PROCEEDINGS OF THE 2ND INTERNATIONAL SYMPOSIUM ON FRONTIERS IN OFFSHORE GEOTECHNICS, PERTH, AUSTRALIA, 8–10 NOVEMBER 2010
Frontiers in Offshore Geotechnics II Editors Susan Gourvenec & David White Centre for Offshore Foundation Systems, University of Western Australia
© 2011 by Taylor & Francis Group, LLC
Cover photo credit © Norske Hydro
CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2011 Taylor & Francis Group, London, UK Typeset by MPS Limited, a Macmillan Company, Chennai, India Printed and bound in Great Britain by Antony Rowe (a CPI Group Company), Chippenham, Wiltshire All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by:
CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail:
[email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl
ISBN: 978-0-415-58480-7 (Hbk + CD-ROM) ISBN: 978-0-203-83007-9 (ebook) © 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Table of Contents
Preface
XIII
Committees
XV
Reviewers
XVII
1 Keynotes A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot, T. Lunne & T.I. Tjelta Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat, H. Dendani, A. Puech & J.-F. Nauroy Geotechnics for subsea pipelines D.J. White & D.N. Cathie
3 33
59 87
Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph
125
New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin, E.C. Clukey, J. Garnier & R. Phillips
155
Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert, J.D. Murff & E.C. Clukey
189
2 Geohazards and gas hydrates Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney
203
Deepwater Angola part I: Geohazard mitigation A.J. Hill, J.G. Southgate, P.R. Fish & S. Thomas
209
Deepwater Angola part II: Geotechnical challenges A.J. Hill, T.G. Evans, B. Mackenzie & G. Thompson
215
Shallow gas hazard linked to worldwide delta environments S. Kortekaas, E. Sens & B. Sarata
221
Analysis of submarine flow slides in fine silty sand P.V. Lade & J.A. Yamamuro
227
Hydrate dissociation around oil exploration infrastructure A.K. Sultaniya, J.A. Priest & C.R.I. Clayton
233
An investigation of past mass movement events in the West Nile Delta S. Thomas, L. Bell, K. Ticehurst & P.S. Dimmock
239
Deformation of seabed due to exploitation of methane hydrate reservoir J. Yoneda, M. Hyodo, Y. Nakata, N. Yoshimoto & R. Orense
245
© 2011 by Taylor & Francis Group, LLC
V
3 In situ site characterisation and pore pressure measurement A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils D. Borel, A. Puech & S. Po
253
Enhancement of the ball penetrometer test with pore pressure measurements N. Boylan, M.F. Randolph & H.E. Low
259
Laboratory free falling penetrometer test into clay S.H. Chow & D.W. Airey
265
Offshore sediment overpressures: Overview of mechanisms, measurement and modeling B. Dugan, T.C. Sheahan, J.M. Thibault & T.G. Evans
271
Angolan deepwater soil conditions: GIS technology development for sediment characterization P. Enjaume, M. Hamon, K. Epalanga & B. Mackenzie
277
Strength measurement in very soft upper seabed sediments P. Kelleher, H.E. Low, C. Jones, T. Lunne, S. Strandvik & T.I. Tjelta
283
CPT in polar snow – preliminary observations A.B. McCallum, A. Barwise & R. Santos
289
Parametric study of a free-falling penetrometer in clay-like soils M. Nazem & J.P. Carter
293
The future of deepwater site investigation: Seabed drilling technology? J.J. Osborne, A.G. Yetginer, T. Halliday & T.I. Tjelta
299
Mini T-bar testing at shallow penetration A. Puech, M. Orozco-Calderón & P. Foray
305
Piezometer installation in deepwater Norwegian Sea T.I. Tjelta & J. Strout
311
Luva deepwater site investigation programme and findings T.I. Tjelta & A.G. Yetginer
315
Investigations into novel shallow penetrometers for fine-grained soils Y. Yan, D.J. White & M.F. Randolph
321
Seabed drilling vs surface drilling – a comparison A.G. Yetginer & T.I. Tjelta
327
4 Soil characterisation and modelling Rheological behaviour of soft clays P.E.L. de Santa Maria, I.S.M. Martins & F.C.M. de Santa Maria
335
A three-dimensional finite element study of the direct simple shear test J.P. Doherty & M. Fahey
341
Repeated loading and unloading of the seabed H.J.E. Hu, K.K. Tho, C.T. Gan, A.C. Palmer & C.F. Leung
347
A new interpretation of the simple shear test H.A. Joer, C.T. Erbrich & S.S. Sharma
353
Physical modelling of the crushing behaviour of granular materials H.A. Joer & S.S. Sharma
359
New evidence for the origin and behaviour of deep ocean ‘crusts’ M.Y-H. Kuo, M.D. Bolton, A.J. Hill & M.J. Rattley
365
Soil unit weight estimated from CPTu in offshore soils P.W. Mayne, J. Peuchen & D. Bouwmeester
371
© 2011 by Taylor & Francis Group, LLC
VI
Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola M.J. Rattley, A.J. Hill, S. Thomas & B. Sampurno Simplified calibration procedure for a high-cycle accumulation model based on cyclic triaxial tests on 22 sands T. Wichtmann, A. Niemunis & Th. Triantafyllidis Understanding cyclic loading behavior of soil for offshore applications J. Yang
377
383 389
5 Shallow foundations Observations of shallow skirted foundations under transient and sustained uplift H.E. Acosta-Martinez, S. Gourvenec & M.F. Randolph
397
Numerical study of grillage foundations on sand under combined VHM loading M. Banimahd, A. Maconochie & J. Oliphant
403
The vertical bearing capacity of grillage foundations in sand M.F. Bransby, P. Hudacsek, J.A. Knappett, M.J. Brown, N. Morgan, D.N. Cathie, R. Egborge, A. Maconochie, G.J. Yun, N. Brown & A. Ripley
409
Behaviour of skirted footings on sand overlying clay C.T. Gan, K.L. Teh, C.F. Leung, Y.K. Chow & S. Swee
415
Numerical study of piping limits for suction installation of offshore skirted foundations and anchors in layered sand L.B. Ibsen & C.L. Thilsted
421
Shallow foundation performance in a calcareous sand B.M. Lehane
427
A numerical study of the vertical bearing capacity of skirted foundations D.S.K. Mana, S. Gourvenec & M.F. Randolph
433
The effect of torsion on the sliding resistance of rectangular foundations J.D. Murff, C.P. Aubeny & M. Yang
439
Foundation design challenges of the MCR-A skirted gravity platform L. Tapper, C. Humpheson & B.M. Lehane
445
Constructing breakwater with prefabricated caissons on soft clay S. Yan, X. Feng & J. Chu
451
6 Piled foundations Simplified analysis of laterally loaded pile groups F.M. Abdrabbo & K.E. Gaaver
459
Behavior of piles under combined lateral and axial loading M. Achmus & K. Thieken
465
Investigations on the behavior of large diameter piles under cyclic lateral loading M. Achmus, J. Albiker & K. Abdel-Rahman
471
BP Clair phase 1 – Pile driveability and capacity in extremely hard till T.R. Aldridge, T.M. Carrington, R.J. Jardine, R. Little, T.G. Evans & I. Finnie
477
Photoelastic investigation into plugging of open ended piles J. Dijkstra, E.A. Alderlieste & W. Broere
483
Soil-pile interaction during extrusion of an initially deformed pile C.T. Erbrich, E. Barbosa-Cruz & R. Barbour
489
BP Clair phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till T.G. Evans, I. Finnie, R. Little, R.J. Jardine & T.R. Aldridge
495
© 2011 by Taylor & Francis Group, LLC
VII
Pile driving experiences in Persian Gulf calcareous sands K. Fakharian & I.H. Attar
501
FLAC3D analysis on soil moving through piles E.H. Ghee & W.D. Guo
507
Cyclic loading of barrettes in soft calcareous rock using Osterberg cells C.M. Haberfield, D.R. Paul, M.C. Ervin & G.A. Chapman
513
Shaft capacity of drilled and grouted piles in calcareous sandstone B.M. Lehane
519
Numerical analysis of mudmat contribution to capacity of piled offshore platforms L.S.D. Lorenti, M.A. Ismail & B.M. Lehane
525
Simplified numerical model for analysis of offshore piles under cyclic lateral loading M.M. Memarpour, M. Kimiaei & M. Shayanfar
531
Centrifuge modelling of rapid load tests with piles in silt and sand C.T. Nguyen, H. van Lottum, P. Hölscher & A.F. van Tol
537
Field measurements on monopile Dolphins A. Sadeghi-Hokmabadi & A. Fakher
543
