only in the specific kind of subsoil for which they were elaborated. The use of empirical methods for calculation of settlements in the soft soil therefore often gives low credibility for the results. The settlements for the final design should always be estimated from parameters obtained in laboratory or field tests and one- or two-dimensional calculations.
5.3.4
Prediction of settlement in one-dimensional consolidation
The estimation of settlement in one-dimensional consolidation requires knowledge of the increase in vertical stress developed in the subsoil as a result of the load from the embankment. The stress increase is calculated by the theory of elasticity. The state of stress in the subsoil under an embankment can be estimated by Gray's approach (1936) presented in Fig. 5.20. in which the stress components ( ~ - vertical, 6h - horizontal, I: - shear stress) are calculated as follows: ~v = q (13 + x c~/a-z (x-b)/R22)/~
(5.17)
cyh = q ([3 + x c~/a + z(x-b)/R22 +2zln(R1/R0)/a)ht
(5.18)
I: = -q (z cz/a-z2/R22 )/n
(5.19)
200
Analysis of subsoil deformations
~"
I
..J
~-!-~1
I
q/unit
J
QreQ
\
\\RI R~ ~ //
X\
/
/R2
Fig. 5.20. Vertical "embankment" loading (Gray, 1936).
On the basis of this solution, Osterberg (1957) published a diagram to estimate the influence factor I for evaluation of the vertical stress increase beneath an embankment with unlimited length (Fig.5.21). In the case of an embankment with limited length 1, the influence factor can be determined by using Fadum's (1948) approach and the chart for vertical stresses beneath the comer of a uniformly loaded rectangle (Fig 5.22). If a point within a loaded area is considered, the area should be divided into four rectangles with comers at the considered point. To obtain the vertical stress at the point, the stresses beneath the comer of the four rectangles should be calculated and added. The settlement calculations for one-dimensional consolidation can be made for one or more subsoil strata. In the case of many strata, the total settlement is defined as the sum of all settlements. It is assumed that the whole subsoil stratum is taken into account in the calculations (see Chapter 9). The final time-dependent deformation Sf is calculated as the sum of the initial settlement Si, consolidation settlement S and settlements due to secondary compression S. The simplest way to determine the consolidation settlement is by means of: So = A~' v H / M
(5.20)
or
So=a~H
(5.21)
201
Analysis of "Final" Deformation b/z =oo 0.5 c~:::==:==~:: 3
0./-,
O.B
u.0
u 0.3 ,2 o u C 0 .._.,
bl b 2 0 2
0.2
C
0.1
'
Gv=I-q 1 -
I=II+I2 6v 0 i
0.01
,--7
I i iIIli
0.1
a/z
10
1
Fig. 5.21. Influence factor I for vertical stress due to embankment loading (Osterberg, 1957).
0.261
I
0.22
m=t/z n=b/z
0.18
~v =ql
I
I I IIII"
rn= o o ~
2.5 1.8
0.6 I
0.14 0.10
0.5
n are/i///f"--~
Note" m and interchangeable
0.~ 0.3 0.2
0.06
Fig. 5.22. Vertical stress beneath the corner of a uniformly loaded rectangle (Fadum 1948).
0.1
0.02 ~.~
~
0.1
1
10
Analysis of subsoil deformations
202 where M = S~ = H = ~ = Act"V =
modulus of compressibility consolidation settlement thickness of soft subsoil consolidation strain effective vertical stress increment
The above formula can be used for homogeneous subsoils with small moduli variation within the actual stress interval, e.g. heavily overconsolidated soils, and with relatively small thickness H compared to the loaded area, (i. e. relatively small change in A~' v with depth). In other cases taking moduli variation into account, the consolidation strain e~ is determined according to the formula: Ae eo =
Cr
1+e 0
where e0 = Ae = c" 0 = cr p = 6",r = Cr = C~ =
(Y'
~
log
1 +e 0
Co
P +~ cr 0
1 +e 0
O'vf
log ~
(5.22) cr p
initial void ratio decrease in void ratio initial vertical effective stress preconsolidation pressure final effective vertical stress recompression index compression index
The final relative compression at varying moduli can also be determined by means of Larsson's (1981) proposition with the estimation of three different moduli. (~ p -(~'0
Eo = ~ + M0
0" L -(3" p
~
+ My
1 M'(~',r ~'L ) +1) ~ ln( M' ME
(5.23)
The necessary explanations are presented in Fig. 3.20. The method can be used for single layer or multi layer subsoils. Based on extensive investigations, performed by the Swedish Geotechnical Institute on organic soils, an empirical nomogram was established for estimation of the
Consolidation Analysis
203
moduli values for some types of peat depending on the degree of decomposition and water content (see Chapter 5.2) A simple calculation of secondary settlement S~ describing the time-dependent movements due to secondary compression after pore pressure equalization can be based on a coefficient of secondary consolidation C~ or Co~, which is relevant for the actual void ratio or water content after primary consolidation. The settlement S~ is then expressed as: S~ = Ca log(tf/tp ) H/(1 +e0) or
(5.24)
S~ - C~e log(tf/tp ) H where tf = time of the end of the period covered by the prognosis tp = time for the end of primary consolidation The coefficient of secondary consolidation is determined in incremental oedometer tests or it may be estimated from empirical relations such as those presented in Chapter 5.2.
5.4
CONSOLIDATION ANALYSIS
5.4.1
Type and scope of analysis
When constructing embankments on organic soils, it is often necessary to utilize the shear strength increase in the subsoil during construction by stages. Such cases involve an accurate estimation of the subsoil deformation rate and the excess pore pressure dissipation in the consecutive stages. The position and thickness of each subsoil layer and the prediction of the effective stresses with time makes it possible to estimate the shear strength increase and to select a preliminary construction time schedule with consideration to stability. The same requirements for an accurate prediction of the consolidation process apply when preloading and surcharging is used in order to stop subsequent settlements. The consolidation analysis of soft soils can be made with methods based on onedimensional theory, in which the settlement St at time t is calculated as: St = U Sc
(5.25)
Analysis of subsoil deformations
204 where U = degree of consolidation So = total consolidation settlement
Due to a significant variability of the soil parameters during the deformation process, the consolidation prediction should be based on a method which accounts for the variation of the parameters with time and preferably also for creep effects. In the case of construction of embankments on deep soft subsoil, especially when the subsoil consists of several layers, the consolidation prediction should be made with methods which take into consideration the interaction of several consolidating layers with different characteristics, as well as the changing geometry and stresses due to large strains in the soil. In construction of embankments on soft soils, the very slow consolidation process causes great difficulties in designing a practical construction schedule. In order to accelerate the consolidation process, vertical drains of different types are often used. The prediction of the deformation process can then be carried out, assuming one-dimensional state of strain and vertical and axi-symmetrical pore water flow.
5.4.2
Empirical prediction of the consolidation course
In practice, the estimation of the settlements is often sufficient to design the embankment when the organic soil is a surface layer with limited thickness. In such a case, the preliminary calculation of the degree of consolidation U can be made with empirical formulae for the specific kind of subsoil for which they were elaborated. The preliminary estimation of consolidation course in peaty subsoil can be made by using Carlsten's diagrams prepared on the basis of experience from Swedish peats (Carlsten, 1988). To obtain simple diagrams for estimation ofthe rate of settlements in peat, the consolidation equation was simplified to the following form: 0.52 (wN)~ -( U = 1 - 0.6 e
qO.5 -)t (5.26)
The equation (5.26) is valid for the case where there is free drainage at both the top and the bottom of the peat layer. Limits for the investigated peats in the empirical data base are: thickness of peat 2 - 6m; water content 900 - 1500 %; applied load q <50 kPa. The corresponding analysis, performed for the case of an impermeable bottom layer, gave the following equation:
205
Consolidation Analysis 0.13 (w~) ~ -( U:
1-
)" t
H 2 q0.5
0.6 e
(5.27)
Equations 5.26 and 5.27 have been transferred by Carlsten to the diagram presented in Fig. 5.23. The diagram shows the case of a permeable bottom layer, but can also be used for an impermeable bottom layer ifa fictive height of twice the real height is used. Sometimes, back analysis of an observed deformation process is used to predict the further course of the process, especially in connection with stage-loading and preloading. In such cases, the settlements are measured during the consolidation process and the predicted course of consolidation can be checked and improved. For this purpose, the method developed by Asaoka (1978) can be used.In this method, the final settlement is estimated with a graphic procedure. A
1400
i:: 1200 E 1000
10}8
2
/ I / f, / l I I /," ! I 1 i ! i/,r/I ,' 1
o 800 ,.600 / , O '6 /-.00 O
,6,5 [4 ,3
I
/
Z
#
/
Thickness of pe:lt, lnh)~
I 1
I 1
1 1
1 1
I
l ! t
I~I /, lollt',l\ , , IIIVl\
t 1 /i--2qili,"
! //,),~ 1/Y/,ti'Ji,@~
,.'/7,4t'./,I~ 1 /,!l)[ll~
I .,'L ~IIi I,',IDF_A , !
t" lApplied ......
I 1
1
I I I !
L
~\ -\\ ~,~, IN . [
I C i
(oad (kPa)
i
"99,.
-\
~\'~85 i\90",I ,8o[-, 1\
: [tt"
[
0.52. w 0"75 ].t [ U=l-0.6.e H2'q 0.5
-4.... "
I I !
4,, ,...
50
100
150
200 250 Time. (days)
Fig. 5.23. Diagram for calculation of consolidation in peat (Carlsten, 1988).
206
Analysis of subsoil deformations Time, t 0 to
t/1
to+z~t to+2Z~t
to+4~t
So $I s2
-~ s3
54
"cJ > i_
s
-- ~ $4 $3 $2
~'~'
-
Sc
$I
u~ r~ C)
0 So
$I $2S3
Si-1
Fig. 5.24. Estimation of consolidation parameters according to the Asaoka (1978) method.
Using the observed settlements at various times, the consolidation curve is determined as S i = f(t i ), where t i = t0+iAt (Fig. 5.24). The points on the consolidation curve are then used in the linear relationship Si= f(Si.1). The intersection point ofthe line thus obtained and the bisector of the coordinates is the total settlement S Gand the curve inclination gives the consolidation parameter ~1" Settlement within the time t can be calculated from the formula: S t = So (1- exp((lnl~ 1/At)O )
(5.28)
Asaoka's method is based on Terzaghi's consolidation theory, where all parameters are constant and independent of time and deformation. Therefore, the parameters in Asaoka's method change during the consolidation process owing to creep effects (~1' > 131)"This is illustrated by the observation of the settlements of a stage constructed embankment at the Antoniny site in Fig. 5.25.
Consolidation Analysis 1.2 Si (m) ~. o
207
s]E o.__g_~
/
~LIst2 g~
tg p =o.9~
O.8 SI I =0. 0.6
0.2 0.0 0.0
I
I
I
I
0.2
0.4
0.6
0.8
1.0 1.2 Si-1 (m}
Fig. 5.25. Change in consolidation parameters in Asaoka's method during consolidation o. organic subsoil at the Antoniny site (Wolski et al, 1988).
5.4.3
Prediction of one-dimensional consolidation at small strains
Terzaghi's linear, one-dimensional consolidation theory (1924) may be used for relatively thin compressible layers and homogeneous soil. Due to the assumptions adopted in this theory, such as constant relationship between void ratio and effective stress, constant permeability and small strains, its applicability is limited to relatively stiff thin layers with small changes in parameter values. The usual form of Terzaghi's equation is:
~U 8t
~2U Cv
(5.29) 3x 2
where u = excess pore water pressure % = coefficient of consolidation = ( k M) / ( g Pw) t = time; x = vertical space coordinate
208
Analysis of subsoil deformations
The degree of vertical consolidation U v c a n be calculated from the relationship between U v and T v represented by curves in Fig. 5.26. For soft subsoils in which the coefficient of volume change m v and permeability k are changeable and are known functions of depth, the consolidation prediction can be made with Schiffman and Gibson's (1964 ) equation:
a~u
1 dk au
%mv(X) au =
~-
ax 2
k
(5.3o)
dx c3x
k(x)
31:
where k = permeability m v - coefficient of volume change Yw = unit weight of water. I
I
I
I
I
I
II
i
I
I
: .'.'.':.'.
";::
. : :'.;
:"
:.'.'.'.-.':
i 0 I/)
c 0.60o u
l'~
~ .' ...' : .': ..~
Curve I
Curve 2 Curve 3
: : ; ''"."i:
: : ". ".'.'.'."
::
'.
:.'..::
I
I
I 1
::
. ".'." :.'.'.'"
Half-closed tayers
i z
T M / / / ,,///z,
1.00 0.001
I
Open layers
: : : .'.'.'..."
,0.80 O~
I
Zv- ;
> 020-
Z)
I
l
l
l
i
i
i Jl
0.01
I
I
Time
J
I
t I
factor,
II
0.1
I
I
I
1
11 I II 1
Tv
Fig. 5.26. Relationship between time factor T v and degree of consolidation U.
209
Consolidation Analysis
5.4.4
Prediction of one-dimensional consolidation at large strains When the course of one-dimensional consolidation of thick and soft soil layers is to be predicted, the change in soil parameters, as well as the change in load and geometry because of the large strains and total deformations, has to be accounted for. In this case, the consolidation prediction among other things involves the application of a reduced or convective coordinate system (Fig.5.27). Reference plane (x=O .,,
a)
Reference
Reference plane (z=O)
plane (~':O) b)
l"(x t)
No
initial depth subsoil
Fig. 5.27.
B
Zo
volume of solids
~(xo t) present depth subsoil
Coordinate systems used in consolidation analysis. (a) Lagrange coordinate system at time t=0. (b) Convective coordinate system at time t. (c) Reduced coordinate system at time t.
Approaches to the consolidation prediction with large strain analysis for onedimensional state of strain and pore water flow were presented by Gibson and Schiffman (1981) and by Yong and Ludwig (1984). These methods were originally designed for the extreme cases of consolidation of slurries.Gibson and Schiffman proposed the following equation in a reduced coordinate system:
_(y~ _1) d~ Yw
k
~ + ~ b e~ f
de (1--~-e) ~z
where y~ - unit weight of solids e = void ratio z - reduced coordinate.
~z
k
de" ~et + D e
yw(l+e) de
~z
Dt
=0 x
(5.31)
210
Analysis of subsoil deformations
Yong and Ludwig (1984) proposed a consolidation equation with the application of a convective coordinate system (Fig.5.28) as"
1
de Du
=0 l+e dcr' Dt
(5.32)
X
where = convective coordinate u = excess pore water pressure Du = material derivative Dt x
Unit
area
>,,
.....
IZ
V/,,~'S 0 LIDS//,//,,~"'-.~
,.,-..
0 Z
=
Vs
4-, .,.._, + r--
N~,
ii -i-,, N ,,._.,
0
VOIDS Soil
skeleton
Fig. 5.28. Relationship between convective position ~ and reduced coordinate z.
An accurate description of the consolidation process, taking time effects on the compressibility of the soil into account, leads to the following equation in a convective coordinate system (Szymanski 1991):
Consolidation Analysis
.
~
.
%
.
.
211
1 1 0~J
l+e
deP Du dcr" Dt V
+~ x
d~1
=0
(5.33)
dt
where deP = change in void ratio due to primary consolidation de ~ = change in void ratio due to secondary consolidation deP
= C r
log(t~'p/G'vo) + C o log(t~' v / t~'p )
de ~ = eo - C~ log(t / tp)
(5.34)
(5.35)
e ~ = initial void ratio tp
= time w h e n e - eo
C~ = recompressionindex Co = compression index C~ = coefficient of secondary consolidation CY'vo= initial effective vertical stress c'p = preconsolidation pressure ~'v = effective vertical stress The application of differential equations to predict settlements and excess pore pressure dissipation requires these equations to be solved by means of numerical procedures because of the non-linear nature of their coefficients. Numerical solutions can be based on a finite difference scheme with the application of the fmitestrain consolidation analysis or a piece-wise linear approach. The finite consolidation analysis developed by Gibson et al. (1981) is based on the reduced coordinate system. The governing equation requires recalculation at each step for the current void ratio value and the boundary conditions defined in terms of void ratio. In the piece-wise linear iterative analysis developed by Yong and Ludvig (1984), the derivation for finite difference consolidation is performed with respect to a convective coordinate system. With the piece-wise linear iterative approach, non-linear soil properties and non-homogeneous material can be accommodated in the analysis. This approach is based on updating the excess pore pressure explicitly.
Analysis of subsoil deformations
212
Calculation of the consolidation process with these types of analysis requires the specific gravity of solids, the relationship between void ratio and effective stress and the coefficient of permeability estimated from oedometer tests. The application of the large-strain consolidation analysis to layered subsoil requires taking into account the boundary problem which appears at the level between compressible layers, because of the differences in pore pressure and pore pressure gradient values obtained from numerical calculations for each different layer at the boundary level. One way of solving this problem is to use common imagined boundaries with imaginary mesh points (Cargill, 1982). This iterative numerical procedure is complex when the soil consists of several thin layers. A more sophisticated method without the necessity of solving boundary problems in the numerical calculation for layered subsoil can be applied by using the implicit scheme in fimte consolidation analysis. This procedure, based on equation 5.33, has been used e.g. to predict the deformation process in the subsoil under a test embankment constructed at the Antoniny site. Results obtained from numerical computations based on the original Terzaghi procedure and Yong's equations for this embankment are shown in Fig.5.29. "
I
i
!
9\ x
~ ~
....
"~-.~,. ~ ~..~.~
E
I
- - - - - CONMULT without creep
0.4
.....,
i
- - - - Terzaghi's met,
~ ~176
0.8
i
.---9 observed value
CONMULT
with creep ~\
~ .....
.
c 1.2 E l.n
1.6
2.0 - ~tja 0
:: ..-.. ~.-..: ..-.;..:..:.,~,
120
240
360
480
,
,
600 720 840 Time. (days)
Fig. 5.29. Measured and calculated settlement of organic subsoil at the Antoniny site.
Consolidation Analysis 5.4.5
Layered
213
soils
Terzaghi's equation can be used in layered soils with a modified calculation procedure. In this case, it is not possible to solve the equations analytically and numerical methods have to be used, Helenelund (1951) proposed the following graphical method: The soil is divided into layers in such way that the coefficient of consolidation c v within each layer may be assumed constant. For each layer the boundary conditions must be fulfilled. Each layer is assumed to have a thickness Ax. The pore pressure is assumed to be u at time t and u" at time (t+At). With designations according to Fig.5.30, the consolidation equation can be written:
(U'i-U i )/At = c v ( 1 / A x ) [ ( U i + l - U i ) / A x - (u i -ui. 1 ) / A x ] =
(5.36)
Cv [(Ui+l + ui.1 - 2u i )/(Ax) 2]
,=
QI t_
tn (b i_
13.
I I I
I
i-1
0 I..
, t!
L
I
i.'~l I t*txt
I I I I I
ui-1 !
0(,J x LU
i I &X I
&X
i
I
&X
Vertical coordinate, x Fig. 5.30.
i*1
I
,,.._ v
Designations for graphical solution of the pore pressure dissipation in one-dimensional consolidation, ( H e l e n e l u n d 1951)
By selecting c v At/(Ax) 2 = AT v it is thus possible to proceed from the pore pressure at time t = 0 and step-wise calculate the remaining excess pore pressure after a certain time for consolidation and thereby estimate the consolidation process for the entire soil strata. With consideration to required accuracy and simplicity, a value of AT v = 1/4 is usually selected for calculation by hand.
Analysis of subsoil deformations
214
Then u" i = 1/4 (Ui+l+Ui.1) + 1/2 u i =1/2 [(l/2)(Ui+l+Ui ) + (l/2)(Ui+Ui_ 1 )]
(5.37)
Hence u i arithmetic mean of the pore pressures at the boundaries towards the surrounding layers. Example: A homogeneous clay layer. Load increase = q. Select t = 1 year = 3.15 9107 s and c v = 10 8 mVs AT v = 1/4 gives Ax=~]10 "8 x 3.15 x 107/(1/4) = 1.12 m The graphical construction is made according to Fig. 5.31.
At :0 U=Uo
I I I o a,X--t12m~ Fig. 5.31.
a•
AX
ere
Example of a graphical solution according to Helenelund's method when c v is constant, (After Hansbo 1984).
When both the % and the k-values vary, the thickness zXxi of a layer with coefficient of consolidation %i is determined by the selected time At according to Ax i = ~/cViAt/AT v . The thickness of the layer is thus directly proportional to the square root of the coefficient of consolidation. Furthermore, at both sides and in the vicinity of the boundary between two layers, the demand for continuity of the velocity of the water flow must be met,
ki (~U/~x)i = ki+l (~U/~x)i+l
(5.38)
Consolidation Analysis
215
How this can be solved graphically is shown in Fig. 5.32, where Cvl/Cv2 = 1.8, kl/k2 = 2, Cv2/Cv3 = 2.25 and k2/k3 = 1.5 etc. The graphical construction requires extra construction lines to determine the inclination of the pore pressure isochrone between the midpoints of the layers. At =0. U=Uo I I I I
(J
I
I I I I I
.c_ 0 i..
I I
r-~
az l:2ax 9
klCvl
az2=l.5ax ,
az3=ax.,
k2cv2
, k3cv3 ,
Fig. 5.32. E x a m p l e
of a numerical
solution
when
ax
. k3cv3
,,,~r
etc
B
c v and k vary (Hansbo
1984).
From the variation of c v and k we obtain Ax2=~/2.25. Ax3= 1.5 Ax3 and mx 1~ 118"Axe= 2.0-Ax3 etc. In the construction, tan c~ denotes (3u/3x)l and tan ~=(3u/3x)2 at time At. By using the extra construction k 1 tan c~ = k 2 tan 13 is obtained and hence the boundary conditions between the layers are fulfilled. In this graphical method, it is also possible to account for the variation ofk and c v during the consolidation process by letting Ax vary so that at all times AT v = [At/(Ax) 21 (kM/g Pw)=1/4
(5.39)
After each step, the increase in effective stress and the relative compression for each layer is calculated. New values of moduli, permeabilities and coefficients of consolidation are estimated and the whole situation may be reevaluated with new geometries of the problem, taking into account changes in the loading situation and associated changes in pore pressures that have occurred during the load step. The calculations in the next step then proceed from this new situation. Performing this kind of step by step calculation by hand is very time-consuming and therefore the number of steps that can be made and the number of variations in
216
Analysis of subsoil deformations
parameters and boundary conditions that can be taken into account is limited. Therefore, the calculations should preferably be carried out in a computer, where any number of steps and variations can be included. Such calculation programmes have been developed by Magnan etal. (1979) and Mesri and Choi (1985). The CONMULT-programme developed by Magnan et al. (1979) has been revised at SGI to take new models of soil compressibility and empirical observations into account (Larsson 1986). In these types ofprogrammes, also creep effects occurring during and after the dissipation of excess pore pressure can be taken into account. In the SGI model, this is done by modifying the Terzaghi equation to M
~)u
C)
gPw ~x
~t
C)Uet
~U
( k - - ) - ~ ~ x ~t
(5.40)
where uet is the additional excess pore pressure that develops due to the creep effects. The programmes can also be used to predict swelling at unloading and secondary swelling or compression thereafter. The effect of partially unsaturated soil can also be taken into account. The CONMULT-programme and the SGI model has also been used to predict the deformation process under the test embankment constructed at the Antoniny site. Results from these calculations are shown in Fig. 5.33.
ui
3
STAGE STAGE
Q 100
2
V
2O0
~00
3oo
~
E o Ii
LI.I ~
600 TINE.days Settlement in calcareous sod (3.0-7.8m)
~
fir)
500
:~
Marker at
3.0m
LLI
I~0.
Z_ ~.. 0.8
~
~
~
' - ~
-
~
Settlement in peaty soil ( 0 -3.0 m }
Z W ._J I-- 1.2 pW ffl 1.
9
"" --- ....~ ~
Measured
....
Calculated
without
-----
Calculated
with
. . . . . . . .
Total settlements
creep
creep
Fig. 5.33. Measured and calculated settlements for the test embankment at the Antoniny site (Larsson 1986).
Consolidation Analysis
217
The application of large-strain theory to predict consolidation in layered subsoil requires taking into account the boundary problems which appear at the level between compressible layers by means of common imagined boundary with imaginary mesh points (Cargill, 1982, Szymanski and Lechowicz, 1987). This iterative numerical procedure is complicated in design calculations of consolidation course in soil subsoil which consist of several thin layers. The adaption of more sophisticated method without the necessity of solving boundary problems between compressible layers in the numerical calculation for layered subsoil can be done by using the implicit scheme in finite consolidation analysis. Then, the goveming equation (5.32) can be written as:
0c
=0
ocj
(5.41)
where (l+e)k C=~ Yw
dS' (5.42) de
According to this approach the ~-th plane is subdivided into sets of rectangles of the sides A~j and At (Fig. 5.34). The representative mesh point P described by coordinates (~i, t,) can be expressed as" i ~i-- j~l ~ J
(5.43)
and tn =nAt
(5.44)
where: i and n are integers. Denoting the value of u at point P by Up = U(~i, tn) = Ui,n
(5.45)
218
Analysis of subsoil deformations
[111
/
.~mesh points
surface
soit tayer III
P(T t n) :
Ti
soit layer IT soit
tayer I
f =ot = 0
t n =n• t
t
Fig. 5.34. Graphical scheme of the subsoil discretisation in one-dimensional state.
the original differential problem (eq. 5.41) can be approximated by: Ui+ 1,n+ 1 - Ui,n+ 1 Ci + 89 " Ui,n+l - Ui,n
Ui,n+ 1 - Ui- 1,n+ 1 - Ci- 89
Qi+l,n - ~i,n
At
w h e r e : n = 0 , 1 , .... N a n d i = l ,
~i,n- ~i-l,n
(5.46)
~i+%,n - ~i- 89
2, .... I
Numerical soluitions of this equation from Ui, n to Ui,n+ 1 involves three unknown v a l u e s ofUi,n+1. Thus equation 5.41 defines Ui,n+ 1 implicitly and a tridiagonal system of simultaneous equation must be solved at each time step with using a straightforward Gaussian elimination method. The equation (5.41) has to satisfy:
Consolidation Analysis
219
9 the initial condition (5.47)
u(~,t = O) = O; for all 9 and the boundary conditions
U(~max, t) -- O;
for all t
u(~ = 0,t) - 0 or
(5.48) (5.49)
~u (~ = O,t) =0;
for all t
(5.50)
These conditions are approximated and added to equation 5.46 and next are solved together. During simulation the value of u(~i,t~) is modified according to loading schedule: Ui, n "= Ui, n + m~ n
(5.51)
where AS, = increase in total stress in subsoil :=
= substitution operator
The variation in value o f A~i during consolidation process is described by relation 1 + ei,n+1 (5.52)
A~i,n+l = A~i,n 1 +ei, n
The total thickness of compressible subsoil for each t is calculated as sum of all Two and three - dimensional water flow cannot be neglected in cases where the width of the loaded area is small compared to the thickness of the compressible soil. The problem becomes mathematically complex, but can be solved by finite element calculations, provided that an accurate soil model can be created. The consolidation equation can generally be written Ou/c3t- c v (c~2u/Oz2) + c h (OZu/Ox2+OZu/Oy2)
(5.53)
220
Analysis of subsoil deformations
If the consolidation process in the soil has been calculated under the simple assumption that the load has been applied instantaneously (t=0, broken line Cl) the consolidation process for a gradually applied load (solid line C) can then be constructed according to Fig. 5.35. Also in the case of partly unsaturated soil, the consolidation curve calculated under the assumption of full saturation may be graphically corrected, as shown in Fig. 5.36. If the soil is loaded on the surface, the increasing pore pressure leads to an instant compression of the gas bubbles (Boyle's law) and thereby an initial settlement corresponding to an apparent consolidation. As the pore pressure decreases, the gas bubbles grow in volume to resume their original size after complete dissipation of excess pore pressures. If the degree of consolidation is judged by the settlements only, there is thus an apparently faster consolidation process (broken line in Fig. 5.36).
/•ql o
Time, t
t2 -1/2 tl
v
I
100
Fig. 5.35.
Graphical construction of the consolidation process at gradual application of the load according to Terzaghi. A
0'
Fig. 5.36.
t
Time, t
Graphical construction of the consolidation process for partly unsaturated soil according to Terzaghi. U0-apparent consolidation at time t=-o due to instant compression of the gas in the pores.
Consolidation Analysis
221
The application of large-strain theory to predict the settlement of loaded subsoil and excess pore pressure dissipation in two-dimensional state requires numerical solution of diffemtial equation based on finite difference scheme (Fig. 5.37), (Szymanski, 1994). In this case the goveming equation can be expressed according to Samarski (1977) procedure as:
EHBANKHENT
-%
u= 0
t
"I"
T
~
"IF"
"11"
"1-
+ -+- ~- -+- -!- + ~
+
+
+
+
+
"I"
PCf~ii. ~hijJ
+ -I
[oyer 1
-I- + -!
0 tt r r o o
[oyer 2
4"+++-t-++-I
§ § +-t
++
u=O or
r
+ ~ ~4 = blU-Uol
Fig. 5.37. Subsoil discretisation and boundary conditions in two-dimensional state.
~u = Lu
(5.54)
~t
where 1
de
l+e
do'
(5.55)
Lu = (L;v + L;h ) u
L~v =
a;v ~9
L~h = oq~ h
I
(5.56)
(5.57) a;v (5.58)
Analysis of subsoil deformations
222
kv (5.59) %
13 = ~ %
(5.60)
Defining the value of u at time t as: u " = u(t--nAt) and A" is the differential operators of operator L, the governing equation can be written as"
E~ n-
A2t A ; f ) v = L n u n
At n" 2 A;h~)ut=V an
(5.61)
U n 1 = U n + m tUt
where E = unitary operator A;v and A;h = differential operators v and u t = auxiliary variables f~ = factor L = operator The differential operators A;v and A;h were expressed as" Ui+l, j - Ui,j 0~i +0.5, j "
A;vu =
A;vij
Uij - Ui.1, j " ~i-0.5, j"
O.5(A~vi.lj
+
A vi. j A~vi,j)
Ui, j+l - Ui,j ~i, j+0.5 "
(5.62)
Ui,j - Ui, j-1 " ~i, j-0.5 "
1
A~h u = 0.5(A~hi, j + A~hij. 1)
(5.63)
223
C o n s o l i d a t i o n A n a l y s i s - Vertical D r a i n s
where - discrete area ofU(~v,~h )
uij
A~hij = distance between nodal points in horizontal direction A~vi,j = distance between nodal points in vertical direction
~i+0.5j = 1//z(~ij + ~i+lj)
5.5
CONSOLIDATION ANALYSIS OF SUBSOIL WITH VERTICAL DRAINS
An embankment construction by stages or with surcharging may require an acceleration of the consolidation by means of vertical drains (see Chapter 9). The consolidation analysis is then based on vertical and axi-symmetrical pore water flow. The consolidation process in axi-symmetrical state can be calculated with the methods presented by Barron (1948) and Hansbo(1972, 1979). In these methods, linear stress-strain relationships and constant soil parameters are assumed in the deformation process, as well as constant stress during the consolidation process and constant vertical surface displacements throughout the drained area (Jamiolkowski et al 1983). The effect of secondary compression is also neglected. With these assumptions, the equation of consolidation becomes: 1 OH
~2u cv
+ ch (
c)x2 where cv = ch = u = x = P = t =
~2U +~ )
p ~)p
c)192
OU =~
(5.64) ~t
coefficient of vertical consolidation coefficient of horizontal consolidation excess pore pressure vertical space coordinate radial space coordinate time.
If the degree of consolidation, U, is defined as the degree of dissipation of the excess pore pressure it can, according to Carillo (1942), be expressed as:
u=u + Uv-UUv
(5.65)
Analysis of subsoil deformations
224 where U = degree of total consolidation Uh= degree of horizontal consolidation U v = degree of vertical consolidation.
The degree of vertical consolidation U v is calculated according to Terzaghi's theory (Chapter 5.4) and the degree of horizontal consolidation, U h is calculated from (Barron 1948) Uh= 1 - exp(-8Th/P) where Th =
Th =
Ch
(5.66)
time factor
t/D 2
n2
~t-
3 1 1 (ln n - ~ -~. . . . ) n2-1 4 n2 4n 2
(5.67)
D n = ----
(5.68)
(5.69)
where D = diameter of dewatered soil cylinder (see Chapter 9.3.3) d = drain diameter The degree of horizontal consolidation, U h , can be estimated from the relationship between U h and T h represented by curves in Fig. 5.38. Equation (5.66) by which the degree of horizontal consolidation is defined, was obtained without regard to the disturbance of the soil around the vertical drain and the well resistance. However, in the course of installing drams, a disturbance of the soil structure occurs, which causes a decrease in modulus and may change the permeability in the immediate vicinity of the vertical drain (smearing effect). Furthermore, the drain discharge capacity is limited, and the well resistance may slow down the consolidation process.
Consolidation Analysis
-
225
Vertical Drains
J
I
I
I
I
I
I
n=lO0 c-
n= 40
0.20
/n=20
E o 0.40
n:15
0 ..... ....,
0
n=7
0.60
u
0
0.80
I,_
0
1.00
0.001
I"1"
D d
n=5
r.= C":, t 0.01
0.04
0.10
0.40
1.0
Time factor, T h Fig. 5.38. Relationship between time factor T h and horizontal degree of consolidation U h.
The smearing effect and the well resistance were taken into account for the first time by Barron. Hansbo (1981) has presented a recommendation for taking these factors into account. These methods are presented in detail in Chapter 9. An analysis of the computed values of the consolidation rate in an organic subsoil (Fig. 5.39) shows that the results of the calculations are only insignificantly influenced if the smearing effect and the well resistance qw are taken into account (Koda et al. 1989). The insignificant influence is the result ofthe small extent ofthe disturbance zone, the high discharge capacity of the vertical drains and the considerable variability of the permeability coefficient during the consolidation process. The changes in the soil parameters and the effect of secondary compression in the deformation process have greater influence on the results than the decrease of permeability caused by the smearing effect and well resistance. The variability of the consolidation parameters in the deformation process was included in Schiffman's method (1980). However, in this method, the secondary creep of the soil skeleton under constant effective stress is disregarded. The consolidation analysis of a deep soft subsoil requires that not only the variability of parameters but also the large strains in the soil be taken into account (Wolski et al. 1987). In this case, the consolidation equation in one- dimensional state of strain and axl-symmetrical pore water flow into a convective coordinate system becomes:
Analysis of subsoil deformations
226
1
1
I
I
I
I
-
peat
O.l~
s~nd 0.8
E Ic
9 1.2
E
1.6
Barron-Hansbo method [--.--- without smear effect [ and constant qw
2~!--'--/
/
~
changing qw
with smear effect
I
t
2/.0
0
1.
.I
I
/..80
|
720 Time, (days)
Fig. 5.39. Calculated settlement of organic subsoil with vertical drains at the Antoniny site.
~)
k v 3u
oq
ka u
+ --~--
~ where u = e = Yw = = 9 =
7w ~9~
k a oqu + ~
7w P
excess pore pressure void ratio unit weight of water convective coordinate radial coordinate
7w ~)P
1
de Du
+
=0 (l+e) dcy' Dt
(5.70)
SwellingAnalysis
227
The numerical solution of the consolidation equation can be made with a finite difference scheme with application of the piece-wise linear or finite-strain approaches. The computation procedure requires knowledge of the specific gravity of solids and the relation between void ratio and effective stress and void ratio and the vertical and horizontal permeabilities estimated from laboratory tests (Fig. 5.40).
Permeability, 0.1 1 ,, • 6 o e=f(kv) o1' x x e=f(6')
-
9 e=flkll
b)
o I..
x x
O O
o
3
X
;
~.eu 9 9
e t
X O UX OeO 9 X.x x x 9 / o~
..I
5.6
O
x
100 0
9
'
OC
9
Xx x
e
"10
Fig. 5.40.
o ~
Ol :) " ) c) xXx PEAT . ~ ~ ~ oxik'e' 9
4
o"
k (x10-9 m/s) 10
oOoO GO
(: 3 ;: X x x
O
1
areous soil 10 Effective
I 100
stress, 6"' (kPa)
1000
Example of calculation characteristics for consolidation analysis of a subsoil with vertical drains in terms of large strains. (a) Variation of vertical permeability. (b) Variation of void ratio. (c) Variation of horizontal permeability.
SWELLING
ANALYSIS
When the settlement of an embankment must be finished at the end of the construction period, surcharging is usually applied (Chapter 9). In this case, a temporary preloading with a load in excess of the permanent fill is used. The surcharge is then removed in order to unload the subsoil. If the effective stress m the soil is decreased sufficiently, the soil starts to swell. When the subsoil has consolidated for a load greater than the previous preconsolidation pressure, the secondary compression will proceed at a relatively high rate in organic soils. However, a small reduction in the load causes a halt of this compression for a short period or a small swelling before compression starts again, but at a reduced rate. If, on the other hand, the load reduction is large enough, compression
Analysis of subsoil deformations
228
stops completely and the soil may begin to swell with time. The procedure for determining the magnitude of the surcharge required to stop or reduce further compression and the time of surcharge removal is presented in detail in Chapter 9. Design of embankments on soft soil with application of surcharge comprises, besides deformation and consolidation analyses, an evaluation of the deformation properties at unloading. The total relative increase in subsoil thickness (swelling) ~w can be calculated from (Larsson 1986): e,w = C,e log((b Cyp')/cru')
(5.71)
where C~e= modified swelling index (C~e = C/( 1+e)) b = load factor when swelling overcomes secondary compression Crp = preconsolidation pressure or'= effective pressure after pore pressure dissipation Application of this equation to the prediction of subsoil swelling S~w (S~w = e,w" Hp, where lip = subsoil thickness at the time of surcharge removal) requires estimation of parameters C,e and b from oedometer test or use of empirical correlations with the plasticity index Ip (Fig 5.41). Larsson (1986) found that in soft soils the swelling index C e varies between 0.007 and 0.012 and the load factor b is about 0.8. Also in organic soils, the load factor appears to be approximately 0.8, (Ingan~ts 1978). a) 1.5
~5
s
.pJ
1.0'
s
s
s
f
S
+
ss.O"
§ '
/
/
,
o Fig. 5.41.
1.0,
"" "U ....
5o 160 "Plasticity index, Ip (%)
q.._
I--
0
u 0.5
+ .'"
0.5'
d3
i
b)
%
o Clay 9Coarser soil ,,
50
Plasticity index, Ip
lOO
(%)
Swelling parameters versus plasticity index Ip. (a) Modified swelling index Cs,. (b) Load factor b. (Larsson, 1986).