Behaviour of driven tubular steel piles in calcarenite for a marine jetty in Fujairah, United Arab Emirates J. Thomas, M. van den Berg, F. Chow & N. Maas CPT-Based design method for axial capacity of offshore piles in clays B.F.J. Van Dijk & H.J. Kolk
549 555
7 Foundations for renewable energy Evaluation of pile capacity approaches with respect to piles for wind energy foundations in the North Sea M. Achmus & M. Müller
563
Installation of suction caissons for offshore renewable energy structures O.J. Cotter, B.W. Byrne & G.T. Houlsby
569
Lateral behaviour of large diameter monopiles at Sheringham Shoal Wind Farm L. Hamre, S. Feizi Khankandi, P.J. Strøm & C. Athanasiu
575
Centrifuge modelling of offshore monopile foundation R.T. Klinkvort & O. Hededal
581
Gravity based foundations for the Rødsand 2 offshore wind farm, Denmark L. Krogh, J.H. Lyngs & J.S. Steenfelt
587
Geotechnics for developing offshore renewable energy in the US M. Landon Maynard & J.A. Schneider
593
Engineering issues for fixed offshore wind turbines on Lake Michigan Mid Lake Plateau, USA P.J. Lang, J.A. Schneider, K. Smith & T. McNeilan
599
Centrifuge model tests on piled footings in clay for offshore wind turbines B.M. Lehane, W. Powrie & J.P. Doherty
605
Design of monopile foundations in sand for offshore windfarms M. Saue, T.E. Langford & N. Mortensen
611
Experimental evaluation of backfill in scour holes around offshore monopiles S.P.H. Sørensen, L.B. Ibsen & P. Frigaard
617
An investigation of the use of a bearing plate to enhance the lateral capacity of monopile foundations K.J.L. Stone, T.A. Newson, M. El Marassi, H. El Naggar, R.N. Taylor & R.J. Goodey
623
© 2011 by Taylor & Francis Group, LLC
VIII
Optimizing site investigations and pile design for wind farms using geostatistical methods: A case study B. Stuyts, V. Vissers, D.N. Cathie, C. Jaeck & S. Dörfeldt
629
Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a high-cycle accumulation model T. Wichtmann, A. Niemunis & Th. Triantafyllidis
635
Cyclic accumulation effects at foundations for offshore wind turbines H. Wienbroer, H. Zachert, G. Huber, P. Kudella & Th. Triantafyllidis
641
Study on soil-structure interaction of suction caisson by large-scale model tests B. Zhu, D.Q. Kong, L.G. Kong, R.P. Chen & Y.M. Chen
647
8 Jack-up units Simplified VH equations for foundation punch-through sand into clay J.-C. Ballard, P. Delvosal, P.H. Yonatan, A. Holeyman & S. Kay
655
Characterisation of undrained shear strength using statistical methods B. Bienen, M.J. Cassidy, M.F. Randolph & K.L. Teh
661
Centrifuge modelling of spudcan deep penetration in multi-layered soils M.S. Hossain, M.F. Randolph & Y.N. Saunier
667
A probabilistic approach to the prediction of spudcan penetration of jack-up units G.T. Houlsby
673
An assessment of jackup spudcan extraction O.A. Purwana, H. Krisdani, X.Y. Zheng, M. Quah & K.S. Foo
679
3D FE analysis of the installation process of spudcan foundations G. Qiu, S. Henke & J. Grabe
685
Undrained bearing capacity of deeply embedded foundations under general loading Y. Zhang, B. Bienen, M.J. Cassidy & S. Gourvenec
691
9 Anchoring systems Trajectory prediction for drag embedment anchors under out of plane loading C.P. Aubeny & C.-M. Chi
699
Setup following keying of plate anchors assessed through centrifuge tests in kaolin clay A.P. Blake, C.D. O’Loughlin & C. Gaudin
705
Seismically-induced displacements of a suction caisson in soft clay A.J. Brennan, S.P.G. Madabhushi & P. Cooper
711
SEPLA keying prediction method based on full-scale offshore tests R.P. Brown, P.C. Wong & J.M. Audibert
717
Set-up of suction piles in deepwater Gulf of Guinea clays J.-L. Colliat & D. Colliard
723
Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up time J.-L. Colliat, H. Dendani, H.P. Jostad, K.H. Andersen, L. Thorel, J. Garnier & G. Rault
729
Numerical FEM and laboratory study of the bearing capacity factor Nc for plate anchors L.N. Equihua-Anguiano, M. Orozco-Calderón, P. Foray & M. Boulon
735
Caisson capacity in clay: VHM resistance envelope – Part 2: VHM envelope equation and design procedures S. Kay & E. Palix Installation and in-place assessment of drag anchors in carbonate soil M.P. O’Neill, S.R. Neubecker & C.T. Erbrich © 2011 by Taylor & Francis Group, LLC
IX
741 747
Caisson capacity in clay: VHM resistance envelope – Part 1: 3D FEM numerical study E. Palix, T. Willems & S. Kay
753
Numerical investigation of the behaviour of suction caissons in structured clays S. Panayides, M. Rouainia & A. Osman
759
Cyclic moment loading of suction caissons in sand B. Zhu, B.W. Byrne & G.T. Houlsby
765
10 Pipelines and risers Multidirectional analysis of pipeline-soil interaction in clay R.G. Borges & J.R.M.S. Oliveira
773
Geotechnical challenges for deepwater pipeline design – SAFEBUCK JIP D.A.S. Bruton, M. Carr & F. Sinclair
779
Large deformation finite element analysis of vertical penetration of pipelines in seabed S. Chatterjee, M.F. Randolph, D.J. White & D. Wang
785
Implementation of geotechnical techniques in the analysis of pipeline response G. Cumming & N. Brown
791
Lateral soil resistance to an untrenched pipeline under the action of ocean currents F.P. Gao, S.M. Yan, E.Y. Zhang, Y.X. Wu & X. Jia
797
Vertical cyclic testing of model steel catenary riser at large scale T.E. Langford & V.M. Meyer
803
Kupe gas project pipeline – optimisation of discrete rock berm design shore approach B.L. Larsson
809
Model test studies on soil restraint to pipelines buried in sand R. Liu, S.W. Yan & J. Chu
815
Pipe-soil interaction on clay with a variable shear strength profile D.R. Morrow & M.F. Bransby
821
Sweeping behaviour of shallowly-embedded pipeline during cyclic lateral movement T. Takatani
827
Advanced nonlinear hysteretic seabed model for dynamic fatigue analysis of steel catenary risers I.H.Y. Ting, M. Kimiaei & M.F. Randolph
833
Mobilization distance in uplift resistance modeling of pipelines J. Wang, S.K. Haigh, N.I. Thusyanthan & S. Mesmar
839
Theoretical, numerical and field studies of offshore pipeline sleeper crossings Z.J. Westgate, M.F. Randolph, D.J. White & P. Brunning
845
Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE® D.J. White, A.J. Hill, Z.J. Westgate & J.-C. Ballard
851
11 Trenching, ploughing, excavation and burial Influence of object geometry on penetration into the seabed A. Ivanovi´c, R.D. Neilson, G. Giuliani & M.F. Bransby
859
Investigation into the effect of forecutters on plough performance K.D. Lauder, M.J. Brown, M.F. Bransby & J. Pyrah
865
State-of-the-art jet trenching analysis in stiff clays J.B. Machin & P.A. Allan
871
Numerical modelling of soil around offshore pipeline plough shares W. Peng & M.F. Bransby
877
© 2011 by Taylor & Francis Group, LLC
X
Anchor–chain–rockfill–soil interaction: Evolution of design methods H. van Lottum & H.J. Luger
883
Development of a jet trenching model in sand J.-F. Vanden Berghe, J. Pyrah, S. Gooding & H. Capart
889
12 Design and risk Structural factors affecting the system capacity of jacket pile foundations J.Y. Chen, R.B. Gilbert, J.D. Murff, A.G. Young & F.J. Puskar
897
The new API Recommended Practice for Geotechnical Engineering: RP 2GEO P. Jeanjean, P.G. Watson, H.J. Kolk & S. Lacasse
903
Comparison of ISO 19901-2 and API RP 2A seismic design criteria for a site in the Caspian Sea, Turkmenistan Z.A. Lubkowski, J.E. Alarcon & Z.A. Razak
909
Offshore geotechnics – safe and sustainable J. Peuchen & J. Haas
915
Author Index
921
© 2011 by Taylor & Francis Group, LLC
XI
© 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Preface
The International Symposia on Frontiers in Offshore Geotechnics (ISFOG), hosted by the Centre for Offshore Foundation Systems (COFS) at the University of Western Australia (UWA), provide a platform for academics and practitioners to discuss the emerging challenges in offshore geotechnical engineering and present recent practice and research. Offshore design and construction presents unique challenges to geotechnical engineers. Many of the challenges that are routinely encountered today have persisted for decades and continue to be gradually overcome with advances in technology and analysis. Meanwhile, new challenges are faced as the industry moves to harness resources in deeper waters, harsher environmental conditions and new frontier regions with previously-uncharacterized seabed conditions. The inaugural ISFOG was hosted by COFS in September 2005 and arose from discussions with colleagues from around the world which highlighted a consensus that there was a niche for a new specialist offshore geotechnical conference. That symposium was attended by 182 delegates from 22 countries and the proceedings contained 7 keynote papers and 127 peer-reviewed general papers that reflected the state-of-the-art of practice and research (Gourvenec and Cassidy, Eds, Taylor & Francis, ISBN 041539063X). Immediately afterwards and over the years following the symposium, we received extremely positive feedback from many of the participants, who had found the event to be both valuable and enjoyable. As a result, in the latter part of 2007 we canvassed the International Advisory Committee (IAC) of the inaugural ISFOG – a 30-strong team of leading academics and practitioners – for their views on a second ISFOG. They offered their unreserved support for a repeat event and it was agreed that a 5-year gap would be appropriate between symposia. The original Australian Research Council grant that led to the establishment of COFS expired in 2005. In some respects the first ISFOG was a chance to mark the achievements of COFS during its original tenure, at what might have been its zenith. However, COFS has grown since 2005 and is now more active and larger than ever. We therefore felt able to organise a 2nd International Symposium on Frontiers in Offshore Geotechnics, inviting friends and colleagues, old and new, to visit Perth again in November 2010. On advice from the IAC, we retained the single session format of the inaugural ISFOG, which kept participants together throughout the event, allowing continuity of discussions and avoiding clashes between concurrent presentations. Emphasis on the poster sessions and general reports offers exposure to the papers that inevitably do not receive full oral presentation given the constraints of the single session format. Key themes of the inaugural ISFOG are still as relevant 5 years on and therefore remain unchanged for the second ISFOG – these include geohazards, site investigation techniques and foundations for renewable energy, as well as anchoring solutions for deep water and pipeline geotechnics. Greater emphasis has been placed on in situ soil testing techniques and pore pressure measurement, and the characteristics of unusual seabed deposits found in frontier regions. The keynote papers reflect the key stages of an offshore project. The first three keynotes cover the assessment and interpretation of offshore geohazards, in situ site characterisation and pore pressure measurement, and the behaviour of West African clays – West Africa being a critical frontier region with particular challenges; two further keynotes describe state-of-the-art design considerations for piled foundations and pipelines; a further keynote describes recent advances in centrifuge modelling and the final keynote examines risk and reliability in offshore geotechnics. The keynotes alone contain many man-years of accumulated experience. We are sincerely grateful to the keynote authors, particularly those from industry who have generously found time to contribute their expertise to these comprehensive papers. The papers collected in these proceedings include the 7 keynotes and a further 117 peer-reviewed general papers that represent the current state-of-the-art in offshore geotechnics. These provide an invaluable resource to all those working in offshore construction, design and research. Each of the papers has been peer-reviewed by at least 2 reviewers, drawn from COFS and from academic institutions and industry within Australia and around the world. We are indebted to the reviewers for all their efforts, which have ensured that the papers are of a very high standard. We also thank Divya Mana and Santiram Chatterjee of COFS for their editorial assistance whilst assembling these proceedings. Stephanie Boroughs and Monica Mackman of COFS have provided administrative support throughout the organisation of this second
© 2011 by Taylor & Francis Group, LLC
XIII
ISFOG, and we acknowledge in particular Stephanie’s tireless work corresponding with the authors and reviewers to compile the proceedings. The assistance of the Local Organising Committee is also acknowledged, in particular Noel Boylan of COFS who co-ordinated our sponsorship programme. Finally, COFS is grateful for the support for ISFOG provided by the following organisations and individuals: the Australian Department of Innovation, Industry, Science and Research through the International Science Linkages Programme, our industry sponsors, the members of our International Advisory Committee and the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE) under whose auspices the ISFOG symposia are held. Susan Gourvenec and David White July 2010
© 2011 by Taylor & Francis Group, LLC
XIV
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Committees
LOCAL ORGANIZING COMMITTEE Susan Gourvenec (Chair)
Centre for Offshore Foundation Systems (COFS), UWA Stephanie Boroughs (Sec.) COFS, UWA Monica Mackman (Sec.) COFS, UWA Britta Bienen COFS, UWA Nathalie Boukpeti COFS, UWA Noel Boylan COFS, UWA Mark Cassidy COFS, UWA Liang Cheng School of Civil and Resource Engineering (SCRE), UWA Fiona Chow Chair, Australian Geomechanics Society (AGS), WA Chapter James Doherty SCRE, UWA Sarah Elkhatib Arup Martin Fahey COFS, UWA Ian Finnie Advanced Geomechanics
Andy Fourie SCRE, UWA Christophe Gaudin COFS, UWA Andrew Grime Arup Jim Hengesh COFS, UWA Shazzad Hossain COFS, UWA Yuxia Hu SCRE, UWA Mehrdad Kimiaei COFS, UWA Barry Lehane SCRE, UWA Mark Randolph COFS, UWA Yinghui Tian COFS, UWA Dong Wang COFS, UWA Phil Watson Advanced Geomechanics David White COFS, UWA Long Yu COFS, UWA Hongxia Zhu COFS, UWA
INTERNATIONAL ADVISORY COMMITTEE Susan Gourvenec (Chair) David Airey Peter Allan Marcio Almeida Knut Andersen Charles Aubeny Fraser Bransby Nick Brown Byron Byrne Tim Carrington John Carter David Cathie Johnny Cheuk Chris Clayton Ed Clukey Jean-Louis Colliat Don DeGroot Jason DeJong Earl Doyle Carl Erbrich Trevor Evans Albert Griffith
© 2011 by Taylor & Francis Group, LLC
Australia Australia UK Brazil Norway USA Australia Australia UK UK Australia Belgium Hong Kong UK USA France USA USA USA Australia UK Saudi Arabia
Andy Hill Phil Hogan Jacques Garnier Bob Gilbert Harry Kolk Richard Jardine Andy Lane Colin Leung Jayme Mello Roger Moore Don Murff Julian Osborne Andrew Palmer Alain Puech Matthew Quah Richard Raines Roderick Ruinen Marc Senders Dan Spikula Jørgen Steenfelt Tor Inge Tjelta Shuwang Yan
XV
UK USA France USA Netherlands UK Australia Singapore Brazil UK USA UK Singapore France Singapore USA Netherlands Australia USA Denmark Norway China
© 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Reviewers
Tony Abbs Martin Achmus Hugo Acosta-Martinez David Airey Tom Aldridge Marcio Almeida Knut Andersen Senthil Arasu Charles Aubeny Jean-Christophe Ballard Edgard Barbosa-Cruz Britta Bienen David Bonjean Nathalie Boukpeti Noel Boylan Fraser Bransby Andrew Brennan David Bruton Byron Byrne Tim Carrington Mark Cassidy Santiram Chatterjee Liang Cheng Johnny Cheuk Fiona Chow Patrick Clancy Ed Clukey Richard Dean Andrew Deeks Jason DeJong James Doherty David Edwards Clarence Ehlers Sarah Elkhatib Ed Ellis Carl Erbrich Martin Fahey Ian Finnie Andy Fourie Christophe Gaudin
© 2011 by Taylor & Francis Group, LLC