Swelling Analysis
229
Swelling, like compression, is time-dependent. At unloading, the pore pressure in the subsoil drops. The rate of swelling depends on excess negative pore water pressure dissipation, related to the natural pore pressure. This primary swelling may or may not be followed by secondary swelling, mainly depending on the degree of load reduction (Mesri et al. 1978, Juarez-Badillo 1988). If the unloading is small, a seeondary compression takes place after the pore pressure equalization. As various layers in a soil profile may have different degrees of consolidation, the effect of unloading o~en becomes uneven and some layers may swell at the same time as other layers undergo compression. The time-swelling-compression relation can be calculated according to classical Terzaghi theory (1931) supplemented with analysis of secondary compression taking into account swelling index C~, coefficient of secondary swelling C~ and coefficient of secondary compression Ca. In these calculations, it should be observed that C~ varies with the deformations. C a has considerably lower values in soils that have been unloaded and then swelled. If the compressions return to the values that were reached before unloading, the coefficient C~ also resumes its original value. Fig. 5.42.
Surcharge ._.1
Finat toad v
tog time
f(Cs,k) L/) t-
-I,-,
f(C s~) ~~V
SS r
\
f(Co~)
Fig. 5.42. Scheme for calculating the vertical movement due to temporary surcharging.
Analysis of subsoil deformations
230
5.7
DEVELOPMENT TRENDS IN DEFORMATION AND CONSOLIDATION ANALYSIS
In the construction of embankments on deep soft subsoils, especially narrow embankments such as dykes, it is often useful to extend the deformation and consolidation prediction to a plane strain state and two-dimensional water flow. In these cases, when large horizontal displacements may occur, the deformation prediction can be carried out with a two-dimensional consolidation analysis based on a constitutive soil model. One such method is the application of Biot's consolidation theory, in which the elastic properties for the soil skeleton as well as Darcy's law for liquid flow are assumed (Biot, 1941; Verruit, 1977). Based on the assumptions in Biot's theory ,the numerical solution by means of a finite element method is relatively simple. Design calculations can be performed using the piece-wise linear approach with simplified stress-strain relation. This method is chosen because of the possibility of determining parameters necessary for the calculations and the facility of computation methods. The numerical calculations can be performed with the following equations: 9 piece-wise linear stress-strain relation for the soil skeleton, ~'ij = G (W i ,j + Wj, i) + (K- 2/3"G) Wk, k 8ij where cr'.j = Wi,j = Wj = 8ij =
(5.72)
effective stress tensor; component of the displacement gradient; displacement vector; Kronecker'sdelta,
K=E/(3(1-2v))
(5.73)
and G=E/(2(I+v)) where E = modulus of elasticity; v = Poisson's ratio.
(5.74)
Development in Deformation and Consolidation Analysis
231
9 Darcy's law for the pore water flow, qj = k u , j
(5.75)
where: qj = component of the specific discharge vector of pore water u,j = gradient of pore water pressure k = coefficient of permeability 9 Terzaghi's formula, (5.76)
(y~j= cy'~j + uSij where ~j = total stress tensor, 9 the de Josselin de Jong storage equation, ()8o
Ou = n[3
~)t
- qi,j
(5.77)
~t
where ~o = volume strain n = porosity [3 = compressibility of the pore water. Results obtained with this method in an organic subsoil under an embankment are shown in Fig. 5.43. The application of more precise soil models to consolidation prediction in organic subsoils makes it possible to include non-linear elastic or elasto- plastic soil characteristics (Szymanski et al 1991). If the load is small, the non-linear Duncan-Chang model can be used with parameters obtained m triaxial compression tests (Duncan, 1980). For plane strain conditions, the hyperbolic stress-strain relation can be modelled by a series of linear increments. The relation is usually expressed in terms of Young's modulus E and bulk modulus K as follows:
Analysis of subsoil deformations
232
10Sl i a)
II
. 9. i .
b) '0 i
: ~-
"
9 o
"""" "'" '
~
gyttjo
o. . . . . . . . . . .
-
' I
'
. . . . . $ond. . . . . . . . . . . . . . . . . . . calcutated - - - measured
----- catcutated ^u..... measured
.
.
.
.
.
.
.
.
I
t
I .
.
.
I .
.
.
Fig. 5.43. Measured and calculated results of consolidation in organic subsoil at the Antoniny site. (a) Excess pore pressure. (b) Displacements.
A%]
3K
A'Cxy
9K-E
I
(3K+E) (3K-E)0 [Aex ] (3K-E) (3K+E) 0 lacy ]
o
o
(5.78)
E[A%J
where A G x A~y and A1:xy= stress increments Ae~ Aey and AV~y = strain increments The non-linear character of the stress-strain relation can be applied in equation (5.78) by varying the values of Young's modulus E and bulk modulus K according to the stress variation during the consolidation process. An elasto-plastic Cam-Clay model may also be used (Wroth and Houlsby, 1980). The total strain increment is then calculated as the sum of two components: Aaij = aij (e) + aij (P) where eli(e) = elastic strain increment aij(P) = plastic strain increment
(5.79)
233
Development in Deformation and Consolidation Analysis
For computation of the plastic deformation, the evaluation of a yield envelope and an associated flow rule is necessary. The modified Cam-Clay model yield surface is given by the ellipse: q~ - M~(pop - p~) = 0
(5.80)
where q = deviatoric stress M = constant parameter po = equivalent isotropic pressure p = mean normal stress. Laboratory tests performed on soft mineral soils (Tavenas and Leroueil, 1977; Larsson and S~tllfors, 1981) show that the shape of the yield envelope for natural soils differs greatly from the Cam-Clay model. Even if the yield envelope appears to have a more or less elliptical shape, this ellipse is not centered around the isotropic line but rather around a line close to the K0-1ine (K 0 - the coefficient of earth pressure). Investigations on organic soils performed by Lechowicz and Szymanski, (1988) show that, as for soft clay, the shape and position of the yield envelope vary during consolidation, depending on the magnitude and the direction of the principal stresses, and are centered around the changing K0-1ine (Chapter 3.5). In calculations of deformation performance in organic soils, an anisotropic model should therefore be used. Such models and calculation methods have been developed by e.g. Runesson (1978) and Magnan (1987).
5.8 REFERENCES Asaoka, A. (1978). Observational procedure of settlement prediction. Soils and Foundations, Vol. 18, No.4, pp.87-101. Azzouz, A.S., Krizek, R.J. and Corotis, P.B. (1976). Regression analysis of soil compressibility. Soils and Foundations, Vol. 16, No. 2, pp. 19-29. Berry, P.L. and Poskitt, T.J. (1972). The consolidation of peat. Geotechnique, Vol. 22, No 1. pp. 27-52 Barron, R.A. (1948). Consolidation of fine-grained soils by drain wells. Trans. ASCE Vol. 113, pp. 718-754 Biot, A.W. (1941). General theory of three-dimensional consolidation, Journal of Applied Physics, No. 12.
234
Analysis of subsoil deformations
Cargill, K.W. (1982). Consolidation of soft layers by finite strain analysis. U.S. Army Engineer Waterways Experiment Station. Miscellaneous paper GL-82-3, Vicksburg, MS. Carlsten, P. (1988). The use of preloading when building roads on peat. Proceedings of the 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn. pp. 135-143. Carillo, N. (1942). Simple two and three dimensional cases in the theory of consolidation of soils. Journal of Mathematics and Physics, Vol. 21, No. 1. Drozd, P.A. and Zajac, W.N. (1968). Razcziot osadld nasypiej na bolotach, Gidrotechnika i Melioracja, No 3. Duncan, J.M. (1980). Hyperbolic stress-strata relationships, Proc. of the Workshop on Limit Equilibrium, Plasticity and Generalized Stress- Strain in Geotechnical Engineering, Montreal, Canada, pp. 443-460. Edii, T.B. and Dhowian A.W. (1979). Analysis of long-term compression of peats. Geotechnical Engineering, Vol. 10, No. 2, pp. 153-178. Fadum, R.E. (1948). Influence values for estimating stress in elastic foundations. Proceedings of the 2nd International Conference on Soil Mechanics and Foundation Engineering, Vol.3, pp. 77-84. Flaate, K. (1968). Setninger i torvjordarter. Innlegg ved NVF -Konferansen, Voksenaasen. Foott, R. and Ladd, C.C. (1981). Undrained settlement ofplastic and organic clays. Journal of the Geotechnical Engineering Division, ASCE. Vol. 107, No GT8. pp. 1079-1094. Gibson, R.E., Schiffman R.L. and Cargill K.W. (1981). The theory of one dimensional consolidation of saturated clays, Part II. Finite non-linear consolidation of thick homogeneous layers, Canadian Geotechnical Journal, Vol. 18, No. 2, pp. 280-293. Gray, H. (1936). Stress distribution in elastic solids, Proceedings of the 1st International Conference on Soil Mechanics and Foundation Engineering, Vol.2, p. 151. Hansbo, S. (1979). Consolidation of clay by band-shaped prefabricated drains. Ground Engineering, Vol. 12, No. 5. Hansbo, S. (1981). Consolidation of fine - grained soils by prefabricated drains. Proceedings of the 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol.3, pp. 677-682. Hansbo, S. (1984). Deformationer och s~tttningar. Handboken BYGG, Geoteknik. Liber Frrlag. Stockholm.
References
235
Hanrahan, E.T. and Rogers, M.G. (1981). Roads on peat. Observation and design. ASCE. Journal of the Geotecnical Engineering Division, Vol. 107. No GT 10, pp. 1234 - 1265. Helenelund, K.V. (1951). Om konsolidering och s~ittningar av belastade marldager. J~irv~igsstyrelsens geotekniska sektion. Meddelande 3. Helsingfors. Hobbs, N.B. (1986). Mire morphology and the properties and behaviour of some British and foreign peats. Quaternary Journal of Engineering Geology, Vol. 19, pp. 7-80. Ingan~is, J. (1978). Vertikaldrgnering i organisk jord, Delrapport, Dnr. 1- 187/77, Swedish Geotechnical Institute, Link6ping. Jamiolkowski, M., Lancellotta R. and Wolski, W. (1983). Precompression and speeding up consolidation, S.O.A. and General Report, Proceedings of the 8th European Conference on Soil Mechanics and Foundation Engineering, Helsinki, Vol. 3, pp. 1201-1226. Janbu, N., Bjerrum, L. and Kjaernsli, B. (1964). Veiledning ved losning av fundamenteringsoppgaver, Report NGI, No 16, Oslo. Juarez-Badillo, E. (1988). Postsurcharge secondary compression equation for clays. Canadian Geotechnical Journal, Vol. 25, No. 3, pp. 594-599. Koda, G., Szymanski, A and Wolski, W. (1988). Behaviour ofgeodrains in organic subsoil. Proceedings of the 12th International Conference on Soil Mechanics and Foundation Engineering, Rio de Janeiro. Vol. 2., pp 1377-1380. Kogure, K., Yamaguchi H., Ohira, Y. and Ishioroshi, H. (1986). Physical and engineering properties of a peat ground, Proc. Advances in Peatlands Engineering, Ottawa, pp. 95-100. Landva, A.O. and La Rochelle P. (1982). Compressibility and shear characteristics of Radforth peats. ASTM Special Technical Publication 820, Toronto. pp. 157-191. Larsson, R. (1981). Drained behaviour of Swedish clays. Swedish Geotechnical Institute, Report No. 12, Link6ping. Larsson, R. and S~illfors, G. (1981). Hypothetical yield envelope at stress rotation. Proceedings of the 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol.1, pp. 693-696. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Report No. 29, Link6ping. Larsson, R. (1990). Behaviour of organic clay and gyttja. Swedish Geotechnical Institute, Report No. 38, Link6ping.
236
Analysis of subsoil deformations
Lechowicz, Z. and Szymanski, A. (1988). Deformation analyses of organic subsoil in anisotropic stress conditions, Archiwum Hydrotechniki. Vol XXXV, pp. 125133. Magnan, J.P., Baghery, M. and Tavenas, F. (1979).Etude num6rique de la consolidation unidimensionnelle en tenant compte des variations de la perm6abilit6 et de la compressibilit6 du sol, du fluage et de la non-saturation. Laboratoires des Ponts et Chauss6es. Bulletin de liaison. No 103 pp. 83-94. Magnan, J. P. (1987). Prediction and behaviour of embankments - settlements and improvements. State of the Art Report. Session 2A International Symposium on Soft Soils. Mexico City.
Mesri, G. (1973). Coefficient of secondary compression. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 99, No SM 1. Mesri, G and Godlevski, P. M. (1977). Time- and stress-compressibility interrelationship. Journal of the Geotechnical Engineering Division, ASCE, Vol. 103, No. GT5, pp. 417 - 430. Mesri, G., Ullrich, C.R. and Choi, Y.K. (1978). The rate of swelling of overconsolidated clays subjected to unloading. Geotechnique, Vol. 28, No.3, pp. 281307. Mesri, G. and Choi, Y.K. (1985). Settlement Analysis of Embankments on Soft Clays. ASCE. Journal of the Geotechnical Engineering Division. Vol. 111, No.GT4, pp. 441-464. Niesche, H. (1977). Niektore uwagi o warunkach osiadania torfu. Sympozjum nt. Nasypy na gruntach organicznych, pp. 45-64. Ostromecki, J. (1956). Method and nomograms for evaluating the subsidence of peat soils under the influence of drainage. International Comm. on Irrigation and Drainage, 4th Congress. Osterberg, J.O. (1957), Influence values for vertical stresses in semi- infinite mass due to embankment loading, Proceedings of the 4th International Conference on Soil Mechanics and Foundation Engineering, London. Vol. 1, p. 393. Ozden, Z.S. and Wilson, N.E. (1970). Shear strength characteristics and structure of organic soils. Proceedings of the 13th Muskeg Reseearch Conference, Fredericton, New Brunswick, National Research Council, Technical Memorandum No. 99, Ottawa. Poulus, H.G. (1972). Difficulties in prediction of horizontal deformations of foundations. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 98, No. SM 8, pp. 843-848.
References
237
Runesson, K. (1978) On non-linear consolidation of soft clay. Chalmers Umversity of Technology. Department of Structural Mechanics. Publ. 78:1. Thesis. Gothenburg. Samarski, A.A. (1977). Tieorja raznostnch schiem. Nauka, Moskwa. Schiffman, R.L. and Gibson R.E. (1964), Consolidation of nonhomogeneous clay layers. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 90, No. SM5, pp. 1-30. Schiffman, R.L. (1980). Finite and infinitesimal strain consolidation. Journal of the Geotechnical Engineering Division, ASCE. Vol. 106, No. GT 2, pp. 115-119. Steinbrenner, W. (1934). Tafeln zur Setzungberechnung. Die Strasse. Vol. 1. p. 121. Szymanski, A. and lechowicz, Z. (1987). Numeryczna prognoza konsolidacji uwarstwionego podloza slabonosnego. Materialy na konferencje "Komputery w geotechnice", Rydzyna. pp. 66-73. Szymanski, A. (1991). The factors determining the deformation analysis of organic subsoil under embankment. Warsaw Agricultural University. Szymanski, A., Wolski, W. and Kr61, W. (1991). Two-dimensional consolidation analysis of organic subsoil in terms of large strain. Proceedings of the 9th European Conference on Soil Mechanics and Foundation Engineering, Florence, pp. 273-276. Szymanski, A. (1994). The use of constituive models in consolidation analysis of organic subsoil under embankment. Proc. of the Intern. Workshop on Advances in Understanding and Modelling the mechanical behaviour of Peat. Balkema, Rotterdam. pp. 231-240. Tavenas, F. and Leroueil, S. (1977). Effects of stresses and time on yielding of clays. Proceedings of the 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, Vol. 1, pp. 319-326. Terzaghi, K. (1924). Die theorie der hydrodynamistischen spannungs- erschemungen und ihr erdbautechnisches answendungsgebeit, Proceedings of 1st International Congress of Applied Mechanics Vol.1, Delft, Netherlands, pp. 288- 294. Terzaghi, K. (1931). The influence of elasticity and permeability on the swelling of two-phase systems. Colloid Chemistry, ed. J.Alexander, Chemical Catalog Co., New York pp. 65-88. Wolski, W., Szymanski A. and Koda G. (1987). Large deformation analysis of organic subsoil with vertical drams. 4 Colloque Franco-Polonais de M6canique des Sols Appliqu6e, Grenoble, Vol. 1, pp. 565-576.
238
Analysis of subsoil deformations
Woiski, W., Szymanski, A., Mirecki, J., Lechowicz, Z., Larsson, R., Hartl6n, J., Garbulewski, K. and Bergdahl U. (1988). Two stage constructed embankments on organic soils. Swedish Gcotechnical Institute, Report No. 32, LinkOping. Wolski, W. (1989). Some aspects of application of band-shaped vertical drains in organic soils. De Mello Volume, pp. 545-551, Editora Edgard Blticher LTDA, Sao Paolo. Wroth, C.E and Houlsby G.T. (1980). A critical state model for predicting the behaviour of clays. Proceedings of the Workshop on Limit Equilibrium, Plasticity and Generalized Stress-Strain in Geotechnical Engineering, Montreal, Canada, pp. 592-627. Verruit, A. (1977). Generation and dissipation of pore-water pressure. Finite elements in geomechanics. J. Wiley, New York. Yong, R.N. and Ludwig, C.A. (1984). Large-strain consolidation modelling of land subsidence. Symposium on Geotechnical Aspects of Mass and Materials Transportation, Bangkok, Thailand. Yong, R.N., Lechowicz Z., Szymanski A. and Wolski, W. (1988). Consolidation of organic subsoil in terms of large strains. Proceedings of the 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn, pp. 261-267.
23 9
PART11: Design and Construction Methods
240
Chapter 6
Methods of Construction J. Hartldn, Swedish Geotechnical Institute
6.1
GENERAL
Construction on soft organic soil gives rise to special problems. Most obvious are the large deformations that may occur during and atter the construction period, both vertically and horizontally. The settlements otten appear very quickly but may also continue for very long time periods due to creep. The low strength often causes stability problems, and consequently the load sometimes has to be placed in stages or, alternatively, the soil must be improved through prior treatment. It is important to realize that organic soil is not a single type of soil but a number of soils with very different behaviour. In one case, it may be appropriate to choose preloading, as for fibrous peat, while in another case this solution turns out to be inpracticable due to low permeability, as with organic clay. In the latter case, soil stabilization may be the most economical solution. Even for a special soil type, such as peat, the properties may vary immensely, as between fibrous and decomposed states. As in all building projects, the choice of construction method is a matter of finding the optimum solution. Economic and technical aspects must be considered. The solutions will of course vary according to demands on the standard, as well as available construction time and geotechnical conditions. The choice of method differs depending on the soil layers beneath the organic soil. If there are sot~ compressible layers underneath, their properties may have a large influence on settlements and stability. When dealing with stability and settlement problems there are several options to choose from, such as 9 adjusting the load from the embankment and traffic to the soil or transfering the load to a more stable level of soil: Load adjustment 9 replacing the soil, totally or partially, by better material: Soil replacement 9 improving the soil properties through some form of prior treatment: Soil im-
provement
Choice of Method
2 41
Preliminary investigations of the geotechnical properties of the soil in question provide a basis for the choice of construction method. These investigations must provide information on the depth of the organic soil, the type of organic soil and presence of compressible layers undemeath. A suitable construction method should be chosen in cooperation between the geotechnical engineer and other persons involved in the designing process. There after, a more detailed geotechnical investigation should be made. The basic conditions of the project vary according to several factors, for instance, whether a new road is to be constructed or an existing road is to be improved. In the latter case, consideration must be taken as to how the existing road is constructed. The design must integrate the longer time perspective so that savings during the construction period are not lost through higher maintenance costs.
6.2
CHOICE OF METHOD
It is always important to bear in mind that organic soil is not one unique type of soil but behaves quite differently through aspects such as its origin. The main difference in behaviour is the permeability which affects the rate of consolidation. A rapid consolidation results in a rapid strength increase, provided that the stability (safety factor) is high enough. Figs. 6.1 and 6.2 presents some commonly used methods and gives basic concepts for the different methods.
Methods of Construction
242
[ Load adjustment I Profile /--
.~f..
lowering
Pressure
preliminary~ decided J" level
9. - . . . . . . . - . ' . . . - . ' . . . . . - ~ , ~
-- ~ - ~----- . ': "--- .": ~ "-
. . . ~ ' : :2~J.:-a~ ~2 L .~,:. -"-".: ~ '
Light weight fill ...:.: ::::
~ O.....:-':" O
'
~:~
Compensation foundation
..:..... :.......: .,7,~
"b a no o'ooo 00~ 0
berms
~o:~
"...
"
"
0 0 0
~ ) ~-~ ~ . . ~ ~ 0 0 O
O
m
'
'
~
I Replacement I
Excavation ( shallow depth )
Progressive displacement (comb. with surcharge and sometimes blasting) /~ - S u r ~
"
~"-
"
-~,
........>~ !.::.i.
:.'" 9 :" ".'" ~':. .:-~:'~.::~:-[':.':~.r.:'t-'.:T.:,."
"./;.'/:
.;,~. :/.':.',~.."/"
-/'..-.;K."./'.:/'- ~.-'.f. :~I.'/:
"F..',~'.',~:
Fig. 6.1. Methods of construction; load adjustment, replacement.
:-:~::~.'./
243
Choice of Method
Stage construction
]
Preloading with vertical drains
Preloading / L ~
.
/ .
~ .
.
9 .
"" ~
.
9
.
.
.
.
Surchar.qe
". 9 " ' . ' . .
". ".
9
\
". ". i "."
.
\\
/
X
/
/--
.
.
/..''-
I Cotumns
.
9 9
.
.
.
.
.
.
.
Surcharge .......... o
~
.
9 .
.
.-.. .
.
.
.
\
: \\ 9
N.
.
]
Lime, cement or other material ,/- -- S"urc"h~rge" -- -~ \
/~ .,-
.
"
/././,
/.
/
/
.
o
,z/.
.
.
9
.
~,.,4 ~ ~
\
.
.L :i.
./, .i../..,~, .I
I
Reinforcement
Geotextile
..~"
.
9A 9
' Other methods
..
.
9
....i i..-..:....-.:-:~:~ ....
Timber
mat
~:i-:-.. :; i. ..! i..i..;. :.'~_
Piling with continuous concrete slab
" ."
9 9 9 "/-
V'.
V-
'/-/.
"/-/-
9
" / - " / . " / - "/. " / - " / -
/ .
"]
Fig. 6.2. Methods of construction; Stage construction, columns, other methods.
244
Methods of Construction
In Table 6.1 a crude division into suitable methods is suggested, related to different soil types. However, as each case is unique, it is always necessary to make sure that the method is applicable. The different methods are presented in the following chapters, together with the limitations. Table 6.1.
General applicability of construction methods to different kinds of soft subsoils. Limitations are presented in Chapters 7-11.
TYPE OF S O I L
GYTTJA
PEAT 03
o
03
METHOD
OF
CONSTRUCTION LOAD ADJUSTMENT Profile lowering Pressure berms Lightweight fill
,
0
GYTTJA-BEARING CLAY
SOFT CLAY
03
o ,.~ o
03
P P P P P P U UU
P P U
P P P
S S S
S S S
S
S
S
U P S
S
S
S
STAGE CONSTRUCTION Preloading S P U
P
U
U
P
S
S
U
P
S
P P P*
P P P*
P P S
REPLACEMENT Excavation shallow deposits Progressive displacement
Preloading + vertical drainage
P P P
LIME/CEMENT COLUMNS OTHER METHODS Corduroys Geotextile Embankment piling
P P P P P P U P* P*
S = suitable P = possible U = unsuitable - = not possible *) Embankment piling and pile-supported concrete deck.
Choice of Method
245
An example of current Swedish practice for road construction in areas with soft soils with organic content is described in Table 6.2, from which the importance of soil type is evident. Table 6.2.
Methods used for road construction on soils with organic content in Sweden,
Soil conditions
Construction methods
Remarks
Fibrous peat on firm soil layers
Preloading (stage construction) or excavation and backfill
Depending on road standard, thickness of peat layer and available construction time
Pseudo-fibrous and amorphous peat Dy and gyttja
Soil replacement (Excavation and backfill or progressive displacement)
Preloading in exceptional cases. Floating structure sometimes for low cost roads
Organic clay
Soil replacement, lime/cement columns or pile foundation (embankment piling and pile-supported concrete deck)
In exceptional cases vertical drainage
Soft clay
Leight-weight fill or compensation foundation replacement, preloading with or without vertical drains, lime/cement columns or pile foundation
Depending on thickness of clay layer
Recently superlight fill material, such as expanded polystyrene (EPS) has been used more and more frequently in road embankments in Sweden. The use of EPSmaterial in combination with preloading with conventional fill material sometimes proves to be the most favourable solution regarding technique and economy. Preloading and thin floating embankments reinforced/supported by geotextiles is a technique that is under rapid development.
246
Methods of Construction
6.3
REVIEW OF BASIC CONCEPTS OF EMBANKMENT CONSTRUCTION ON ORGANIC SOILS
6.3.1
Load adjustment
The optimum method of utilizing the natural soil properties is to adjust the load. The simplest, and often the most efficient way of adjusting the load is to lower the embankment height. Due to geometrical requirements, this simple method is not always applicable. Another way of adjusting the shear stresses induced in the soil is to reduce the inclination of the embankment slopes or to construct pressure berms. The use of pressure berms produces a more deep-seated critical slip surface. Pressure berms are therefore most efficient when the shear strength increases with depth or when the depth to firm layers is limited. The method requires more material and also more space along the embankment. Pressure berms solve the stability problem, but the settlements may still be troublesome. For this reason pressure berms may be less suitable for structures sensitive to settlements. The use of lightweight material in embankments can solve the stability problems as well as reduce the settlements. The lightweight fill must have low density, good beating capacity, high long-term durability, low compressibility and be nonpollutant in respect e.g. to groundwater. The material must also be easy to handle when constructing the embankment. In areas with organic soil the groundwater table often is very high, sometimes up to the ground level, but may be subjected to seasonal variations. Therefore, it is important that the embankment is not too light, otherwise it may be damaged by uplift forces. Some organic soils have a very low shear strength and high compressibility. The bearing capacity of such soils only allows very low embankment heights, and large settlements will occur in any case. Support of the embankment with piles may then be necessary. Usually the top layer is very weak and it is then necessary to construct a piled continuous concrete deck.
6.3.2
Soil replacement
Replacement of organic soil by non-cohesive soil or blasted rock can be carried out with excavation and backfill or byprogressive displacement. Excavation often leads to a better result while progressive displacement sometimes leads to costs for maintaining the embankment some years after completion. The use of excavation has its limitations due to economic and technical restraints when the excavations have to be deep.
Basic Concepts
247
Progressive displacement means that the soft subsoil is displaced by producing a continuous shear failure through the weight of the embankment. There is often a need for a surcharge to produce this shear failure. A surcharge also helps to speed up the settlement development in order to produce most of the settlements before the embankment is put into operation. The surcharge should in any case be untouched for up to 3-6 months. Sometimes the progressive displacement is complemented with excavations and blasting.
6.3.3
Soil improvement
Using stage construction increases the effective stress in the soil which will result in an increase in shear strength and stiffness. It is essential for the method that the soil have a sufficiently high permeability and/or that the compressible strata are sufficiently thin to permit dissipation of the excess porewater pressures within the available construction time. During stage construction it is important to make a careful follow-up of settlements. Information from settlement measurements can lead to a change of design.
Incremental loading with surcharge is one kind of stage construction. When the calculated consolidation has occurred, the surcharge is removed. With a correct design of the surcharge, it is possible to produce an embankment free from long-term settlements, even if the subsoil has a high compressibility. The method has proved to be very useful, especially on fibrous peat with a low degree of decomposition and with a total thickness of less than about 4-5 m. When the subsoil consists of amorphous peat, gyttja or gyttja-bearing clay, the method is not as useful because of the low permeability and consequently slow consolidation of these materials. The method also requires a careful follow-up. Embankments constructed by using incremental load and surcharge can be combined with vertical drains. The applicability of preloading depends on the type of organic soil. A fibrous peat has a high permeability so there may be no need for vertical drams. When the subsoil consists of amorphous peat on clay, the method with vertical drains may be an efficient alternative. Settlements will normally be considerable and it is therefore necessary that the surcharge be so high that the reduction of load due to settlements is taken into account. As a consequence of the initial height there may be a need for pressure berms to ensure the stability. The applicability of vertical drains in organic soil can be discussed, but the method has proved successful in practice. To avoid problems with long term (creep) settlements, the load of the final embankment should be at most about 80 % of the preconsolidation pressure achieved by preloading.
Lime/cement column stabilisation has often been considered to be less useful in
Methods of Construction
248
organic soils. However, in recent years, the method has been tested in organic soils, often with good results. Quick lime, gypsum, cement or fly ash mixed with the soil increases the shear strength and decreases the compressibility. The advantage of lime/cement column stabilisation in comparison with, for instance, vertical drams is an "instantaneous" increase in shear strength, which may lead to a reduced or eliminated need for pressure berms. Lime/cement columns can also be considered as drams and will consequently reduce the time of consolidation. The settlements will be reduced approximately in the order of 30 % to 50 % when using lime or lime/ cement columns, compared to those occurring using preloading and vertical drains. When designing an embankment with lime column stabilisation, laboratory tests should be made to clarify the effect of the stabilising agent on the soil properties. The method demands a follow-up during installation and preloading period.
6.3.4
Other techniques
In recent years geotextile reinforcement of embankments has been used and is often combined with a lightweight fill. A geotextile can act as a separating layer between the gravel in the embankment and the soft subsoil. It also provides better possibilities for compacting the embankment. Whether the geotextile always works as a reinforcement is not clarified. There are many other methods, such as stone columns, that can be chosen. However, these methods are often related to very specific soil conditions.
249
Chapter 7
Load Adjustment P. carlsten, Swedish Geotechnical Institute
7.1
PROFILE LOWERING
7.1.1
Introduction
In cases where settlements are large or stability is poor, the costs of improvement may increase dramatically. An effective and inexpensive alternative to improvement is to lower the entire road profile so as to reduce the soil loading. This possibility should be considered as early as possible in the project study phase so that the planners can make changes to the design. Therefore it is important to establish close cooperation between geotechnical engineers and road planners. Often, road culverts or bridges determine the embankment height, but between these a lower road profile would be an excellent solution. Naturally, the demands on the standard of the road must be taken into account in this context. When dealing with dykes and dams, the height of the embankment is critical and obviously profile lowering is no alternative.
7.1.2
Design considerations
It is necessary to have sufficient knowledge of ground conditions so that both stability and settlement calculations can be made. Other initial data naturally include the alignment of the embankment and its chosen profile and cross- section. Settlements and stability are calculated for the given embankment. If it is found that the settlements are large or the stability unsatisfactory, the cost for a number of improvement alternatives must be calculated. The road planner is then brought in and given firm data for making a cost comparison. Discussions can then take place to establish an optimal solution. This may combine profile lowering with normal improvement methods. Considerations of cost must not take precedence over stability requirements for the road. However, it is always possible to discuss with the road planner the settlements that can be accepted with regard to the usage of the road.
250 7.1.3
Load adjustment Limitations
The limitations to this method are many. Adaptation to bridges and culverts has already been mentioned. In addition, there are minimal requirements on pavement thickness and limitations on maximal gradient and curvature. As stated earlier, the possibility of lowering the embankment is obviously not at hand when dealing with dams and dykes. 7.2
PRESSURE
7.2.1
Introduction
BERMS
Pressure berms are used on roads and for dykes where stability is unsatisfactory. The method is based on building a counterweight that prevents the embankment from sliding outwards, see Fig. 7.1. The height and width of the pressure berm depend on the density of the material as well as the shear strength and depth to firm bottom. In general, any material may be used, which makes the method simple and inexpensive. The use of pressure berms is very common in building objects where there are problems with surplus soil and rock.
l
•
Fig. 7.1. Pressure berms as counterweight.
Pressure Berms
7.2.2
251
Design considerations and dimensioning
Initial data: The use of pressure berms demands knowledge of the soil's shear strength, its variation with depth and depth to firm bottom. It is also advantageous to know the density of the material that will be used for the pressure berm. Otherwise, calculations may be made for various altematives. The density of pressure berms, made from sand or gravel, is normally between 1.7 and 1.9 t/m 3.
Dimensioning: In order to dimension pressure berms, a calculation is made of the most dangerous slip surface for the embankment. This gives the following safety factor:
F=
Mresisting
(7. la)
M overturning Where Mresisting is the resisting moment and Moverturningis the overturning moment. If the safety is considered unsatisfactory, normally when F<1.5, pressure berms are built which give a negative pushing force, z f~ is the mean shear strength. For other definitions in Eqv. 7. lb, see fig. 7.1.
F=
Mresisting M overturning- M pressureberm
=
q~fu" R" S
(7.1b)
Xp.Wp + Xv.W T - X c- W c
The effect of the pressure berms is that the most dangerous slip surface will be moved and will have a larger radius, which increases the safety against sliding. If the safety is unsatisfactory, the pressure berm is enlarged until the most dangerous sliding surface offers a reliable level of safety. In general, the height of the pressure berm, hpb , can be calculated from the relation: ~f. 5.52 ql - ( g" 0pb" hpb ) = (7.2) F where ql is the applied load from the embankment and traffic. It is appropriate to give the loading difference from the level of the embankment to the level of the
252
Load adjusm~,ent
pressure berm instead of the height of the pressure berm, see Fig. 7.2. This method offers a safety margin for unknown ground levels. If ditches are planned, these must also be included in the calculations since they may present a risk to stability. Loading difference
Fig. 7.2. Loading difference.
The loading difference, ld, is then defined as the difference in levels of the embankment, hemb, and pressure berm, hpb , respectively: ld = hemb-hpb The width of the pressure berm can be estimated from nomograms in Figs. 7.3 and 7.4 (Ekstr6m et al. 1963). The nomograms were originally constructed by Odenstad (1960). The width of the pressure benn is primarily governed by the depth to firm bottom or firm material. Other factors that affect the shape of the pressure berm are the width of the load (embankment) and the magnitude of the load in comparison to the shear strength of the subsoil. In Fig. 7.3 the quotient q~all/ql is plotted against the quotient b 2/D, where: ~,11 = the allowed shear stress in the soil, i.e. the shear strength divided by the safety factor: q~all-" Zfu/F q~ = the total load from the embankment and traffic b 2 = the necessary width of the pressure berm D = the depth to finn bottom layers Only the unbroken line in the diagram may be used. As shown, this line is composed of two curves, which are derived from equilibrium analyses of the earth pressures and of circular cylindrical slip surfaces, respectively. To the left of the intersection ~all/ql = 0. 122 and b 2/D=2.5, the most dangerous slip surface is circular and to the right of the intersection the width of the berm is governed by the equilibrium analyses of the earth pressures. The calculation for circular slip surfaces is performed according to the Swedish circle method (cf Chapter 4.3.3). Active, J , and passive earth pressure, Jp, are calculated according to equations 7.3a and 7.3b, respectively. The method of calculating earth pressure may be questioned, but the equations 7.3a-c were used by Odenstad. T is the resisting force beneath the pressure berm.
253
Pressure Berms
Tall ql
,k" q2
111~4Jl11111!
;-
q
-
0.20 \ 0.15
\ \
Sufficient width of pressure berm, b 2, assuming that the most dangerous slip surface is derived from equilibrium analyses of the earth pressures and of circular slip surfaces, respectively.
0.10
0.05 o
1
2
3
4
b9
5
6
D
Fig. 7.3.
N o m o g r a m for estimation of the width of the pressure berm. The shear strength is a s s u m e d to be constant with depth. (After Ekstr6m et al, 1963)
Ja = ql- (2" ~-" 1:all)
(7.3a)
Jp- 2. ~-"
(7.3b)
q~all
T - b2" q~alI
(7.3c)
The equilibrium equation gives Jp + T - J a Pressure berms with inclined surface Figure 7.4 shows a nomogram to assist calculation of width and inclination of pressure berms when the shear strength increases in proportion to the depth below ground surface.
~f~ - t o f u+ k f . d where" tofu = shear strength at ground surface kfu - constant d = depth beneath ground surface
254
Load adjustment .
.
.
.
.
.
.
.
.
.y...q
,
,
,
--i
i
,
|
,
!
,
1
,
!
I
1
r
,
1
0.5
k_ b
Tail =l'o+k d
l, ~
-~
/'~'k._.D~......................
z_L) "T'o =10
7o.~
:5
~.,.,-.,~-/y'YTI//////l///.~
~/////////////////////~.2
0,5
u-- u 11 ro 0.8:
kD== 1 . 0 ~
...~C
0.4
0.6 0.4
1=Ii.0
0.285
dThe i _ ogram is not watid in the'e~ shaded area since a plane n = _.k. J~ sheer surface is more dangerous b k
0 5 29 '
Fig. 7.4.
'
, 5tO ` '
,
, 100 , ,
,ql , To
, 150 ,
0.2 0.3 O285 1
5.52
l
1
t
t
10
1
L
I
I
I
IS
I
1
R~ To
L
1
[
20
Nomogram for estimation of the width and inclination of the pressure berm. The shear strength is assumed to increase linearly with depth. (After Ekstr6m et al, 1963)
The loading difference is calculated in the same manner as before for constant shear strength with depth. The inclination of the surface of the pressure berm mainly depends on the increase in shear strength - the greater increase the steeper the inclination. The nomogram is only valid for a horizontal ground surface. Using the diagram starts with calculating the allowable shear strength; 1:ll = 1:o+ k- d. The axes in the diagram are q~/~o and k/o, respectively. where: the total load from the embankment and traffic
ql
=
F
= safety factor
9o = allowed shear stress at ground surface = ~of./F k
= constant = k ~ / F
c
= the increase in load in the transversel direction
The diagram consists of a curve in the form of a parabola and a number of mostly horizontal curves. The latter curves are valid when the shear surfaces touch the firm bottom. The diagram is used as follows: From the known values 1:o, k, D 1 (depth to firm bottom) and ql the values kD1/~ o and ql/~o are calculated. On the abscissa axis the value of ql/1:o is found. From this point the vertical direction is followed until either the parabola curve or the calculated value kD1/1:0 is reached. The horizontal direction is then followed to
255
Pressure Berms
the ordinate axis and the value k/or is read. From this value the inclination n of the pressure berm can be calculated from the equation: k
P
n =
(7.4) k
where p is the density of the material in the embankment. When the soil stratification is very complicated the simple solution given in Eq. 7.2 calmot be used. Other stability calculations have to be made (cf Chapter 4). Modem computer programs can consider very complicated shear strength profiles. This means that the nomograms 7.3 and 7.4 mostly are used for rough estimations. The final design will normally be based on computer analysis.
7.2.3
Limitations
The method is very economical if surplus soil can be used for the pressure berms. What may limit its use is the size of the road area available. The maximum effect of pressure berms is obtained when the depth to finn bottom layers is limited and when the shear strength increases with depth, since pressure berms increase the depth of the most dangerous slip surface. When using pressure berms, the settlements in the embankment increase owing to the load-spreading effect.