Guido Gottardi Susan Gourvenec Jürgen Grabe Andrew Grime Ole Hededal James Hengesh Andy Hill Matt Hodder Phil Hogan Shazzad Hossain Guy Houlsby Yuxia Hu Hans Hugel Mostafa Ismail Ana Ivanovic Christophe Jaeck Richard Jardine Eric Jas Hackmet Joer Steve Kay Lindita Kellezi Feizi Khankandi Yoshiaki Kikuchi Mehrdad Kimiaei Melissa Landon Thomas Langford Kok Kuen Lee Barry Lehane Colin Leung Nina Levy Tom Lunne Jon Machin Chris Martin Jayme Mello Richard Merifield Vaughan Meyer Neil Morgan Damian Morrow Don Murff Tejas Murthy
XVII
Andrzej Niemunis Steve Neubecker Tim Newson Elio Novello Conleth O’Loughlin Michael O’Neill Julian Osborne Ashraf Osman Andrew Palmer Jeff Priest Richard Raines Mark Randolph Mike Rattley Oliver Reul David Richards James Schneider Marc Senders Shambhu Sharma Tom Sheahan Dan Spikula Doug Stewart Kevin Stone Kar Lu Teh Luc Thorel Yinghui Tian Manh Tran Jean-Francois Vanden Berghe Alastair Walker Dong Wang Phil Watson Zack Westgate David White Nobutaka Yamamoto Jun Yang Gulin Yetginer Long Yu George Zhang Hongxia Zhu
1 Keynotes
© 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans BP Exploration Operating Company, Sunbury-on-Thames, UK
ABSTRACT: The development of economic and safe oil and gas facilities in geotechnically challenging and geohazard-prone areas calls for inter alia early engagement of geospecialists, fit-for-purpose data acquisition, geological understanding, engineering pragmatism and innovation and perhaps above all, multi-disciplinary teamwork. Jeanjean et al. (2005) described some of the strategies and practices developed by BP that embody these basic principles and this paper updates and expands these ideas and concepts with special reference to the investment sharing and planning opportunities afforded by programmes of multiple projects. 1
INTRODUCTION
centralised Geohazard Assessment Teams (GATs) and regional ground modelling. This paper develops the geohazard engineering risk management themes and methods described by previous authors, with particular reference to the opportunities for investment sharing and learning from BP’s multiple-project development programmes in deep water offshore Angola and Egypt. The paper is in two parts: Part 1 (Section 2) is a review of the some of the main engineering challenges faced by operators such as BP in geohazardous environments and Part 2 (Section 3) describes some of the ways that BP is tackling these challenges in Angola and Egypt as part of its asset management process.
In their keynote paper to ISFOG 2005, Jeanjean et al. (2005) described the challenges faced by oil and gas operators when developing offshore facilities in deep water frontier areas and environmentally and geologically difficult settings. The main theme of the paper was geotechnical and geological risks faced by projects operated by BP in geohazard-prone areas in the Caspian Sea, West Nile Delta (WND) and the Gulf of Mexico (GoM), with particular reference to the lessons learnt from the Mad Dog and Atlantis fields that are located along the Sigsbee Escarpment in GoM. The underlying message of the keynote was that geohazards do not necessarily preclude safe offshore developments provided operators are proactive and invest sufficient resources and planning in the geohazard management process. Some of the principles and best practices described in the paper include the use of multidisciplinary geospecialists, integrated and phased field-wide geophysical surveys and geotechnical surveys, development of new fit-for-purpose survey tools and use of project-specific engineering analyses to inform risk assessments. From about 2003 the focus of BP’s new deepwater projects shifted from the GoM towards the Caspian Sea, Angola and Egypt West Nile Delta (WND). Each of these areas has its own specific geotechnical and geohazard challenges that require some form of rigorous management. Significantly, BP’s developments in Angola and Egypt involve rolling programmes of projects that have provided systematic learning and investment leveraging opportunities in geohazard risk management, integrity management and value engineering not necessarily available for single projects. Evans et al. (2007) and Moore et al. (2007) introduced some of the multiple-project investment concepts that have been applied in the WND, including the use of © 2011 by Taylor & Francis Group, LLC
2
OFFSHORE GEOTECHNICAL ENGINEERING CHALLENGES
2.1 The geotechnical maze The ever-increasing geotechnical challenges faced by offshore operators in pursuit of hydrocarbons in frontier areas have been widely reported in the past decade: Power and Clayton (2003), Kvalstad (2007), Jeanjean et al. (2003, 2005), Evans et al. (2007), Evans et al. (2010) and Hadley et al. (2008) and many others. However, as discussed by Evans et al (2007), offshore exploration and development policy is rarely, if ever, dictated by shallow subsurface and geotechnical considerations. Rather, the shallow geo-specialists are expected to work as part of a bigger team to: (1) ensure that shallow-subsurface issues do not compromise project-life safety and function (loss prevention) and (2) offer cost-effective geotechnical solutions. These two goals are particularly onerous in frontier areas where the challenges for geotechnical and foundation engineering are especially labyrinthine, as illustrated on Figure 1. The chaotic nature of this figure shows
3
Figure 1. Geotechnical challenges.
how offshore geotechnical engineering challenges can arise from the interactions of one or more negative factors that can result in increased costs and risks to exploration, development and production. For the purpose of this paper the geotechnical challenges are assumed to have eight root causes:
seabed-founded structures. Offshore developments, in water depths of 500 m to 1500 m, are now relatively commonplace and a number of projects are underway in over 2000 m of water. Within five years it is expected that development will extend beyond 3000 m of water. The costs of geotechnical operations in such water depths are high, especially in remote and harsh conditions. Consequently, despite the obvious technical advantages of early geotechnical data for supporting exploration and development, it is often difficult to justify spending money on such activities until there has been some exploration success. Engineers therefore rarely get the early geotechnical insight that is actually needed for frontier developments where there are no ‘offset’ data.
1 2 3 4 5 6 7 8
Water Depth Harsh environments Remoteness Geological Complexity Development Scale Data Acquisition Constraints Structural Complexity Inexperience and Uncertainty Each of these root causes is discussed in the remaining parts of Section 2.
2.3
In an unpublished review of literature on submarine geologies and environments Brunsden (2010) confirmed the complex morphological, geomorphological and geological legacies which geo-specialists and engineers need to understand and resolve to support offshore exploration and development. The main purpose of this review was to summarise conventional wisdom on landslide controls and processes. The main findings of the review were:
2.2 Water depth, environment and remoteness – (Root Causes 1 to 3) The first three root causes listed relate to access difficulties; geographical, subsea and subsurface. The logistical problems of working in frontier areas away from established centres of industry and support infrastructure are clear and will have a significant affect on the efficiency and cost of even routine offshore operations. Deep water and rough weather can restrict access further and cause additional problems for geophysical and geotechnical data acquisition and for the installation of foundations, anchors and other © 2011 by Taylor & Francis Group, LLC
Complex geology (Root cause 4)
– Offshore depositional centres are largely determined by plate tectonics and they are subject to repeated sedimentation and landslide processes.
4
Table 1.
Offshore geological and geotechnical controls.