7.2.4
Construction aspects
It is very important that the pressure berms be built at the same time as the embankment. The calculated loading difference must never be exceeded. Construction of the road embankment must therefore take place in stages. When filling with an end tipper, the front of the heap must be placed in steps or flattened to avoid longitudinal slides.
7.2.5
Calculation example
A 200 m long road section is to be built. The uppermost soil consists of about 4 m silty gyttja. The shear strength is approximately 13 kPa (corrected values in accordance with SGI Information 3). Beneath the gyttja is a metre-thick layer with firmly stratified non-cohesive soil and beneath this a semi-firm clay down to 15 m. The planned road can be built with acceptable safety against ground failure. However, relatively large settlements would occur. The embankment height varies between 1.2- 1.8 m.
Load adjustment
256
In order to reduce the settlements as far as possible during construction it was decided to use preloading, giving an embankanent height of 3.0 m. A stability calculation showed that safety against ground failure for this embankment height was unsatisfactory. It was then decided to improve the stability by the use of pressure berms during the period when there was an overload. An estimate of the necessary height for the pressure berms was calculated: The height, hi, of the pressure berm close to the embankment can be calculated from: zf~. 5.52 ( h- P emb" g + qtraffic ) F hi =
(7.5)
Ppb" g
where: h
= height of embankment (m)
P emb
=
Ppb
= density of the pressure benn (t/m 3)
density of the embankment
(t/lll 3)
%amo = traffic load, normally 10 kPa ~f~
= shear strength in the subsoil (kPa)
F g
= safety factor = acceleration of gravity
The shear strength, Xfu, is 13 kPa; the safety factor, F, is chosen as 1.5; the height of the embankment is 3.0 m and the density of the fill is 1.8 t/m 3 . The traffic load is 10 kPa. This gives the following calculation: 13.5.52 (3.1.8 "10 + 10) 1.5 hi =
= 0.90 1.8 910
The height of the pressure berm is 0.9 m, which means that the loading difference is 3 - 0 . 9 = 2.1 m.
257
Pressure Berms Example of use of the nomogram in Fig. 7.3 qTalI
= Xfu/ F = 13 / 1.5 = 8.67 kPa
ql
= 3.0-1.8.10+10=64kPa
qSall/ql =
0.135
The value of ~all/ql is used in the nomogram and this gives the value b 2/D=2.2. Since the total depth, D, is 4 m this yields the necessary width b 2 b e= 4 . 2 . 2 = 8.8 m This value ofthe width ofthe pressure berm is theoretically necessary (cfFig. 7.3). In reality both the embankment and the outer part of the pressure berm are inclined. Usually, therefore, the width b 2 is measured as the distance between the middle of the both slopes. A more accurate stability calculation (done with the computer program SSTAB) showed that the loading difference could be increased to 2.20 m and the pressure berm should be made at least 6 m wide, see Fig. 7.5. E
q C~ J
l"= 20 kPa
.~
i
lOkPa ~ ~
i,. /
\\
Fig. 7.5. Calculation of pressure berms.
258
Load adjustment
Example of the use of the nomogram in Fig. 7. 4
= 10 kPa J
10 kPa
'
Id h=4m
?=1.
Dl=5m
=10+1x5:15 kPa
'Tofu :lOkPa
9
kfu = l k P a l m
Fig. 7.6. Embankment.
The height, h I , of the pressure berm close to the embankment can be calculated from equation 7.5. 10" 5.52 (4"1.8-10+10)
1.5
h 1
=
=
2.51
1.8. 10 This means that the loading difference, I.d., is 4-2.51 = 1.5 m The n o m o g r a m is used as follows: k~ 1 K . . . . . F 1.5
't o .
'tof~ . . F
.
10 . 1.5
0.67
6.7
259
Pressure Berms
k. D 1
0.67.5 =
1:0
=0.5 6.7
q~
4 . 1 8 + l0
I:0
6.7
= 12.3
In the nomogram the value k / cy = 0.365 is given and entered in the equation: k n
~,
=
k The result of the calculation is that n = 0.365 9 18/0.67 = 9.8, which can be rounded to 10. This means that the pressure berm must be started at a loading difference 1.5 m below the crest of the embankment and that the surface of the pressure berm must have an inclination of 1"10.
~1.5m
/./.
/.
/./././.
/././.
/.
/.
~1:10
/./.
/.
/.
/.
/.
/.
/.
/.
/.
/.
/./.
/././
Fig. 7.7. Results of calculation with the nomogram in Fig. 7.4.
7.3 7.3.1
LIGHTWEIGHT
FILLS
Introduction
If the settlements or the stability of an e m b a n k m e n t of conventional fills (P - 1.7-2.0 t/m 3) are unsatisfactory, the embankment may partly be replaced by lightweight fill. Lightweight fill is normally defined as a material with a density
Load adjustment
260
lower than 1.0 t/m 3. However, there are materials with higher density which are used in the same manner as lightweight fill. The material must be chemically long-lasting and stable, as well as mechanically durable and resistant to frost. In addition, the material must not cause corrosion of concrete and steel in the embankment or exert a damaging effect in any other way on the environment. There are two types of lightweight fill, residues and industrially produced material. The type of lightweight fill used must be decided from case to case. Below are presented a number of lightweight fills and their individual properties.
7.3.2
Different lightweight fill materials
Slag produets: These have a bulk density of about 1.0 t/m 3. However, there are certain types of slag, such as blast furnace slag, which absorb water with time and whose density may thereby increase to 1.5 t/m 3. Fly ash is normally placed at least 0.3 m above the highest water level, for environmental reasons. Bark: Bark should preferably be from fir or pine and have a density of between 0.8-1.0 t/m 3 when moist. Bark should not be used indiscriminately since sap may leach out and damage the groundwater and environment. Most suitable is bark from floated timber or timber stored in water and with about 1 year's storage in a moderately high stack. Fresh or dry bark is difficult to compact and should be avoided. Damp bark laid and compacted causes settlements during the first few months and it is therefore necessary to build the road with an overload. In Norway, it is considered that bark has such good strength that it can be partly used as a subgrade and thereby reduce the pavement thickness.
Wood chips: Wood chips have a density of 0.8-1.0 tim 3. They are most often used below groundwater level in order to utilize the flotation of the material. It is important that the chips be kept moist the whole time and that they be properly compacted. It is advisable to lay some form of reinforcing bed (corduroys or geotextiles) beneath the chips to avoid horizontal deformations. Peat bales: Peat is excavated and compressed into bales measuring about 1 x 0.7 x 0.5m. The density is 0.8-1.0 t/m 3. The bales are placed below groundwater level so that
Lightweight Fills
261
their flotation is utilized. The method with peat bales has been tested by Hanrahan (1964). Peat bales were placed in an embankment that had suffered extensive damage due to settlement. Parts of the embankment material were replaced by peat bales to make the embankment lighter. Samples were taken from the peat bales after these had been lying in the embankment for 8 years. Hanrahan gives the following data on the peat bales: Table 7.1.Data on peat bales.
At factory
Dry. densi.ty, t / m 3 Bulk densi .ty, t / m 3 Water content, %
0.123 0.170 37.5
After 8 year s u b m e r g e n c e
0.133 0.761 473.0
The bales had a bulk density of 0.761 t/m 3 after 8 years of submergence and consequently they were still capable of exerting an upthrust of about 0.24 t/m 3.
Sand-peat mixtures: This type of lightweight material is obtained by adding an excavated peat to sand in order to obtain a composition with bulk density Pm between 1.1-1.3 t/m 3 (Golebiewska, 1985). This density value is reached when the ash content of the mixture, PM, varies between 55-70 %. The sand-peat composition can be constructed by placing the sand and the peat in layers: - sand about 0.2 m, and - peat calculated from the relation:
h -h P
Ps
- PM
PM - PP
Pds
Pap
where hp
-
thickness of peat layer
h = thickness of sand layer; Ps, PM, P P - ash content of sand, mixture and peat respectively 9d~- dry density of sand P dp =
dry density of peat.
(7.6)
Load adjustment
262
The peat and the sand layers, which are placed on top of each other, are mixed and compacted by a bulldozer (Wolski & Golebiewska, 1980). It is important to note that to obtain satisfactory compaction of the composition built up, the excavated peat should be partly dried to achieve a water content of less than 300 %. Moreover, it must be remembered, that for embankment construction, using a sand-peat mixture, a peat with a degree of decomposition exceeding 25 % and CaCO 3 content less than 5 % should be used.
Expanded clay (Lightweight aggregates): Expanded clay is clay that is heated at very high temperatures, forming porous balls of 12-20 mm fraction. The density varies between 0.5 and 1.0 t/m 3 depending on how much water the material absorbs. When placing fills of expanded clay, a geotextile is used to separate the layers.
Expanded polyst.vrene: Expanded polystyrene (EPS) has a bulk density of 0.02 t / m 3 and is used in the form of blocks of maximum size 5xlx0.5 m. If laid in several layers, the EPS blocks must be placed so that the vertical joints are staggered. The uppermost surface of the blocks is covered with a reinforced concrete slab at least 0.1 m thick. When using EPS, it is necessary to observe the highest water level since this material has a very high flotation, which can cause difficulties. If the flotation is to be utilized, the water level should not vary by more than 0.2-0.3 m. Foamed concrete:
Foamed conrete is a porous material produced on site by mixing cement, water and a foam agent to increase the percentage of air in the mixture. Normally foamed concrete with densities of about 0.6 t/m 3 is used in connection to roads. To achieve even lighter concrete, pellets of EPS are sometimes added. If sand is added a higher density and thereby an increasing compressive strength is achieved. Foamed concrete is especially suitable in connection to repair works of existing roads.
7.3.3
Design considerations
Initial data and dimensioning: The most important parameter in a lightweight fill is the density. As shown above, this is generally below 1.0 ton/m 3 for residual products and somewhat lower for industrially manufactured products, in particular EPS, which is extremely light.
Lightweight Fills
263
Naturally, it is also very important to know how the subgrade will behave when a load is applied. Thorough preliminary studies should always be carried out to determine the stratigraphy and properties of the soil. A lightweight fill must be designed so that the loading is light enough for the road embankment to have satisfactory stability and so that settlements are acceptable in regard to the purpose of the embankment. In principle, a road embankment using lightweight fill is constructed as shown in Fig. 7.8.
Lightweight
fit[
Pavement base course ~
Supporting fi[[
Fig. 7.8. Road embankment with lightweight fill. 7.3.4
Construction
aspects
The load on the soil is the sum of the weight of the pavement and the lightweight fill. The minimum thickness of the pavement depends either on the insulation properties of the lightweight fill material or the bearing capacity of the lightweight fill. The high pore volume of lightweight fill materials means that they have excellent insulation properties. They prevent the transport of heat from the subgrade and therefore during cold autumn days icing on the road surface may easily occur. It is consequently necessary for the pavement to be sufficiently thick for heat to be stored in it, thereby avoiding this problem. In locations with a warmer climate, it will instead be the bearing capacity of the lightweight fill material that determines the thickness of the pavement. For a road embankment with lightweight fill to be sufficiently stable at the edges when traffic is allowed on the road, supporting fills of heavier material are necessary along the edges of the embankment, see Fig. 7.8. The gradient and the thickness of the supporting fill will depend on the fill material chosen, as well as the height of the road embankment. In general, two types of foundation can be used for lightweight fills; compensated foundations and preloaded foundations. In compensated foundations, the natural soil is removed and replaced with a lightweight fill material so that the pressure against the lower edge of the fill remains unchanged, see Fig. 7.9. It is important to observe the level of the ground water.
264
Load adjustment
J
Fig. 7.9. Compensation foundation.
While it is possible to utilize the flotation of certain lightweight fill materials this may also cause problems, with the road simply "floating away". It is furthermore necessary to note the variations in ground water level. In compensated foundations, it is often possible to allow a certain increase in pressure against the bottom edge of the fill. However, this must be determined m each individual case according to the soil characteristics and especially the purpose of the road. In preloading, the road embankment is built with lightweight fill directly upon the sub-base and an overload is applied, see Fig. 7.10. The embankment is then allowed to settle so much that the lightweight fill sinks below the ground water level.
Before unloading //
/"
~ " . .
,-~-,-,-~,--,-~.-,,-,-
"
Surcharge Base course
,, Supporting.flit .-, \
---./~\.
L!ghtweight fi!t i~:. ". ".(Wood chip,~)'"i"i."i:"~"~~,,~,,~,.~,~.,~,,~,~
Groundwater
After unloading Base
course
~~
;ii
tabte
/
Fig. 7.10. Preloading technique.
The load is then removed, after which the subgrade no longer imposes an extra loading owing to the flotation in the lightweight fill material. This method is used especially with wood chips or bark.
Lightweight Fills
265
Lightweight fill can also be used in roads damaged by settlements. The old embankment fill masses are then removed and replaced by lightweight fill material, while the profile is adjusted at the same time, see Fig. 7.11. Before
After ~:'iLightwei~ht
f i tt "' . i....':':h~.~>,,.
Fig. 7.11. Adjustment of road damaged by settlements.
This method avoids imposing any extra loading on the soil, which would only make the road embankment settle further. Soft lightweight fill materials, i.e. those not produced in the form of blocks or slabs, are laid 0.5 - 1.0 m thick. Each layer is then compacted thoroughly with a crawler tractor, for example. When laying peat bales or EPS blocks, it is advisable to smooth the surface first with a layer of sand, for example, so that the bales or blocks can be laid as closely as possible. Foamed concrete is produced directly on site. Normally the maximum thickness of each layer is 0.6 m. After one day the concrete has hardened and a new layer can be placed on top of the old one. No compaction is needed for foamed concrete. When using lightweight fills to eliminate settlements, it is always extremely important to follow up the development of the settlements. It is then possible at an early stage to modify the actions used if these do not produce the intended results. In addition, it is necessary to determine when an overload, if any, can be removed.
Transitions: Another area of application for lightweight fills is in transitions from a nonstrengthened section of road to a heavily strengthened section, where the reinforcement is intended to reduce settlements. If it is expected that the transition between the strengthened and unstrengthened road section will produce a hazardous step in the road surface, the transition can be "softened" by compensating with lightweight fill, (cf Chapter 7.3.6).
266
Load adjustment
Lever of road (year O)
Level of road (year 10)
.
.
.
|
I"
Embankment piles Fig. 7.12. Transition between strengthened and unstrengthenedroad section.
7.3.5
Limitations
The choice of lightweight fill depends very much on cost. Residual products are naturally less expensive than specially manufactured materials if they are available at a reasonable distance from the worksite. Mostly, lightweight fill is used for shorter sections (especially transitions) since the method is rather expensive. If problems with settlements are expected over a long section, other methods are often more suitable (for example, profile lowering, preloading, lime colunms or vertical drams).
7.3.6
C a s e history from lightweight fill in a transition to a bridge
The example below relates to a 1200 m section of road 156, between Kungsbacka and Gothenburg, built on organic soil. The road lies on an embankment with a maximum height of 1.5 m. Almost in the middle of the 1200 m section the road crosses a fiver on a bridge. The bridge is built on piles driven to a firm bottom. The soil consists of soft gyttja-bearing clay down to 6 m depth on top of clay down to approximately 12 m depth. The area has been in agricultural use for a long period and the top layer (= 1.0 m) has been transformed into a dry crust. According to compression tests, the clay is weakly overconsolidated.
267
Lightweight Fills Shear strength (kPa) 00 50
~,
Density, p (t / m3 ) 010 I I I I 20
"C_.
Water content (%) Liquid limit,w L (%) 100 00,, I
I
I
I
/
/ ! I
5
-
I
5-
5-
I
I
E _c"
\ \x 7///
~,-
r;I
cl
4~10
10
Sensitivity
100
II
101
1
I
I
1
1
t
I
Fig. 7.13. Clay parameters.
Stability:
To ensure acceptable stability down to the fiver, it was decided that the road should partly be built up with lightweight material (EPS). Stability calculations were made to dimension the amount of EPS that was needed. These simple calculations are based on circular slip surfaces. Since the load has a limited width (= 15 m), the derived safety factors are on the safe side. A more accurate calculation has to take into account the shear strength along the ends of the cylindrical element that slides (see Fig 7.15). Settlements:
Compression tests were performed on three sections and showed that the soil is weakly overconsolidated, see Fig. 7.14. With an embankment height varying between 0.5 - 1.5 m, the total settlement is in the order of size 0.15-0.50 m. This would have been regarded as acceptable if there had not been a bridge in the middle of the section where the settlement is 0 m. This means that alter a time a step will form in the transition from the bridge to the road. If an attempt is later made to adjust the step by laying extra surfacing, this will only increase the loading of the soil, and consequently lead to increased settlements. Therefore, it is important to unload the road adjacent to the bridge.
Load adjustment
268 Effective vertical stress 0
I
I
i
1
,
I
I
i
~
f
6' (kF~) I
100 I
9
r'~
10 l
i
\\~6",' i O~ " ~'~
Fig. 7.14. Effective stress and preconsolidation pressure.
By calculating the settlements for a number of different loadings, it is possible to determine how far the soil can be loaded in order to obtain a gradual transition, see Table 1. Table 7.2. Relation load/settlement.
Load (kPa) 22.8 15.7 12.4 2.0
Settlement (cm) 23 15 10 0
Here it was decided to lighten the road embankment stepwise using expanded polystyrene. Since it is now known how much the soil can be loaded, it is possible to calculate how much expanded polystyrene will be needed. The road embankment at the bridge, hemb,is 1.20 m high. This gives a loading of 22.8 kPa with a heavy fill (Phil- 1.9 t/m3). To arrive at the load 15.7 kPa with the embankment height 1.20 m it is necessary to replace X m of gravel with EPS (Plight- 0.1 t/m3).
Lightweight FilLy
269
[9~11 " g" (hemb.- X)] + (Plight" g" X ) = q [1.9- 10 ( 1 . 2 - X)] + (0.1 910. X ) -
15.7 => X = 0.40 m
In this example, a roadbase of at least 0.6 m is required, which means that when the loading is to be about 2 kPa a depth of 0.55 m must be excavated and replaced with expanded polystyrene.
Results of design calculations: The length of the transition must be decided in each individual case. This decision should take into account the speed limit on the road, the type of traffic for which the road is intended and any other information that may be of importance. In this example, it was decided to make the transition 18 m long, see Fig.7.15. As always, the solution is affected by both stability and settlement aspects.
Slip surface 1 2 3 /~
Safety factor 2.02 1.92 1.65 1.70
L-B Concrete sLab on piles
A-A
B-B
i
.....~ < , :,:,!.:.:.;'~,,,'~ ......
i~xpanded polystyrene
Fig. 7.15. Transition.
I i
! a
9 slab i
[
270
Load adjustment
7.3.7
Case history from use of lightweight fill in repairing an existing road
The example relates to a 500 m section of road E65, at B6rringe monastery (between Malm6 and Ystad). The road was built in the late 1960"s over an area with peat, gyttja and gyttja-bearing clay. The maximum depth of organic soil was about 8 m. The foundation of the road was built by means of excavation and refill with good material, see Fig. 7.16.
/ GYTT]A /,
Fig 7.16. Bbrringe monastery; Geotechnical conditions prior to treatment. However, the excavation was not wide enough and in some sections there were also some organic soil left under the embankment. Over the years large deformations have occured, due to long time creep settlements and bad stability. Adjustments of the road surface had been made with base-coarse material and asphalt. When investigations were made in 1991 the asphalt layer was about 1 m. By adding load the stability becomes worse and a 30 m longitudinal crack was observed in the road. To avoid the recurrent adjustment of the road it was proposed that the next adjustment should be combined with an unloading of the embankment according to Fig. 7.17.
I I
~Foamed concrete
Fig. 7.17. B6rringe monastery; Principle of unloading.
The unloading was performed by using foamed concrete and a design based on the assumption that the density of the foamed concrete, after a maximum absorption of water, should be less than 0.6 t/m 3. The design was based on stability analyses
Lightweight Fills
271
where the safety factor should be increased from the existing 1.2 up to the required 1.5. The unloading was performed according to Fig. 7.18. Stage 1: Excavation of one half of the road.
Stage 2: Casting of foamed concrete.
Stage 3: Excavation of the othet half of the road.
Stage 4: Casting of foamed concrete.
Stage 5: Road completed.
Fig. 7.18. B6rringe monastery; Performance of unloading at section 1/780-11880.
272
Load adjustment
The work was performed in two stages. In stage I an excavation was made in the fight half of the embankment. The traffic was passing by on the left part of the road. Foamed concrete was cast in four layers and after it had hardened it was covered by base coarse material and asphalt. The traffic was then altered to the fight part of the road. In stage II excavation and casting of foamed concrete was made for the left part of the embankment. In total the unloading, casting of foamed concrete and preparation of the road surface for the 500 m section of the road was finnished in about 3 weeks.
7.4 REFERENCES Carlsten, P. (1989). V/igbyggnad p~ torv. Handbok. V/igverket. Publ. 1989:53. BodAnge. 35 p. Carlsten,P (1995). Construction methods for roads in peatland areas. European conference on soil mechanics and foundation engineering, 11, Copenhagen, MayJune 1995. Proceedings, vol. 8. The interplay between geotechnical and engineering geology, pp. 8.13-8.18.
Carlsten, E, Eriksson, L., Bergman, L., Andersson, R. & Hansson, L. (1995). Skumbetong i viig-och markbyggnad. AnvAndning, projektering, produktion samt erfarenheter. Statens geotekniska institut, SGI. Viigledning 6. LinkOping. 43 p. Construction of roads on compressible soils (1979). Organisation for Economic Co-operation and Development (OECD). Paris. 148 p. Ekstr6m A. (1974). SRS geotechnical strengthening measures in road and street construction. Ekstr6m, A. & Melhus, B. (1977). Geoteknik 1. Hermods Skola. Malm6./174/p. Flaate K. & Rygg N. (1962). Sagflis i vegfylling pA myr (Statens Vegvesen, Norge Medd nr 16)Norske Vegtidskrift nr 12, 1962. Frydenlund, T.E. & Aab6e, R. (1994). Expanded polystyrene - a lighter way across soft ground. International conference on soil mechanics and foundation engineering, 13, New Delhi, Jan. 1994. Proceedings, vol. 3, pp. 1287-1292. Gandahl R. (1971). NAgra svenska erfarenheter frLri anvAndning av bark i viig. Frost i jord, No. 5, pp. 5-13. Statens Vegvesen, Oslo, Norge Golebiewska, A., (1985). Construction of embankments using peat (in Polish) Katedra Geotechniki. Hanrahan, E.T. (1964). A road failure on peat. Geotechnique, Vol XIV, No 3. Harbruck, D.I. (1993). Lightweight foamed concrete fill. Transportation Research Record No. 1422, pp. 21-28.
References
273
Hartlrn, J. (1985). Pressure berms, soil replacement and lightweight fills. Soil improvement methods. International geotechnical seminar, 3, Singapore, Nov., 1985. Proceedings, pp. 101-111. Hartlrn, J. & Carlsten, P. (1992). Improvement of soft soil. New technology for foundation engineering, NTFE '92. International geotechnical conference, Hanoi, Oct. 1992. Proceedings, vol. 2.15 p. Knutson A.(1973). Praktisk bruk av bark i vegbygging, NRRL rapport nr 47, pp. 515, Oslo. Larsson, R., Bergdahl, U. & Eriksson, L. (1984). Evaluation of shear strength in cohesive soils with special reference to Swedish practice and experience. Statens Geotekniska Institut. SGI Information; 3. Linkrping. 32 p. Olofsson, T. (1987). L~ittldinker som l~ittfyllning i v~tgbankar. V~igverket. V~ig- och Brokonstruktion. Geoteknik. Publikation; 1986:78. Bod~,nge. 12 p. Oppbygging av fyUinger (1994). Statens Vegvesen. HS,ndbok 176. Oslo. 124 p. Skuggedai, H. & AabSe, R. (1991). Temporary overpass bridge founded on expanded polystyrene. Statens vegvesen. Veglaboratoriet. Intern rapport No. 1511, pp. 9-11. Swedish National Road Administration (1990). Cellplast som l~ittfyllning i v~igbankar. V~igverket. V~ig- och Brokonstruktion. Geoteknik. Publikation 1990:49, 22 p. Wolski, W., Golebiewska, A. (1980). Compaction of peat and peat-sand mixtures. International conference on compaction, Paris, April 1980. Vol 1, pp. 287-289.
274
Chapter8
Replacement P. Carlsten, Swedish Geotechnical Institute
8.1
GENERAL
When building roads or dykes on organic soil or any other soil, it is necessary to ensure that the stability is acceptable both during construction and also when the road is in use by traffic. A road embankment shall also be practically free from settlement to limit maintenance costs. For dykes, it may be possible to accept settlements to some extent as long as the dyke is still above the highest water level. One way of fulfilling these conditions on stability and permissible settlements is to replace the compressible soil with gravel or blasted rock. Replacement can be performed by excavation and backfill or by progressive displacement. Excavation leads to a better result, while progressive displacement sometimes leads to maintenance costs for the road some years after it has been taken into service. The use of excavation has its limitations, mainly depending on how deep the excavator can reach. To obtain a satisfactory result, it is important that all soft soil be excavated. Compared to preloading, for instance, replacement offers a better possibility for choosing the level of the road and normally no extraordinary costs for maintenance arise.
8.2
EXCAVATION AND BACKFILL
8.2.1
Description of the method
When the soft soil is excavated, it is normally replaced with non-cohesive material. The method is based on creating equilibrium conditions between the backfill and the remaining organic soil. Stability is calculated by comparing the active earth pressure, P~, from the backfill to the passive earth pressure, Pp, from the organic soil. The safety factor shall be 1.7 and this will be achieved if the depth and width of the excavation are properly chosen and adapted to the prevailing conditions. The extent of the excavation also depends on the choice of backfill material.
275
Excavation and Backfill
The backfill material must not contain peat, gyttja, dy, wood remnants or similar material. For dykes, the demands on the backfill material are not as hard as for roads. Depending on the conditions in the particular case, the following shall be considered: 9 If the backfill is placed under water, the material should be blasted rock or non moisture-susceptible friction material. 9 If the backfill is placed in a pit with a muddy bottom layer, the material should preferably be coarse blasted rock or coarse-grained moraine containing boulders. In some cases other backfill material, such as sandy gravel, can be used. If this kind of material is used, consideration has to be given to large settlements and problems with maintenance. 9 Normally, soft organic soils are highly compressible and the excavation must be made to full depth. Sometimes, it may be very difficult to make an excavation in peat to full depth, since the walls of the pit tend to collapse. In this case, it may be better to stop the excavation some 0.5 m above the bottom of the peat layer and compact the rest with the backfill material. There are of course alternative methods of excavation. Sometimes, the soft soil is not excavated to full depth. By partly excavating the soft soil, it is possible to reduce the time needed for consolidation (the length of the drainage path is shortened). Sometimes, the soft soil is excavated only beneath the side slopes of the embankment (cf. Fig. 8.1). This method has been used especially in Sweden for improvement of existing roads on peat and has the advantage that the road can be used by traffic
Existinq embankment .New e,mbqnkrnent
....
!
,
a ......
. . z ..
: " !: : : l : :
~!
I
:
~
..... a
I
Fig. 8.1. Example of excavation beneath the slopes of an existing embankment.
Replacement
276
during construction. On the other hand, it obviously has disadvantages since the good quality material is placed beside the actual road and there is also a risk of serious problems with transverse differential settlement. Nowadays, this latter method has been more or less abandoned and the improvement is instead achieved by preloading (cf. Chapter 9.2.5). To separate the material of good quality m the embankment from the backfill material in the excavation, a geotextile is placed on top of the backfill.
8.2.2
Design considerations
Initial data and dimensioning: The extent of the excavation is governed chiefly by the shear strength and the thickness of the soR soil. The properties of the backfill material are important, as is the height of the embankment and the inclination of the embankment slopes. The calculation methods described in this chapter are intended for preliminary decisions. For a more detailed calculation of stability, see Chapter 5. According to the excavation manual issued by the Swedish National Road Administration the main concern is the stability of the future road. The safety factor shall be F > 1.7, i.e. 1.7" P~ _< Ppl and (8.1)
1 . 7 " P _< Pp2 + T
Load
/ I I
I
Pp 2
J'.l:".'.'l.'.':l.
1
/
i
i
i
i
t
/
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q
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". T ' . "
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Fig. 8.2. Comparison of active and passive earth pressure.
,'l-
".'.1
Excavation and Backfill
277
T is the resisting force along the bottom of the excavation. The most dangerous slip surface must not be allowed to pass through the crest of the embankment. To calculate the necessary width of the excavation through equilibrium analyses of the earth pressures, there are diagrams for normal values of the friction angle and slope inclination (Swedish National Road Administration, 1991). When the sos soil is excavated to full depth the following rules of thumb are given.
Rule of thumb, no. 1 Non cohesive material (~) = 35 ~ in refill and embankment.
A. Inclination of slope 1:2 x-2-H
J
f
f
f
J
f
I
J
H _> 1,5m
f
D <3,5m
Fig. 8.3. Rule of thumb, no. 1.
If H < 1.5 m and/or D > 3.5, there is a need for backfill material with a higher friction angle
Replacement
278
B. Inclination of slope 1:3 30 = 350 n=3.0
E
20
e.
o
~
co ~
/
/
LU
..
0
1
2
3
4 5 6 7 Height of embankment, m
8
9
10
Fig. 8.4. Excavation width vs height of embankment. (D= depth of soft soil)
of thumb,
Rule
no.
2
Blasted rock (~ = 42 ~ in refill and embankment. 20" = 42 ~
f
E x~ .c: "O . m
~ o
~-~ 9 } n = 2.0 .I
~
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10"
UJ
0
0
1
2
3
4 5 6 7 Height of embankment, m
8
9
10
Fig. 8.5. Excavation width vs height of embankment. (D= depth of soft soil)
279
Excavation and Backfill
If the most dangerous slip surface does not pass through the crest of the embankment (Cf Fig. 8.2) and the criteria described above (Eqv. 8.1) are fulfilled there will be no problems with lateral displacement. E m p i r i c a l rules employed in various countries:
The OECD report on "Construction of roads on compressible soils" (1979) gives a scheme of empirical rules from various countries, developed for determining the excavated cross-section beneath road embankments. The rules have been developed according to the local conditions in each country and do not normally consider the properties of the backfill or subsoil. For this reason, the OECD report states that the more general application in some cases can lead to problems. A check on the stability during excavation is of course necessary. If this calculation gives unsatisfactory values of the safety factor, the excavation can be performed in terraces (Fig. 8.6).
.N_
x O ' / / / / /
\ \ \ \ \ \\
/ / / / / / / @~r
///////////
\ \,~x,\~;
/// // ~\ \
Fig. 8.6. Excavation with terraces.
An alternative is to keep the pit open to a minimum during excavation. This means that excavation and backfill are performed in parallel as shown in Fig. 8.7.
x
x
X___ __,,
x _\ __
__~_._
Fig. 8.7. Excavation and backfill. 8.2.3
Limitations
The excavation cannot normally be made deeper than 5-6 m, depending on the capacity of the excavator. The method is also costly in comparison to preloading, for instance. When the excavation is made below water, compaction of backfill may be
Replacement
280
1. ORGANIC SOILS
W,"R,~/,,7,,"~,,~,,'7",,'-Rx2~,~
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.
.
.
.
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2. COMPRESSIBLE SOILS
Shoulder
Soft soil
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Depending on the shoutder width line a or line b witt be determinant for the excavation width
3. ALL COMPRESSIBLE SOILS ,
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Any soft soil
Minimum width 4. PEAT 7\ I D -~ 1.5rn
Fig. 8.8.
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~
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Empirical rules employed in various countries for determining the width of excavation of compressible soil beneath road embankments.
ProgressiveDisplacement
281
difficult and consideration should be given to using material which will not require compaction. This material shall be non moisture-susceptible.
8.2.4
Construction aspects
Depending on the conditions, it is possible to choose between a normal excavator and a dragline. The dragline has the advantage that it can be placed at a distance from the pit and the vehicles removing the soil from the site need not get close to the excavation. An important factor in minimising costs and maintaining production is the availability of conveniently located spoil areas for depositing the unsuitable material. Close control of the excavation operation is necessary to ensure that pockets of soft material are not overlooked. These may lead to post-constructional problems such as differential settlement, lateral displacement or even instability.
8.3
PROGRESSIVE DISPLACEMENT
8.3.1
Description of the method
Progressive displacement is a method of constructing an embankment by placing an overload on the ground surface. A continuous shear failure is produced in the soft soil and as a consequence the soft soil is squeezed away and replaced by material from the overload. Usually, a surcharge is applied to increase the force and to compensate for loss in embankment height due to settlement and lateral displacement. This method can be applied when the subsoil has a low shear strength, which is normally the case in organic soil (with the exception of peat with a low degree of decomposition). If the subsoil is stiffer there is an altemative method, where the embankment is constructed in its full length. The displacement is then created from bursting in the subsoil. In assessing the feasibility of the displacement method for a particular situation, consideration must be given to the influence of the displaced soil on the adjacent structures. Relatively large areas are likely to be affected and structures at distances from the edge of the embankment of five times the depth of treated soil are known to have been influenced. The method has proved most successful with high embankments and when the subsoil is very soft and too deep for normal excavation. With well-planned and carefully executed schemes, the method has produced a satisfactory solution even
282
Replacement
for roads with high standard. Plan
Longitudinal section
!, !, !,
Lever of future road
:_i_..$.ur.c.h.a.r.ge" :...i...i.._...,~ :.'..'..............-.........
....
/.
/
/
/
/
/
/
/
/
Fig. 8.9. P r o g r e s s i v e displacement.
Displacement methods are currently used mainly in the United States, Canada and Scandinavia. Good results have been reported from these countries, particularly when suitable rock fill has been employed and where the construction period is sufficiently prolonged to allow the movements to be largely completed. Where a stronger layer, or a dessicated crust, overlies the very soft compressible material, it may prove necessary to remove this by excavation methods. Alternatively, a blasting technique may be employed to achieve the same effect (cf. Fig. 8.10). Cross section
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,, ...-._7
Progressive Displacement
8.3.2
283
Design considerations
1 Initial data
Density The weight of the embankment is calculated from an assumed value of the density of the fill. If no exact values are available, the following values can be chosen: 9 Multi-gradedfill
- 1.9t/m 3
9 Fill consisting only of boulders and coarse blasted rock
- 1.7 t/m 3
Friction angle When calculating stability and earth pressure, the friction angle is chosen with respect to the material in the embankment fill. The friction angle can be estimated as: 9 Gravel, cobbles, boulders and coarse-grained moraines
- 35 ~
9 Blasted rock with large boulders
- 42 ~
Shear strength When checking stability, earth pressures and displacement depth, the shear strength is chosen from measured values. Using bursting, the shear strength of the soil will normally be reduced. The reduction is judged for each case. As a rule, the reduction can be assumed to be about 20%.
I Dimensioning Displacement is created by causing a failure in the soil through a surcharge on the ground. The necessary height of the surcharge can be estimated from H >_0.4.~f~ where H is the embankment height in metres and q:fu is the shear strength in kPa. This corresponds to an embankment load exceeding the failure load by at least 20 %. To make the displacement easier, it is possible to use bursting or excavation in front of the fill and consequently reduce the resistance against displacement. Bursting and excavation are also used to control the direction of the displaced masses. When bursting is used, the shear strength is normally decreased. The decrease is judged for each case, but as a rule a 20 % reduction in shear strength can be assumed.
Replacement
284
H1H0 s 9. / . . . / . .
/.../.../
S- surcharge Hproj-height of future road H- total height of embankment and surcharge Fig. 8.11.
The necessary embankment height H, including surcharge, as a function of the shear strength t fu"
The displacement depth is influenced by soil stratigraphy and properties of the fill. If the subsoil consists of a homogeneous cohesive soil, especially clay, the displacement depth can be derived from a trial calculation. If the ground water table is assumed to be at the original ground surface, the depth can roughly be estimated from: Of" hi + P'f" h2 - 0.55"qTfu h
Z
~..
Ps where h z = Displacement depth, given as an assumed distance from the level of the ground surface after displacement, m ~f~ = Undrained shear strength, kPa P'~ - The density of the original subsoil beneath the ground water table, t/m 3 P'f = The density of the fill below the ground water table, t/m 3 Pf = The density of the fill above the ground water table, t/m 3 If there are layers of non-cohesive material in the cohesive soil, the fill may stop at these layers. The displacement depth may also be influenced by measures taken during construction, such as bursting and excavation. The lateral displacement is normally limited to the vertical plane through the foot of the embankment. If the top layer is very soft, this vertical plane passes through the intersection between the slope of the embankment and the lower level of the soR material.
Progressive Displacement
285
Original ground surface
: PiPi :
~j'
7 . . .
/...
i ~.
. . "
jo! h 2 . ~ displacement !1 ~ t h
/.
..
/.
..
/.
../.
.
.
/.
9 . / .
.
./..
./..
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~. I
.../.../.../.../.../.../.../.../.../.../.../.../...
Fig.8.13.Lateraldisplacement.
9 /.
.
-/
Replacement
286
8.3.3
Limitations
As mentioned before, it is necessary to consider the effect on adjacent structures from the progressive displacement. Other factors that affect the design considerations are the availability of a suitable fill (such as crushed rock), the construction time required, and the maintenance costs resulting from long-term deformations. The shear strength in the subsoil cannot be too high since this creates a large resistance in the soil against failure. For this reason, the method is very difficult to use in peat with a low degree of decomposition. The entanglement between the fibres in the peat results in a high shear strength and the shear strength also increases with effective vertical pressure. The method is preferably used for high embankments. For low embankments, other methods may be better from both economic and functional aspects. Normally, the fill does not reach firm bottom layers, but leaves a layer of compressible soil beneath the fill. Though this layer may be very thin, it may take a long time to reach full consolidation in this layer. The progressive displacement shall be combined with preloading, i.e. the embankment with surcharge shall be in place for at least 4-6 months. During this preloading, the settlements should be measured continuously.
8.3.4
Construction aspects
The following factors affect the result of progressive displacement: 9 The higher the embankment, the higher will be the probability of failure in the subsoil. 9 A narrow embankment is more likely to reach firm bottom layers, due to smaller movements in the soil. 9 When the embankment is very wide, the displacement is mostly longitudinal. Sometimes the volume of the displaced masses may be so large that there is a need for bursting or excavation in front of the fill. 9 Ensure sufficient preloading time. 9 The settlements shall be measured during the preloading period. Sometimes extra time for preloading will be needed due to adjustments of the surcharge. 9 Materials forced upwards will change the level of the ground surface in the vicinity and cause changes in dewatering conditions. 9 Progressive displacement in water causes the bottom of the lake or river to heave so that the water becomes muddy. This will have an effect on currents, river channels, bird life and fish life.
Progressive Displacement
Damaged tel ephone cables
287
/
Accumua l ted water~
Damage totrees
Tn '
L " 5D
Fig. 8.14. Effects on the environment close to the construction area.