Origin
Type
Cause
Effect/Factor
Natural
Inherited quasistatic conditions
Past geological & environmental processes
Preconditioning – structural relief/seabed ruggedness – slope geomorphology & geomorphology – sediment fabric and structure – sediment stress-strain history – sediment mechanical properties
Inherited transient Equilibration of conditions Aggravating processes arising from past geological & environmental processes or events Future processes Ongoing or new slow Aggravating geological and environmental processes Future episodic Random short-duration Triggering events natural phenomena Man-made Future processes or events
Episodic or progressive offshore activities
Triggering
– fluid flow – swelling – strain softening – physiochemical changes – sedimentation at slope crest – slope toe erosion – uplift from diapirism – earthquakes – storm waves and currents – tsunamis Project-life operations: Exploration, e.g. – exploration well drilling – rig anchoring – geotechnical investigation Construction/Installation, e.g. – development well drilling and anchoring – installation of gravity structures – pipe laying – pile driving Hydrocarbon Production, e.g. – operational loadings on facilities – fluid and gas injection – fluid withdrawal and subsidence
complex interactions between inherited characteristics and future processes, events or activities that may precondition, aggravate or trigger geohazard incidents that threaten projects. A good understanding of the geological past and project-life future is important since damaging geohazard events, such as submarine slides, may arise during the lifetime of a project due to ongoing geological processes alone, or in combination with shorter duration triggers. Geological history also has an important influence on the physical properties of marine sediments. Chandler (2000) and Cotecchia and Chandler (2000) describe how the mechanical behaviour of a clay is controlled by the ‘soil structure’ that it develops during deposition solely as a result of one-dimensional consolidation and by natural post-depositional geological processes such as mechanical unloading, creep, diagenesis and tectonic shearing. The following subsections discuss some of the sedimentological and post-sedimentological processes that can affect the physical behaviour of marine clays in areas such as Egypt and Angola. Soil Structure. A challenge for the offshore geotechnical engineer is to understand the physical characteristics of shallow marine sediments and how these may affect their behaviour when disturbed by natural events or by exploration and development activities. Marine sediments of most interest to engineers
– Terrestrial, hemipelagic, contourite and shelf sediment sources provide soils at variable rates and in relatively definable packages. – Structural controls are determined by plate tectonic history, fabric, structural relief, sedimentary and basin architectures, surface morphology and residual accumulated strains are the results of a chequered inheritance. – Mass movements are much larger than onshore but the seabed often exhibits extraordinary freshness that may reflect long periods of inactivity between episodic events and/or our inability to resolve slow sedimentological or geomorphological changes with current geophysical imaging technology. Leroueil (2001) describes how geological and geomorphological inheritances and future aggravating or triggering factors are the major controls for terrestrial and submarine landslides. Locat and Lee (2000) and Locat (2001) present similar ideas in connection with instabilities along ocean margins which can be equally applied to geohazards in general. The main natural and man-made controls on seabed instability and other geohazards in the context of offshore exploration and development are summarised on Table 1. A significant challenge for those responsible for planning and engineering offshore developments is to unravel the © 2011 by Taylor & Francis Group, LLC
Examples
5
in deep water are predominantly clays and Burland (1990), Chandler (2000) and Cotecchia and Chandler (2000) describe how such soils have significant geotechnical inheritances that condition their response under mechanical loading. Burland (1990) used the inherent compressibility and strength characteristics of reconstituted clays, called intrinsic properties, as a reference for interpreting the in situ state and physical characteristics of inorganic natural non-carbonate clays. He performed one dimensional compression tests on clay slurries and by plotting a normalising parameter called the Void Index , Iv, against the vertical effective stress σv , derived an Intrinsic Compression Line (ICL) as a datum for the in situ stress states for natural clays. Iv is defined as: Figure 2. Normally consolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).
deposited slowly in still water will have more random open structures with Iv values that may lie above the SCL, Point’C’. Highly flocculated clays deposited in seawater with higher than average salt concentrations and siliceous and carbonate oozes would also be expected to plot above the SCL. When natural clays are loaded by foundations in the field, or in one dimensional compression tests (oedometer tests) in the laboratory, the rate of loading is sufficiently high to disrupt the inter-particle bonding and fabric of the clays. The corresponding compression lines are relatively steep compared to the natural SCL and fall towards the ICL, also shown on Figure 2. It is therefore particularly important to detect and characterise clays with natural states that plot above the SCL since they are especially brittle and compressible and therefore more susceptible to progressive failure and possibly liquefaction under external loading. Post-sedimentation Effects. All naturally sedimented clays are altered to some degree by environmental, geological and physio-chemical processes, or other ageing effects. As described previously, many large-scale post-sedimentation changes can occur in submarine environments and these can be expected to alter the stress states and mechanical properties of sediments significantly. However, even in the most benign environments, soils will experience postsedimentation changes due to one or more of the following processes:
In this expression e0 is current in situ voids ratio, e∗100 is the voids ratio of a reconstituted sample of the same soil compressed one dimensionally to a vertical effective stress, σv , of 100 kPa and C∗c is the compression index of the reconstituted soil measured between σv of 100 kPa and 1000 kPa. The asterisk denotes an intrinsic property of a reconstituted soil prepared from a slurry mixed at a water contents of between 1.25 and 1.5 times the liquid limit. According to Chandler (2000), Iv is a useful normalising parameter for different soil types because C∗c is defined uniquely by the one-dimensional compression procedure, the soil mineralogy and the pore water chemistry. Burland also established an Iv compression line, for marine clays in their natural state, called the Sedimentation Compression Line (SCL). The distance between the SCL and ICL in Iv-σv space is a measure of the sensitivity and brittleness of the natural clay. The SCL represents the average condition for natural marine clays with sensitivities of between about 2 and 9 and soils that plot on the idealised SCL have sensitivities of about 5 (Cotecchia and Chandler 2000). Sedimentation Structure. Mitchell (1976) defined soil ‘structure’ as a state that reflects the arrangement of soil particles, or soil fabric, and interparticle bonding. As shown on Figure 2, the ICL and SCL provide a simple and convenient framework in for comparing the in situ stress states or ‘structures’ and the mechanical behaviours of natural and reconstituted clays (Chandler 2000). The structure that clay has inherited during sedimentation depends on its mineralogy and the depositional conditions and it is not easily changed by further burial. The natural stress states of most marine clays lie on or around the SCL, Point ‘A’ in Figure 2. Clays that are deposited relatively rapidly from dense suspensions in high energy environments such as hyperpycnal flows, tend to have more orientated compact structures with Iv’s closer to the ICL, Point ‘B’. Debrites, contourites and turbidites may also have these characteristics. Conversely, hemipelagic clays that are © 2011 by Taylor & Francis Group, LLC
– Ageing – Creep./Secondary Consolidation – Unloading and swelling (Mechanical Overconsolidation) – Diagenesis – Physio-chemical changes including bonding and cementation, – Biological activity (Ehlers et al. 2005, Kuo et al. 2010) Such soils will exhibit gross yield, a threshold stress state beyond which soil stiffness and strength fall significantly (Hight et al. 1992, Cotecchia and Chandler 2000). Mechanical behaviour pre- and post-yield will be fundamentally different, so the identification of
6
Figure 4. Normally consolidated and mechanically over-consolidated clays with diagenetically altered structures – after Cotecchia and Chandler (2000).
Figure 3. ‘Aged’ normally consolidated and mechanically overconsolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).
and slopes and (2) the interpretation of insitu tests that are used to infer these mechanical properties (Randolph et al. 2005). In addition soils with natural stress states close to or above the SCL and/or with significant post-sedimentation structures that do not fit well in to normalised stress history frameworks such as SHANSEP that require consolidation beyond gross yield (Ladd & Foot 1974). Burland (1990) indicates that the peak strength and brittleness of such soils are underestimated using the SHANSEP test procedure. The detection, identification and characterisation of atypically or unusually sensitive soils are therefore important in offshore geotechnics. The natural soil structure framework described by researchers like Burland (1990), Cotecchia and Chandler (2000) and De Gennnaro et al. (2005) can be used with other soil frameworks such as Critical State Soil Mechanics to predict the mechanical behaviour of natural soils affected, for example, by ageing, desiccation, ice-loading, erosion and mass wasting. However, BP and other offshore operators are increasingly encountering post-sedimentation processes that are more difficult to characterise using traditional methods and which may require different, possibly bespoke, soil mechanics approaches, as suggested by Gens (2010). Four examples of effects that have posed additional challenges during BP’s exploration and development activities are:
gross yield stress states in shear and compression is also important for modelling soils correctly in offshore foundation design and geohazard assessments. Aged normally consolidated clays and clays that are over-consolidated due to mechanical unloading but which are unaffected by diagenesis or biological activity, would be expected to yield close to their natural sedimentation compression lines; the SCL for most natural marine clays. The oedometer com pression lines and vertical yield stresses, σvy , for an aged normally consolidated clay and a mechanically over-consolidated clay are shown schematically on Figure 3. The vertical yield stress for an over consolidated soil, σvy , will be coincident with the . geological preconsolidation stress of the soil, σvc The ratio of geological preconsolidation stress and /σv is traditionthe in situ vertical effective stress σvc ally termed the over-consolidation ratio (OCR). The ratio of the yield stress and in situ vertical effective stress σvy / σv is the called the yield stress ratio, YSR. Conversely, as shown on Figure 4, many normally consolidated clays will have been affected by diagenesis and will have diagenetically altered postsedimentation structures that will result in yield to the right of the SCL. Some over-consolidated soils that have retained diagenetic characteristics will also yield to the right of the SCL above the preconsolidation pres sure, σvc . In this case the YSR is greater than the OCR. Oedometer compression curves for normally consolidated and over-consolidated clays with diagenetically altered structures are shown schematically on Figure 4. The natural in situ states of soils condition their responses to mechanical loading and disturbance and are therefore important geotechnical controls. Grossyield in compression and shear and the post-yield brittleness are important characteristics in offshore geotechnical engineering since they affect: (1) in situ operational shear strengths and stiffnesses of soils, and consequently the behaviour of foundations, anchors © 2011 by Taylor & Francis Group, LLC
– – – –
Shallow Gas Gas Hydrates Salt Diapirism Hydrocarbon Migration Products
Shallow Gas. Gas-charged sediments, largely of biogenic (bacterial activity) and petrogenic (thermally altered) origins, are commonplace offshore. The presence of gas, in the form of free bubbles, dissolved gas or gas hydrates can change the shear strength and compressibility characteristics of sediments under static and cyclic loading. Unlike unsaturated soils, gassy
7
soils encountered offshore contain large amount of dissolved gas that can exsolve. The degree of saturation of offshore gassy sediments is generally above 90% and free gas is generally in the form of discrete bubbles much larger than the pore spaces between individual soil particles (Sultan 2009). Shallow gas per se is a ubiquitous drilling hazard and shallow gas hazard assessments are obligatory for hydrocarbon exploration, development and production operations in most parts of the world. For example, in the GoM, the requirements for assessing shallow gas are specified by MMS (1998) and in the UK sector North Sea they are covered by guidelines prepared by UKOOA (1997). Gas also has important influences on the mechanical properties of sediments. These are less well understood but have potentially negative effects on offshore wells and structures too. Sands are especially vulnerable since the difference in pore water pressure and pore gas pressure is small, so high gas pressures equate to low effective stresses and reduced strength, as well as increased compressibility. However, the effects of gas on the physical properties of clays with low free gas concentrations, typically encountered offshore, have been the subject of limited research and are consequently less predictable. Research by Wheeler (1988) showed that the effects of gas on the mass undrained shear strength of clay depend on the initial effective and total stresses and the volume fraction of gas voids, and that these effects may be detrimental or beneficial. For example, Wheeler’s model suggests that the undrained shear strengths of gassy soils in triaxial compression at shallow depths in deep water, which have high initial total stresses and low initial effective stresses, may be less than the comparable strength of the saturated clay. Conversely, gassy clays with low initial total stresses and high initial effective stresses such as deep sediments in shallow water are estimated to be stronger than the saturated clay matrix. Both effects are predicted to increase with increasing gas saturation. However, Wheeler’s research was limited and it is clear that the mass strength of gassy clay depends, amongst other things, on the degree of soil de-structuring caused by the formation of free gas (Hight & Leroueil 2003, Sultan et al. 2009), and the direction and sense of loading. Wheeler’s soil model may provide useful upper and lower bound modifying factors for the effects of gas on undrained shear strength but the industry’s capabilities of inferring the operational properties of gassy soils from laboratory and in situ tests and for modelling their behaviour in routine geotechnical engineering are still very limited. Gas Hydrates. Gas hydrates are crystalline solids, physically resembling ice, that form when gas molecules are encaged by hydrogen-bonded water molecules. The main factors affecting hydrate formation and stability are temperature, pore pressure, gas chemistry and pore-water salinity and they can occur in the pore spaces of marine sediments under the right conditions.A change in any of these parameters such as a reduced pressure or an increased temperature could © 2011 by Taylor & Francis Group, LLC
Figure 5. Bottom simulating reflector (BSR).
affect the system equilibrium and reduce the stability of hydrates, resulting in decomposition, dissociation and dissolution, and the release of free gas and water (Sultan 2007). Many gases of low molecular weights can form gas hydrates but in marine conditions the gas is generally methane of biogenic origin. Marine gas hydrates are therefore common in organic-rich sediments at depocentres, especially where deposition has been rapid. They are therefore of relevance to most offshore hydrocarbon developments in relatively deep water where gas has infiltrated shallow sediments. Figure 5 is a seismic section from an offshore location which shows high acoustic amplitude traces below a bottom simulating reflector (BSR) that is indicative of possible gas-charged soils below a stable zone of gas hydrate-bearing sediments. A joint industry project (JIP) carried out on behalf of West African offshore operators (Sultan 2007), concluded that cementing action of gas hydrates could inhibit the normal compaction processes, resulting in sediments with more open structures and higher voids indices than would otherwise be the case. Hydratebearing sediments would therefore be expected to have higher yield stresses, elastic moduli and peak strengths, and to be more brittle, than comparable hydrate-free sediments. Sultan (2007) describes the potential physical collapse of such soils and the development of excess pore pressures, leading to loss of strength, increased compressibility and possible hydraulic fracture. The threats posed by gas hydrates to offshore exploration and production are still uncertain and speculative and views on this subject are sometimes controversial. Although problems have been reported (Nimblett et al. 2005), it is generally agreed that well drilling in hydrate-bearing sediments is possible with good planning and close control of well bore pressures and temperatures. However, experience of hydrocarbon production in hydrate-prone areas is limited so there is more uncertainty about interactions arising from longer-term offshore operations, for example raised temperatures around hot production wells. Therefore, for the present, the industry tends to avoid producing in areas where there is direct evidence or strong indirect evidence of hydrate-bearing sediments.
8
Figure 6. Salt dome.
soil overlying salt. These observations were based on the results of preliminary large strain numerical analyses performed by Imperial College/GCG (2007, unpublished) and reflect conditions often encountered offshore in regions such as Angola. Further generic and project-specific research work is clearly needed to support developments in areas where salt effects are unavoidable. Hydrocarbon Migration Products. Hill et al. (2010a, 2010b) describe areas of persistent hydrocarbon seepage in deep water Angola where the shallow predominantly clayey sediments contain beds/layers with anomalously high acoustic impedances, Figure 8a. These seismically reflective zones often have a surface expression and locally extrude above the seabed, as shown on Figures 8b and 8c. Investigations have shown that these atypical soils comprise sequences of hard carbonate-rich claystone beds with variable amounts of heavy hydrocarbons that have the consistency of thick oil and asphalt (bitumen) at room
Salt Diapirism. Hill et al. (2010a, 2010b) describe how salt diapirism in deep water Angola has uplifted older sediments which has resulted in strong soils close to the seabed that can cause installation problems for suction-installed foundations and anchors and increased seabed slopes that can lead to pipeline routing difficulties and instabilities. The seismic section and schematic shown in Figure 6 indicate how the overburden can be severely distorted by salt movement resulting in large changes to the depositional stress states of cover sediments. These changes may affect the behaviour of foundations such as suction-installed caissons that are designed using empirical methods developed from field tests in areas well away from such influences, or from centrifuge or physical models that replicate these far-field conditions. This point is illustrated in Figure 7, which summarises some of the observations made about the possible combined effects of uplift and subsequent erosion on a normally consolidated © 2011 by Taylor & Francis Group, LLC
9
Figure 7. Effects of salt uplift and surface erosion.
temperature when recovered in geotechnical samples, Figure 8d. Sediments that have been modified by hydrocarbon seepage over geological timescales are, unsurprisingly, relatively common in areas of interest for oil and gas exploration and development. These soils are inevitably highly variable and are difficult to characterise geotechnically. Hard beds arising from authigenic mineral cementation can clearly pose problems for foundation and anchor installations that will probably warrant location-specific geotechnical investigations and bespoke engineering solutions. Sediments contaminated with hydrocarbons may pose even greater engineering challenges and arguably open up a new branch of soil mechanics that requires unconventional engineering thinking and new geotechnical solutions, as suggested by Randolph et al. (2005) and Gens (2010). 2.4
Figure 8. Hydrocarbon migration products. © 2011 by Taylor & Francis Group, LLC
Development scale (Root Cause 5)
Offshore oil and gas developments range from discrete bottom-founded shallow water structures with relatively small seabed footprints at locations that are generally fixed early in the development cycle, to floating or subsea-tieback developments in deepwater that cover large areas of the seafloor and which are constantly evolving. For example, the total footprint area of the three subsea developments currently planned by BP in the WND amounts to about 1400 km2 . The footprint of a typical BP Angolan project is about 750 km2 . Scale is therefore a major challenge for deepwater geospecialists who are required to define the geotechnical conditions and geohazards for a large number of structures spread across the seabed with sufficient flexibility to guide, or accommodate, inevitable changes in scheme layout. For logistical and economic reasons, it may not be possible or practical to acquire geotechnical data at every structure location and it is often necessary to infer local design conditions from regional geotechnical data; a methodology described in some construction industry standards such as Eurocode 7 (2004). Eurocode 7 classifies structures in to three geotechnical risk categories, according to structural
10
parameter selection and geotechnical design reliability assessments.