9 It is advantageous for the progressive displacement method from an economic viewpoint if there are excess masses in the vicinity of the treated area. It is also valuable if the volume of the masses is so large that extra surcharge masses are available at low cost. It is often difficult to calculate the exact volume of the masses required. 9 Ensure that there is a sufficient, safe distance to adjacent structures. The claims due to settlement damage may be considerable. Note that structures within a distance of five times the depth of the treated soil may be affected. 9 Culverts must be placed outside the area affected by the displacement. 9 Excavations or piling close to the displacement area may be unsuitable or even impossible. For this reason, progressive displacement will affect future contruction works in the area. 9 In materials with high sensitivity, shear failures may become uncontrolled unless care is taken. 9 Displacement to depths greater than 15 rn should be avoided. The following demands should be placed on the filling material: 9 The masses shall be coarse-grained to create good binding in the fill. Preferably the fill shall consist of coarse, blasted rock. Moraine with a large content of stones and boulders, stones and boulders from boulder ridges, or gravel containing stones and boulders can be used. 9 If the alternative solution is chosen, i.e. if the embankment is constructed to the full length and the displacement is created through bursting, the demands on the fill are not as stringent. One condition, though, is that the material shall be freedrained.
288
Replacement
Transitions: For transitions to areas with better beating capacity, progressive displacement is normally complemented with excavation. If and when there is a transition between two different embankments built by progressive displacement, it is very difficult to replace all soft soil down to firm bottom. Normally, some material will be left and there is a need for extra surcharge in this area. It is advisable to avoid this problem by making the displacement take place in only one direction and supplementing it with excavation where the soil is shallower. Transitions from progressive displacement to other construction methods, such as embankment piling, are not advisable.
Construction documents: The following aspects shall be detailed in the documents presented to the contractor. 9 Demands on the fill material. Demands on placing the fill. 9 Temporary transport roads to the site. Temporary diversions for traffic on existing roads. Restrictions with respect to existing buildings. 9 The site where the contractor is to obtain the fill material. The direction in which the displacement is to be managed. 9 Location of a place to store excavated material. 9 Preparatory works, such as bursting or excavation. 9 Performance, filling, surcharge and necessary preloading time. 9 Necessary measures for controlling and facilitating the progressive displacement. Since areas in the vicinity ofthe site for displacement will be affected, it is important to observe effects on buildings and other structures in the neighbourhood. 9 Excavation of material that has been lifted up on the side of the embankment. This is normally not necessary unless the embankment is located m an urban district. 9 Transitions 9 Time schedule 9 Inspection programme. Includes follow-up of quantity of fill material, sounding to check the effect of the displacement and levelling of the embankment with surcharge.
Progressive Displacement 8.3.5
289
Case history
The example concerns a progressive displacement performed in Sweden for road 50 between J6nk6ping and Orebro, approximately 200 km west of Stockholm. At this site, the soil stratigraphy was as follows: On top there was a 2-3 m thick layer of peat with a high degree of decomposition and 1-3 m of gyttja. Beneath this organic material was 2-6 m soft clay. The total depth to firm bottom layers was about 8-10 m.
+25
Level of future road
7
i
'
SurchQrge. ,
Exc~176l TI I x
7
/
* IO
0 I
0/100 I
0/200 I
0/300 Length
I
Fig. 8.15. Longitudinal section of the progressive displacement.
For stability and settlement reasons, it was decided that the embankment should be pressed down to firm bottom layers using the progressive displacement method in combination with excavation. The replacement was made through excavation between the sections 0/010 and 0/030, where the excavation depth was gradually increased to 4 m. In the section between 0/030 and 0/300, the excavation was to 4 m depth and the rest of the replacement performed through progressive displacement.
Replacement
290
Between sections 0/300 and 0/310, the excavation depth was decreased gradually from 4 to 1 m. For reasons of stability, masses from the excavation had to be transported away. Refilling was performed in direct connection with the excavation. The refill material consisted of blasted rock and the fill was placed with a surcharge of at least 3 m over the level of the future road. The necessary height of the embankment including overload can be calculated from the following formula: H>0.4.%
Length or
'
"
9~
i.~
9 N.
9
10 -15 m
~> H t o t a I
"
~"
9 \/
Level of f u t u r e r o a d / ; \ " i . z, a
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.
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soil
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Safety margin for placing the fill. The fill is subsequently transported to the front by caterpillar.
The shear strength in the clay was about 9 kPa and consequently the necessary load was 3.6 m. The surcharge was placed with the same crest width as the future road (7 m) and its length was 30 m. This temporary surcharge was moved forward gradually as the progressive displacement was achieved. After removing this temporary surcharge, the terrace was covered with fine-grained rock material and a new surcharge was made from basecourse material. This surcharge was to have a crest width of 6 m and the level was to be at least 2.5 m above the level of the future road.
Progressive Displacement
291
Surcharge of blasted, rock to c r e a t e failure in the \soft soil
. Temporary surcherge
\
from base course material
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:..'..'.-.':
.'.-~'.':'.-:
Fig. 8.17. Diagram of progressive displacement. The documents also contained detailed instructions conceming the direction in which the organic material was to be squeezed out. Instructions regarding suitable machines (caterpillars) to place the surcharge were given. To make the displacement easier a dragline was used. The dragline removed loose pressed-out masses close to the front of the fill. One condition was that the excavation pit in front of the fill had to be at least 5-6 rn long and at least 4 m deep. After finishing the progressive displacement, the surcharge functioned as a preloading for 3 months. In the instructions to the contractor, it was stated that this preloading period could be prolonged by up to 6 months and that the surcharge was not to be removed before the Road Administration had given their consent. The following control measurements were made: 9 The consumption of fill material was measured and compared to the (theoretically) calculated value. If the difference was too large, construction was to be stopped and an investigation made to clarify the reason. One possible cause of differences could occur in connection with the fill stopping at a layer of non-cohesive material. In this case, measures such as bursting or extra excavation were to be taken. 9 Levelling was to be carried out in sections, every 20 m, and was to start as soon as the large movements close to the front had stopped. The results were communicated to the Road Administration in order to make a decision on the need for an extra surcharge.
292
Replacement
8.4 REFERENCES Carlsten,P (1995). Construction methods for roads in peatland areas. European conference on soil mechanics and foundation engineering, 11, Copenhagen, MayJune 1995. Proceedings, vol. 8. The interplay between geotechnical and engineering geology, pp. 8.13-8.18 Carlsten, P. (1989). V~igbyggnad ph torv. Handbok. V~igverket. Publ. 1989:53. Bod~,nge. 35 p. Hartl6n, J. (1985). Pressure berms, soil replacement and lightweight fills. Soil improvement methods. International geotechnical seminar, 3, Singapore, Nov., 1985. Proceedings, pp. 101-111. Holtz, R.D. (1989). Treatment of problem foundations for highway embankments. National Cooperative Highway Research Program. Synthesis of Highway Practice 147. Washington, DC. 72 p. Oppbygging av fyllinger (1994). Statens Vegvesen. HS,ndbok 176. Oslo. 124 p. Swedish National Road Administration (1991). U r g r / i ~ g f6r v~igbank. Allm/in teknisk beskrivning. V~igverket, V/ig-och Brokonstruktion. Geoteknik. Publikation 1991:06. Bodange. 53 p. Swedish National Road Administration (1979). Nedpressning av v/igbank. Statens V~igverk; TU 139.39 p.
293
Chapter 9
Staged Construction W. Wolski, Department of Geotechnics, Warsaw Agricultural University
9.1
GENERAL
Staged construction consists in the filling of an embankment at a controlled rate, so as not to cause failure but to permit an increase in shear strength due to consolidation. In such a way, the obtained strengthening of the foundation soil should be sufficient to support safely the required load. Thus, in staged embankments, the precompression technique is used, which according to Johnson (1970) is defined as: "compressing the soil under an applied stress prior to placing or completing the structure load". In the case of the staged embankment, the first stage of embankment (preloading with first stage) compresses the subsoil prior to the filling of the second stage. In the case of roads or dykes, two or three stages are often used. When settlements after the end of embankment construction have to be minimised, surcharging is used. This is a temporary preloading with load in excess of the permanent fill. Where there is a high ground water level the settling embankment is gradually submerged. Because of uplift the effective load is then decreased. A schematic explanation of the staged construction is shown in Fig. 9.1. The staged construction technique is frequently used in conjunction with the installation of vertical drains in order to accelerate the consolidation process and to reduce the period before the next stage of the embankment construction can be commenced. The staged embankment is considered a very useful technique of construction on most organic soils. But, in spite of an apparent simplicity, it needs: 9 a good quality soil investigation 9 a design well-adjusted to the foundation behaviour, taking into account the relationship between the stress state developed in the subsoil and stability 9 careful monitoring of the subsoil behaviour during the construction period.
294
Staged Construction
"0 0 0 _A
Possible load decrease due to uplift / ~, f
I
...... i
I
........I.......-
nd stage
I Surcharge
Permanent fill Time
r
C
E @
Fig. 9.1. Staged construction scheme.
9.2
PRECOMPRESSION TECHNIQUE
9.2.1
Introduction
Staged embankments are constructed by the use of precompression technique in which each stage is utilised as preloading before the next stage is placed. Each stage should be constructed according to earthwork standards, using layers. Unlike temporary preloading fills, used for improving soft, soils before a structure is built, each stage of the embankment should be made of an appropriate soil, with proper compaction. Depending on the site conditions, it may be necessary to construct several layers of the first stage as a working mat - without compaction, to support the hauling vehicles and the equipment for compaction. When the top layer is composed of peat it is important to keep the surface intact, which can be done by removing vegetation with very light equipment to avoid disturbance of the surface mat. In some cases, e.g. where there is a top layer of decomposed peat, it would be advisable to use a geotextile as a separating layer (see Chapter 11) to prevent local failure. The first stage of the embankment is usually wider than the next one, thus providing loading berms which improve stability condition in the subsoil (Fig. 9.2).
Precompression Technique
295
/
1II,,'~I//
j~
F
~"
/
Surchorge
2nd sta.qe Ist
stage
\
x
~'~
Lomdingberm
l 1~~_
I//,~,WlX,.~<.
Fig. 9.2. Cross-section of staged embankment.
9.2.2
Design considerations
The design of the staged embankment consists of: 9 dimensioning (height and width of embankment stages, inclination of slopes) .
planning the construction schedule (duration of periods between construction stages) depending on consolidation rate. The height and the slope inclination of the first stage of the embankment will depend on the shear strength of the organic subsoil. The stability analysis may be performed as for embankments placed in one stage (see Chapter 4). The time when the second stage of the embankment can be commenced, as well as its maximal allowable height, depends on the shear strength increase during the period elapsing since completion of the first stage. The shear strength increase can be anticipated on the basis of the empirical equations given in Chapter 4. For more important projects the shear strength tests should be performed. Since both aforementioned methods are based on the distribution of the effective stresses in the subsoil, the important part of the design is the consolidation prediction. When postconstruction settlements are to be minimised, e.g. a road embankment in the vicinity of a bridge abutment, the surcharging method is chosen. The use of a heavy surcharge can be assumed to reduce the postconstruction settlements due to both primary consolidation and secondary compression.
Surcharging to eliminate primary consolidation settlement: This is performed according to the schematic diagram given in Fig. 9.3. The broken line represents the predicted consolidation settlement So for the final height of the embankment (load qf), which has a final consolidation settlement equal to Sof. The solid line shows So versus time during and aider surcharging (load q,r)" The final consolidation settlement for the embankment with surcharge So(f+~) is also shown. When the solid line (settlement with surcharging) reaches the level of the final settlement of embankment S~e this is the time for surcharge removal t~r. At this time the
296
Staged Construction
consolidated layer will reach an average degree of consolidation: (9.1)
Up - Scf/So(f+sr )
Possibte load decrease
{:I 0 __1
Surclnarget:'[[;l~"!......I "due to uptift .............................................
qsr
qf
Final embankment
Time t
tsr
=,..__ v
i
Scf
~
q
tO 0 1:3 0 U1 E 0 0
I
Surcharge removal
Sc(f+sr)
Fig. 9.3. Surcharging to eliminate primary settlement.
However, the surcharge which is designed to obtain the settlement So equal to he predicted S~f(Fig. 9.3), has to fulfil an additional requirement. This is explained in Fig. 9.4. Let cr'~f equal the consolidation stress under the final loading and 6"v~ equal the consolidation stress under the surcharge fill. The cy'w profile at the time of surcharge removal t~ depends on the degree of consolidation U(f+sr), which was then achieved. In Fig. 9.4 the or' line corresponding to U(f+~r)= 50 % shows that a portion of the layer may undergo further consolidation after removal of the surcharge load. This is not the case when the O'v~ line corresponds to Uff+~O= 70 %. Therefore, to eliminate further primary consolidation settlement, following the removal of the surcharge, this should be kept in place until O'v~> O'vf throughout the entire soft layer. To fulfil this requirement, the degree of consolidation at the midpoint of the compressible stratum at the time of surcharge removal can be computed according to the following proposition by Johnson (1970):
Precompression Technique
297
CI
n wilt settle ,//~ after surcharge f l ' / / A movot
c-
I |
-7~176176
13.
n
\N .4--"
ffvs at
o
~
x
Sand
Fig. 9.4. Stresses at the time of surcharge removal.
CYvf log (1 + ~ ) CYv o Uff+~,:) = , (~ vf (T'vs log (1 + + ) (~' G' VO
(9.2)
VO
where CY'vs = the vertical effective stress under the permanent load with surcharge after excess pore pressure dissipation ~'vo = initial vertical effective stress Cy'vf = final vertical effective stress under the permanent load (see notations given in Figs. 9.3 and 9.4.). A solution of this formula, that is the value for U(f§ is given in Fig. 9.5. From this value the moment for surcharge removal can be indicated. The above requirement leads to the conservative design in which the actual precompression settlement will be greater than designed Sol. It is important to emphasise that the design of preload should be considered as preliminary and the decision to remove the surcharge should be based on field observations.
Staged Construction
298 ~,
100 I
.f, g o s0F
~
~.o
r~m
30
L
~176t
"5 ~
0
i
0.2
i
0.4
l
I
I
,
0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0 Surchorge to embankment Iood rcttio. ~vs/6'vf
Fig. 9.5. Diagram for estimating the surcharge to eliminate primary settlement (after Johnson, 1970).
Surcharging to compensate for secondary compression: This is particularly important for embankments on organic soils. Effects of the secondary compression may result in a significant settlement during the economic life of the embankment. This is especially evident when vertical drams are used, which drastically reduce the time required for primary consolidation. For a preliminary estimation of the reduction in secondary settlement due to surcharge, in normally consolidated soils, the diagram given in Fig. 9.6 can be utilised (Ladd, 1976). The notations used in the diagram are as follows: G'v~ = vertical effective stress after excess pore pressure dissipation under the load with surcharge ~'vf = final vertical effective stress under the permanent load only aider pore pressure dissipation C ae = modified coefficient of secondary compression tp = time at the end of primary consolidation t~r = time at which the surcharge is removed R = surcharge ratio According to Ladd (1976) for the first cycle of secondary compression the reliable surcharge ratio is obtained when the "maximum reduction" line is used. The time at which the surcharge is removed t~r should be close to the tp value, i.e. tfftp> 1. A preliminary assumption of the surcharge value can be based on the secondary behaviour of the Swedish clays described by Larsson (1981). It was found that
Precompression Technique 100
299 ~\,
,
80
\\
,
,
Minimum
.c_ .x2_ 60 c-
C 0
20 ~o
~
~
~
" -... /Maximum ~
0
0
10
20
Surcharge
Fig.9.6.
Averooe
[Reduction k, ----7='-"~, 30
/.,13
50
ratio R s = ( % s - % f ) / S v f , (%)
Reduction in rate of secondary compression due to surcharging (after Ladd, 1976).
secondary settlements will take place when the effective stress in the soil exceeds 0.8 ~'p (where C'p is preconsolidation pressure). Hence surcharging ought to result in the stress G'v~= ~' f/0.8, where C'vf = effective stress in subsoil under final load. This is valid only in limited depth, where change of additional load distribution is small by depth. For a more detailed analysis, the method given by Johnson (1970) can be used. The general layout of the surcharge design for partial compensation to secondary compression is given in Fig. 9.7. The degree of consolidation required under surcharge fill loading to produce primary consolidation plus the desired amount of seeondary compression can be expressed as follows:
U(f+sr)-Up (1 +
log
tso
)
(9.3)
where Up = degree of consolidation required under surcharge loading to produce settlement equal to primary consolidation C~ = coefficient of secondary compression = strain at the centre of the compressible stratum, caused by primary consolidation under permanent loading t~o = time determined by the useful life of the embankment or by the amount of secondary compression for which compensation is desired tp = time corresponding to primary consolidation under permanent load.
300
Staged Construction
/due touplifi Possible lo
[qf
decrease
,,,
t,tog time
=.._ v
tsc \/o.. \ v~;,
\R
~ S c f
t/)
Sc s r
~' s/e r e ~ v o~~
C
N
E lla U3
-\ \
\
c~s
~Sc(f.sr)
Fig. 9.7 Compensation for secondary compression by temporary surcharging.
The value of the required surcharge load can be estimated, assuming that the settlement S~, at the surcharge removal time t~ is" Scsr -- U(f+sr) " Sc (f+sr)
(9.4)
where S c(f+sr) = total consolidation settlement under permanent load plus surcharge. *Equation 9.4 is solved using successive approximation. Upon the removal of the surcharge swelling usually occurs, which depending on the magnitude of the unloading may be followed by slow rate settlement due to secondary compression in the overconsolidated state (C~), and subsequently by settlement due to secondary compression in normally consolidated state. The value
Precompression Technique
301
of swelling and its rate can be computed according to Terzaghi's theory from formulae: Ssw = Cse H-log
(Y"v(f+sr)
(9.5)
(Y'vf where C~e= modified swelling index, C](1 +Co) C ~ - swelling index H = thickness of the compressible layer CY'v(f+~r) = effective stress under permanent load and surcharge 6"vf= final effective stress under permanent load. Usually for compensation of secondary compression one cycle (t~o/tp=10) is used. On the basis of long-term observations of the post-construction settlement of an expressway built on peat, Samson (1985) observed that the swelling value is a function of the surcharge ratio, and the swelling time is approximately equal to the duration of surcharge.
9.2.3
Designparameters
The effective application of the precompression technique requires a detailed evaluation of the stress history, shear strength, compressibility and consolidation characteristics of the foundation subsoil and accurate location of drainage boundaries. In particular, the following parameters of organic subsoil are required: r
VO
cy p
= initial vertical stress preconsolidation pressure
Co = compression index C r = recompressionindex cv
= coefficient of consolidation
C a = coefficient of secondary compression ~f~ = undrained shear strength and its increase during consolidation c" and r = the effective shear strength parameters. For details, see Chapter 2 and 3.
Staged Construction
302
9.2.4
Limitations
An important feature of the staged construction technique is the relatively long time required to perform the works. Also, considerable time is needed for subsoil investigation. These facts, as well as the comprehensive construction monitoring programme, which is an integral part of staged construction technique, may prevent its use. On the other hand, there are no limitations in the application of the precompression technique from the soil conditions point of view; even a fibrous peat with a high water content need not be considered an obstacle when the precompression technique is used. It can therefore be concluded that a decision to use the staged construction technique is chiefly influenced by factors of economics and time.
9.2.5
An e x a m p l e - t h e Dalar@v~igen road
In order to illustrate the precompression technique, the construction of a road on organic soil, as presented by Carlsten (1988), is briefly described. The 850 m long section of the Dalarrv~igen road in Sweden was built on a peat bog, with a fibrous peat layer 2 to 3 m thick, underlain by a very thin layer of gyttja and organic clay (about 0.1 m). Under the organic soils there was a sand layer 0.5 to 2.0 m thick on top of another compressible layer of slightly overconsolidated clay and silt about 3.0 rn thick. The soil profile of the organic subsoil is given in Fig. 9.8. Undrained shear Degree of Water content strength kPa Description humification 1000 1500 2.5 5 7.5 10 Topsoil i
i
i
i
I
I
i
I
E
i
X
H3 I
i
X
X
X
Fibrous peat
H2
X
X
X
H4
rm
X
X
X XX
Peat/Gyttja
Fig. 9.8. Soil profile at the Dalar6v~igen site (after Larsson, 1986).
Precompression Technique
303
The embankment, 24 m wide at the crest, was constructed using the precompression technique. In the first stage, a fill about 1 m high was built. After about 5 0 days, when almost all excess pore pressure had dissipated, the second stage fill about 1.5 rn high including surcharge, was added. After almost one year the surcharge was increased by another 0.5 m, but only for halfa year; then a layer about 1 m thick was removed. This simplified construction schedule and the results of settlement and pore pressure measurements are given in Fig. 9.9.
60 50
Load,
Excess pore pressures
Excess pore pressure,
"----'0.91m'] below
kPa
. . . . 1.41m t . . . . . 1.91m
ground
surface
Z,030
~..
Load
/h-
20 10- ~ ~ . - - ~ . . . : _ a _ _
. . . .
..
----
.~.
9
~ 400
i
__.__
:.____. ~ .. =_ ____:~
3'0
~
Road ready for use ~
" " ~
..
.,.=.~,.~,,..
12o
~:.
'~..-4-..~._%~. ," "qI \ - ~"~ ./.
2~o
~',~, 9~o
~95o
3~o
~-
Time, days (log scale)
800 1200
Settlement, m m
Fig. 9.9.
Measured settlement, excess pore pressures and load at DalarSv~igen site (after Carlsten, 1988).
The measurements show that the surcharging reduced and delayed the post- construction settlements. The embankment with road being used by traffic settled during the first four years 10 to 20 mm. The total settlement during the 18 months of construction was 1.2 m, of which 1.0 m occurred in the peat layer and 0.2 m in the clay layer. Atter removing the surcharge a minor swelling was observed (Fig. 9.10).
304
Staged Construction E E
~:-~o 1 ~, ~o ~
w
93 o
'
"
~2o
'
'z~o
w
9
1920
Time, days Fig. 9.10. Swelling and settlement versus logarithm of time at Dalar6v~igen site (after Carlsten, 1988).
9.3
VERTICAL DRAINS
9.3.1
Introduction
To accelerate the rate of consolidation beneath the embankment, vertical drams are used which shorten the drainage path of the pore water (Fig. 9.11). Because in general, the coefficient of permeability for horizontal direction kh is higher than that in vertical direction kv, the effect of the consolidation acceleration is essentialy increased.
9
9
.
9
9
Organic
I
oit
9.'Send
I 9~
9
9
.
I
9
"i"
I
.
It
~
',:1" . - - 1 ~ ~ 1
~
I P~
~
9
I
I
verticat drains
" .
.
. ". "."
Fig. 9.11. Vertical drains under an embankment. In the bottom of the fill there normally is a 0.5 m thick drainage blanket.
There are two general types of vertical drains: 9 sand drains, and 9 prefabricated, band-shaped drains. Sand drains can be defined as columns of sand, 150-600 mm in diameter, installed in the soil by different techniques: displacement and non- or limited displace-
305
Vertical Drains
ment techniques. When using displacement technique, the drains are installed with the aid of a mandrel consisting of a closed, hollow pipe which is driven to the bottom of the layer to be consolidated. The mandrel is then filled with sand under air pressure and gradually lifted out of the ground. The hinged valve located at the end of the mandrel opens and the sand under pressure is squeezed out, forming the sand drain. When using the non-displacement technique, the drains are installed with the aid of a hollow continuous flight auger. When the auger reaches the desired depth, sand is introduced under pressure inside the auger and is left in the soil when the auger is rotated in the opposite direction. Water jets can also be used for sand drain installation. The cycle time for these installation methods is 5 to 20 minutes. Sand drains can also be enclosed in fabric. Prefabricated drains are most commonly in the shape of a band (strip) consisting of a plastic core surrounded by a synthetic (geotextile) sleeve (Fig. 9.12). The plastic core with grooved channels or protruding studs (rarely of mesh-type material) provides a flow path along the drain, and supports the sleeve, which in turn serves as a filter separating the core with its flow channels from the soil. Most often the bandshaped drain has dimensions similar to those designed by Kjellman*, i.e. about 100 mm wide and about 4 mm thick. Prefabricated band-shaped drains are installed in the ground with a mechanical rig equipped with a mandrel protecting the drain (Fig. 9.12) during penetration into the soil. The cycle takes 1 to 5 minutes.
Cross-section of mandrel
E .
Fig. 9.12. Prefabricated band-shaped drain.
*) W. Kjellman developed in the late 30's band-shaped drains made of cardboard strips with longitudinal grooves (Hansbo, 1979)
Staged Construction
306
In the last decades prefabricated band-shaped drains have mostly replaced sand drains, mainly because of speed and simplicity of installation, resulting in economic advantages, and also because of less disturbance of the soil compared to sand drains. Hansbo (1986) states that "prefabricated drains have an obvious advantage over the sand drains in that their quality and characteristics can be well defined and the risk of necking eliminated". Only prefabricated band-shaped drams will be considered in the following text, but principles are mainly the same for sand drains.
9.3.2
Installation of vertical drains
Band-shaped drains are generally inserted into the soil foundation usmg the equipment schematically shown in Fig. 9.13. The installation rig (most often a piling machine) is equipped in the drain delivery arrangement. The main part of this is the mandrel, which protects the drain during penetration into the soil. The cross-section of the mandrel is mostly rhombic, and rarely rectangular or circular.
Upper main
rotter
/
.
//
Drain
stQtic
route
guie /Z
~ l/
III \11 III
co ,o,-- 41 Ill
Drain
IlL/'--Drain
IW III
Mandrel
Lower m o i~ roller
////////////////////////////,5. 9 .
. .
.
, .
.
.. .
,
.
'
.
.
9
.
.
.
.
.
Fig. 9.13. Equipment for installation of the prefabricated vertical drains
(after Rixner et al., 1986),
307
Vertical Drains
There are two methods of installation: static and dynamic. In the former case, the mandrel is pushed into the soil by its own weight, in combination with dead- weight or the weight of the rig. In the latter case, vibrators similar to those used to install piles or a conventional drop hammer are used. It is not yet clarified which of the methods is better, but in the case of dynamic installation it should be carefully considered whether or not the excess pore pressure induced by installation will affect the soil properties (Hansbo et al., 1981). In organic soils static installation seems preferable. Before the mandrel is inserted into the ground an anchor is placed at the end of drain. After the mandrel is withdrawn, the drain is cut above the ground surface (working mat) leaving extra length for a drainage blanket, if this has yet to be built. In situations where the surface layer consists of dense or frozen soil or other materials difficult to penetrate, the predrilling technique may be appropriate. Before the vertical drain installation, the standard site works preparatory to the embankment construction, should be carried out. It is advisable to grade the ground surface with a slope less than 5 % to aid proper installation of the drain. During installation of the vertical drains in organic subsoil, usually a drainage blanket is used as a working mat to support the installation rig. It is considered preferable to build the drainage blanket prior to the installation the drains. This ensures free drainage between the drains and drainage blanket, and minimises disturbance of the surface soil layer, which is particularly important in cases where it is composed of peat.
9.3.3
Design considerations
When vertical drains are used, the vertical and horizontal (radial) drainage occur simultaneously in the soil. The degree of consolidation is then the result of the combined effects of vertical and horizontal drainage and can be expressed by the Carillo equation: U = 1- (1-Uv) (1-Uh)
(9.6)
where U
= degree of consolidation
Uv,U h= degree of consolidation due to vertical and horizontal drainage, respectively. The degree of consolidation is treated below as the ratio of the settlement at any
Staged Construction
308
time to the total primary settlement that is expected to occur. This is the average degree for the entire layer analysed. Because the evaluation of U~ has been discussed in Chapter 5, only consolidation due to horizontal (radial) drainage will be considered. The basic solution for radial drainage with sand drams was given by Barron (1944). The Barron solution was adopted for prefabricated drams, and modified for practical purposes in consideration of soil disturbance due to drain installation and well resistance by Hansbo (1979). The disturbance of the soil around the dram due to the installation process results in a decrease in the horizontal permeability. The well resistance depends on the discharge capacity qw ofthe dram, which may be too limited in the case of long drains. A solution of the Barron-Hansbo consolidation equation which is presented in Chapter 5 gives the following relation for evaluating the consolidation time t required to achieve a predetermined degree of consolidation (the variables of the equation are shown in Fig. 9.14): D2~t t=~ l n 8%
1 ~
(9.7) 1-Uh
where D = diameter of the soil cylinder, influenced by the drain, which is 1.05 times the spacing when drains are placed in a triangular pattern, and 1.13 times the spacing for a square pattern (see Fig. 9.14) ~t = factor including the effect of drain spacing = ~tl, effect of soil disturbance = la 2 and effect of drain resistance = ~t3, factor ~ = lal+Ja2+~3 c h = coefficient of consolidation in horizontal drainage U h = average degree of consolidation due to horizontal direction The effect of the drain spacing g l can be calculated with the relation: ~t1 = ln(D/dw) - 0.75
(9.8)
where" d w - equivalent diameter of drain, which according to Hansbo (1979) is equal to 2(a+b)/rt, and to Rixner et al. (1986) dw = (a + b)/2 a, b = dimensions of band-shaped drain.
309
Vertical Drains d s
Radial drainage
only
~ /
U
/ / /
-Vertical discharge capacity
?,,"?, / /
_.! kh
/
k
/
Square pattern" D = 1.13 S .[~
/
/ / / / / / / / / /
_
J
,
,
Triangular pattern' D = 1.05 S
/ /
Undisturbed soil Zone of soil disturbance Drain
Impervious boundary
Fig. 9.14. Scheme of vertical drain, with drain resistance and soil disturbance; relationship of drain spacing to drain influence zone.
When the effect of soil disturbance has to be considered, the factor ~t2 i s la2 = ((k h / k ) - l) ln(d~/d w )
(9.9)
where k h = permeability coefficient in the horizontal direction in the undisturbed soil k = permeability coefficient in the horizontal direction in the disturbed soil d s = diameter of the disturbed zone dw = as before. The effect of the well resistance can be estimated by the relation:
'3
=
xz(L-z) (kh/%)
(9.10)
where z = distance from the open end of the drain L = length of the drain when open at one end (half length, when open at both ends) qw = discharge capacity of the drain, at the unit hydraulic gradient (i = 1,0) k h = as before.
Staged Construction
310
2/..
"~'
VI JE ..i-, tO
II i-
'~
E 15
!11
r .D
r I
d E ~ 12
/
/
/
;/7 o~ r j'
/
I
,I
/
/
//
/
i
Y',,,// ' /" /
/
I
I"
/ 7- / /
.;
/ /,. ~--/
/
.-
J
~~"
I ! // / / / . , ' ~ / 7 ~ ~'~~ l l//I //,~,:~'/r !I ! i I / / / .~ -jl 6 ! !//I//'//'.,f,~'~ ~ .J'-
.
.,f ~
..,I'
9
0
0.5
fl =D2)u/8ch,
E
1.0
1.5
in years
2
c~
.
0
0
1
.
.
.
.
.
.
.
2 fl " Ch, in m 2
Fig. 9.15. Diagram for estimating vertical drain spacing (Hansbo, 1989).
3
Vertical Drains
311
The ideal case, when soil disturbance and well resistance are ignored, may be accepted for preliminary design. However, the negative effect on the consolidation process of soil disturbance in particular cannot usually be neglected. By experience, for seemingly homogeneous soils, the following disturbance parameters can be assumed: dJd w = 2 and kh/k~= 2. On these assumptions, the maximum drain spacing required to obtain a certain degree of consolidation, one of the steps of staged construction design, can be estimated from the diagram given in Fig. 9.15 (Hansbo, 1989). When a more accurate prediction of the consolidation performance is needed, the effects of soil disturbance and/or drain resistance should be considered. For evaluation ofthe soil disturbance factor ix2 (Eqn. 9.9) the value of ratio d~/dw = 1.5-3 is assumed (Hansbo 1979), whereas permeability of the disturbed zone can be reduced up to 5 times (kh/k = 5) according to the macrofabric of the soil (varvic). Instead of computing drain spacing with factor Ix2 evaluated as above, a reduced value of the equivalent diameter dw can be introduced in Eqn. 9.8 (Hansbo 1979). The well resistance effect has a significantly smaller value because the discharge capacity of good quality band-shaped drains is high enough for drain resistance to be neglected, particularly when dram length is smaller than 10 rn (drain closed at the bottom) (Jamiolkowski et al., 1983). In CIRIAReport (Holtz et al., 1991) the authors state: "As long as qw (discharge capacity) is greater than 100-150 m3/year, there should be no significant increase in the consolidation time". When designing a vertical drain system in organic soils, it is important to consider the deviations from the assumption used in the Barron-Hansbo method. A very high variation of parameters during the consolidation process and the influence of the dissipation of pore water pressure on the rate of secondary compression (Ftirstenberg et al., 1983) can invalidate the applicability of the linear consolidation theory for prediction of consolidation performance.
9.3.4
Design parameters
The prediction of the consolidation rate in the case of vertical drains requires an evaluation of the same soil properties as for the precompression technique. Additionally, the estimation of the permeability coefficient in the horizontal direction k h as well as permeability coefficient in the disturbed zone k might be advisable. Particularly for more complex soil conditions with high sensitivity or distinct macrofabric (e.g. fibrous peat and gyttja) it is recommended to estimate the value of k from laboratory or field test, performed under test embankments. The ratio ka/k is generally considered to range from 1 to 5, exceeding the upper limit in sensitive or in macrofabric soils. For preliminary design it is suggested that k h/k = 2 (Hansbo, 1989) be used.
312
9.3.5
Staged Construction
Limitations. Antoniny case history
There are three main reasons why the application of band-shaped vertical drains may be limited or ineffective in organic soils. They are as follows: 9 the relatively high permeability, which can make the effect of vertical drains practically negligible 9 the secondary compression dominating the deformation process, leading to significant deformation of subsoil after the completion of the primary consolidation accelerated by vertical drains 9 large deformations under load, which can lead to harmful axial deformations e.g. buckling (kinking) of drains. The answer to the question of how far the aforementioned properties may affect the performance of vertical drains in organic soils cannot be explicit. Due to the great variety of organic soils and different construction conditions, each case has to be analysed separately. Observations of test embankments on amorphous peat and calcareous gyttja at the Antoniny site in the Notec fiver valley, Poland (Wolski et al., 1988), provide some experience concerning the performance of band-shaped drams in organic soils. This may be helpful when a decision regarding the use of vertical drams in organic soils of a similar type is undertaken. Two test embankments, of the same shape, were constructed in three stages. Under one of the embankments, vertical prefabricated drains (Geodrain type) were installed in a 1.2 m square grid. A cross-section of the embankment with vertical drains is given in Fig. 9.16. Properties of the organic subsoil are given in Fig. 9.17. The construction schedule (Fig. 9.18) was decided on the basis of the stability analysis using vane shear strength measured prior to each stage. Deformations, both vertical and horizontal, were measured with monitoring equipment (Fig. 9.16) installed under each test embankment. Settlements of subsoil under the embankment with vertical drains, both measured and calculated, are given in Fig. 9.18. The results of calculations made according to the Barron-Hansbo formula disregarding the smear and well resistance effects gave relatively good agreement with the measured values. This is probably due to the smooth installation and short length of the drains. Horizontal displacements under the embankment with vertical drains, measured with inclinometers, are given in Fig. 9.19. The maximum measured horizontal displacement was about 0.55 m, which related to the maximum settlement, equal to 1.9 m, gave a ratio close to 0.3. The maximum horizontal displacement was observed in the middle of the organic subsoil, which is slightly at variance with the shape de-
313
Vertical Drains
1
5.0 J, 3.6 1
f
4.5
Calc. soil
Gyttja
3.9
f
~<~J~.
35.0m 11.0m
~
9"
3.9
~'
~'
4.5
.Stage i"..~--,-~ .2 ~
3.6
~
f
5.O
r
oo_-'.
~
~ o
~
.r
oo
.".. :"..".'.'.;.'..i....-:"'"" Monitoring equipment
~
,_,_~"
.... "
~"Geodrains"'":"
"
A-settlement gauge o-magnetic settlement gauge =-hose settlement gauge f-inclinometer n. BAT piezometer II open standpipe
Fig. 9.16. Cross-section of the Antoniny test embankment (Wolski et al., 1988).
Soil description
Reduced Water content Density vane Preconsolidationpressure % P, Ps (tIm3) 1"fu (kPa) 613(kPa) 0 200 4130 1 2 3 0 5 10 0 10 20 30 v
i
.
i
!
'
t
u
!
v
OX
_.
Black, very calciferous " Wp wL
f
"~ amorphous n
3
Dark brown.very calcifer.
E .c" 4.
Yellow-white Lacustrine marl
.g, Grey gyttja - bearing 6 - ! ~, calcareous soil 77.8
Gr'een- grey calcareous gyttja Dense sand
-
9
0
-
ii,i , , , , i o
-
0-4o t--,Io
I,'-'I
9
I,,,-ie
;
,e
X
0
X
0
X 0 0
II,i I O -
"\1
:o
lii,-io
-
X X-
0
X
-
0
X-
0
X
-
0
"'tt
X-
I
0
j
!
10
i
I
200
Sensitivity (o)
!
10
t
20
6'v, (kPa)
Fig. 9.17. Soil properties at the Antoniny site (Wolski et al., 1988).
30
Staged Construction
314 - - - Acting toad. kPa ----- Thickness. m
E &5
Stage 3
.x4 ~. 8c .u_ C
~3
Stage 2
C QI
E2
"". . . . . . . . .
St.1
2O [.__3
C
R1
,,
E m0
~x 0.4 ,_ ,,,,
=.
U.O
1.2
lw.....
i
89
Time.years
"~x \\~, -~-" " , ,
observed values Barron- Hansbo
---
"~~z.z~,
$ ~ 1.6
Fig.9.18.
i
x
C o n s t r u c t i o n s c h e d u l e and s e t t l e m e n ts of the test e m b a n k m e n t with vertical drains at the A n t o n i n y site.
o--o prior stage 1-o--oprior stage 2- +--+prior stage 3O---O1 year after stage 3 loading" - - 2 year after stage 3 loading I-5
I-7
..~_j
1-6 t
+
,
~
,2
.3
.
,3
4
.4
,s
15
i6
6
i ,7
.7
8
~ ,8
0.2 0 {m)
0.4 0.2 (m)
0
J
l
/
t
-
-
/
Stage 3 Stage 2
I
1
,1
,2
f f
7! .2 3
u
Stage 1 E
C~
Peat
4
Iq 5
j
!. 6 Calcareous soil / gyttja
0.6 0.4 O.2 0 (m}
Sand
Fig. 9.19. Horizontal d i s p l a c e m e n t s u n d e r test e m b a n k m e n t with vertical drains at the A n t o n i n y site (Wolski et al., 1988).