complexity, the ground conditions, type of load, and the consequences of failure. Small and relatively simple low-risk structures constructed on uniform ground are designated Category 1. Large or conventional structures which are supported on unexceptional ground and have well defined loads and normal risks fall in to Category 2. Structures that are unusually large or complex, or which are associated with abnormal loads, ground conditions or risks are classified as Category 3. Eurocode 7 advice is that it may be acceptable to use regional (offset) geotechnical data for the detailed design of Category 1 structures but that location-specific data are needed for Categories 2 and 3. Many offshore structures fall in to Categories 2 and 3 but for the reasons alluded to earlier it may be overly constraining or uneconomic to apply Eurocode 7 advice rigidly in deep water. However, the concept may be extended by improving the reliability of the regional database through informed ground modeling and applying risk-based design to manage the residual geotechnical uncertainties. Probabilistic-based geotechnical design methods using offset geotechnical data have been discussed and presented by various authors including Gilbert and Gambino (1999), Clukey et al (2000), Jeanjean et al. (2005) and Griffiths et al. (2009). These methods generally use normal or log normal soil property distributions that are best suited for areas with relatively uniform geology and soils that have similar sedimentation and post-sedimentation histories. They are less suitable for use in areas with complex geologies or numerous geohazards. The approach may also be unconservative in uniform conditions if mechanical behaviour of the soil mass is dominated by local weak or brittle seams, layers or zones that have gone undetected. The SHANSEP normalization procedure is often used to infer the undrained shear strengths of clays of similar origins and compositions over large areas. The method is proposed in ISO 19901-4 (2003) to achieve consistent shear strength profiles and has been used successfully for geotechnical engineering and geohazard assessments in Norway (Kvalstad 2007) and in the Gulf of Mexico (Jeanjean et al. 2005). Burland (1990) and Chandler (2000) indicate that the procedure may be reliable for relatively insensitive clays that lie below the SCL. However, for the reasons described earlier and discussed by Le et al. (2008), they propose that the procedure may be unreliable for more structured clays. Jardine et al. (2005) also suggest that the procedure may be unconservative for plastic clays that develop brittle shear bands that suppress dilation, Reconsolidation to the in situ effective stresses as performed in CAU tests results in less sample compression than SHANSEP. However, this technique may also cause some ‘destructuring’ of some highly structured clays resulting in underestimates of undrained shear strengths and sensitivities. It is therefore important to recognize the significance of soil structure when preparing or ‘conditioning’ laboratory test samples and to consider factoring this into geotechnical © 2011 by Taylor & Francis Group, LLC
2.5
Data acquisition constraints (Root Cause 6)
There are no hard and fast accepted strategies for geophysical and geotechnical data acquisition in deep water. Rather acquisition plans are invariably a compromise between budget and technical requirements, and are often constrained by an industry-wide shortage of survey vessels, equipment and know-how to carry out the work. Data Acquisition Costs. Shallow geological and geotechnical characterization of the shallow subsurface, including geohazards, is generally achieved by a combination of geophysical surveying and intrusive geotechnical investigation. The precise scope of work is defined by the area, line density and resolution of the geophysical survey and by the numbers and locations of in situ tests and samples. A range of acquisition strategies is possible as illustrated on Table 2. One extreme, Strategy 1, would be to spend very little on location-specific data until the scheme layout has been finalised and to rely on regional or offset data for planning and design during the early stages of the project. The other extreme would be to carry out fully calibrated area-wide high density ultra high resolution surveys at the beginning of a project as described for Strategy 3. The delayed approach is most likely to be suitable for single projects in mature areas with uniform well understood geology and few geohazards, whereas multiple projects in geologically complex frontier areas with numerous geohazards may justify the significant front-end investment implied by Strategy 3. A phased approach somewhere between these two extremes is generally adequate for most deep water projects. The amount of money spent on shallow subsurface investigations should ideally reflect positive and negative risks, i.e. the opportunities or benefits of investing the right amount wisely and the threats from not spending enough, or spending it wastefully. The UK Institution of Civil Engineers UK indicate that the largest element of technical and financial risk on construction projects is normally in the ground (ICE 1991), and that the costs overruns and delays arising from inadequate geotechnical investigations can effect 18 to 50% of construction projects. For underground structures with significant soil interfaces such as tunnels, up to 85% of projects could be affected (USNCTT 1984). The amount spent on geotechnical investigations for onshore construction projects in the UK is typically between 1% and 5% of the capital cost but geotechnical cost overruns can be up to 50% of the project value (AGS 2003). There are few comparable statistics for offshore projects but industry surveys have shown that offshore geotechnical risks may be underestimated and that associated installation problems arise frequently in delays and cost overruns (Clayton & Power 2002). There is also little published about the costs of offshore
11
Table 2.
Data acquisition strategies.
Mature Areas/ Less Geohazards ⇔ Frontier Areas/ More Geohazards Increasing Front-end Uncertainty ⇔ Increasing Front-end Cost Purpose
Strategy 1
Strategy 2
For Concept Selection
Use exploration seismic data, Medium-density grid of 2D AUV & offset data & regional 2D UHR survey lines calibrated knowledge with geotechnical cores, boreholes & insitu tests
For Preliminary Medium to high-density design (FEED) AUV & UHR surveys with selective geotechnical calibration For Detailed Targeted geotechnical survey Design when development fixed
Strategy 3 High-density 2D AUV & 3D UHR blanket survey calibrated with geotechnical cores, boreholes & insitu tests
Additional selective ‘infill’ AUV & UHR surveys and geotechnical calibration
Possibly no additional acquisition
Selective location-specific geotechnical investigations
Selective location-specific geotechnical investigations
Table 3. The cost of data acquisition – Influencing Factors. 0.1% Capex ⇐
⇒ 1.5% Capex
– Shallow water – Mature areas – Simple geology – Few geohazards – Uniform soil conditions – Compact developments
– Deep and ultra-deep water – Frontier areas – Complex geology – Many geohazards – Highly variable soils – Laterally extensive developments – New unproven technologies
– Routine proven offshore technologies – Limited well understood – Many poorly understood soil-structure interactions soil-structure interactions – Financial risk only – Societal, Environmental and Financial Risks – 1st Party risks only – 3rd and 1st Party Risks Figure 9. Optimum spend on data acquisition.