Vertical Drains
315
scribed by the Poulos diagram (Fig. 5.13). This is probably due to stiffness at the boundary of the organic layer. The horizontal deformations developed mostly during the application of the load. A comparison of the horizontal displacements of the subsoil under an embankment with and without vertical drains given in Fig. 9.20 shows a difference in the distribution of displacements. Those measured under the toe of the drained embankment slope were smaller than in the case of the embankment without drains. This effect can be explained by differences in the pore pressure distribution, developed under the embankment with and that without vertical drains (Wolski, 1989). When analysing the first of the limitations mentioned in the beginning of this paragraph (relatively high permeability) reference may be made to the soil behaviour under the Antoniny test embankments with and without vertical drains, presented in Fig. 9.21 as deformation curves (settlement versus logarithm of time). It can be concluded that, in spite of a relatively high initial permeability of amorphous peat (k = 2 910-8- 8.10 -8 m/sec), the consolidation was accelerated from 2 to 5 times. It is worth mentioning that band-shaped drains accelerate the consolidation also in fibrous peat, as reported by Hansbo (1989). It is a matter of economic analysis whether vertical drains should be used in such situations. However, it must be emphasised that in the staged construction even a 50 % reduction ofthe consolidation period may essentially improve the time schedule of the construction works. To evaluate an importance of the second of the aforementioned limitations (secondary compression) the following can be stated. The results of more than two years of observation of the settlement presented in Fig. 9.21 (third stage) show that after a lapse of one year, the increase in settlement of subsoil with vertical drains is, for practical purposes, insignificant. It can be therefore assumed that the secondary compression need not be considered an essential limitation on the use of vertical drains in amorphous peat and gyttja, particularly if surcharging is used. The third objection to the application of vertical drains in organic soils, buckling, was also investigated at the Antoniny site in the Notec river valley (Koda et al., 1989). It was found that, in spite of a rather high relative compression of subsoil, exceeding 20 %, the configuration of the drain strip deformation can be classified as a "bending", not "kinking", type according to Lawrance and Koerner (1988). This may be entailed by large horizontal displacements, which occur in very soft organic uniform layers. The detected shape of bending makes the decrease in discharge capacity insignificant. It is worth mentioning that a possible limitation which may be entailed by a decomposition of the drain sleeve due to chemical reactions of the organic soils depends on the quality of the sleeve material. Good quality drains which have sufficiently resistant sleeve material do not exhibit any essential deterioration during the
316
Staged Construction
t2~
---- 4
E
-~ 3-
t
~ 1 f-""-::i:: 0 , 0 ----. ---
26o
'
without drains with drains
&o
'
20
0
Time
days
66o
86o
Time, t 1
i Sand
10 0
20
0
cm
Time, t 2
....-
!
j Peat
, '/i /
! Calc. I s oil
Ifrond
~,
10 0
20
0
20
0
cm
Fig.9.20. Horizontal displacements under slopes of embankments with and without vertical drains, Antoniny site (after Wolski, 1989).
usual period of operation (2-3 years). This was explicitly shown by the long-term tests at the Antoniny site. Discharge capacity of two types of Geodrains, with paper and polyester sleeves, was measured about 250, 500 and 1000 days after installation in the subsoil (Koda et al., 1986, 1989, 1993). The results, shown in Fig. 9.22,
317
Vertical Drains
01.,
i
!
i
i
i
I
Time, days
10 =
1
I
I
,
,
.
I
I
I
100
I
I
i
I
,
,,,,,4
,
,
~
I
I
I
i
J
J
m
f
m
I
I
!
I
I
I
I
I
2nd stage
\
0.Sm settlement in 120 days, with drains in 400 days. without drains
0.6
1000
!
~ ,
~
o.z.-
?
'
0.2
I
st stage
~
n,0"2
i
\
\
~" E 0.8 E
~ I~
I L
__
I
.
= 9
i
I
=
I
I
I
I
I
I
I
I
I
I
i
I
I
I
I
I
I
I
I
I
1
1
I
I
1
I
I
I
I
I
I
t/3
~'P I
3 rd stage
0.2 0.4
0.6 08.
--
II
----I
without drains with drains I
I
I I
I I I
I
~ ~~ ~ . . ~ . . ~ I
I
I
i l
1111
1
I
I
I I
Ii
Fig. 9.21. Consolidation performanceof the embankmentswith and without vertical drains, Antoniny site.
indicate that during almost three years of operation the discharge capacity qw of drains with paper filters decreased dramatically to less than 10 % of the initial value, while qw of drains with polyester filters was reduced by only 50 to 70 % and was therefore still much higher than the lowest permissable value of 1O0 m3/year.
Staged Construction
318
2500 "C
~ . .
~ 2000
er 1500
'
(~) 10 days ~ 250 days 500 days 1000doys
.
""- -
U
8 o (..)
ter filter
1000
o
u 500 r-
D
0 0
100 200 Total I~eral stress (kPa)
300
Fig. 9.22. Influence of stress and time on discharge capacity of vertical drains in peat (Koda et al. 1989).
9.4
CONSTRUCTION MONITORING
9.4.1
Introduction
Monitoring during the staged construction of embankments on organic soils aims at controlling the loading rate in each stage as well as creating a basis for the decision when the next stage can be commenced. In the case of surcharging, the time of the surcharge removal can also be decided. The aforementioned aims of monitoring can be achieved on the basis of reliable measurements provided by field instrumentation. An important component of the field measurements is evaluation of the shear strength increase, determined by vane tests or possibly by flat dilatometer or piezocone tests. When the precompression technique is associated with the use of vertical drains, the installation of drains should be recorded in detail, e.g. lengths of drams, as well as the rate of installation, verticality and soil resistance during the mandrel penetration should be monitored. Construction monitoring of the staged embanbnents comprises observations which make it possible to confirm the design assumptions. For this reason, monitoring should be regarded as an essential extension of design. In some periods of construction, the piezometer as well as the settlement gauge readings should be taken daily.
Construction Monitoring 9.4.2
319
Instrumentation
Selection of field instruments and their arrangement in the embankment subsoil is based on the requirement that the observational data should enable verification of the design analysis and, if necessary, the introduction of corrections to the designed construction. To provide the required observational data from the construction monitoring of a staged embankment three kinds of instruments are used: piezometers, settlement gauges and inclinometers. An example of the instrument arrangement in the foundation soil under a test embankment for staged construction is shown in Fig. 9.23.
2nd
......., J~
x
x
x
x
x x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x j'
3
Peat
x
x
x
x
x
x
x
x
x
x
x
x
x
x
x
4
F
6
f
1-piezometer,
%
3
- Gyttja .: .-...'.'. : .."
Ii
.i
" II
9. s6;~d . . . . . .: :" 9-'"' -" ' .-" ~
4-hose
/.
2-open
settlement
standpipe.
gauge,
3-multiple
5-settlement
point
platform,
settlement
gauge,
6-inclinometer
Fig. 9.23. Instrument layout for the stage constructed embankments.
Piezometers provide pore pressure values which in turn enable us to compute profiles of a vertical consolidation stress 6"v = CYv-u, which is needed for the evaluation of the shear strength increase. It is useful to place at least one piezometer in each stratum beyond the embankment toe; this makes it possible to measure normal variation of pore pressures during construction. In the case where vertical drains are used, piezometers placed before the installation of the vertical drains measure the initial ground water condition as well as the excess pore pressure caused by drain installation. In case a significant increase of pore pressure is observed, the shear
Staged Construction
320
strength used in stability calculations has to be corrected. Open standpipe piezometers allow boundary pore pressure to be obtained. Settlement gauges, particularly of the multiple point type in conjunction with pore pressure observations make it possible to verify the coefficients of consolidation c v and ch as well as compression indices Co. These in turn enable correction of the consolidation analysis. In order to follow the changes of the embankment profile, the hose settlement gauge is most appropriate (see Chapter 2); for the same purpose, settlement platforms are used. Inclinometers are usually installed beneath the toe of the embankment, and provide a profile of horizontal displacements. It is also advisable to place inclinometers beneath the toe of the second stage of embankment, particularly when the height of this stage is large. The inclinometer readings are of great importance. They can warn against impending failure if they reflect larger movements than were predicted. The spacing between cross-sections with monitoring equipment depends on the variability of soil conditions along the embankment route. In general, one wellequipped cross-section (as in Fig. 9.23) should be located at each section of the embankment with similar foundation conditions. It is advisable also to locate settlement gauges, preferably hoses, at intervals of 100 m along the embankment route.
9.4.3
Interpretation of the monitoring results
A successful application of the staged embankment technique depends on the monitoring of foundation soil behaviour. Observational data allow a decision concerning construction progress and, if necessary, changes in the embankment crosssection, e.g. an addition to the temporary berm. The interpretation of the monitoring results has two main goals: 9 to evaluate consolidation behaviour, and 9 to assess foundation stability. To reach these goals, two principal techniques of analysis are used: 9 direct estimation, e.g. an increase of the settlement rate indicates an impending failure, or pore pressure measurement which providesdatafor the estimation of vertical stress, needed for stability analysis (see Chapter 4); 9 verification of design parameters,e.g. - the estimation of a horizontal coefficient of consolidation back-calculated from piezometer and settlement data. Interpretation of the observational data for staged embankments is extensively discussed by Ladd (1988), who, among others, describes a case history which can be treated as an example of the direct estimation technique. Following is a short description of the case.
Construction Monitoring
321
A test embankment of a rectangular shape, close to a square, was built on 10 meter thick soft organic silty clay (Ip = 33 %, IL = 1) with constant initial strength cu - 12 kPa, and preconsolidation pressure slightly decreasing with depth. During construction of the first stage (height of about 3 m), foundation instability comprising failure of the north slope and a cracking oflae east and south slopes was observed. This led to the flattening of the slopes, which reached 1:8, and to the removal of slightly more than 0.5 m of the fill (Fig. 9.24).
4.0 E &
Flatten East Slope after Cracking
Flatten South Slopes /
3.0
S~
e-
I
"" , ~ ~
ll
2.0
~e/
I05
..._.
E o.~o C
Failure
11'0
11"5
120
1 89
"~T~---"~'~---+_~---..-._ ~ + Construction day ~ ~ ~ ~ -.~""',,--.-!._.~. Center
"v,,~ . ~ ~ Nort.hStope_~. " ~ ' ~
130
"'b,,,~ "x,,South Stope .. "-q'<'--~'-.o-5-o..o
~ 0.20
Fig.9.24. Settlements during first stage filling, which reflect instability of the subsoil (after Ladd et al., 1969).
The settlement data obtained from measurements of the settlement platforms located at the north slope show high settlement rates, in comparison to those locatedbeneath the centre of the embankment, during the week preceding the failure. As Ladd (1969) explains, there were "obvious signs of excessive undrained shear". During the following weeks, similar increases of settlement rate - but on a smaller scale - of platforms located on the east and south slopes were observed. They were followed by cracking. The failures of both slopes were prevented by flattening, which is strongly reflected in the settlement plots. The verification of design parameters technique is greatly advantageous when used in order to evaluate the coefficient of consolidation for horizontal flow towards
322
Staged Construction
vertical drains. As stated in 9.3, the installation of vertical drams causes disturbance of the soil, particularly as regards its permeability characteristics. Therefore, in order to correct the consolidation course prediction, it is important to evaluate the "effective" coefficient of consolidation. This can be done on the basis of readings from piezometers placed at the midpoint between vertical drams. As proposed by Ladd (1988) who followed the Asaoka (1978) technique (see Chapter 5), the measured excess pore pressure u~ (measured value minus equilibrium piezometric water elevation multiplied by the unit weight of water) should be plotted on a logarithmic scale versus time. The field effective coefficient of consolidation in the horizontal direction can be calculated using:
D2 g
Cat -8
In(uel/Ue2)
(9.11)
t 2 - t~
where D and g are defined as in equation 9.7. U~land Ue2 are excess pore pressure measured in times t 1 and t2, respectively. It should be emphasised that the value obtained, although close to the effective one, contains errors due to the assumption of the constant boundary conditions and the omission of drainage in the vertical direction.
9.5
CONSTRUCTION ASPECTS
The staged construction technique, besides ordinary earthwork specifications, requires a detailed description of monitoring procedure and, if vertical drains are used, a detailed description of the drains, their material and installation. Monitoring, a particularly important part of the staged construction process, consists in a detailed evaluation of the field performance of each stage of the embankment. By comparing the observed performance with the predicted performance, rational decisions can be made as to when to start the next stage, when to remove the surcharge load, and the allowable rate of filling. Therefore in the specifications for staged embankments the required increase in shear strength, the allowable increase in pore pressure, and the predicted rate of consolidation should be used for each stage and for selected areas of foundation subsoil. When the precompression technique is associated with speeding up consolidation through the use of vertical drains, the contractor should furnish all the necessary equipment and materials and perform all operations necessary for the installation of drains in accordance with drawings and with the requirements of the detailed specifications.
Construction Aspects
323
The specifications for prefabricated band-shaped vertical drains usually concern materials for sleeve and core as well as the quality of the assembled drain; they also include a description of the installation operations. The sleeve material should conform to specifications given in appropriate standards, e.g. minimum permittivity, and should be tested according to methods used for geotextiles. The quality of the assembled drams is usually expressed in terms of discharge capacity. For details concerning testing methods and requirements, the reader is referred to Christopher and Holtz (1984), Holtz et al. (1987) and Rixner et al. (1986). In the specifications it should be stressed that before selecting a drain whose behaviour in field conditions similar to those in the analysed project is unknown, full-scale tests are strongly advisable (Hansbo, 1986 and Holtz et al., 1989).
9.6
DESIGN EXAMPLE FOR STAGED EMBANKMENT WITH THE USE OF VERTICAL DRAINS
9.6.1
Introduction
The example explains the geotechnical design procedure for a river dyke, constructed by stages on very soft organic subsoil. The design calculations in this example (final settlement, consolidation course and stability analysis) have been carried out using the classical methods presented in Chapters 4 and 5. According to these methods, the calculation parameters, except for coefficient of consolidation and shear strength, were assumed to be constant for the first and second construction stages. Also some other simplifications have been used. In the case of more important embankments, particularly those on deep organic soils, the procedure presented in this example should be treated as a preliminary one. For a more accurate analysis it is necessary to take into account the variability of parameters during the course of consolidation and to use more precise methods based on numerical calculations considering large- strain analysis. This enables the prediction of the variability in elevation of particular subsoil layers during the deformation process, which is helpful for the correct estimation of the shear strength increase and consequently of the embankment stability in each construction stage. The designed dyke, which will be constructed of sand, should have the final height hf= 2.75 rn above the terrain level, during an economic lifetime assumed to be t - 50 years. The geotechnical characteristics of the subsoil at the embankment site are given in Figs. 9.25, 9.26 and in Table 9.1.
324
Staged Construction
Reduced vane Precon soli dat i on Water content (%) Density, ~ (t Im 3) Tfu (kPa) pressure,61; (kPa) 0 200 400 1 2 3 5 10 10 2O 30
#
"6
g_ ~3
f
./
/
I I
t.-----4-tl-
\
\
x~4
/
@
/
f
~o
k
o
Fig. 9.25.
/
/
o 5 o
/
\
\
Initial soil properties.
I
ld65
d
Verticat effective stress, gv (kPa) 10, 20, 40, 80,
0
4 --
----
160
Peat 8yttja
10-7
U
10.8 Fig. 9.26.
I
i
~
I
Coefficients of consolidation versus vertical effective stress.
Design Example for Staged Embankment - Vertical Drains
325
Tab. 9.1. Parameters of foundation soil. i
Parameter
Symbol
Unit
Peat
Gyttja
Density of soil Liquid limit Plastic limit Plasticity index Undrained shear strength Initial void ratio Compression index Recompression index Swelling index Coefficient of consolidation (initial*) Coefficient of secondary compression Coef.of sec. comp. after surcharging
p wL Wp Ip ~fu e0 Cc Cr CS cv Ca C~
t/m 3 % % % kPa " m 2/s
1.10 313 189 134 8.5 5.5 2.5 0.28 0.20 1.5.10 -7 0.15 0.03
1.40 103 54 49 9.0 2.8 O.8 0.08 0.05 2.0.10 -8 0.044 0.010
* values taken for the first stage embankment; for the next stages cv was taken from Fig. 9.26.
Taking into consideration that the embankments on organic soils usually settle more than 1/3 of their height, in this example it is conservatively assumed that the initial height h o = 4.7 m. The stability analysis of so high an embankment results in an insufficient factor of safety, and therefore construction in stages has to be chosen. As the desired time of completion of the construction is two years; two stages, lasting one year each, are used in the design (Fig. 9.27). The first stage is 2.5 m high, the second is 2.2 m.
5 ..--.,.
E 4 IE
3
I
.u_ 2 tI--
stage
2nd
1
oiF 0
I
1 st stage
o
&t1
I I
at 2
_1 "-1
26o
.6o
soo
66o
Time, {days) Fig. 9.27. Schedule of the embankment filling.
I
760
Staged Construction
326
9.6.2
Stress distribution under the embankment axis during the first stage
T
a= 7.5 m
0
b= 10.0
,,s
,,% F
\
ql = 45 kPo
x xPeQtx x x
o
x
x
x
,z Tfu=8.5
x
- &.7,:,-"
Tfu =9.0 kPa
Sana: Fig. 9.28.
Cross-section of the dyke, first stage.
S T E P 1. Initial effective vertical stress:
O'vO = zp'g where p ' = density of submerged soil (cf. Table 9.1 and Fig. 9.28) g = 10 m/s 2 Thus z (m)
9" (t/m 3 )
0.0
cy -r (kPa) 0.0
1.1 0.5
5.5 0.1
1.5
6.5
0.1 3.0
8.0 0.4
5.5
18.0 0.4
8.0
28.0
J
=1.27
Design Example for Staged Embankment- Vertical Drains
327
Obtained values o f initial effective vertical stress (effective overburden) a'vo as well as preconsolidation pressures (3" p, according to the oedometer test results, are plotted in Fig. 9.29.
0
0
20
/..0
Vertical
stress,6v(kPa)
60
80
100
120
t52.2
,,,-..,,
E 3
,,013"
~b > 6"vo
u~ /.. o c
U
Cvo:zf'g ff,//5,,'6va " "
1 a . o \ y / / ' . , ' / , / / / / / 5 , ~.61.2 ,,NC"
Fig. 9.29. Vertical stress distribution due to the first stage of loading.
STEP 2. Stress distribution caused by the first stage embankment load computed at the embankment centre line, according to Osterberg's formula given in Chapter 5"
(3" 9v l (a) = (3" , vo + m O " v l
where AC3vl - 2 I ql I
= f(a/z,b/z) - the influence value from diagram given in Fig. 5.21.
ql
= hi " Pl " g - 2.5 91.8 910 = 45 k P a
328
Staged Construction
Thus z (m)
a/z
1.5 3.0 5.5 8.0
5.00 2.50 1.36 0.94
9.6.3
b/z
6.67 3.33 1.82 1.25
t
I = f(a/z, b/z)
Ao vl (kPa)
(3"'vl
0.50 0.49 0.48 0.47
45.0 44.1 43.2 42.3
51.5 52.1 61.2 70.3
(kPa)
Prediction of the immediate and consolidation settlements at the first stage of embankment loading
S T E P 3. The immediate settlement
estimated according to the procedure explained in Chapter 5, from formula 5.7: - Sil ,
II
Sil = (4 I~ ql B)/E. where I v = 0.036 - the influence factor from the equation I v = (1 -v 2 )fl, derived from equation 5.7 (Chapter 5), for L = 10 B, and v = 0.5 ql = 45 kPa - unit load of embankment B = (17.5 + 10)/2 = 13.75 m - reduced embankment width, E u = undrained modulus. Undrained modulus is computed according to the empirical formula (equation 5.6): E-
(~f~ 215 ln(F))/Ip
where zf~ = undrained shear strength, from Tab. 9.1 F = 1.27 - safety factor computed for the first stage of loading according to Bishop's method (Fig. 9.28) Ip
=
plasticity index from Tab. 9.1.
Design Example for Staged Embankment - Vertical Drains
329
Then:
for the peat layer
E u = (8.5-215-1n(1.27))/1.24 = 352 kPa,
for the gyttja layer E u = (9.0.215.1n(1.27))/0.49 = 944 kPa, and weighted average undrained modulus
E u = 722 kPa
Thus, immediate settlement is" Sil = (4.0.036.45.13.75)/722 = 0.12 m, in this: 0.07 m for the peat layer, and 0.05 m for the gyttja layer (depending on Eu). The above computed immediate settlement resulted mainly from the horizontal displacement which at depth 1.5 m evaluated from nomogram in Fig. 5.15, is equal to Shl= 0.13 m.
STEP 4. Settlement due to primary consolidation- Sol , computed according to Chapter 5 using formula 5.21" Cr
(Y'p
Sol = ~
H log ~
l+eo
Cc
+~
Cr'vo
H log
(Y vl
l+eo
p
where Co, Cr, e o- from laboratory tests, given in Tab. 9.1.
In the analysis, the subsoil has been divided into four strata. The thickness of each stratum does not include the immediate settlement. The primary consolidation is assumed to begin immediately after loading. Hence Z
9 (Y v0
o" (Y'vl(a) Hi (kPa) (k~a) (kPa) (m)
Cc /(1 +e0)
Cr/(1 +e0)
0.0-1.5 1.5-3.0
4.0 7.2
18.5 17.5
49.9 51.8
1.5 1.5
0.385 0.385
0.043 0.043
3.0-5.5 5.5-8.0
13.0 23.0
23.0 24.0
56.7 65.8
2.5 2.5
0.211 0.211
0.021 0.021
(m)
Sci (m) 0.29 0.30 0.59 m - peat 0.21 0.23 0.44 m - gyttja
Total S cl-- 1.03 m
Staged Construction
330
9.6.4
Consolidation performance at the first stage of loading
To enable the construction of the second stage of the embankment, it is necessary to achieve the average degree of consolidation for the whole subsoil U = 0.8. STEP 5. The consolidation time of soft subsoil, to achieve U = 0.8, without use of the vertical drains, is computed from formula 5.29: Assuming only vertical drainage (U h = 0), the average time factor T v, for the whole subsoil, from nomogram (Fig. 5.26), is 0.57. The consolidation time for the peat layer, assuming only upward drainage (the coefficient of consolidation for gyttja c v = 0.001728 m 2/day is drastically smaller than for peat, Cv= 0.01296 m 2/day, so gyttja can be treated as impervious): t~ = (TvH2)/Cv= (0.57"3.02 )/0.01296 = 396 days The consolidation time for a gyttja layer with upward and downward drainage: t 2 = (Tv(0.5 H) 2 )/Cv= (0.57.(0.5.5.0) 2 )/0.001728 = 2062 days Because the second stage loading is to be applied when the degree of consolidation is U = 0.8, and no later than after one year, vertical drains should be installed to reduce the consolidation period. S T E P 6. Estimation of the vertical drain spacing according to Chapter 9.3: ] i
The vertical drain spacing is computed to reach U = 0.80 in the gyttja layer, which is of smaller (therefore controlling) permeability, within one year. The degree of consolidation U vfor gyttj a is" Tv=(365.0.001728)/(0.5.5.0) 2= 0.10, and from nomogram (Fig. 5.26) Uv= 0.35 The horizontal degree of consolidation U h is computed from Carillo's equation 9.6 (when U v = 0.35): Uh=l -((1- U)/(1-Uh) ) = 1 - ( ( 1 - 0.80)/(1- 0.35)) = 0.70 The optimal vertical dram spacing is estimated from the Barron-Hansbo equation (9.6), in which: equivalent diameter: dw = (2.(0.004 + 0.10))/rr = 0.066 m, for strip
Design Example for Staged Embankment- VerticalDrains
331
drain dimensions a = 0.004 m and b = 0.10 m, and consolidation coefficient in the horizontal direction %= 2%, which is often assumed for preliminary computations, hence:
Ch-- 2Cv= 2 (2" 108 m e/s) - 0.00346 m 2/day - 1.26 m 2/year Taking the above values into consideration and noting that: Uh= 0.70, the following Barron-Hansbo formula is obtained: D2 t -
D (ln( ~
8:0.00346
tma x --
365 days, and
1 ) - 0 . 7 5 ) ln( ~
0.066
)
1-0.70
The diameter of the cylinder of influence of the drain D, may be obtained using the successive approximation method: D
t
(m)
(days)
(days)
2.00 1.60 1.80
463 272 361
> 365 < 365 -- 365
- too large - too small - correct spacing
To obtain the consolidation degree ofgyttja layer U - 0.80 after one year, vertical drains in the square pattern with spacing S (see Fig. 9.14) should be installed: S = D/1.13 = 1.80/1.13 = 1.59 m The dram spacing can also be estimated by means ofHansbo's nomogram (based on the assumption dJdw=2 and kh/k =2) given in Fig. 9.15. D=l.70 m, and consequently the drain spacing in a square pattern: S = D/1.13 = 1.70/1.13 = 1.51 m Drain spacing S = 1.50 is selected.
9.6.5
Total settlement at the end of the first stage
Due to significant differences in permeability and consequently in the rate of consolidation of the peat and gyttja layers, the settlements should be calculated separately.
332
Staged Construction
STEP 7. Total settlement of peat layer after one year of loading:
Using formulas as in STEP 5, it is estimated that the average consolidation degree U = 1.00, will be achieved in tp = 110 days. Therefore, the primary settlement of the peat layer will be completed in 110 days and will equal: S~l= 0.59 m After this period, the secondary settlement will progress until the second stage loading. The thickness of the peat layer subjected to the secondary compression will be: H" = H
-
Sil
-
Sol = 3.0 - 0.07 - 0.59 = 2.34 m, and
the settlement due to secondary compression of the peat layer in one year, according to Chapter 5, using the formula 5.24: S~l : (C a/(l+ep )) H' log(tf/tp ) where C~ = 0.15, from laboratory tests, given in Tab. 9.1; ep
= 4.07, void ratio at the end of primary consolidation;
H' = 2.34, thickness of peat layer. Hence S~l = 0.030.2.34.1og(365/110) = 0.04 m The thickness of the peat layer before the second stage of loading is as follows: H ' = H - Stl = H - (Sil + So1 + S~I ) H ' = 3.0- (0.07 + 0.59 + 0.04) = 2.30 m STEP 8. Total settlement of the gyttja layer after one year of loading:
In the gyttja layer, after one year when the average degree of consolidation has reached U = 0.80: then U v = 0.35, and U h = 0.70 (from Carillo's equation 9.6). The primary settlement of the gyttja layer will be equal to:
Design Example for Staged Embankment - Vertical Drains
333
Sol = 0.80"0.44 = 0.35 m
The thickness of the gyttja layer before the second stage of loading is equal to: H ' = H - Stl = H - (Sil + Sol ) H ' - 5.0 - (0.05 + 0.35) = 4.60 m
9.6.6
Shear strength increase due to the first stage of loading
The increase in shear strength of the soil is a function of the effective vertical stress and stress history (see Chapter 4).
STEP 9. Vertical stress distribution before the second stage of loading: Due to the settlement a lower part of embankment will be sunken in the ground water. The height of the submerged part of embankment at the end of first stage of loading will equal to" h~(~l = 0.07 + 0.59 + 0.04 + 0.05 + 0.35 - 0.5 = 0.6 m Because the embankment have sunken under the ground water level, the uplift effect should be considered as a continually developing value (iterative procedure) during a significant part of the first stage of loading. However, for simplification, in this example the uplift effect will be considered only at the end of the first stage of loading. Therefore the effective stress at the end of the first stage of loading is equal to: ~'vl(e) -- ~'vl(a) - A~'vl(u)
where CY vl (a) = assumed (without uplit~) effective stress, under the first stage of loading
A~'v~(u~= decrease in the effective stress due to uplift effect, caused by the embankment settlement below ground level (after a lapse of one year): A~'vl(u~= 2 1 ('Pw g h~(u~)
Staged Construction
334 Thus z
a/z
b/z
I=f(a/z,b/z)
(m) 1.15 2.30 4.60 6.90
-Aa'vl(u )
ovl(~~ (kPa)
~'vl(e) (kPa)
6.0 5.9 5.7 5.3
51.5 52.1 61.2 70.3
45.5 46.4 55.4 65,0
(kPa) 6.52 3.26 1.63 1.09
8.70 4.35 2.17 1.45
0.50 0.49 0.47 0.44
Computed values are given in Fig. 9.30
00
20
Vertical stress, 6'v (kPa) 60 80 100
40
120
9
"- Sancl Peat 3 E. d 4 c-
o 5
.C I---
6
7 8
9
.
l=:::
~a
"."
-~vl
.
.
~, :~.5 V
m.O
I~v1.1~8- :
=used in ~'~
!
Ls2. \
stability \ ~, \ catculation~ ~-" \ 9 / "~ . . . . Gyttjd ~.\-P-vl-~ 9
\ \
\65.0
\61.1 \\
~ \\ \ 70.3
Fig. 9.30. Vertical stress distribution before the second stage of loading.
STEP 10. Shear strength increase" The shear strength increase due to consolidation under the applied load of the first stage embankment is estimated according to the empirical formulae 4.22 and 4.27.
Design Example for Staged Embankment - Vertical Drains
335
zf~ - K s O"vl(e ) where K=
m S (ESL) no
E S L = (Cr'p)o/(Y'vl(e) In which S- normalized undrained shear strength in the normally consolidated state at E S L = 1. F r o m n o m o g r a m Fig. 4.21 the following values were obtained"
9
under centre line:
- for peat
rno = 0.15, S = 0.50,
- for gyttja
mno = 0.10, S = 0.45;
9 under slope" - for peat
rno = 0.10, S = 0.45,
- for gyttja
m o - 0.10, S = 0.40.
(or p )o- initial preconsolidation pressure, from the diagram in Fig. 9.29, ~'vl(~ - vertical effective stress before the second stage of loading.
In the peat stratum after a lapse of one year, U = 1.0, then: cy vl(e)
45.5 91.0 = 45.5 kPa,
In the gyttja stratum after a lapse of one year, U - 0.80, then: cr vl(o~ 5 5 . 4 - 0.8 = 44.3 kPa. Thus the shear strength:
Staged Construction
336 9under the centre line:
peat gyttja
(G'p)~ (kPa)
G'(kPa) vl(e)
ESL
I,(
'l:fu (kP a)
18.0 23.5
45.5 44.3
0.392 0.527
0.435 0.422
19.8 18.7
l~a)O
~'vl(e)
ESL
K
~f~ (kPa)
18.0 23.5
24.7 29.7
0.728 0.790
0.436 0.391
10.8 11.6
9under the slope:
peat gyttja
(kPa)
Using the shear strength values estimated above, the safety factor obtained for the second stage embankment from Bishop's method for the critical circle is equal to 1.22 (Fig. 9.31). 0
a=6.6m
~
:
~
-
f
b= 3./., ,
.
I
\
. ~q2 =
;_:~=_~:~13z .
;, S9 a n. d
.
9
Fig. 9.31.
. "
Cross-section
.
"
.
'
.
'
-
,
,
.
~ -
u =18.8 kPa ',
9 .
9 .
"
.
9 .
\
"
. . .
./
I
9
. . .
"
..
,"
,"
.
'
.
"
,
of the d y k e at the s e c o n d s t a g e of l o a d i n g .
"
,,
,"
,'
.
Design Example for Staged Embankment- Vertical Drains
337
Stress distribution in the subsoil under the embankment centre line, due to the second stage of loading
9.6.7
STEP 11. Vertical stress distribution due to the second stage of loading: The computation procedure is similar to that used in STEPS 2 and 9.
CY'v2(a) -- l~'vl(e ) -t- A~'v2 where
1~ v2(a) = assumed (without uplift) effective stress, due to the second stage of loading (~'vl(e) = vertical stress due to the first stage loading A(Y'v2 = stress increase due to the second stage of loading, ACr'v2 = 2 I Aq2 where I = f(a/z,b/z) - influence value from diagram given in Fig. 5.21. Aq2 = h 2 9p . g = 2.2" 1.8.10 = 39.6 kPa The calculations were carried out taking into account a decrease in subsoil thickness due to settlement after one year. The stress state at the beginning of the second stage of loading is (cfFig. 9.32):
z (m)
a/z
3.65 4.80 7.10 9.40
1.81 1.43 0.93 0.70
b/z 0.93 0.71 0.48 0.36
I---f(a/z,b/z) 0.47 0.42 0.39 0.33
ACY'v2
O'vl(e )
(kPa)
(kPa)
O'v2(a) (kPa)
37.2 34.8 30.9 26.1
45.5 46.4 55.4 65.0
82.7 81.2 86.3 91,1
Staged Construction
338
0 0
I..
20
'
..
Vertical stress, 6v (kPa) 40 60 80
1
,-Peat
~
.
9
"
9 ,
~
.
.
/
"
/'/
| /
.
.
",
9
. .
.
,
.
-
o
.
.
.
~'82.7
3
~g
!
~
4
IJ r
5-
r"
120
100 9
6v2 =6"vl + a 6 v 2 -
Gyttja
Fig. 9.32. Vertical stress distribution due to second stage of loading.
9.6.8
Prediction of the total s e t t l e m e n t St2 u n d e r the s e c o n d stage of loading.
St2 = Stl 4- Si2 + 5c2 4- Ss2
where Stl -- 1.10 m - total settlement after one year of consolidation due to the first stage of loading Si2, S~2, S~2 - immediate, consolidation and secondary settlements entailed by the second stage of loading.
Design Example for Staged Embankment - Vertical Drains
339
S T E P 12. Immediate settlement (see STEP 3): ii
Si2 = (4 1v Aq' 2 B 2)/E2 where Iv
- 0.0868 - influence factor from calculation (Chapter 5)
Aq' 2 = 2 1 Aq2- 2-0.47.39.6 = 37.2 kPa B2
- (10+3.4)/2=6.7m
E2
- (l:f~ 215 ln(F 2))/Ip
where ~f~ = 19.4 kPa for peat, and 18.8 kPa for gyttja (see STEP 11) F 2 = 1.22 (see STEP 10) Ip (see Tab. 9.1),then: for peat layer
Eu2 = (19.4.215.1n(1.22))/1.24 - 669 kPa,
for gyttja layer
E 2 = (18.8-215-1n(1.22))/0.49 = 1642 kPa, and
weighted average undrained modulus E 2 - 1318 kPa
Thus Si2 = (4.0.0868.37.2.6.7)/1318 = 0.06 m, in this: 0.03 m for the peat layer, and 0.03 m for the gyttja layer (depending on E ) . Thus computed immediate settlement was mainly due to the horizontal displacement, which evaluated from nomogram in Fig. 5.15 at depth 1.15 m is equal to Sh2 = 0.05m.
STEP 13. Settlement due to primary consolidation" The increase of stress caused by the second stage of loading is beyond the preconsolidation pressure, settlement is computed according to Eqn. 5.21. S~2 = (Co/(1 + e I )) H" log((S'vZ(a) / (Y'vl(e))
Staged Construction
340 where eI =
3.98 - for peat, and 2.50 - for gyttja layer,
Cr = assumed the same as at the first stage of loading, H' = thickness including the settlement after a lapse of one year of consolidation, ~'v~(~), CY'v2(a)- according to Fig.9.32 Thus
Z
O'vl(e)
~'v2(a)
H'
CJ(1
0.00-1.15 1.15-2.30
45.5 46.4
82.7 81.2
1.15 1.15
0.502 0.502
0.15 0.14 0.29 m - peat
2.30-4.60 4.60-6.90
55.4 65.0
86.3 91.1
2.30 2.30
0.229 0.229
0.10 0.08 0.18 m - gyttja
(m)
(kPa)
(kPa)
(m)
Sc2i
+ el)
(m)
Total S~2 =0.47 m
STEP 14. Secondary settlement: S~2 = (C~/(1+%2)) H' log(tf/tp) where ep2 = 3.29 - for peat, and 2.34 - for gyttja, C~ = from laboratory tests (the same as at the first stage of loading), H'=
thickness of the computed stratum, including the settlement after a lapse of one year of consolidation,
tf = 50 tp, the same as at the first stage of loading, assuming 50 years as the economic life of the dyke. The calculations were carried out for two strata (see STEP 4).
341
Design Example for Staged Embankment - Vertical Drains
Thus C~/(l+ep) peat laver gyttja layer
0.035 0.013
H' (m)
(m)
1.98 4.39
0.12 0.10
Ss2 i
Total Ss2 = 0.22 m
I
!
II
STEP 15. Total settlement St2 when the second stage of loading is completed" Ill
Illl
Ill
II II
II
I
St2 : Stl -at- S'ol + Si2 + 5c2 "+- S s2 where S'Q1 = the rest of primary settlement in second year of consolidation under the first stage of loading for gyttja: S'cl
=
0.44 - 0.35 = 0.09 m (see STEP 8)
Thus St2 = 1.10 + 0.09 + 0.06 + 0.47 + 0.22 = 1.94 m The submerged part of embankment (see STEP 9) will equal /~12(w) :
S t 2 - S t l --
1.94 - 1.10 = 0.84 m
For simplification the uplift effect in the submerged part of the embankment will not be considered as a continually developing process.
9.6.9
Consolidation performance at the second stage of loading.
Equations and procedure as in STEP 6 are used.
STEP 16. Horizontal degree of consolidation" Uh2 - horizontal degree of consolidation after a lapse of one year of consolidation (second stage of loading).
342
Staged Construction For peat
c v = 6.0.10 8 m2/s = 0.0052 me/day, and c h = 2c v = 0.0104 me/day,
For gyttja c v = 1.5"10 -8 mZ/s = 0.0013 mZ/day, and % = 2c v = 0.0026 m2/day. For the peat layer: (1.7) 2
1.7
365 =
(ln( 8"0.0104
1 ) - 0.75) l n ( ~ ) 0.066 1-Uh2
,
thus:
Uh2 = 0.99,
9 For the gyttja layer: (1.7) 2 365 =
1.7 (ln(
8.0.0026
1 ) - 0.75) In (-
0.066
),
thus"
Uh2 = 0.65.
1-Uh2
STEP 17. Vertical degree of consolidation after a lapse of one year of consolidation" 9 for the peat layer: T=
(t c v)/H "2 - (365.0.0052)/2.302= 0.36
from n o m o g r a m 5.26:U~2 = 0.69 9 for the gyttja layer: T v = (t c v)/(0.5 H') 2 = (365-0.0013)/(0.5.4.60) 2 = 0.09 from nomogram5.26:Uv2 = 0.35.
STEP 18. Mean degree of consolidation after a lapse of one year of consolidation: 9 for the peat layer:
U 2 -- 1 - ( l - Uv2)(1 - Uh2) -- | - ( 1 -
0.69)(1- 0.99) = 1.00
Design Example for Staged Embankment - Vertical Drains
343
9 for the gyttja layer:
U2-- ] - (] - Uv2)(1 - Uh2 ) = ] - (l - 0 . 3 5 ) ( 1 - 0.65) = 0.77
9.6.10
Settlement of the subsoil after a lapse of two years from the beginning of consolidation
STEP 19. Settlement and thickness of the subsoil after a lapse of two years: I St2-- Stl-t- S'cl + Si2 + Sc2 where St1 = settlement at the end of the first stage of loading,
Stl
-- 1 . 1 0 r n
S'ol = 0.09 m (see S T E P 15) Si2 = immediate settlement at the second stage of loading,
Si2 = 0.06 rn So2 - consolidation settlement at the second stage of loading.
9 for the peat layer, when U 2 = 1.0,
SQ2i = 0.29.1.00 = 0.29 rn
9 for the gyttja layer, when U2-- 0.77,
SQ2i = 0.18-0.77 = 0.14 m Total $ 2 = 0.43 m
Thus S t 2 - 1.10 + 0.09 + 0.06 + 0.43 = 1.68 m. The thickness of the subsoil strata after a lapse of two years of consolidation: 9 Peat layer:
H"H"-
9 Gyttja layer"
H " = H'- S'cl - Si2 - 8c2
H ' - Si2 - SG2 2 . 3 0 - 0.03 - 0.29 = 1.98 m
H " = 4 . 6 0 - 0 . 0 9 - 0.03 - 0.14 = 4 . 3 4 m
344
Staged Construction The total thickness after a lapse of two years of consolidation:
H " = 1.98 + 4.34 = 6 . 3 2 m Thickness, (m) Load. (kPa)
8C
t
1st stage
2 nd sta~e _ . . . . . . .
[
I I
I
.J
0
~6o ,..-..
36o
26o
i
I 1
I 2
T ime, ( year s) 8~
(days)-
0.5"
E
,,_,.
c 1.0-
m
E
- "
" r ' -
_e 1.52.O-
Fig. 9.33. Settlement of embankment versus time.
('N
~
q
P3
9
'i:':
.... :", 0
03
X
X
X
.
.
.
.
.
.
" 2od l~
~
x Peat x x X
_,_,
X
1st stage
~
=
O0
0.66__~
r
m
C3
ud
-
03
8yttjo_-
..--.~ r
i ~na:: :! ,
9
.
.
.
.
.
.
.
. . .
.
.
.
9
.
.
., . .
Q
Fig. 9.34. Cross-section of the dyke one year after second stage of loading.
S T E P 20. Settlement and thickness after a lapse of 50 years"
The settlement after a lapse of 50 years, from the beginning of consolidation:
St5 0 -- Sil 4- Sol + S'sl-F Si2 4- Sc2 + Ss2 St50 = 0.12 + 1.03 + 0.04 + 0.06 + 0.47 + 0.22 = 1.94 m
Design Example for Staged Embankment - Surcharging
345
Therefore the crest of the dyke will be elevated above the terrain during the period of 50 years more than: 4.70 - 1.94 = 2.76 m, which is in agreement with the desired value of 2.75 m. The thickness of the subsoil after a lapse of 50 years from the beginning of consolidation: Hs0 = H - St50 = 8.00 - 1.94 = 6.06 m
9.6.11
Final remarks
The analysis presented above concerns deformation and stability in the conditions of the construction period. The shape of the dyke obtained should now be analysed in the operating conditions (steady seepage, sudden drawdown etc., see Chapter 4). In case of an insufficient safety factor, a flattening of slopes or pressure berms will have to be constructed.
9.7
DESIGN EXAMPLE FOR THE STAGED EMBANKMENT WITH SURCHARGING
9.7.1
Introduction
A road embankment is to be built in the immediate vicinity of the dyke analysed in Chapter 9.6, with the same soil conditions, The crest of the road embankment should be placed 1.8 m above the terrain and because of the road requirements the settlement during at least 10 years must be minimised. Therefore, in order to eliminate primary consolidation settlement and to delay the settlement due to secondary compression beyond the period of 10 years, staged construction with surcharging has to be applied. According to the requirements, the embankment should be ready for pavement works two years from the beginning of construction. The embankment is assumed to be constructed in two stages. The first loading stage is assumed to be analogous to that used in the dyke design (Chapter 9.6). The second load will be partially removed after one year. A principle aim of this example is to estimate the required surcharge. For the first stage of the embankment, all computation results obtained in the dyke design, (STEPS 1 - 10 in Chapter 9.6) are utilised. Taking into consideration the results obtained in the design of the dyke (Chapter 9.6) it can be assumed that the total settlement will not exceed 30 % of the final thickness of the embankment. Therefore, for further computation, the final height is assumed to be hf = 3.5 m.
346
Staged Construction
7.Sin
{
o:6.6
{b:S.t.{
.--"-~ su,-d',a,.m 22--
~
" "
9 ...
:.
.to.F, ".." lst's'ta~le' "'. "". ~j...~ .... . . . . . . . . ~
Peat
Gyttja
:"S6nd.
9.
"
".
9. .
9" . :
. : ii . .
" "" "
'..'.'
." ..' 9 ". "" "- :
Fig. 9.35. Cross-section of the road embankment to estimate the required value of the surcharge.
9.7.2
Required height of the embankment at the second stage of loading
In order to achieve a delay of the secondary compression settlement, the final height o f the embankment has to be temporarily increased by surcharging.
STEP 1. Estimation of the surcharge load" The surcharge load can be approximately evaluated from the diagram given in Fig. 9.6, using the surcharge ratio
R = (~'w- ~ in which the stresses can be approximately replaced by the respective height o f the embankment. Referring to Table 9.1, the ratios needed for the diagram are obtained: C~ ~ 9 for peat
0.03 =
Ca
C~ ~ 9 for gyttja Ca
- 0.20 0.15
0.01 = ~ = 0.044
0.23
Design Examplefor Staged Embankment- Surcharging
347
Making conservative use of the peat value, the ratio R = 0.29 for the average t/tp is obtained from the diagram. Hence the required height of the embankment with surcharging is: hs+f= R
9h f + h f
where hf= 3.5 m - final height of the embankment (after removal of the surcharge), thcn
hs+f - 0.29"3.5 + 3.5 - 4.52 m The above value can also be roughly verified using the simple relationship given by Larsson for Swedish clays (see Chapter 9.2.2)" ~'w - CY'vf/0.8 where ~'w - effective stress due to surcharging together with the final load, Cy'vf - effective stress due to final load. Replacing, as before, stress by the height of the embankment, the minimum height of the embankment with surcharge is: hs+f= 3.5/0.8 - 4.38 m
On account of the approximate character of the methods used, the value hs+f = 4.70 m, is conservatively accepted in the subsequent computations.
9.7.3
Prediction of the settlements under the second stage embankment (with surcharge)
The height of the second stage of embankment with surcharge is the same as in design of the dyke embankment (Chapter 9.6). Therefore, the vertical stress distribution given in Fig. 9.32 can be used. Consequently, as in the case of the dyke (see Chapter 9.6), the total settlement of the second stage embankment, before the surcharge is removed (after a lapse of one year) is: St2-- 1.68 m
Staged Construction
348
9.7.4
Vertical stress distribution after removal of the surcharge
In order to estimate the swelling due to discharging, we need to know its effect on the vertical stress.
STEP 2. Estimation of vertical stress distribution: (3",vf = (Y,vf+s(a)- A6'v2(u ) - ~ ' v s
where o'~f
= effective vertical stress after removal of the surcharge,
o vf§
= ~v2" according to STEP 11,
-ACr'v2(,) = stress decrease after the second stage of loading due to uplift caused by embankment settlement below the ground water level before the surcharge removal (see STEP 9). -AO"v2(u ) = 2 1 (-Pw g Ah2(~)) ~ ( u ) -" St2 - Stl = 1.68 - 1.10 = 0.58 m
Thus z (m)
a/z
b/z
I=f(a/z,b/z)
-AO'v2(u) (kPa)
0.99 1.98 4.15 6.32
14.24 7.12 3.40 2.23
3.43 1.72 0.82 0.54
0.50 0.49 0.47 0.44
5.8 5.7 5.4 5.1
-O'vs = 2 I (-qsr) where -q~ = 1.2.1.8.10 = 21.6 kPa z+h 1
a/z
b/z
I=f(a/z,b/z)
3.49 4.48 6.65 8.82
1.89 1.47 0.99 0.75
0.97 0.76 0.51 0.39
0.47 0.44 0.40 0.35
(m)
I
20.3 19.0 17.3 15.1 I
Design Examplefor Staged Embankment- Surcharging
349
Thus Z (m)
1~"vf+s(a)
(kPa)
-n(3' 'v2(u) (kPa)
(3"'vf+s(e) (kPa)
-O'vs (kPa)
O "vf (kPa)
0.99 1.98 4.15 6.32
82.7 81.2 86.3 91.1
5.8 5.7 5.4 5.1
76.9 75.5 80.9 86.0
20.3 19.0 17.3 15.1
56.6 56.5 63.6 70.9
00
I
20
.
--~
1
40
I
1
60
I
I
I
Vertical stress, 6v(kPa) 80 100 .
i
!
I
120
I
.
Embankment .
.
.
.
.
.
L6"vf. s
Peat
56.6 56.5
E
r v u 5 tI---
6vf
0
_~v~.9 ~-~v2 75.5
,
6"vf+,
Gyttja
8
I
99
1
.
I
,
I
9
,70.9 \
Decrease due 182.7 to uplift after [81.2 2nd stacje
i
.3
~i.o \ ,~.9~.~,
9
Fig. 9.36. Vertical stress distribution after removal of surcharge.
! 9
9
o
350
Staged Construction
9.7.5
Prediction of the swelling behaviour after removal of the surcharge STEP 3. Calculation of the swelling value (Eqn 9.5)"
S~w - - (C~/(1 + e 2 )) H " log(C~'vf+
a(e)/~'vf)
where C
= swelling index, given in Tab. 9.1 ( C ~ - 0.12 - for peat, and C gyttja),
e2
-
0.04 for
void ratio after a lapse of two years of consolidation (e 2 = 3.29 - for peat and e2= 2.22 - for gyttja),
H" -
thickness of the subsoil strata after a lapse of two years of consolidation
(H"=
1.98 m - for peat layer, and H"= 4.34 m - for gyttja layer).
Thus z (m) 0 - 1.98 1.98 - 6.32
o'w (kPa)
r (kPa)
H" (m)
C~/(1 +e2)
Sswi (m)
56.6 63.6
76.9 80.9
1.98 4.34
0.028 -0.008 0.012 -0.006 Total Ssw= -0.014 m
STEP 4. Calculation of swelling time according to the Carillo equation (9.6): Coefficients of consolidation assumed according to Fig. 9.26, for the reduced stress due to unloading 9 peat:
c v = 3.5-108mZ/s = 0.0030 m2/day, and %= 2.%=0.0060 m2/day
9 gyttja: Cv= 1.5.10SmZ/s - 0.0013 mZ/day, and %= 2.%=0.0026 mZ/day Thus 9 in peat: i f t ~ - 730 days, then:
U h - 0.98,
if t~ = 1095 days, then:
U h - 0.995, U v - 0.95, and
U v - 0.75, and
U = 0.99,
U = 1.0,
Design Examplefor Staged Embankment- Surcharging
351
and 9 in gyttja: if t~ = 1460 days, then:
U h = 0.96,
U v = 0.61, and
U = 0.98,
if t~ = 1825 days, then:
U h = 0.98,
U v = 0.78, and
U = 1.0.
Time of swelling in peat is 3 years, and in gyttja 5 years.
9.7.6
Prediction of the secondary compression behaviour of the subsoil
After the swelling period the delayed secondary compression will start. In the initial period the secondary compression will develop in preconsolidated soil. This will be followed by the settlements due to secondary compression in normally cons o l i d a t e d state.
STEP 5. Time of secondary compression in "OC" part: S~w = (CaV(1 +e2) ) H " l o g ( t ' / t ) where 9 in peat:
C~ ~ = 0.03, e 2 = 3.29, H " = 1.99 m, S w = 0.008 m, t~ = 1095 days
9 in gyttja: Ca ~ = 0.015, e 2 = 2.30, H " = 4.35 m, S~w- 0.006 m, t = 1825 days
Thus 9 in peat:
t'~ = 4106 days = 11.5 years, and
9 in gyttja: t ' -
4618 days = 13.0 years.
After a lapse of the periods computed above, the road embankment will settle due to secondary compression.
Staged Construction
352
Thickness
4 --60-
....'Ea
40-
2
El O
"J 20-
1
O:
C
--.]
Load
Decreased due to uplift I .
,
.
.
.
.
.
.
...
,._ .=== . . . . .
_=
I
1 st stage 100
0.'I
I
,
i
0.5
Primary consolidation
,
,
,,
1
Surcharging
I L I
'i
2nd I stage I
I I i I
I I
11000 !
2
,
I
1
5
I
~j
.....
I
I I I I ,
i
,,,
I I
10
I
I I
0.5-
I
I
I IjO0~30
Time
I
112'0
I I
100 (years)
t
I I I I I Secondary I Secondary
t
ISweUing ~ompress~n Icompressaon
..-..
E
I~
~1 .
I
"~ 1.0E
!
.
.
.
in.OC"
I
i
in .NC"
I
...., ~q
1.5About 0.02m heave
2.0Fig.
9.37.
Subsoil
settlement
performance.
9.7.7 Final remarks Due to the assumptions made, the computation presented m this example has to be treated as preliminary. Therefore, during the construction period, careful monitoring of the foundation behaviour is required, because changes in the construction progress or of the cross-section of the embankment may be necessary.
References
9.8
353
REFERENCES
Asaoka, A. (1978). Observational procedure of settlement prediction. Soils and Foundations, Jap. Soc. of Soil Mech. and Found. Eng., Vol. 18, No. 4, pp. 87-101. Barron, R.A. (1944). The influence of drain wells on the consolidation of finegrained soils. Diss. Providence, US Eng. Office. Carlsten, P. (1988). Peat - Geotechnical properties and up-to-date methods of design and construction. State-ot-the-art report. Swedish Geotechnical Institute, Varia 215. Christopher, B.R. & Holtz, R.D. (1984). Geotextile Engineering Manual, prepared for Federal Highway Administration, National Highway Institute, Washington D.C. Fiirstenberg, A., Lechowicz, Z., Szymanski, A. & Woiski, W. (1983). Effectiveness of vertical drains in organic soils. Proc. of the 8th Eur. Conf. on Soil Mech. and Foun. Eng. Helsinki, Vol.2, pp. 611-616. Hansbo, S. (1979). Consolidation of clay by band-shaped prefabricated drains. Ground Eng. Vol. 12, No. 5. Hansbo,S., damiolkowski, M. & Kok, L. (1981). Consolidation by vertical drains. Geotechnique, Vol. 31, pp. 45-66. Hansbo, S. (1986). Preconsolidation of soft compressible subsoil by the use of prefabricated vertical drains. Annals des travaux publics de Belgique. No.6. Hansbo, S. (1989). Full-scale investigations of the effect of vertical drains on the consolidation of a peat deposit overlaying clay. De Mello Volume, Rio de Janeiro, pp. 159-165. Hansbo, S. (1989). Vertikaldr~inering av jord. Metodblad, Byggforskningsr~tdet, Stockholm. Holtz, R.D., Jamiolkowski, M., Lancelotta, R. & Pedroni, S. (1991). Prefabricated vertical drains. CIRIA report, Butterworth-Hainemann, London. Holtz, R.D.,Jamioikowski, M., Lancelotta, R. & Pedroni, S. (1989). Behaviour of bent prefabricated vertical drains. Proc. of the 12th Int. Conf. Soil Mech. and Found. Eng., Rio de Janeiro, Vol.3, pp. 1657-1660. Jamiolkowski, M., Lancellotta, R., & Wolski, W. (1983). Precompression and speeding up consolidation, S.O.A. and General Report, Proc.of the 8th Europ. Conf. on Soil Mech. and Found. Eng. Helsinki, pp. 1-26. Johnson, S.J. (1970). Precompression for improving foundation soils. ASCE, Jour. of the Soil Mech. and Found. Div. Vol. 96 SM 1, pp. 111-144.
354
Staged Construction
Koda, E., Szymanski, A., & Wolski, W. (1986). Laboratory Tests on GeodramsDurability in Organic Soils. Archiwum Hydrotechniki, Vol. XXXIII, No. 4 pp. 491-498. Koda, E., Szymanski, A. & Wolski, W. (1989). Behaviour of Geodrains in organic subsoil. Proc. of the 12th Int. Conf. on Soil Mech. and Found. Eng., Rio de Janeiro, Vol.2, pp. 1377-1380. Koda, E., Szymanski, A. & Wolski, W. (1993). Field and laboratory experience with the use of strip drains in organic soils. Can. Geot. Jour., Vol. 30, No. 2, pp. 308-318. Ladd, C.C., Aldrich, H.P. & Johnson, E.G. (1969). Embankment failure on organic clay. Proc. of the 7th Int. Conf. on Soil Mech. and Found. Eng., Mexico City, Vol. 2. Ladd, C.C. (1976). Use ofprecompression and vertical sand drains for stabilization of foundation soils. M.I.T. Dept. of Civil Eng. P76-4. Ladd, C.C. (1988). Stability evaluation during staged construction. The twentysecond Karl Terzaghi lecture, Jour. Geot. Eng. Div. ASCE. Lawrance, C.A. & Koerner, R.A. (1988). Flow Behavior of Kinked Strip Drams. Proc. of the Symp. sponsored by the Geot. Eng. Div. of the ASCE, Geotechnical Special Publication No. 18. Larsson, R. (1981). Drained behaviour of Swedish clays. Swedish Geotechnical Institute, Report No. 12, Link6ping. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Report No. 29, Link6ping. Rixner, J.J., Kramer, S.R. & Smith, A.D. (1986). Prefabricated Vertical Drains, Vol. 1, Engineering Guidelines, FHWA Report RD-86-168 Washington D.C. Samson, L. (1985). Postconstruction settlement of an expressway built on peat by precompression. Can. Geot. Journal, Vol.22, No. 3, pp. 308-312.
Wolski, W., Szymansld, A., Mirecki, J., Lechowicz, Z., Larsson, R., Hartlen, J., Garbulewski, K. & Bergdahl, 13. (1988). Two stage-constructed embankments on organic soils. Swedish Geotechnical Institute, Report No. 32, Link6ping. Wolski,W. (1989). Some aspects of application of band-shaped vertical drains in organic soils. De Mello Volume, Rio de Janeiro, pp. 545-551.
355
Chapter 10
Lime and Lime/Cement Columns P. Carlsten, Swedish Geotechnical Institute
10.1
DESCRIPTION OF THE METHOD
Lime and lime cement columns are columns of stabilised clay 0.5-0.6 m in diameter. In lime stabilisation, finely milled, burnt lime is mixed with soft clay using a lime column machine (see Fig. 10.1). In lime/cement columns Standard Portland cement is added to the lime. Normally the proportions of lime/cement in percent per weight are 50/50 in lime/cement columns. In clays that react positively to mixing the soft clay is converted into a firm clay resembling a dry crust. The soft clay outside the stabilised zone is practically unaffected. The shear strength and compression modulus of lime and lime/cement columns are considerably higher than in the unstabilised clay. Columns improve the bearing capacity of sot't clay according to the spacing between the columns. Owing to the higher compression modulus of the columns, consolidation settlements beneath a loaded surface will be smaller than for an unstabilised clay. This description is based on Swedish experience. In other countries other diameters of the lime and lime/cement columns are used, in addition to different techniques for installation. The clay mixed with lime and cement in the columns is not homogeneous. When mixed with lime and cement, lumps of stabilised'clay are formed. The shear strength in the joints between the lumps is lower than within the lumps. The mixing technique also leads to the formation of layers in the columns approximately at fight angles to their longitudinal axis. The permeability of lime and lime/cement columns is considerably higher than that of unstabilised clay. The irregular structure of lime/cement columns leads to the following results: 9 the shear strength varies in different directions within the column 9 the column acts as a vertical dram and speeds up the consolidation settlement For organic soils with high organic content (> 6 %) and high water content (> 120 %) lime has little or no stabilising effect, see Chapter 10.3.3. This means that the use of lime or lime/cement columns in peat or gyttja should be avoided. For these kinds of soils other methods of construction are more suitable.
356
Lime Columns
Lime and lime/cement stabilisation of other organic soils has been tested with good results. However, there may be big differences between one type of clay and another. For this reason it is important to test the reaction on the organic clay both in laboratory and in field tests. The most common application is for layered organic and mineral soil or pure mineral soil. Research is being performed to test whether other chemicals can be used for stabilisation of soft clays. For instance lime-cement mixtures in different proportions have been tested (Almberg et al, 1995). The investigations show that the addition of cement to the quick-lime results in a good stabilisation effect in organic clay with a water content of 100-140 % unlike stabilisation with lime only. In Japan cement is used as a stabilising agent with good result. The text in this chapter is to a large extent taken from a handbook published by the Swedish National Road Administration (1986) and from a handbook (Carlsten and Ekstr6m, 1995). The handbooks are based on experience from research at the Swedish Geotechnical Institute and on the Design Handbook published by Lime Column AB, Sweden (Broms, 1984).
10.1.1
What happens inside a lime/cement column?
9 The burnt lime reacts with the water in the clay, so that the lime is slaked. The reaction is accompanied by heat generation. Slaking is a rapid process and is terminated within 1 hour aider the lime is mixed in. 9 In ion exchange, which follows slaking of the lime, the dissociated divalent calcium ions change places with the clay mineral's univalent ions of calcium, sodium and ammonium. Since each calcium ion can bind two clay particles, the structure of the clay becomes coarser. The ion exchange and structural conversion of the clay begin as soon as the lime is added. The strength thereby starts increasing immediately, but it may take several months for the ion exchange to be completed. After a rapid initial increase, the strength develops at a gradually slower rate until the process is concluded. 9 In the third stage, the lime that has not been consumed by ion exchange reacts with the silicates and aluminates in the clay. Compounds are formed with strong binding power and low solubility. These act as a binder between the clay particles and thereby increase the firmness of the lime column. This hydraulic stabilisation of the lime and clay mix continues for several years. A condition for these chemical reactions taking place is that the pH in the columns is at least 8 aRer mixing in the lime. A low ground temperature at the ground surface may delay and even halt the chemical process.
357
Description of the Method
I
-T __7
-11 -.U
li
,'~ I"1"1 ~.1 ot::_- - -.~'.~: ",J "I ol ' L
mU=-L-_- _ - .
Rotary table Mixi_.___._nng toot
Fig. 10.1. Lime column machine.
9 When cement is added to the mixture the cement reacts with the water and this results in a higher strength. 9 During the first months stabilisation with lime normally gives relatively lower strength compared to stabilisation with cement. In the longer run the shear strength many times becomes higher in soil stabilised with lime only. Some aspects on choice of suitable admixtures to different kinds of soils are given in 10.3.3. 10.1.2
Installation
of lime/cement
columns
When installing lime/cement columns, the shaft with the mixing tool is driven down to the bottom of the planned location for the column at about 100 mm/revolution. This process has the effect of churning up the clay within the borehole. The
358
Lime Columns
admixture is then mixed with the clay at the same time as the mixing tool is withdrawn under rotation in the opposite direction. The admixture is forced out with compressed air at a prescribed quantity per metre through a hole m the upper part of the mixing tool. The rate of withdrawal during lime mixing is normally 25 mm/ revolution. When mixing cement in to the columns their is a higher demand on mixing and therefore lime/cement columns normally are performed with 15-20 mm/ revolution. Columns can be installed at a maximum inclination of 30-70 ~ and to a maximum length of about 20 m. The admixture feed is stopped 0.2-0.5 m below the ground surface so that the lime is not blown into the open air. Where there is a dry crust thicker than 1 m, the admixture feed is normally stopped 0.5 m above the bottom edge of the crust. Since the lime/cement column increases the bearing capacity of soft clay and reduces and hastens consolidation settlements under loaded surfaces, the method is suitable for foundation reinforcement in: 9 road embankments 9 equalising settlements in road embankments when building piled structures
(transitions) 9 foundations for culverts and light bridges 9 stabilising excavated and natural slopes 9 reduction of earth pressure against sheet piling in clay
10.2
REQUIREMENTS FOR FIELD AND LABORATORY INVESTIGATIONS
10.2.1
Field Investigations
In addition to the determination of the thickness, relative firmness, possible layers of non-cohesive soil and stone and strength and deformation properties of the soft soil layers, the following aspects in particular must be studied: 9 composition, thickness, firmness and content of tree roots and stones in the soil, if any 9 filling of the ground surface and its composition and thickness 9 occurrence of fixed obstacles to installation of columns, such as overhead and buried cables 9 ground water and pore pressure conditions Samples must be taken in the soft soil layers for determination of the strength and
Field and Laboratory Investigations
359
deformation properties of the natural soil. The samples are also used for studying how the soil reacts to mixing with lime and/or cement.
10.2.2
Laboratory investigations
The laboratory investigations primarily involve the properties of the natural soil and as a rule also the reaction of the samples to the admixture. In Swedish practice the soil samples are classified through the following routine investigations: 9 type, bulk density, water content, liquid limit In the case of organic clays, the organic content is also determined. Undrained shear strength and sensitivity are determined with the fall-cone method. If the settlements in an embankment on lime/cement columns are to be investigated, it is necessary to know the compression moduli and preconsolidation pressure of the clay. These parameters are determined through compression tests.
Investigation of soil~admixture samples: Using the results of the classification tests on soil samples, the soil profile is divided into characteristic layers with similar properties. Primarily, layers of organic and inorganic postglacial clay and glacial clay are differentiated. During installation of lime/cement columns, the quantity of the admixture fed per unit length is constant along the entire column. This means that the required quantity of admixture per metre of the column is determined by the characteristic layer where the growth in strength is lowest. The strength is investigated at different times after mixing. Usually, tests are performed on three to five occasions. Five stabilised samples are thus required from each characteristic layer. The effect of the admixture is studied through tests on five occasions after mixing, normally after 10, 20, 30, 90, 180 and possibly 360 days. The tests provide a basis for evaluating the suitability of the lime/cement column method compared to other strengthening methods, as well as for dimensioning column spacing and setting up a time schedule for column installation and loading of the stabilised area. The following properties are determined each time a test is performed: 9 bulk density, water content, shear strength according to unconfined compression tests. If planning also comprises the settlements within the stabilised area, compression tests on laboratory samples are normally carried out 90 days after adding the admixture.
360
Lime Columns
The unconfined compression tests reflect the average shear strength of the sampies. The beating capacity of lime/cement columns beneath a road embankment on a horizontal ground surface is normally judged from the results of these unconfined compression tests.
10.2.3
Test columns
In large projects or where there is doubt about the effect of mixing in lime, for example, in soils with organic contents - a more reliable basis for planning can be obtained by the installation of lime/cement columns as a complement to laboratory tests. The test columns provide information on possible problems with installation, such as hard layers in the soil, and also on the performance of the lime/cement columns. Test columns should be installed in groups of 6-10 so that at least 3 colunms can be examined on 2 different occasions after installation. The lime/cement content in the test columns is chosen on the basis of laboratory tests or earlier experience.
10.3
DESIGN CONSIDERATIONS
10.3.1
Introduction
Reinforcement with columns in road embankments is primarily aimed at increasing the bearing capacity of the soil. The lime/cement columns are therefore dimensioned with regard to the stability of the embankment. The high compression modulus and permeability of lime/cement columns as compared to unstabilised clay provide a further advantage in that settlements in the embankment are smaller and are concluded faster than if the clay had been unstabilised. In order for the road to remain on the planned level when opened to traffic, the embankment is filled to an excess height corresponding to the calculated settlement plus a temporary surcharge. Owing to the interaction between the proportion of lime/cement columns per surface unit and settlement of the embankment, dimensioning involves successive trial calculations of stability and settlement before the total thickness of the embankment fill is decided. In general, the calculation process is similar to dimensioning of embankments on vertical drains. Lime/cement columns are often dimensioned on the basis of shear strength in soil/admixture samples determined with unconfined compression tests. If test columns have been installed, dimensioning of the column reinforcement is primarily based on the performance of the test columns. The longer the time that testing is
Design Considerations
361
allowed to continue, the more reliable will be the data for technically optimal dimensioning of the column reinforcement. As a minimum requirement, the shear strength should be measured at least 3 months after mixing. Since the strength of lime/cement columns increases with time after mixing, the extent of the strengthening will depend on how soon the bank is filled to its full height atter installing the columns. In addition, the season when the installation takes place must be considered, since the growth in strength of the upper sections of the lime/ cement columns is delayed by low temperatures. However, measurements have shown that the strength of lime/cement columns increases faster if these are loaded immediately after installation. It is therefore appropriate to fill out, with a normal safety factor, the part of the embankment that corresponds to the bearing capacity ofunstabilised clay. This is done about two weeks after installing the columns. During construction, the strength of the lime/cement columns is measured at time intervals in accordance with programmes designed for this purpose.
10.3.2
Demands on the admixtures
The lime should be unslaked and burnt with grain-sizes between 0-0.2 mm. The content of CaO sl,ould be > 80 %. The cement should be Standard Portland cement with grain-sizes between 0-0.2 mm. There is also a demand on the liquidity of the admixtures. The unslaked lime shall have a liquidity > 70 and the cement shall have a liquidity > 40. Liquidity indicates the quantity of the total sample, expressed in per cent by weight, passing through when the sample is shaken on a screen with a mesh of 0.5 mm. The test method is described by Van lmse VW. ~Messung der Fliessfahigkeit yon Zement, Zement Kalk - Gips Nr 3, 1972~. The test is performed by the manufacturer of the lime and the cement respectively.
10.3.3
Choice of admixture for different kinds of soils
On the basis of mixing tests in the laboratory the following recommendations are given for seven different kinds of soils. In the text is given a value of the "stabilisation effect", S~. So~ is calculated as the quotient between the shear strength in stabilised clay and the shear strength in unstabilised clay. There are obviously big regional and local differences in stability effect, which motivates that mixing tests in the laboratory should be performed on each object. Furthermore the stability also increases with time. The given values on S ~ should be looked upon as normal values to use in design for lime and lime/cement columns. If the construction time is short it could be difficult to reach this values.
362
Lime Columns
Peat
In a few cases mixing tests have been performed on peat that has been stabilised with cement. Normal values on S~can be 5. Often peat is present in stratigraphies with other kinds of soft and compressible soils, such as clay and gyttja. Normally the stabilisation effect is not good enough and therefore the peat is excavated.
Gyttja On tests performed in laboratory lime gives very low stabilising effect when added to gyttja. When cement or lime/cement mixtures are added the effect is considerably better. Anyhow the stablisation effect normally is to low and also the gyttja layers are excavated.
Gyttja-bearing clay When lime is added to gyttja-bearing clay the stabilisation effect could be about 5. Cement and lime/cement gives considerably higher effect and S~will be about 1020. The increase of shear strength with time is relatively slow. Lime/cement is recommended as admixture when gyttja-bearing clay is to be stabilised. Becuase of the slow increase in strength the embankment should not be constructed until 3 months after the installation of the columns. A load that corresponds to the bearing capacity of unstabilised soil can be put m place at an earlier stage.
Sulphide-rich clay It is difficult to draw any general conclusion on how sulphide-rich clay reacts with lime and cement. There are big regional and local differences in stabilisation effect. Lime/cement is recommended as admixture to sulphide-rich clayes since it gives higher shear strength than lime. As is the case for gyttja-bearing clay the increase m strength with time is relatively slow. The embankment should not be constructed until 2-3 months after the installation of the lime/cement columns.
Clay, clay with silt layers and silty clay Clay, clay with silt layers and silty clay are highly suitable soils for mixing with lime and lime/cement. The stabilisation effect differs with the amount of admixture and the time from mixing. Normally S~will be at least 10-20. The higher the content of silt the more suitable is a mixture with lime/cement. S//t Only a few tests have been performed in Sweden. When silt shall be stabilised it is recommended to use a mixture of lime and cement.
Design Considerations
363
Stability calculations
10.3.4
Embankments on horizontal ground surfaces or ground surfaces with slight lateral inclination: Provided that no movements occur in the natural soil where the ernbanbnent is to be built, the stability of embankments on lime/cement columns can be calculated under the simple assumption that full interaction exists between the lime/cement columns and intervening unstabilised clay. This condition is assumed to be fulfilled when the stresses in the natural soil are within the elastic range. In general, it may be assumed that this is the case for ground inclinations shallower than 1:7-1:5. The stability of the embankment is calculated with circular cylindrical slip surfaces for an average value of the shear strength of lime/cement columns and unstabilised clay. In heterogeneous soil stratigraphy, however, dimensioning may be governed by plane or composite slip surfaces. The calculation model assumes that the columns are placed beneath the embankment on the active portion of the slip surface. When strengthening is introduced on the passive side of the slip surface, the lime/cement columns are installed as slabs or blocks. The calculated safety factor against base failure in the completed embankment must be at least 1.5. The calculation principle is illustrated in Fig. 10.2 for an embankment on clay with constant shear strength throughout the whole clay layer. X
Traffic
load
'~
....--"
9 " .
,
Fig. 10.2.
C
C [ C
.
,
~
.
~.~
,
C
Stability calculation on soil stabilised by lime/cement columns (from Swedish National Road Administration, 1986).
364
Lime Columns
The distance between centres of the columns is calculated with the moment equation: [a 9q:ooI + (l-a) "[clay]" R2" O~ -b R 2" ~ " '[clay -- W " x " F
(10.1)
Designations: '[clay('[fu) =
undrained shear strength of the clay
'[col
= =
shear strength of the lime/cement colunms
r
distance between centres of the columns
A
=
area of the columns, 0.2 m z for a column of 0.5 m diameter
a
=
A]c 2 , proportion of lime columns per unit of area
W
=
resultant of the load, including the effect of an inclined ground surface
X
=
lever arm of the load
~ , 1~
=
centre angles in radians
F
=
safety factor
When evaluating the shear strength of the lime/cement columns, attention is normally paid to the fact that this varies along the column, since the effect of the admixture differs with the layers in the soil profile, see Fig. 10.3. When classifying into characteristic layers, the variation in the shear strength of the unstabilised clay is also taken into account. The mean shear strength in a slip surface is calculated with the relation: c
Z [a" "[col q- ( l - a )
9'[clay] " ~n q" "[clay" ~clay
%~,, =
where '[clay
--
shear strength of the clay
'[col
--
shear strength of the lime/cement columns
~m~.~ -- mean shear strength along the slip surface a
= proportion of lime/cement columns per unit of area
13
= centre angle of each sector of the slip surface
(10.2)
Design Considerations
365
0A
/
::-!i~-i.: : .
i--i . . : - : i -
:.-.
...
:-:-..:-.-...-...-~-..
:i
-..
" " " ..-..
|
"
-
'""
9 :""
Fig. 10.3. Stability calculation (from Swedish National Road Administration).
Embankments along a slope:
In strengthening with lime/cement columns for embankment fills on soil exposed to creep deformations, full interaction between column and unstabilised clay cannot be assumed. Creep deformations occur in natural clay slopes with a low safety margin against failure. Under such conditions, load absorption in the lime/cement columns is ensured by combining these into slabs placed at fight angles to the slope. In highly stressed areas the slabs are combined into blocks. The lime/cement columns must overlap each other by at least 50 mm if they are to interact. The principle for installation of lime/cement columns as rigid screens beneath an embankment along a clay slope is illustrated in Fig. 10.4.
10.3.5
Settlement calculations
Magnitu de of settlements: The loading of a surface strengthened with lime/cement columns is supported both by the columns and by the unstabilised clay between and immediately outside the columns. The compression modulus of the lime/cement columns will be considerably larger than that of the unstabilised clay. Consequently, the settlements will be very much smaller for a loading on a surface stabilised with columns than if the clay is unstabilised. The settlements within the volume containing lime/cement columns
366
Lime Columns 0 o(
o o o o oooooooox:Z)oooi.
o o o o o o o o
ooo
.
.
.
oo
0 0 () 0 0 0 0 O 0 0 0 0 0 ( X ) ( ) O 0 0 0 0 I oOC
0 0 0 0 0
0 0 0 0 0 0
() o o o o o c ~ ) c x 3 c x 3 o
I
9
"
'
"
"
'
'
i
"
"
"
"
"
'
1
j
i
I I Fig. 10.4.
..J
r!
i'ii!!.!!. Illlll Stabilisation of an embankment along a slope with lime/cement columns (Swedish National Road Administration, 1987).
are influenced by the following factors: 9 the relationship between the compression moduli of the lime/cement columns and the unstabilised clay 9 the proportion of the surface stabilised with lime/cement columns 9 the degree of consolidation of the clay 9 the creep load of the columns 9 the time between installation of the columns and application of the load 9 permeability in unstabilised soil and column respectively. The distribution of the loading between the lime/cement columns and unstabilised clay is calculated under the idealised assumption that the same settlement occurs in columns as in unstabilised clay at each level (equal strain theory). This means that the loading on the unstabilised clay is successively transferred to the lime/cement columns and that the entire loading is transferred to the bottom surface of the lime/cement columns, as outlined in Fig. 10.5. This assumption is probably on the
367
Design Considerations
safe side. In Fig. 10.5 the values ql and % are shown, ql stands for the load theoretically carried by the lime/cement columns and q2 for the load carried by the unstabilised clay. If the clay layer continues under the column block, the settlements in the layer are calculated with a load distribution according to the 2:1 method, whereby the loading is considered transferred to the bottom surface of the columns. The layer may be considered as drained by the lime/cement columns.
xy
~d
~'
~f
~tc
,1
\ /
q=ql+q2
2
/ Fig. 10.5.
Distribution of load between lime/cement columns and unstabilised clay (Swedish National Road Administration).
The compression modulus of the lime/cement columns increases with time after installation. Investigations indicate that the compression modulus may vary between 50 and 150 times the shear strength of the column. For organic clay the multiplier usually is smaller. Approximate values for the compression modulus and its increase with time are obtained with compression tests at different times on samples mixed with lime in the laboratory. Owing to different mixing methods and stress conditions, the compression modulus of lime/cement columns develops in a different manner compared to samples in the laboratory. Settlement calculations should therefore be documented as maximum and minimum values for the settlements. From measurements during construction, it is possible to determine whether significant deviations from the predicted settlements have occurred. A basis for decisions on corrective action is thereby obtained, for example, when a temporary surcharge is to be reduced or if the loading requires adjustment.
368
Lime Columns
The load-absorption capacity of the lime/cement columns depends partly on the thickness of the fill. Beneath low fills, the columns should therefore be installed closer together than required for stability of the fill in order for the settlements not to be uneven. Distribution of load- lime~cement column and unstabilised soil:
The creep load is the load above the proportionality limit for the loading vs deformation curve of the lime/cement columns, see Fig. 10.6.
Load. Gco[
Creep toad 6co[ creep
col 1
Deformation
I
E
Fig. 10.6. Assumed stress-strain relation of stabilised soil.
According to investigations, the creep load for a lime/cement column is not less than 65 % of its failure load. The failure load, ~pf,ip is calculated from the empirical formula: (Ypfail= 2
"1:col +
3 9~H
(10.3)
where O H is the horizontal pressure of the soil against the lime/cement column. ~H can be set equal to the original vertical stress of the soil because of the lateral pressure caused by the expansion of the column when lime/cement is mixed in. Equation (10.3) is a total stress approach with ~ = 30 ~
Design Considerations
369
The calculated maximum load borne by the individual column, ql creep, will thus be q~creep --
0,65
9a" Crpfail
where A a = c2
A = cross-sectional area of the lime/cement columns c = distance between centres of the columns, when columns are placed in a square grid Since the shear strength of the columns increases with time after mixing with lime, the creep load of the column also increases. The creep load will also vary with the depth below the ground surface. The equal strain theory presupposes that the load on lime/cement columns and unstabilised clay is in proportion to their respective modulus, MooI and Molay. The modulus in the column may beconsidered constant as long as the creep load is not reached (Cf Fig. 10.6), but the modulus in the unstabilised clay will vary with the vertical effective pressure according to Fig. 10.7. This means that the calculation of distribution of load on lime/cement columns and unstabilised clay must be made in iterations. The load on the unstabilised clay, q2, is obtained as the difference between the total load, q, and the load on the lime/cement columns, ql-
q2=q-ql The settlement is calculated by dividing the clay layer into characteristic layers. The settlement in the columns is calculated from Eqn 10.4. M o1is the modulus in the lime/cement column. Ah
ql (10.4)
S1= ] ~ ~ "
a
Meo1
370
Lime Columns
Vertical effective pressure, kPa
61~'6~ 6~6~ 1 [tn{M'(6'-6[] +I}] s Mo M L N ' ML o
L.
E o o
o > o o cr
Fig. 10.7 Assumed stress-strain relation of unstabilised clay.
while the, settlement in the unstabilised clay is calculated from the following equations 10.5a-c. The choice of equation will depend on the vertical effective pressure (Cf. Fig. 10.7).
9 if O'vo
+
q2
< d'" p
1-a Ah
q2
1 -a
Mo
(10.5a)
S2=Z~
q2 9 if O'p _< (Y'vo + ~
1-a
-- (fL
Design Considerations
371 q: _
VO
p
S2-ZAh.
(~'
1-a
vo
p (10.5b)
M0
ME
q2 9
ifc'
~_ (3"L
+
VO
1-a
I (j, .(~" S 2 - 2; All"
M"(c'
1
+~. M'
+ (~'L - (~p +
P vo Mo
ME
vo
q2
+
1 -a
In
- ~'L) +
(10.5c)
ME
In Eqn. 10.5a, 10.5b, 10.5c the parameters are as follows (cf Chapter 5 ) vo = in situ effective stress, kPa ~ p = preconsolidation pressure, kPa ~'L = the pressure at which the relation load-deformation is transformed from a linear to a logarithmic relation, kPa M ~ = Compression modulus below preconsolidation pressure, kPa My = Compression modulus between pressures c p and ~'L, kPa M"
= Modulus number
The calculated settlement in the columns is then compared against the calculated settlement in the unstabilised clay. If the settlement in the columns is greater than in the unstabilised clay, the calculated load should be reduced for the columns so that S 1 becomes equal to $2. This load transfer can be calculated by a stepwise reduction of ql and a corresponding increase of q2 so that SI=S 2. The settlement that is assumed to occur, Sin, is
372
Lime Columns
equal to S 1 and Sz. If, however, the calculated settlement is less in the columns than in the unstabilised soil, the columns cannot be subjected to any further loading and the settlement that occurs, Sin, will be equal to the settlement in the unstabilised clay, S z. Settlement/time relationship:
Where loads are less than the preconsolidation pressure of the clay, the settlements are concluded rapidly, more or less as in loading dry crust clay. Where loads are greater than the preconsolidation pressure of the clay, the chronological sequence of the consolidation settlements will be generally the same as in vertical drainage with sand drains and prefabricated band drains. The permeability of the lime/cement columns will be greater than that of the unstabilised clay. According to measurements of settlements in embankments on lime columns, the permeability of the lime columns is 400-1000 times that of the unstabilised clay. The columns thus function as vertical drams. In calculations of the time-rate of settlements in a lime-stabilised clay the permeability in the column is assumed to 1000 times that ofunstabilised clay. The permeability of a lime/cement (50/50) stabilised clay is assumed to 400-800 times that of unstabilised clay. For embankments on lime/cement columns with distances between centres of 0.81.8 m, the settlement/time relationship may be calculated approximately with Hansbo's equation for radial flow (Hansbo, 1979). (cf Chapter 9.3.3)
U = 1 - exp
I -2ch 9t
(10.6)
R 2" f(n)
or
I n ( l - U ) . R2 9f(n) t =-
(lO.6b) 2c u
where U = degree of consolidation % = coefficient of consolidation, normally twice cv t
= time for consolidation
Design Considerations
373
R = radius of influence of the lime/cement columns. For columns with distance between centres a, set in a square grid or as isosceles triangles, R =c/~/~=0.5 6c. For lime/cement columns set in an equilateral grid R=0.525c (c = distance between centres of the columns). nz f(n) ~ ~ ~ n2-1
1
1
In(n)- 0.75 + ~ . [ 1- ~ ] + n2 4n 2
n2-1
1 klay*
n2
r E k~oI
9( L D ) 2
(10.7)
where n
= R/r
r
= radius of a lime/cement column
kolay= permeability in the unstabilised clay kooI = permeability in the column L D = length of the columns in single-sided drainage and half the length of the columns in double-sided drainage The quotient kolay/ko 1 is based on experience from measurements on performed lime-column projects.
10.4
LIMITATIONS
The penetration capacity of the mixing tool is limited by hard surface layers such as dry crust clay, surface layers containing stones, thick roots of trees and bushes and fills with such contents, and construction waste. The surface layer must not be frozen to more than 0.1 m depth. Firmer surface layers must be excavated and removed and may where necessary be replaced by sand fills. Layers of firm deposits, especially stones and boulders, may damage the mixing tool and restrict the installation of the columns. Installation of lime/cement columns may lead to raising of the ground surface since the soil volume increases when the admixture is mixed in. The extent of the rise depends on how closely the columns are spaced. In installation of lime columns as blocks, rises of 0.2-0.3 m have been recorded. Inclination in ground level and in rock level can lead to stability problems. The shape of the mixing tool causes a disturbed zone about half a meter below the bottom of the column. The problem is most obvious when the friction material above the rock is very thin (Cf. fig. 10.8).
374
Lime Columns
Fig. 10.8. Disturbed zone beneath the lime/cement columns.
10.5
CONSTRUCTION
ASPECTS
The geotechnical requirements are listed in a Construction Description, Geotechnique, with relevant drawings. The Construction Description is shown on the draft for the colunm installation and the Site Plan. Applicable drawings are specified in a special list. The draft must contain the following information: 9 a plan of the siting of the columns 9 characteristic properties of stabilised and unstabilised soil 9 calculated settlements, magnitude and rate of settlements 9 construction time, surcharge 9 the length of the columns, normally with a tolerance of 0.2 m from specified depth and 0 m for columns to firm bottom 9 inclination of the columns, normal inclination tolerance 0.01 m/m 9 quality of the lime and cement 9 quantity of lime and cement, kg/m 3 9 rate of withdrawal during installation of the columns, normally max 75 rpm and 25 mm/revolution. For lime/cement columns 15-20 mm/revolution 9 tolerance for deviations in plane, normally + 10 % of specified distance between centres, but at least 0.1 m for solitary columns 9 recommendations on installation sequence of the columns
Construction Aspects
375
9 recommendations on the loading sequense over the stabilised soil. Normally a load that corresponds to the bearing capacity of unstabilised soil can be applied within 1-2 weeks after installation. The time before the full load can be applied depends on how the stability effect increases with time 9 demands on preparatory works prior to the columns installation. Excavation of peat and gyttja, removal of obsticles, culverts and fill 9 restrictions on placing of the lime column machines 9 subcontractor's documentation of the work 9 programme for control of achieved shear strength in the columns 9 programme for control of settlements, pore pressures and movements
10.6
REQUIREMENTS FOR FIELD MEASUREMENTS
10.6.1
Determination of the shear strength of lime/cement columns
The number of lime/cement columns that should be respected in each case must be decided with regard to the nature of the project and the reliability required for the construction. For small projects, i.e. less than about 200 columns, where the clay is inorganic and has a water content of less than 80 %, it is more economical to replace column investigations with a number of extra lime/cement columns. The higher the loading in relation to the bearing capacity of the unstabilised clay, the larger the number of columns that should be investigated for a statistically certain value of the bearing capacity of the columns to be obtained. As a general rule, about 1% of all columns should be investigated, with a minimum of 5 columns/test occasion. If the safety factor for the load with regard to the capacity of the unstabilised clay is less than 1.0, about 2 % of all lime/cement columns should be investigated. An example of an equipment for testing shear strength in lime/cement columns is shown in Chapter 2. The number of test occasions and the timing of these should be chosen so that the load can be modified or the column reinforcement extended if the columns prove to have a lower strength than expected. If the area strengthened with columns is to be filled out immediately after installation of the columns, a preliminary investigation of the columns may be made one week after installation and a further investigation 1-2 months later. Normally, the strength of the lime/cement columns is investigated on only one occasion, provided the result agrees with the assumptions used in the calculation.
376
10.6.2
Lime Columns
Inspection of settlements
The timing of the removal of a temporary surcharge is decided with the aid of continuous measurements of the settlements in the ground surface in a similar way to vertical drainage. If the embankment is low, the settlements can be measured by using levelling pegs placed on the ground surface. Where such pegs hinder filling of the embankment, there is a considerable risk that they will be damaged and become unusable. Settlements in the ground surface should therefore be measured with a hose settlement gauge (cf Chapter 2.6). Using ridge pegs on the fill, the settlement within the fill can be evaluated. Since the settlements are formed comparatively rapidly, they should be measured at least once a month, with 6-8 measuring occasions. The times for the measurements are chosen on the basis of the settlement/time relationship. An example of instrument layout is shown in Fig. 9.23 (Chapter 9).
Dimensioning Example
377
10.7
EXAMPLE: DIMENSIONING OF LIME COLUMNS FOR REDUCTION OF SETTLEMENTS AND FOR STABILISATION OF A ROAD EMBANKMENT ON SOFT AND ORGANIC CLAY
10.7.1
Introduction
The example relates to an actual case in Sweden, for road E6 passing Tanumshede approximately 130 km north of Gothenburg, where it was decided to construct a road using the preloading method in combination with lime columns. The preloading is designed to last 18 months and during this period the major part of the settlements will have occurred. The road was constructed over a farming area where the ground surface is mostly horizontal (Fig 10.9). Section .50
3/300
31400
31500
31600
31700
31800
31900
4/000 § 50
._Gyt_.tja bearing d a y +/.,El
~
~ /
Level of future road ,, :. . . . .
~ .
.~ . 3 0
J~'.'"~ . ~ , n. ' ~ - ~ .
+40
+30
9
.2o
+10
.I0
Fig. 10.9. Cross-section of the area.
The soil consists of 0.3-0.7 m of humus-rich topsoil above 1.0-1.5 m of clayey gyttja and gyttja-bearing clay. Below about 1.5-2.0 m depth there is a clay with high sensitivity to depth. At some depths the clay contains shales and thin silt layers but mostly it is a homogeneous clay. Beneath the clay is friction material. The maximum clay depth is about 20 m. The clay is very soft (I: - 6 kPa) beneath the dry crust and the shear strength increases with depth. However, the shear strength is only 12 kPa
Lime Columns
378
at 14 m depth (Fig. 10.10). According to CRS tests, the clay is normally consolidated or only slightly over-consolidated (Fig. 10.11). Density, tO (t/m3)
Shear strength, Tfu (kPo)
~
. . . .
io,
,
1.0
9
-,,
i
I
)
:y 9
I
2D
9
Woter content.w n (%) Liquid limit, WL(%)
oO
i
r
9
Liquid limit, _ WL (%!_
~oo
!
~
0 ....
9
i
:
5
E x:"
~,
; I *
10
'k
lO-
; i
;
;
H
t I
121
~
i -
!u
-
!
,3oo
5
I
i
10
oo
i i
-
s
Sensitivity
~oo
10 b
10
I
I
-
15
-
I
_ 151
15'''''
/
/
-
;~L, . . . . . .
15
T i
i
i
i
15
'
Fig. 10.10. Basic parameters in the clay.
Effective verticQI pressure. 3/611 0 50 6": (kPo) 100 0,~ - ""X
I
i
m
I
r
I
I
1000
,
A
II
~.
lOO
u)
E .E
-
C~ k
-
0 L_ tn
13. ql C3
m
I1
10
:
c~~ ~g
-
I
I
i
I
I
1
I
I
J
10
1
10 Time
after
100 mixing
(days)
Fig. 10.11. Effective stresses and results from using clay-lime mixture.
1000
Dimensioning Example
379
At an early stage of planning, different methods of construction were considered. The methods were compared from both the economic and technical- functional aspects. As mentioned, the clay was very soft and also very compressible. Replacing the clay with better material was not possible. Adjusting the load by lowering the road profile only has a short-lived effect since settlements will occur also with small loads, and this leads to problems with flooding and water in the pavement. Use of lightweight fill was also excluded since the road would have been too light and problems would have occurred when the ground water level was high. Vertical drains could have been used but due to the low shear strength of the clay large pressure berms are necessary for stabilisation. Furthermore, there will be very large settlements when using vertical drains. The methods that were considered acceptable from technical-functional aspects were piling and stabilisation with lime columns. The use of piles would have been very expensive, since it would have been necessary to transfer the load from the embankment to the piles with a continuous concrete deck. The topsoil layers contain organic soil and therefore separate concrete slabs cannot be used. It remained to see whether the lime column method was suitable. As has been stated earlier, the effect of lime in organic soil is not always good, but the laboratory investigation showed that the mix with lime increased the shear strength considerably (Fig. 10.11). The laboratory tests were performed with 82 kg lime per m 3, corresponding to 16 kg lime per meter of column (500 mm). It is essential to perform these laboratory tests to evaluate the effect of lime-stabilisation on the soil, since this effect may vary considerably. Through cooperation between the road engineer and the geotechnical consultant, the lowest possible profile was chosen for the road in regard to the problems with flooding and water in the pavement. On the basis of field and laboratory investigations it was possible to start the dimensioning of the lime column reinforcement.
10.7.2
Dimensioning of lime column reinforcement
Lime columns are designed to have a stabilising effect as well as decreasing settlement and speeding up the rate of settlements with time. The spacing of lime colunms (c) is calculated from their function as drains and at the same time the necessary load (q+Aq) to ensure sufficient settlement during construction time (t=l 8 months) is calculated. The load (q+Aq) demands an increase m the shear strength of the clay to ~m~an- a "~ool + (l-a)" ~ol,y (see Chapter 10.3.4). To ensure this mean shear strength in the clay, the lime columns shall have a maximum spacing (c) that can be compared with the assumed value. In this case, the dimensioning parameters have been chosen according to Figs 10.12-13.
Lime Columns
380
Shear strength,l"fu 00 (kPa} 40
Density, Water cont.,wn (%) jo (tim 3) Liquid limit, WL(%) Sensitivity. St 00~5,1.5 2.5 020,60 100, 00 /.,13 80
5 ' ' '
g
L .
.
.
. ,--'} . .
s
s
10
10
i
.
Effective vertical pressure (kPa) 50 100
.
-\'%(,' '..L/_' ....
s
O
,
10
10
,j., . , . 15
'
15 ,
'
15
10
120 15 . . . .
.115 15 . . . . . . .
Fig. 10.12. Basic parameters in the clay.
C,
Dimensioning Example
381
M o (kPa)
6"c (kPa} 100
_[3
O0
u-,
E5
!_ 1
M L (l<Pa) 5O0
,,
,.,0
,---
u ~
5
5
I0
10
I0
15
15
L
'
'
M
6" L (kPa} ,,,
100 ....1
~
Oi
5 ~
15
2
i
~_0
Fig. 10.13. a) Design parameters of unstabilised clay.
'1' COtdi m (kF~)
oo _
l
I
I
]I
_
i
-
i
_1oo
-
i
-
!
oO
,, Ioc,o
_
5
i
O
M cot dim {1<~)
!
-
-
10
i
_
i
15
-
.~,-, , , /'U
I
,,,
0
i
i
ii
Fig. 10.13. b) Design parameters of clay stabilised with lime columns.
20
Lime Columns
382
Calculation of settlements in stabilised clay: The settlements in the lime-stabilised clay can be calculated according to Chapter 10.3.5. The model described in that chapter demands extensive calculations and is suitable for processing by computer. C a l c u l a t i o n for 15 m o f clay a n d 15 m o f lim e c o l u m s
Table 10.1.
Input data for computation lime column stabilisation 3/440-3/480 XcoI =75 kPa.
Thickness
Eft. dens.
a p
ML
o'L
M'
Mo
"l;co1
m
t]rn 3
kPa
kPa
kPa
-
kPa
kPa
1.00 2.00 2.00 2.00 2.00 2.00 4.00
1.30 0.50 0.50 0.50 0.50 0.50 0.50
50.0 21.0 31.0 38.0 54.0 65.0 76.0
2000.0 226.0 226.0 221.0 183.0 271.0 319.0
100.0 39.0 49.0 54.0 73.0 89.0 96.0
12.0 11.9 11.9 11.9 14.1 15.8 13.1
9000 1375 2100 2100 2500 2750 3250
75 75 75 75 75 75 75
-
Distance from ground level to ground water table = 1 m
-
Diameter of lime columns = 0.5 m
-
Length of lime columns = 15 m
-
The distance between the lime columns is 1.2 m and the lime columns are placed in a square grid
-
Correction factor for creep load = 0.65
-
Coefficient of consolidation for the clay - c h - 2.6 910 .8 m2/s
-
Relation between permeabilities k 1.y/kooI = 1/1000
Lime columns in a quadratic pattem and free drained top and bottom layer. The modulus in the lime columns is assumed to be 75 times the shear strength of the lime columns, i.e. MooI =75.~oo I =75.75=5625 kPa. Below is shown a calculation for a load, q, of 30.4 kPa.
Dimensioning Example
383
0.0 - 1.0 m
b e l o w g r o u n d surface
(10.3)
(3"pfail =
2 9q:p -k- 3
(Ypfail--
2 . 75 + 3 96.38 = 169 kPa
9o H
0.196 a=~=0.136 1.2 2 q l creep
=
0.65.0.136.
q2
q
- ql
=
=
169 = 15 kPa
30.4 - 15 = 15.4 kPa 1.0
9 Settlements in the lime columns
S 1 .
.
15
.
0.136
9 Settlements in the unstabilised clay S z .
0.02 m
.
(10.4)
5625
1.0 15.4 . . . 0.864 9000
0.002 m
S 2 "( S 1 . The settlement in the unstabilised clay is less than the settlement in the lime columns, which means that a bigger load will be transferred to the unstabilised clay. Assuming that the load on the lime columns is ql = 3 kPa, the load on the unstabilised clay is q2 = q-ql = 27.4 kPa. 1.0 9 Settlements in the lime columns
S 1 .
.
3 .
0.136
9 Settlements in the unstabilised clay S 2 .
9
S m - - S 1 - - S 2 --
0.004
-- 0 . 0 0
m
.
.
0.004 m
5625
1.0 27.4 . . . . 0.864 9000
0.004 m
(10.4)
Lime Columns
384
1.0 - 3.0 m
b e l o w g r o u n d surface
(Ypfail-" 2 . 75 + 3 927.47 - 232 k P a ql creep =
0.65 90.136 9232 = 20.6 k P a
q2
- ql
= q
=
30.4 - 20.6 = 9.8 k P a 2.0
20.6
0.136
5625
= 0.054 m
Settlements in the lime columns S 1 =
(10.4)
9 Settlements in the unstabilised clay
q2
9 + ~ _
cr'e _ CY'vo + cr vo
(y,
P
1-a
(lO.5b)
S2=Ah" Mo
ML
9.8 17.66 + 21-17.66 $ 2 = 2.0-
- 21 0.864 = 0.076 m
+ 1375
226
In this case, the result is that the calculated settlement is bigger in the unstabilised clay than m the lime columns $2>S 1. This means that Sin-- S 2. 9
Sin= $ 2 = 0.08 m
Dimensioning Example
385
3.0 - 5.0 m
13pfai I = 2 . 7 5
ql creep
q2
below ground surface
+ 3 956.90 = 321 k P a
= 0.65 90.136 9321 = 28.4 k P a - q-
ql
-
30.4 - 28.4 = 2.0 k P a
2.0 9 Settlements in the lime columns S 1 = ~ . 0.136
28.4 = 0.074 m
(10.4)
5625
Settlements in the unstabilised clay
S2 .
2.0 . . 0.864
2.0 . 2100
0.002
S 2 < S 1. The settlement in the unstabilised clay is less than the settlement in the lime columns, which means that a bigger load will be transferred to the unstabilised clay. A s s u m i n g that the load on the lime columns is q l - 22 kPa, the load on the unstabilised clay is q2=q-ql = 8.4 kPa.
2.0 9 Settlements in the lime columns S 1 .
.
22 .
.
0.136
.
0.057 m
5625
Settlements in the unstabilised clay 8.4 27.47 + ~ 0.864
31-27.47 S 2 = 2.0 9
+ 2100
31 = 0.058 m
226
(10.4)
386
Lime Columns
9 Sm -
S 1 = S 2 = 0.06
m
Table 10.2. Results from computer calculations.
H EFFECT. TOTAL PRESS. PRESS.
"l;co1
m
kPa
kPa
kPa
Mco1
1st calc. ql q2
2nd calc. ql (12
1st calc. $1 $2
Sm
kPa
kPa
kPa
m
m
m
kPa
kPa
1.0 6.38
6.38
75
5625
14.99 15.41 2.47
27.93 0.02
0.00
0.00
2.0 2.0 2.0 2.0 2.0 4.0
27.47 56.90 86.33 115.76 145.19 189.33
75 75 75 75 75 75
5625 5625 5625 5625 5625 5625
20.60 28.42 30.40 30.40 30.40 30.40
m_ 8.45 6.9 9.93 12.09 20.80
0.08 0.00 0.00 0.00 0.00 0.00
0.08 0.06 0.07 0.06 0.05 0.11
9
17.66 27.47 37.28 47.09 56.90 71.61
9.80 1.98 0.00 0.00 0.0 0.00
-21.95 23.49 20.47 18.31 19.60
0.05 0.07 0.08 0.08 0.08 0.16
T O T A L S E T T L E M E N T = 0.42 M
In a similar manner, calculations are made for other distances (c) between lime columns and for other loads. The relation between load and settlement can be drawn as in Fig. 10.14. .,.-..
E
......
Celcutation
P
for the
/
distQnce 31M~O-31Z.K) 8
100 QSO
/
m
---
9
'
'
15 m C[Qy; lime columns spQcing c=1.2m embQnkment : 1.2+1.6 = 2.8 m IoQd is 2.8-19 = 53.2 kPQ
r_C~sAT__~wL
50
" 53 2 kPa
"
'
11)0
LoQd (kPQ)
Fig. 10.14. Relation of settlement vs load in lime-stabilised area.
Dimensioning Example
387
The lime columns shall be installed at a distance of 1.2 m and the load during the preloading period shall be 2.8 m (= 53.2 kPa). When the settlements occur in the subsoil, the load will decrease and the total settlement is calculated as 0.80 m, (see Fig. 10.14). Since the road is planned to have a profile 1.2 m above the original ground level, this means that the unloading will be (2.8-0.8-1.2) = 0.80 m. This magnitude of unloading is considered sufficient to hinder future creep deformations m the lime-stabilised soil.
Settlement~time relationship: The settlement/time relationship is calculated from the equation: -2Ch" t U = 1 - exp
(10.6) R~ 9f(n)
where
f(n)
~
~
1
In(n)-0.75 + ~ . [ 1 n2
-
n2-1
1
-~] 4n 2
+.
n2-11 . . . n2
r2
.
1
. 1000
LD2
1
The relation between permeability m the unstabilised clay, k~l,y , and the permeability in the lime column, kool, is assumed to be k,lay/kooI =1/1000. R = c / ~ n = 1.2A/~-- 0.677
n.
R . r
.
0.677 . 0.25
2.71
f(n)= 1.215
U = 1 -exp
f
- 2 . 2 . 6 - 1 0 -8 -t t
-9.34.10 -8.t = 1 -exp
0.6772. 1.215
In (l-U) t
(t = time in seconds)
.._ _
9.34.10 -8
(10.6)
Lime Columns
388
Degree of consolidation,U (%)
Time, (days)
3O 50 60 70 75 80 85 90 95 99
10.7.3
Stability
44 86 114 149 172 199 235 285 371 571
calculation
Stability during construction" The stability of the embankment including the overload (q+Aq) during construction can be calculated according to Chapter 10.3.2 or with the help of a computer programme.
~-23.~ r--~kNL~3 1 ' .... " ~ " " " ' " i " ...... "
//
/ / /
//
..... : ~ / " " : " " ' " " " [ " " " " i "
/
/
\~ .... ' ....'""~'-
~'-~s.Po,
\~
r-,~,,N,m~ \ ~
,'%
Fig. 10.15. Stability calculation in lime-stabilised area.
\ -
DimensioningExample
389
angle
shear strength
la 2= ~3 = ~4= [~5=
'~1 = z2 = '~3 = z4 = "ts=
9.30 ~ 3.61 ~ 40.16 ~ 57.20 ~ 3.62 ~
9.42 kPa 23.10 kPa 14.5 + (0.57.z) kPa 5.0 + (0.66.z) kPa 15 kPa
W h e n the shear strength increases with depth, the m e a n shear strength '~mean c a n be calculated f r o m Fig. 10.16.
ol
L Tz
-t"~
2a "[ rneon =Xo+k(D
xo+ k . z
+
180& a ~ . -R}
Fig. 10.16. Calculation of mean shear strength.
q~3mean= 16.85 k P a
q~4mean -- 8 . 1 9 k P a
5 ]~ ~n " q~n "C mean
1
(2) 5
:C~n 1
9.3.9.42 + 3.61.23.1 + 40.16.16.85 + 57.20.8.26 q~mean =
113.9 1374.47 = 12.07 k P a 113.9
+ 3.62-15
390
Lime Columns
Mresisting- "Cm~n'R'S = 12.07.20.6.40.95 = 10179.9 kNm/m
Table 10.3. Calculation of Moverturning
Area (m2) A B C D E F
T (kN/m3) Weight(kN/m) Lever arm(m)
2.55 21.56 6.00 4.80 6.40 0.96
18 18 18 18 18 18
45.86 388.08 108.00 86.40 115.20 17.30
Moment (kNm/m)
16.91 12.45 6.60 5.60 -1.40 -6.20
775.5 4831.6 712.8 483.8 -161.3 -107.1 Z 6535.5
F=
__M-esisting Moverturning
10179.9 =
= 1.55 6535.5
Stability o f the road when in use: The stability shall of course also be calculated for the road when open to traffic. The conditions are somewhat different, i.e. the embankment height is smaller, but on the other hand a traffic load of 10 kPa shall be included. Note that in the example no attention is paid to the fact that the shear strength is increased in the clay during preloading. To calculate the increase in shear strength see Chapter 4.5.2.
Ps x = 23kPa 1'= 13 ~k,. . . . .
z;
c
'a/fJ/
1
"~"\~
~= 15 kPa 1,- 13 kNlm3
/
Fig. 10.17. Stability of road when open to traffic.
Dimensioning Example
391
angle
shear strength
[31= 5.14 ~ ~2 = 4.51 ~ 133= 61.37 ~ 134= 45.25 ~ 135= 4.51 ~
"C1= 8.59 kPa q;2 = 23.10 kPa t:3= 14.5 + (0.57.z) kPa I:4 - 5.0 + (0.66.z) kPa -c5= 15 kPa
('l;3mean=16.97) ('l;4mean=7.45)
5.14.8.59 + 4.51.23.1 + 61.37.16.97 + 45.25.7.45 + 4.51.15 mean
120.8 1594.55 = 13.20 kPa 120.8
Mresisting- qSmean.R.S -- 13.20" 15.4.32.46 = 6598.4 k N m / m Table 10.4. Calculation of M overturning
Area(m 2)
y (kN/m 3) Weight(kN/m)
Lever arm(m)
Moment (kNm/m) p
A 0.41 18 B 11.57 18 C 2.16 18 F Traffic load 10 kPa
Z
7.34 208.22 38.88 103.20
2777.9 M resisting F =
6598.4 =
Moverturning
= 2.37 2777.9
13.27 8.22 2.20 8.56
97.4 1711.6 85.5 883.4
392
10.7.4
Lime Columns
Results from dimensioning
The results from the calculations are given as follows: the sufficient distance between lime columns, sufficient load (q+Aq), sufficient preloading time, calculated settlement and finally the magnitude of the unloading at the end of preloading, are given for each part of the road. The lime columns will be installed in a square grid with distances according to Table 10.5. The magnitude of the load (q+Aq) will vary and is given in the same table. In the transverse direction, the lime columns are to be installed beneath the slope at a distance of 2 times the height of the embankment (including surcharge). Compare with the stability calculations. TablelO.5.
Distance (c) between lime columns, required height of preloading and length of lime columns.
Section
Distance c (m)
Req. embankm. height (in)
Average length of lime col. (in)
3/370-3/390 3/390-3/419 3/419-3/440 3/440-3/660 3/660-3/800 3/800-3/825 3/825-3/835 3/835-3/860
1.6 1.4 1.2" 1.3 1.6 1.2 1.6
1.4 2.3 2.4 2.8 2.5 2.3 1.8 2.3
6 11 15 15 12 11 8
3/860-3/880
1.6
2.3-1.5
8
* Near the culvert in section 3/616 the lime columns are installed at a closer distance. The lime columns shall be installed as far as the bottom of the clay layer or to a maximum depth of 15 m (= capacity of the lime column machine in 1991). The necessary preloading time is calculated to be 18 months, partly because the lime columns will not be installed to the full depth of the clay layer (3/450-3/700). Table 10.6 gives calculated settlements and unloading. The unloading is calculated to be 0.40-0.80 m, which is considered to be enough to avoid future creep settlements. Over the distance 3/450-3/700 full consolidation will not be reached and for this reason future settlements may occur (= 0.1 m).
Dimensioning Example
393
Table 10.6. Estimated settlement and unloading after 18 months of preloading.
Section
3/380 3/400 3/430 3/500 3/700 3/810 3/830 3/860
Planned embankment excluding including overload overload
Estimated settlement
Unloading
(m)
(m)
(m)
(m)
0.4 1.2 1.2 1.2 1.1 1.0 1.0 1.0
1.4 2.3 2.4 2.8 2.5 2.3 1.8 2.3
0.20 0.65 0.80 1.05 0.95 0.80 0.30 0.60
0.80 0.45 0.40 0.55 0.45 0.50 0.50 0.70
Transition to the culverts:
The road will pass over two trenches where two culverts will be placed. In principle there are two solutions to the problems of the foundation of the culverts. a) During the preloading time the water may be led past the construction area. In this way, the settlements in the ground will have finished before it is time to put the culverts in place. b) The water can be led through the embankment (including the overload) through a temporary culvert during construction. In this case, the lime columns should preferably be installed at a smaller distance. Consequently the settlements will be smaller. When the preloading is finished, the temporary culvert is replaced by the permanent culvert. Construction schedule:
9 The area is prepared for lime column installation. Surface layers containing stones, thick roots and other materials that can hinder the penetration of the mixing tool shall be removed. 9 A total of about 8000 lime columns with a total length of about 110 km will be installed. The installation is preferably divided into three separate areas. 9 Directly following the installation of the lime columns, the embankment is filled to 0.8 m height (corresponds to the load the unstabilised clay can support). 9 The shear strength of the lime columns shall be verified with apparatus such as the lime column penetrometer (described in Fig. 2.4 Chapter 2). Tests are made after 30, 60 and 90 days respectively. Preferably tests on different parts of the area should be made on the same occasion.
394
Lime Columns
9 The temporary culverts are put in place. 9 The pressure berms are prepared on both sides of the future embankment. 3 months after installation of the lime columns, the embankment including the overload is put in place. During this stage, settlements will occur. The height of the embankment, including overload from Table 10.6, must not be exceeded. This means that during construction of the embankment the settlements shall be monitored. 9 The time for preloading is scheduled to be 18 months. However, the preloading stage should not be halted until the effect of the preloading is checked and approved. 9 The temporary culverts are taken away and replaced by the permanent culverts. 9 When unloading is completed, the pressure berms can be taken away and the ditches along the road can be finished.
Inspection and follow-up: 9 To follow the settlements, horizontal hoses can be used for measuring with the hose-settlement gauge. The settlements shall be monitored with hoses at maximum 50 m distance. 9 The settlements shall be measured an average of once a month, with the shorter intervals in the early stages of preloading and longer intervals at the end. 9 The follow-up will give information on the need for a supplementary overload and also on the time for unloading. Comments:
In this case, the number of lime columns is very large (= 8000). For this reason, it may be appropriate to install trial lime columns and also to prepare a test embankment. Such a test embankment should be at least 50 m long and shall have the same crest width as the planned width of the preloading. The test embankment should preferably be aligned with the future road. A follow-up shall be made on a number of lime columns to check the increase in shear strength with time and the settlement shall be recorded throughout the duration of the test. Results from the tests may later be used in the dimensioning of the remaining part of the preloading. The described example was designed in the beginning of 1990. Then lime columns were much more common than lime/cement columns. In 1994 more than 90 % of the columns are lime/cement columns. The columns used in on road E6 Tanumshede were 0.5 m in diameter. Nowadays columns with diameter 0.6 m are more common. In the example the preloading time is set to 18 months which was possible since the lime column installation could be made at an early stage of the road construction.
DimensioningExample
395
Today preloading times of about 6 months are most common. The road was opened for traffic in 1993. The settlements have been measured during preloading and were somewhat smaller than the calculated. The meaured time-settlement rate was in good agreement with the calculated.
Case history - Bridge foundation on soft clay stabilised with lime columns A new road near Karlstad was rebuilt for a distance of about 10 km. One section of the road passes over "Kvamtorps/ilven". Earlier the area has been used for agricultural purposes and as meadowland. The suroundings of the stream are rather fiat and as a consequense the width of the stream varied from 5-100 m, depending on the water level. At the bottom of the stream there is a layer of highly decomposed peat and gyttja at most 2 m thick. Beneath, there is soft, silty clay with high sensitivity down to 20 m depth. The clay is over-consolidated by about 20 kPa down to 6 m depth and beneath this level the clay is over-consolidated by about 80 kPa. The shear strength of the clay is about 15 kPa. As an alternative to expensive piled foundations, the bridge was built with a concrete slab on lime-stabilised clay. The bridge has a span of 9 m. By utilising the over-consolidation in the deeper-sited soil layers, the lime columns did not need to be longer than 7 m (Fig. 10.18). The spacing between the columns is 0.8 m beneath the bridge and 0.9 m beneath the access embankments. A preloading was performed both for the future bridge foundation and the access embankments. Through the lime stabilisation, the average shear strength was increased from 15 to 25 kPa (Bengtsson et al, 1991). Rood profite_
~ 60.2 +
/ Embankment )
) )
J j
~ ~ 7----.._... v
I
II
v 48.0 15S (S=0.Sm)
~
10S (S=0.gm)
Fig. 10.18. Longitudinal section of lime stabilisation.
1
396
Lime Columns
During the construction period, the stream was realined and led through a large ditch. The organic soil was excavated and replaced with sand. From a working surface made from sandy material, the lime columns were installed to a depth of about 8 m below ground level (level +48.0). The area stabilised with lime columns was then covered by a preloading to the level + 57.0 (corresponding to the load of the bridge). After three months ofpreloading, the load was excavated and the bridge was built on a 0.2 m thick layer of compacted gravel. Settlements were measured during the construction of the bridge. Most of the settlements occured when the thick bottom plate was cast. In total, the settlements for the bridge from construction of the access embankments were very small (less than 5 mm). It seems that the settlements during construction were mostly immediate (elastic). The road is now open for traffic and because of the choice of construction method, a flexible transition from access embankment to bridge has been achieved. If the bridge had been built on piles, serious problems would have occured in the transition. Besides the quality aspect, the lime column solution was also much cheaper. The cost of lime columns is in this case about 50 % of that for piles. Fig. 10.19 shows the bridge during construction.
Fig. 10.19. Construction of bridge on lime-stabilised soil.
10.8
REFERENCES
Barron, R.A. (1948). Conoslidation of Fine-grained Soils by Drain Wells, Transaction of the ASCE, Vol. 113, No 2346. Broms, B. (1984). Stabilization of soil with lime columns, Design Handbook, Third Edition, Lime Column AB, 51 pp.
397
Balasubramaniam, A. P. & Buenesuceso, B. R. (1989). On the Overconsolidated Behavior of Lime Treated Soft Clay. Proceedings of the Twelfth International Conference on Soil Mechanics and Foundation Engineering, Vol 2, pp. 13351338, Rio de Janeiro. Boman, E (1979). Kontroll av kalkpelare, Nordiskt Geotekniker M6te, NGM 1979. Broms, B.B. (1982). Lime columns in theory and practice, International Conference on soil mechanics, pp 149-165, Mexico. Broms, B.B. (1984). Stabilization of soft clay with lime columns, Seminar on soil improvement and construction techniques in soft ground, pp. 120-133, Singapore. Broms, B.B. (1985). Stabilization of slopes and deep excavations with lime and cement columns, Soil Improvement methods, International geotechnical seminar, 3, pp. 127-135, Singapore. Broms, B.B. (1988). Stabilization of soft clay with lime and cement columns in Southeast, International Conference on engineering problems of regional soils, Proceedings, pp. 41-67, Beijing. Broms, B.B., Boman, P., Ingelsson, I. (1978). Investigations of lime columns at Smistav~en, Huddinge, Sweden, Kungliga Tekniska H6gskolan, Jord- och Bergmekanik. Carlsten, E (1989). Manual till Limeset, Statens geotekniska institut, Varianr 248. Carlsten, P., Triink, R. (1992). Deep stabilisation with lime and lime/cement columns - comparison of performance, NGM -92, Vol. 1, pp. 25-30, Aalborg. Carlsten, P. & Ekstriim, J. (1995). Kalk-och kalkcementpelare. Vitgledning f6r projektering, utf6rande och kontroll. Svenska Geotekniska F6reningen. SGF Rapport 4:95. Link6ping. 103 p. Ekstr6m, J. (1994). Kontroll av cementpelare. Slutrapport B 94:3, Chalmers tekniska h6gskola, G6teborg. Halkola, H.A. (1983). In-situ investigations of deep-stabilized soil, Improvement of Ground, European Conference on SMFE, 8, Proceedings, Vol. 1, pp. 33-36, Helsinki. Hartl~n, J. & Carlsten, E (1992). Improvement of soft soil. New technology for foundation engineering, NTFE '92. International geotechnical conference, Hanoi, Oct. 1992. Proceedings, vol. 2.15 p. Holeyman, A., Mitchell, J.K. (1983). Assessment of quicklime pile behaviour, Improvement of Ground, European Conference on SMFE, 8, Proceedings, Vol. 2, pp. 897-902, Helsinki.
398
Lime Columns
Holm, G. (1979). Lime colunm stabilization - experiences concerning strength and deformation properties, V~ig- och Vattenbyggaren, No. 7/8, pp. 45-48. Holm, G. (1994). Deep stabilization by admixtures, 13th International Conference on SMFE, Proceedings, Vol. 3, pp. 1123-1126, New Dehli. Holm G., Bredenberg H. & Broms B.B. (1981). Lime columns as foundation for light structures. Proceedings of the 10th International Conference on Soil Mechanics and Foundation Engineering, Vol 3, pp. 687-694, Stockholm, Sweden. Imse, W. (1972). Messung der Fliessf~igkeit von Zement. Zement, Kalk, Gips, no 3, pp. 147-149. Kimura, T., Nakase, A., Saitoh, K., Kusakabe, O. (1983). On the effectiveness of relatively shallow soil improvements, Improvement of Ground, European Conference on SMFE, 8, Proceedings, Vol. 1, pp. 375-380, Helsinki. Kujala, K. (1983). Use of gypsum in deep stabilization, Improvement of ground, European Conference on SMFE, 8, Proceedings, Vol. 2, pp. 925-928, Helsinki. Kujala, K., Halkola, H., Lahtinen, P. (1985). Design parameters for deep stabilized soil evaluated from in-situ and laboratory tests, 1l th International Conference on SMFE, Proceedings, Vol. 3, pp. 1717-1720, San Francisco. Kujala, K., Lahtinen, P. (1988). Use of cement for deep stabilisation, Nordiske Geotekniker MOte, NGM -88, Artikler och Poster sammendrag, pp. 215-218. Kujala, K., Nieminen, P. (1983). On the reactions of clays stabilized with gypsum lime, Improvement of ground, European Conference on SMFE, 8, Proceedings, Vol. 2, pp. 929-932, Helsinki. Lahtinen, P.O., Vepsiliiinen, P.E. (1983). Dimensioning deep-stabilization using the finite element method, Improvement of Ground, European Conference on SMFE, 8, Proceedings, Vol. 2, pp. 933-936, Helsinki. Lime stabilisation "88 (1988) BACMI technical symposium on lime stabilisation, 1, 88 p, London. Locat, J. et al (1990). Laboratory investigations on the lime stabilization of sensitive clays; Shear strength development. Mise, T., Nishida, K., Kamon, M., Mashima, M. (1992) Soil Improvement, Current Japanese Materials Research 9, Elsevier Applied Science. Pan, Q.Y., Xie, K.H., Liu, X.L., Lin, Q. (1994). Some aspects of the soft clay ground improved with cement columns, 13th International Conference on SMFE, Proceedings, Vol. 3, pp. 1123-1126, New Dehli. Paus, K. (1979). Kalkpelarmetoden- Produktionstekniska synpunkter och praktiska rhd f'6r olika anv~dningsomr~den, KTH Seminarium 27 november 1979.
399 Sherwood, P. (1993). Soil stabilization with cement or lime. Transport Research Laboratory, State-of-the-art review, 152 p. Swedish National Road Administration (1986). Kalkpelare grtmdfrrs~rkning vid v~igbyggnad. V~igverket, V~ig-och Brokonstruktion. Geoteknik. Publikation 1986:72, Bod~a-age. Suzuki, Y. (1982). Deep chemical mixing using cement as hardening agent. Symposium on Soil and Rock Improvement, Bangkok. Tatsouka, F. & Kobayashi, A. (1983). Triaxial strength characteristics of cementtreatad soft clay. Proceedings of the 8th European Conference on Soil Mechanics and Foundation Engineering, Helsinki. Terashi, M., Tanaka, H. (1981). Ground improved by deep mixing method, International Conference on SMFE, 10, Proceedings, Vol. 3, pp. 777-780, Stockholm. Terashi, M., Tanaka, H. (1983). Settlement analysis for Deep Mixing Method, Improvement of Ground, European Conference on SMFE, 8, Proceedings, Vol. 2, pp. 955-960, Helsinki. Terashi, M., Tanaka, H., Kitazume, M. (1983). Extrusion failure of ground improved by the deep mixing method, Asian regional Conference on SMFE, 7, Proceedings, Vol. 1, pp. 313-318, Haifa. Terashi, M., Tanaka, H., Okumura, T. (1979). Settlement analysis for Deep Mixing Method, Pr~e~ings 8th International Conference on SMFE, Vol. 2, pp. 955-960. Terashi, M., Tanaka, H., Okumura, T. (1984) Engineering properties of lime treated marine soils and D.M. method, Proceedings 6th Asian Regional Conference on SMFE, Vol. 1, pp 191-194, Singapore. Wild, S. et al (1989). Fabric development in lime treated clay soils, Ground Engineering, Vol. 22, No 3, pp. 35-37. /~hnberg, H. & Holm, G. (1987). Om inverkan av h~trdningstemperaturen p~t skjuvhhllfastheten hos kalk- och cementstabiliserad jord., Swedish Geotechnical Institute, Report No. 30. Ahnberg H., Bengtsson, P.-E. & Holm, G. (1989). Prediction of strength of lime columns. Proceedings of the 12th International Conference on Soil Mechanics and Foundation Engineering, Vol 2, pp. 1327-1330, Rio de Janeiro. /~hnberg, H. & Holm, G. (1986). Kalkpelarmetoden. Resultat av 10 ~rs forkning och praktisk anvmadning samt framtida forskning, Swedish Geotechnical Institut, Report No. 31.
/~hnberg, H., Johansson, S.E., Retelius, A., Ljungkrantz, C., Holmqvist, L. & Holm, G. (1995). Cement och kalk for djupstabilisering av jord - en kemisk/ fysikalisk studie av stabiliseringseffekter, Swedish Geotechnical Institut, Report No. 48.
400
Chapter 11
Other Methods P. Carlsten, Swedish Geotechnical Institute
11.1
REINFORCEMENT
11.1.1
Traditional m e t h o d s
Different kinds of corduroys are in use under many of the roads that pass over peat bogs. The corduroys are intended to spread the load from the traffic and the embankment to a larger area. Lighter corduroys, in Sweden, have been made as a fascine bed of flesh twigs and branches from spruce and juniper trees. The branches were laid in the transverse direction to the road and were placed in a manner so that a strong mat was formed. The bed should be at least 0.25 m thick when loaded. To slow down decay, the bed was covered with at least 0.2 m of soil with low permeability, e.g. clayey moraine. Heavier corduroys must withstand bending, which leads to considerable dimensions in both thickness and breadth. Corduroys such as these are only recommended (according to the Swedish Code for Road Construction, 1938) for use on smaller roads with low embankment heights.
Fig. 11.1.
Plank road in muskeg, England, Carbon dating has indicated that this road is between 4000 and 4800 years old. Source: Muskeg Engineering Handbook
Reinforcement
401
Comparing the effects of these reinforcements, it seems that the heavier corduroys has a better effect on spreading the load compared to the lighter corduroys and also compared to the geotextiles of today. The cost of constructing corduroys is high (owing to the labour requirement) and because of this corduroys are not used as much today as in earlier years.
11.1.2
Geotextiles
Introduction and description of the method: Geotextiles (permeable textile products) are usually made from polyester, polypropylene or polyethylene, which may either be woven or non-woven. The most common use for geotextiles is as a separating layer between a sot~ subsoil and a coarse-grained fill. The geotextile can also be used as reinforcement in a road embankment to improve the stability or reduce and equalise distribution of settlements. However, there is no generally recognised calculation method that can help in choosing the right type of geotextile. The calculation methods today are based on the assumption that the geotextile provides a force (directed parallel to the textile), which helps the soil to balance the tensile stresses that occur. By using several layers with geotextiles or geotextiles with very high tensile strength the embankment can be seen as a rigid load. Obviously the geotextile can work excellently as a separating layer between the soft subsoil and the embankment. However, there is no need for a separating geotextile on a fibrous peat, since the fibres themselves create a natural separating layer when the peat is compressed. For this reason, it is important to keep the natural surface mat (from roots and plants) intact. In more decayed peat, however, the need for a separating geotextile increases. When looking at the effect of geotextiles as reinforcement in embankments, it seems that there is a lack of knowledge as to how and when they achieve this function. So far, there has been no rational design for geotextile-built low embankments on organic soils. The role of the geotextile is highly dependent upon the safety factor that would have occurred in the absence of reinforcement. Where there is a high safety factor, the geotextile does not alter the behaviour of the embankment. For low embankments (<1.0-1.5 m), it is for this reason questionable whether the geotextile will have any reinforcing effect at all. However, there should be an effect on spreading a concentrated traffic load, such as wheel load. However, when the safety factor is lower, the reinforcement plays an increasingly important role in equalising the distribution of settlements and reducing lateral displacement. The geotextile is thought to decrease the lateral displacement and has a
402
Other Methods
beneficial effect on stability, as well as preventing local failures. Not even with a very strong geotextile is there any obvious effect on the consolidation settlements. For higher embankments, normally other construction methods, such as those described in Chapters 7-10, are more suitable. Nevertheless, the geotextile acts as a separating layer, which is necessary for successful performance of the embankment. The conclusion is that the choice of geotextile should not be governed only by the strength of the geotextile but instead primarily by its ability to provide separation and filtration. The geotextile is also beneficial at the moment when the gravel in the embankment is being compacted. Reading the case histories on objects where geotextiles have been used to reinforce embankments on soft subground, it is obvious that the present design methods are not ready for use in practice. Researchers who have studied reinforcement of embankments on organic soils have encountered many surprises in their field and laboratory experiments. The mechanism by which the geotextile assists the construction have not been adequately defined for organic soils and also the properties and behaviour of the organic soils are poorly understood and difficult to measure. More research is needed before design procedures can be used with confidence.
Construction aspects: As mentioned before, there is no generally recognised method for calculating the reinforcing effect of a geotextile in an embankment. On the other hand, much effort is placed on arriving at such a method. Below are a number of reviews of what some authors have written on the subject. Rowe et al (1985),(1986) compared calculations carried out with the finite element method to determine the effect of geotextiles in the reinforcement of road embankments on fibrous peat. As is often the case, the results of finite element calculations depend largely on the parameters assigned to the soils. Based on their own assumptions, Rowe et al state that the geotextile has an effect on the behaviour of the embankment and that this effect increases the higher the modulus of the geotextile. Rowe et al also state that the main effect of the geotextile is that it decreases the lateral displacement and has a beneficial effect on stability. Not even with a very strong geotextile is there any obvious effect on the consolidation settlements. The geotextile also has the best effect when the total depth of the peat layer is limited. Douglas (1987) contains a proposal for the reinforcement role of geotextiles used in access roads. According to Douglas, the geotextile can reinforce the road structure, reduce vertical displacements under design axle loads and increase the bearing capacity of the road structure. This is achieved through a combination of two loadcarrying mechanisms.
Reinforcement
403
9 membrane action, where the in-plane tension in the geotextile has a vertical component which adds to the load resistance of the subgrade. 9 alteration of the stress field in the gravel fill, caused by horizontal shear developing at the road base/geotextile interface. Both mechanisms would cause a reduction in the stresses applied to the subgrade, and as a result, the displacements seen at the road surface should be reduced. The road cross-section should therefore be measurably stiffer, according to Douglas.
a)
Imposed pressure,, Granutarbase \~ / Geotextite ~ J
Point of ~'";*" Normal to grav,
~=i~ 9
~_
.
"Subgrade "Subgrade,~ b)
Grovel moves with ig~toe:Jidtept~rrd:
I
_~_
/ Anchorage due to
sheer stress development (Te~~:: in geotextite Tension crocks at edge of geotextite~
on subgrode side of geotextite Fig. 11.2.
Schematics of anchorage mechanisms: (a) theoretical, infinitely wide section; (b) typical section built in practice. (Douglas, 1987).
404
Other Methods
There is a significant difference if the geotextile is installed in a low or in a high embankment. When installed in a high embankment on organic soil, large strains will occur in the geotextile due to settlements in the subsoil. In this case, membrane action can develop, and the geotextile can reinforce the embankment structure, decreasing the lateral spread of the fill. If, however the geotextile is installed in a low embankment, live loads will control the situation. The weight of the embankment will be insignificant in comparison to the live load applied by the vehicles using the road. Under these conditions it appears questionable, according to Douglas, whether a geotextile can indeed reinforce the road section at all. Certain conditions must be met if the geotextile is to reinforce the road section through membrane action. Significant tensile forces must develop in the geotextile and these tensile forces must have a significant vertical component. In order to develop the required high tensile force, the geotextile must be adequately anchored. The anchorage depends on the anchored length, as well as on the surcharge afforded by the fill. Thirdly, the anchorage depends on the friction that can be developed between the gravel/geotextile interface and the subgrade/geotextile interface. In the case of a low embankment, the surcharge from the fill is low and often the anchored length is rather short. It has also been shown that there is little or no shear displacement between gravel and geotextile and the surcharge stresses are small. At the interface between geotextile and subgrade the angle of friction that can be developed is usually low, since the subgrade (organic soil) normally has a very low shear strength. Consequently, the angle of friction that can be developed is usually low, even though there are large shear strains at this interface. All these aspects together show that little tensile force can be generated in the geotextile if the embankment is low. For this reason the geotextile will not be able to provide additional reinforcement to the road structure.
Applications: Rowe et al (1984) describes a follow-up from a portion of an embankment built on a soft organic deposit constructed using geotextiles as reinforcement. The geotextile was primarily selected on the basis of a circular arc stability analysis and on assumptions considered suitable in 1981. One assumption was that the tensile forces required to maintain stability can be mobilised at small strains and that tensile force in the fabric acts tangentially to the failure arc. Rowe states that more recent research casts doubt on these assumptions. Large settlements were observed at the instrumented sections, with a maximum fill thickness of 5.7 m and a maximum settlement of 4.7 m. According to Rowe, a
Reinforcement
405
significant component of the settlement appears to have been due to shear distortion, particularly during the second loading stage. Tension cracks were observed in the fill and lateral movements were evident from "mud waves" and tree movements adjacent to the embankment. The follow-up showed that small geotextile strains developed in the first loading stage, despite the large settlements. At this stage the deformation predominantly was due to compression of the peat. At the 2nd loading stage, when large shear deformations were apparent, large fabric strains (in excess of 21%) developed. It was apparent from the field data that the use of a single layer of even a very strong geotextile did not prevent large shear deformations. Rowe et al have made an analysis of instrumented sections using a filfite element programme. This analysis indicates that the role of the geotextile is highly dependent upon the extent of local shear failure that may occur in the absence of reinforcement. At low load levels (i.e. where there is a high safety factor), the geotextile does not alter the behaviour of the embankment. As the extent of local plasticity increases (i.e. the safety factor is reduced), the reinforcement plays an increasingly important role in reducing lateral displacement and the settlement. The major effect of the geotextile is to reduce lateral spreading and to increase stability. Even a very high modulus geotextile will have relatively little effect on consolidation settlements. Rowe is of the opinion that there is a need for more research before design procedures can be used with confidence. Lupien et al (1983) have performed an experimental programme for comparison of the behaviours of two test embankments built with and without geotextiles. They arrive at the conclusion that geotextiles as reinforcement of the foundation are unnecessary when the peat is not susceptible to bearing failure displacement. Figure 11.3 shows an example from Sweden of a road being widened and reinforced by means of the preloading method. The existing ditch is filled with peat, which is compacted with a bucket, and a new ditch is made at a greater distance from the existing road. A small excavation is made in the road and a geotextile is placed on the bottom of the excavation and under the future preloading. The reason for excavating parts of the existing embankment is that the edges of these unpaved roads tend to contain very poor material. A large quantity of fine-graded material has migrated to the edge of the road through the years. This is also partly the reason for using the geotextile, which acts as a separating layer. The geotextile is also considered to help in keeping the two parts of the road together. It is important to keep the surface mat intact under the preloading. This surface mat will reinforce the embankment and also acts as a natural separating layer when the peat is compressed.
406
Other Methods
,_New road 6.5mwide _, Existing road
....... "..Friction.
-_-
rooter'loll "." . ' ' . "
F |
-] Pretoading ~
New ditch
............. ". "." ." ". ' . -
." ." [ ." ." ." ." : '." . ' - . '
." .'-." .'..'..-'
Fig. 11.3. Diagram of strengthening of existing road. The method has been used at several locations in Sweden and has proved successful. When the preloading has continued for about 3-6 months the surcharge can be removed and the road can be paved. At some of the locations follow-ups have been performed. These show that it is important to apply the preloading in steps (see also Chapter 9). The geotextile does not lend any help if the loading is too fast. It is important that the peat be allowed to consolidate between the loadings. Normally the time for consolidation in a peat with a low degree of decomposition is short (= 1 month). The reinforcing effect of the geotextile has not been clarified, but there is no doubt that there is a need for a geotextile as a separating layer between the original fill and the preloading, and that the geotextile also is beneficial when the gravel in the embankment is to be compacted.
11.2
PILE FOUNDATION
11.2.1
Description of the method and construction aspects
Using piles, the load from embankment and traffic can be transferred to a soil with a higher beating capacity. Piling is very seldom used for building roads or dykes on organic soil. The method is much more expensive than others and usually the load must be transferred to the piles with a continuous concrete slab. The horizontal resistance of peat or other organic soil surrounding a pile foundation is so small that the pile group should be so calculated and constructed that it possesses sufficient stability in itself, without the side resistance of the soil being taken into account. If the pile-group is influenced by horizontal forces, those are handled with inclined piles.
407
Pile Foundation
/'f/,~
7//" ~
JfJ
...............................................
./'f'/" ~ _ / : 1 1
~
fff
Fig. 11.4. Pile foundation with a continuous concrete slab.
Large settlements close to the piled area can cause problems with extra load on the piles from negative skin friction. It is important to notice this and consequently protect the construction piles from this extra load. This can be done either by using protecting piles or methods such as light weight fill in the transition to unstabilised area (cf Chapter 7.3.6). It is important to check the stability in the longitudinal direction to the road, for an embankment close to a bridge, (cf Fig. 7.15). The safety factor in this direction should preferably be in the order of 1.7-1.8. With a lower safety factor there will be lateral movements in the piled area, which can cause severe damage to the piles. There are obviously many practical problems when piling is used in organic soil. The bearing capacity of the soil is low and as a consequence it is difficult to reach the piling area with the piling equipment. Sometimes beds of bark are used to raise the bearing capacity of the topsoil. Other solutions are to use mats oftimber or if necessary piles to obtain a good foundation for the piling machine.
11.3
REFERENCES
Allen, T.M. & Kilian, A.P. (1993). Use of wood fiber and geotextile reinforcement to build embankment across soft ground. Transportation Research Record No. 1422, pp. 46-54. Bergado, D.T., Long, P.V., Lee, C.H., Loke, K.H. & Werner, G. (1994). Performance of reinforced embankment on soft Bangkok clay with high-strength geotextile reinforcement. Geotextiles and Geomembranes, Vol. 13, No. 6-7, pp. 403-420.
40 8
Other Methods
Carlsten, P. (1995). Construction methods for roads in peatland areas. European conference on soil mechanics and foundation engineering, 11, Copenhagen, MayJune 1995. Proceedings, vol. 8. The interplay between geotechnical and engineering geology, pp. 8.13-8.18. Carlsten, P. (1989). V~igbyggnad p~ torv. Handbok. V~igverket. Publ. 1989:53. Bod~nge. 35 p. Douglas, R.A. (1987). Modelling geotextile behaviour in thin access road fills over peat subgrades, Canadian geotechnical conference, Geotechnique in resource development, preprint volume, Regina, pp. 111-120. Douglas, R.A. & Kelly, M.A. (1986). Geotextile "reinforced" unpaved logging roads: the effect of anchorage, Geotextiles and geomembranes, Vol. 4., No. 2, pp. 93106. Giroud, J.- P. & Noiray L. (1981). Geotextile-reinforced unpaved road design. ASCE, Vol. 107, No GT9, pp. 1233-1254. Hartl~n, J. & Carlsten, P. (1992). Improvement of soft soil. New technology for foundation engineering, NTFE '92. International geotechnical conference, Hanoi, Oct. 1992. Proceedings, vol. 2. 15 p. Helenelund, K.V. (1975). Geotechnical peat investigations. Baltic conference on soil mechanics and foundation engineering, 1, Gdansk 1975. Proceedings, Vol. 1, pp. 105-123. Holtz, R.D. (1985). Soil reinforcement with geotextiles, Third International Seminar- Soil Improvement Methods, Singapore, pp. 55-74. Holtz, R.D. (1989). Treatment of problem foundations for highway embankments. National Cooperative Highway Research Program. Synthesis of Highway Practice 147. Washington, DC. 72 p. Jarrett, P.M. (1983). Reinforcement of roads on organic terrain. 7th Panamerican conference on SMFE, proceedings, Vol. 1., pp. 329-342, Vancouver. Jewell, R.A. (1982). A limit equilibrium design method for reinforced embankments on soft foundations, Second International Conference on Geotextiles, Las Vegas, Proceedings Vol. 3, pp. 671-678. Landva, A.O. (1980). Geotechnical Behaviour and Testing of Peat, Ph D thesis, Laval University, Quebec. Lupien, C., Lefebvre, G., Rosenberg, P., Pare, J.J. & Lavalle, J.G. (1983). Use of fabrics for improving the placement of till on peat foundation. Transportation Research Record No. 916, pp. 54-59.
409 Milligan, V. & La Rochelle, P. (1984). Design methods for embankments on weak soils, Proceedings of Conference on Polymer grid reinforcement, Thomas Telford Ltd, London, pp 95-102. Rowe, R.K. (1984). Reinforced embankments: Analysis and design, Journal of Geotechnical Engineering, Vol. 110, No. 2, pp. 231-246. Rowe, R.K., Gnanendran, C.T., Landva, A.O. & Valsangkar, A.J. (1995). Construction and performance of a full-scale geotextile reinforced test embankment, Sackville, New Brunsvick. Canadian Geotechnical Journal. Vol. 32, No. 3, pp. 512-534. Rowe, R.K. & Soderman, K.L. (1986). Reinforced embankments on very poor foundations. Geotextiles and geomembranes, Vol. 4., No 1, pp. 65-81. Rowe, R.K. & Soderman, K.L. (1985). Geotextile reinforcement of embankments on peat. Geotextiles and geomembranes, Vol. 2., No 4, pp. 277-298. Rowe, R.K., MacLean, M.D. & Barsvary, A.K. (1984). The observed behaviour ofa geotextile-reinforced embankment constructed on peat, Canadian Geotechnical Journal, Vol. 21, No. 2, pp. 289-304 Rowe, R.K., MacLean, M.D. & Soderman, K.L. (1984). Analysis ofa geotextilereinforced embankment constructed on peat, Canadian Geotechnical Journal, Vol. 21, No. 3, pp. 563-576 Swedish National Road Administration (1995). Bankp~dning. Allm~inteknisk beskrivning. V~igverket. Publikation 1994:68. Borl~inge. 30 p.
410
Author Index A Aas, G 170 Aboshi, H 107 Adams, JI 100 Mmeida, MSS 175 AlmOn, KE 93 Andrejko, MJ 21 Asaoka, A 205, 206, 207, 322 Azzouz, AS 139, 171, 185 B
Baecher, GB 140 Bailey, WA 146 Baligh, MM 171 Baranski, T 112, 113 Barden, L 105, 129 Barron, RA 223, 224, 308, 311, 312, 330, 331 Becker, DE 169 Bergdahl, U 35, 40, 43, 164, 168 Berre, T 125 Berry, PL 105, 127, 188 Bezirci, MH 120, 121 Biot, AW 230 Bishop, AW 124, 154 Bjelm, L 34 Bjerrum, L 102, 122, 139 Broms, B 356 C Cadling, L 64 Cargill, KW 212, 217
Author index
Carillo, N 223 Carlsten, P 29, 103, 128, 185, 196, 198, 199, 204, 205, 302, 303,304, 356 Casagrande, A 97 Chameau, JL 175 Chang, YSF 73 Chen, RH 175 Choi, YK 216 Chowdhury, RN 175 Christopher, BR 323 Cousins, BF 148, 145, 150 Crapps, DK 47 Crooks, JH 175 D
Dhowian, AW 100, 105, 188 Douglas, RA 402, 403, 404 Drozd, PA 196, 197 Duncan, JM 192, 231 Dyvik, R 123 E
Edil, TB 100, 105, 188 Ekstr0m, A 252, 253, 254 Ekstr0m, J 356 Eriksson, L 58 F
Fadum, RE 200, 201 Felix, B 112 Flaate, K 185, 186, 196 Folkes, DJ 175 Foott, R 193 Fredlund, DG 145, 159 Fredriksson, D 94 FOrstenberg, A 311
411
412
Author index
G Garneau, R 120 Gens, A 171, 172 Gibson, RE 105, 208, 209, 211 Godlewski, PM 189 Golebiewska, A 140, 261,262 Graham, J 126 Granlund, E 14 Gray, H 199, 200 It Hallden, BE 5 Hanna, TH 66, 71 Hanrahan, ET 120, 183, 261 Hansbo, S 16, 19, 20, 21, 29, 89, 90, 99, 120, 214, 215, 223, 225, 305, 306, 307, 308, 310, 311, 315, 323, 372 Head, KH 115, 117, 125, 130 Helenelund, KV 29, 213 Hight, DW 175 Hobbs, NB 8, 9, 11, 12, 22, 25, 26, 28, 29, 87, 188 Holtz, RD 311,323,408 Houlsby, GT 169, 192, 232 Hudson, RR 73 Hungr, O 172, 173 I
Ingan~is, J 228 J
Jamiolkowski, M 58, 144, 163, 223, 311 Janbu, N 62, 107, 116, 156, 157, 158, 194 Jardine, RJ 175 Johansson, HG 34 Johnson, SJ 293, 296, 298, 299 Juarez-Badillo, E 229
Author index K
Kallstenius, T 51 Karlsson, R 16, 19, 20, 21, 29, 88 Kivinen, E 7 Kjellin, B 94 Kjellman, W 122 Koda, E 225, 315, 316, 318 Koerner, RA 315 Kogure, K 185, 186 Konvalov, PA 21 Kortjunov, SS 26 Krahn, J 145 Kulhavy, FH 100 L Lacasse, SM 175 Ladd, CC 122, 140, 161, 162, 163, 164, 193, 298, 299, 320, 321, 322 Landva, AO 21, 22, 23, 24, 25, 26, 28, 29, 40, 60, 61, 62, 63, 65, 73, 122, 125, 139, 143, 178, 183 La Rochelle, P 183 Larsson, R 37, 39, 49, 61, 64, 92, 96, 98, 99, 107, 108, 109, 111, 118, 122, 127, 139, 140, 163, 164, 165, 166, 169, 170, 185, 189, 190, 191, 193, 202, 216, 228, 233, 298, 302 Law, KT 121 Lawrance, CA 315 Lazarev, AV 26 Lebihan, JP 120 Lechowicz, Z 101, 117, 118, 119, 164, 166, 169, 217, 233 Lee, CF 175 Leroueil, S 97, 98, 175, 233 Lo, K 105, 175 Lowe, J 107 Ludwig, CA 209, 210 Lupien, C 405
413
414 M
MacFarlane, IC 26 Magnan, JP 216, 233 Magnusson, NH 13 Marchetti, S 45, 46, 47 Mayne, PW 100 Mesri, G 189, 216, 229 Mitchell, JK 118 Morgenstem, NR 145, 151, 152, 158, 159 Moum, J 95 Mulabdic', M 61 N
Niesche, H 196, 197, 198 O
Odenstad, S 252 Okruszko, H 21 Osterberg, JO 200, 201 Ostromecki, J 196, 197 Ozden, ZS 105, 188 P
Pakarinen, P 7 Paute, JL 112 Pheeney, PE 23, 24, 25, 26, 28, 29 Poskitt, TJ 105, 127, 188 von Post, L 18, 19, 22, 23, 25, 26, 29, 187, 191 Poulos, HG 194 Price, VE 158 R
Radforth, NW 26, 27, 28, 29 Ramalho-Ortigao, JA 175 Rixner, JJ 306, 308, 323
Author index
Author index
Robertson, PK 36, 38 Rogers, MG 183 Rowe, PW 129 Rowe, RK 402, 404, 405 Runesson, K 169, 192, 233 S
Samarski, AA 221 Samson, L 301 Schiffman, RL 208, 209, 225 Schmertmann, JH 47, 48 Schmidt, B 100 Schneider, S 26 Schwab, EF 62, 65 Senneset, K 62 Singh, A 118 Skempton, AW 99 Smith, RE 107 Stefanoff, G 42 Steinbrenner, W 194, 195 Szymanski, A 101, 105, 117, 118, 119, 210, 217, 221, 231, 233 S/~llfors, G 97, 98, 99, 100, 108, 109, 110, 111, 169, 233 T Tallis, JH 14 Tavenas, F 103, 125, 127, 175, 233 Taylor, DW 7, 147 Terzaghi, K 146, 229 Teunissen, JAM 175 Thalme, O 93 Tokheim, O 107 Torstensson, BA 64 Tremblay, M 58
415
416
Author index
V Verruit, A 230 Viberg, L 32 V~gverket 276, 277, 356, 363, 365, 366, 367 W Wahls, HE 107 Wasti, Y 120, 121 Wesley, LD 124 Whitman, RV 146 Wilson, NE 105, 188 Wolski, W 21, 64, 65, 73, 99, 112, 113, 166, 169, 189, 207, 225, 262, 312, 313, 314, 315, 316 Wood, DM 120 Wright, S 146 Wroth, CP 169, 192, 232 Y Yong, RN 191, 209, 210, 211 Z
Zajac, WN 196, 197 Zawadzki, S 21
A /~hnberg, H 356
417
Subject index acidity aerial photo interpretation ash content
25 32, 33 21, 22, 86, 91,261
backfill band-shaped drain bellows hose settlement gauge bending biogenic matter boundary conditions buckling bulk modulus
274 304, 305, 306, 308, 311,312, 315 68, 69 315, 400 4,5 76, 105, 122, 181,211,213, 215, 216, 219, 221,322 70, 312, 315 116, 184, 231, 232
calcareous gyttja calciferous soil carbonate content Carillo equation Carlsten's diagram Casagrande method chemical sediment Chittick apparatus classification
8, 14, 312 20, 92 10, 14, 20, 93 307, 350 2O5 97 5 94 16, 19, 20, 21, 22, 25, 26, 27, 28, 29, 34, 37, 39, 47, 48, 49, 85
coefficient of earth pressure at rest of permeability of secondary compression of volume change colorimeter compression index compressiometer consistency limits consolidation analysis Asaoka equation average degree of Barron-Hansbo equation Carlsten formula coefficient of
99,100 101, 104, 183, 212, 231,304 101,105,106,183,189,229,298,299,301,325 101,185,186, 198,208 92 101,104, 105,183,185, 189,202,211,301,325 85,103, 104 17,22,29, 88 203,230 206 308,330,332 308,312,330,331 204 104,105,110,188,207,213,301,308,320,321,322
418
Subject index
degree of 204, 208, 220, 225, 299, 307 Gibson and Schiffrnan equation 208 large strain analysis 209 layered soils 213 parameters 188 Schiffman and Gibson approach209 settlement 200, 202, 204, 295, 328 strain 202 Terzaghi assumption 207 construction 240, 246 construction monitoring 302, 318 convective coordinate system 209, 210, 211,225 creep 96, 105, 117, 166 creep test 118, 119 deformation analysis parameters degree of humification density of soil design designation dilatometer test direct simple shear test disturbed zone drain resistance drainage blanket dry density Duncan-Chang model dy dy-bearing soil dynamic probing
112, 185, 230 109, 114, 183 22, 191 85 240 24 45, 61, 138, 141, 142, 143 122, 139, 140, 144, 145, 171 309, 311,373 308, 309, 311 307 86 231,232 6, 17, 18, 19, 20, 275 17 35, 42
elastic strain equivalent diameter of drain excess pore pressure
105,232 308 161, 182, 188, 203,211,213,216, 221,232, 297, 303
factor of safety fall-cone test ferrous sulphide field instruments final deformation flood-plain sediments
145, 148, 151, 171 88, 120,140, 142, 143 15, 17, 95,96 319 192,196 8, 17
Subject index
419
flow conditions
130
general survey Geodrain geotechnical classification geotextiles groundwater gyttja
gyttja-bearing soil
32, 33 312, 316 29 401 53 4, 5, 6, 8, 10, 13, 14, 15, 16, 17, 18, 19, 20, 33, 61, 101,124, 126, 140, 183,190, 197, 244, 247, 255, 270, 275, 302, 311,312, 315,325, 329, 330, 331,332, 335, 336, 339, 340, 341,342, 343,350, 351,355,375, 377, 395 17, 244
horizontal drainage displacement hose settlement gauge humification humus
223,307, 308 182, 193 66, 67, 68, 320, 376 22, 26, 191 6, 18,20
immediate settlement initial vertical stress tangent modulus void ratio inclinometer influence factor horizontal displacement
193,328, 339
inventory phase
202, 301,326 115 197, 202, 211,325 71, 72, 78, 312, 319, 320 193, 194, 200, 201 71, 72, 74, 78, 79, 182, 183, 192, 193, 194, 230, 312, 314, 315, 316, 320, 329, 339 194 194 195 182, 193, 194, 402 97, 99, 104, 110, 114, 160, 169, 185, 188, 199 200, 201,301,320, 326, 327, 333,334, 337, 338, 347, 348, 349, 368 32
kinking
312,315
laboratory investigations large strain analysis
85, 359 209
Janbu's chart Poulos' chart Steinbrenner's formula vertical displacement vertical stress
420
Subject index
lateral displacement lightweight fill lime/cement columns admixture field investigation laboratory investigation settlement shear strength stability test column lime column penetrometer liquid limit load adjustment loading berm
113,285 247,259 244,247,355 361 358,375 359 365,377 375 363,377,388 360 44 88,139, 165 244,246,247 74,294
magnetic screw settlement gauge mandrel marl mapping mire modulus number modified coefficient of secondary compression compression index swelling index recompression index monitoring equipment Moum apparatus
69, 70 305, 306, 307, 318 4, 5, 8, 14, 15, 17, 19, 20 32 6, 8, 9, 10, 14, 15 110, 185, 187, 371
Niesche's diagram
198
observational data oedometer modulus test organic clay content organogenous soil Ostromecki nomogram overconsolidation ratio
319, 320
189,229 105, 189, 229 105, 228, 229 105 66, 76, 77, 79, 312, 313,320 95
110, 183, 185 101, 185, 188, 190, 212 8, 47, 245, 377 4, 8, 15, 16, 20, 25, 87, 91 5 197 100, 123, 144, 163
Subject index parameter selection Passon apparatus peat amorphous areas drill fibrous peat
421 183 93 18, 100, 123, 189, 245, 247, 312, 315 6,8,33 52, 53 15, 16, 18, 19, 22, 25, 40, 51, 52, 61, 62, 63, 65, 73, 81, 85, 90, 91, 99, 103, 120, 122, 127, 138, 189, 240, 244, 245, 247, 302, 311,315, 401,402 85 51 35 35 40 42 43, 44 58, 127, 160, 227 211,227, 230 58, 141,318 54, 55, 56, 57, 58, 59, 66, 80, 81,318, 319, 320, 322 50, 51
pseudo fibrous sampler penetration testing cone penetration test standard penetration test dynamic probing weight sounding test permeability piece-wise analysis piezocone test piezometer piston sampler plastic core limit strain plasticity index chart plate load test Poisson's ratio pore pressure porosity precompression technique preconsolidation pressure prefabricated drains preloading pressure berms pressuremeter test progressive displacement
305 22, 25, 88, 90 117, 232 25 90, 99, 139, 164, 193,228, 328 90, 91 62 114, 116, 117, 184, 193,230 36, 53, 66, 80, 149, 213,232, 303 88 294, 311,318, 322 96, 268 3O5 203,227, 240, 244, 245, 256, 264,294, 377, 392, 405 246, 250 62 246, 281
Radforth system
28
422
Subject index
radial coordinate drainage raised bog recompression index reduced coordinate system reinforced soil replacement river dyke root threads Rowe cell sampling sand drain screw auger plate test secant modulus secondary compression compression behaviour settlements settlement due to primary consolidation due to secondary compression final gauges initial measurement total postconstruction sedge fibres shear modulus strength drained effective increase normalized undrained undrained shell soil
223,226 223,308 8, 10, 13, 14, 17 105, 202, 211, 301 209, 211 400 244, 274 323 24 121, 122, 129, 130 50 304, 305, 306, 308 52 62 115, 117 188, 189, 298, 299, 300, 351 188, 351 203,340 329, 339 200, 300, 332, 340 192, 196 66, 67, 68, 69, 70, 320 193 66, 67, 68, 70, 71, 78 192, 198, 331,332, 338, 341 295, 303 23 116 120, 138, 161, 163, 168, 169, 170, 283, 284, 333, 334, 389, 391 126, 145 144 163,253,254, 333,334 144, 164, 165, 168, 335 60, 120, 123, 126, 138, 141, 143, 144, 161, 163, 164, 165, 168, 169 19
Subject index shrinkage limit single-stage embankment site investigation sleeve smear smell specific gravity soil improvement layer sequence model radar square pattern stability analysis Cousin's stability chart finite element analysis Janbu general method Janbu simple routine procedure Morgenstern stability chart Morgenstern-Price method non-circular failure surface simplified Bishop method Swedish circle method Taylor's stability chart three-dimensional staged construction embankment Steinbrenner's formula stress distribution calculation method conditions Fadum's chart Osterberg's chart surcharge ratio time of removal surcharging subsoil deformation surface mat svartmocka
423 88 160, 162, 293 31 305,315,323 224,225,312 25 86,87 247 33 230 34 308,331 137 150 175 156 157 152 158 159 154 151 147 171 244, 247, 293 162, 323, 345 195 199 114 201 201 247 298,301,346,347 227,229, 295 203,223,227,229,293,295,296,298,299,300,303, 315,345,346 181 294,401,405 4,15
424
Subject index
swelling index synthetics
227, 301,304, 350 105, 228, 229 305
tangent modulus tensile strength test embankment topsoil triaxial test triangular pattern
115 25 68, 71, 72, 73, 216, 312, 394, 405 18 112, 124 308
undrained modulus
193,328, 329, 339,
vane shear test field correction factor laboratory vertical displacement drain spacing drainage void ratio
120, 122, 138, 141 63, 139, 140, 141 65, 139 120, 140, 142
water content well resistance wood renmants working mat
182, 194, 402 223,247, 304, 323 310, 330 244, 245, 309 87, 104, 127, 128, 185, 186, 190, 197, 202, 203, 207, 211, 227 23 224, 225, 308, 309, 311, 312 24 294, 307
yield envelope Young's modulus
117, 233 114, 184, 231,232