10% loss would be 10−4 . The probabilities of a 10% loss for intermediate spends would be between 10−4 and 1.0. No attempt has been made to discount the costs in Figure 9 and consequential losses such as reduced production are not considered. Available Resources. The offshore geotechnical industry’s capabilities and resources for supporting the relatively rapid advance of hydrocarbon exploration and development in to deep water are sorely stretched. One of the main constraints for deepwater operators is the limited availability of shallow geophysical survey and geotechnical investigation vessels that can operate in these conditions and the dominance of one or two main contractors in these markets. The development of survey class autonomous underwater vehicles (AUV) (Bingham et al. 2002), has significantly improved the offshore industry’s capabilities for acquiring engineering quality ultra high resolution data (bathymetry, sidescan sonar and subbottom profiling) in deep water. However, at present only three companies offer AUV surveying services on a commercial basis in water depths greater than
geophysical and geotechnical surveys in the oil and gas industry. From BP’s experience these may range from as little as 0.1% of the capital cost for simple fixed shallow water platforms in uniform predictable soils, to 1.5% for large deep water projects in geotechnically difficult areas. Although there are no strict rules about the level of spend on site characterization, it is possible to identify those factors that may influence it, as shown on Table 3. The optimum spend on a project would be that which minimises the overall geotechnical cost, namely the front-end cost of the geophysical and geotechnical investigations plus the risked cost of geotechnical construction problems. A purely hypothetical example for a large deepwater project in a geologically complex area is shown on Figure 9. The optimal spend on investigations for this illustrative case is 1.5% of capital cost (Capex). This is based on the simple premise that: (1) if no investigations are performed the equivalent financial loss would be 10% of Capex and this would occur with certainty (probability of 1.0) during the project life and (2) at the other extreme, if 10% of Capex was spent on investigations the probability of a © 2011 by Taylor & Francis Group, LLC
12
about 500 m. Geotechnical investigation in deep water is still a significant technical challenge, time consuming and expensive, and requires significant investment by contractors. Only one contractor, Fugro, currently provides a full range of geotechnical services in deep water. The world’s fleet of deep water geotechnical vessels is also severely limited with Fugro operating the only three dedicated geotechnical drillships capable of working in water depths over about 1500 m. This shortage can mean that some deepwater basins may not have access to such vessels for years; a situation that is likely to worsen as exploration and production moves in to even deeper water and the technical, operational and safety challenges with conventional geotechnical drillships increase and acquisition costs rise. The market capability for geotechnical laboratory testing, engineering and reporting is slightly better but the numbers of companies offering independently accredited laboratories is still very small and the pool of deepwater geotechnical engineering specialists is tiny compared to that available to clients in the onshore construction business. For operators like BP, the lack of competition and resources reduces programming flexibility and often leads to geotechnical acquisition as a critical path activity. Sole-source generally means additional costs to the operator and in busy periods, poses extra burdens on the sole-provider with greater risk of log -jams and delays in post-field activities such as laboratory testing and reporting. The use of consortia as an alternative to a ‘one-stop’ service can work well but involves the management of a number of interfaces that can pose extra contractual, technical and HSSE pressures on operators’geotechnical staff. In the long term the deepwater oil and gas business would clearly benefit from the emergence of new geotechnical site investigation contractors and the use of quicker and remotely-operated investigative tools and equipment. However, these would require significant investments by contractors that ultimately can only be justified by market forces and a sustainable industry workload. Nevertheless, there has been encouraging progress in recent years; for instance through the use of Fugro’s SMARTPIPE® and SMARTSURF™ Systems (Hill & Jacob 2008; Denis & De Brier 2010), and breakthroughs made by Benthic Geotech Pty with their multi-functional seabed drilling system, PROD (Tjelta 2010). These developments have been achieved with the encouragement and support of one or more operators. Guidelines and Standards. Currently, there is a lack of international guidelines and standards for the performance of marine geophysical and geotechnical investigations, including vessel and equipment audits and best practices for data acquisition, laboratory testing and reporting. As a consequence operators often produce diverse job specifications that may have led to global inconsistencies and inefficiency, especially in the geotechnical investigation contracting business. The lack of best-practice industry guidelines has also hampered the integration of the two branches of © 2011 by Taylor & Francis Group, LLC
Figure 10. Subsea structures.
geoscience that underpin deep water site characterization; geophysics and geotechnics. Advice on offshore site investigation practices has been published by some national bodies Norsok (2004), or specialist groups such as SUT (2003 & 2004) and ISSMGE (2005), but these efforts are largely uncoordinated and are generally focused on shallow water. This industry standards deficiency is currently being addressed by the joint API RG7/ISO and SC7/WG10 committee for offshore foundations with the aim of publishing two new standards; ISO 19901-8-1, ‘Marine Soil Investigations’ and ISO 19901-8-2 ‘Geophysical Site Investigations within the next few years (Jeanjean et al. 2010). Ultimately, it is hoped that a third standard will be published that will provide guidance and encouragement for a multidisciplinary approach to shallow geohazard risk assessments and engineering for offshore developments. 2.6
Structural complexity (Root Cause 7)
Offshore Facilities. Randolph et al. (2005) provide a comprehensive review of geotechnical design practices for offshore facilities and how these have diverged from onshore practices because of the nature of marine soils, and the sizes, loads and performance characteristics of offshore foundations. As explained by Randolph et al. exploration and development in frontier regions has meant engineering in soils that are significantly different to those encountered onshore and which require new geotechnical design approaches. BP and others have reported many cases of unusual soil conditions, including extremely high plasticity structured clays in Angola (Gennaro et al. 2005; Le et al. 2010; Hill et al. 2010b), highly saline clays in the Caspian Sea (Kay et al. 2005), extremely dense sands in Norway (Alm et al. 2004) and exceptionally hard clays West of Shetlands (Evans et al. 2010). Structure Types. Figure 10 shows some typical subsea structures, ranging from relatively light surfacelaid in-field flowlines to large manifolds weighing
13
several hundred tonnes. The foundations for the heavy structures can be much larger than those built onshore. They are also required to resist much larger lateral loads and overturning moments and are often subjected to combined loads in all six degrees of freedom, including significant torsion. Large Strain Behaviour. Operationally subsea structures can generally tolerate relatively large displacements, although settlement may be a controlling factor for heavy structures such as manifolds. Many seabed interactions also involve repeated loads that can result in large cumulative strains and failures. A good understanding of large strain soil behaviour under cyclic and monotonic loading and the potential for progressive failure is therefore important for design reliability. In addition, the ability to model such behaviour accurately increases the potential for cost savings through the use of performance-based design methods that are described later. Low Stress Behaviour. Conversely, knowledge of the compressibilities and shear strengths of soils at very low stresses is also important for deep sea projects. For example, the operational normal total stresses applied by unburied in-field flowlines are generally in the range of 1 to 10 kPa, which is significantly below the range considered in onshore and shallow water engineering. The study of ‘low stress soil mechanics’ problems is difficult using conventional laboratory soils testing machines and often requires bespoke equipment to produce reliable results (Bruton et al. 2007). Uniqueness. Some deepwater structures have interactions with soil that are distinctive by having few onshore analogues (Evans et al. 2007). Examples of behaviour that are difficult to predict and continually pose engineering challenges include: (1) lateral buckling, walking and vibration of surface-laid pipelines that transport multiphase products under high pressures and temperatures, often on undulating seabeds (Figure 11a), (2) the random motions of compliant steel catenary risers (SCRs) that dangle on the seabed (Figure 11b) and flow-induced vibrations of bottom-founded flexible spools (Figure 11c). Design Codes. The background to present-day offshore foundation design practices is discussed by Jeanjean et al. (2010). Foundation design procedures in offshore codes and recommended practices are still largely based on the traditional bearing capacity methods that have been developed for onshore foundations; ISO 19904-1 (2003), API RP2A, (2000) and DNV (1992). These methods generally give safe designs for relatively simple foundations under fully undrained or fully drained conditions but provide little advice about designs under partially drained conditions or with torsional loads, and do not cover structures with more complex load-displacement characteristics such as described earlier. New Methods. The development of new reliable user-friendly geotechnical solutions for offshore engineering is proving difficult. According to Randolph et al. (2005), there are only a few theoretically exact © 2011 by Taylor & Francis Group, LLC
Figure 11. Soil-structure interactions.
plastic solutions and limit equilibrium methods are generally unconservative. Numerical methods based on finite elements have been shown to be useful for improving fundamental understanding of soil structure interactions but are difficult to adapt for routine design and, according to Randolph et al. (2005), may also be unconservative in some cases. ISO 19901-4 recommends the use of large scale load tests, model tests and field instrumentation to reduce residual uncertainties with foundation behaviour. Geotechnical engineering in geohazard-prone areas is potentially even more challenging due to more complex geology and greater natural soil variability that increases the ‘aleatory’ design uncertainties, Jeanjean et al. (2005) and possible interactions arising from active geological processes such as submarine slides. The goal for offshore operators such as BP is to develop reliable, cost-effective and robust, preferably codified, design methods that can be used with confidence by all their design and installation contractors. However, at the present time some design procedures are still under development, often through proprietary research, with very little field validation. The theoretical and empirical bases and expertise behind these methods is often exclusive to academic institutions or specialists contractors and some of the design
14
Table 4. Industry and International Guidelines – Coverage of Foundation Design and Geohazards – after Jeanjean (2010).
solutions are not readily assimilated into standard structural computer codes. Well Geotechnics. The commentaries above relate to offshore facilities but years of experience suggest that the geotechnical engineering of wells also needs serious attention. The top section of a well, or tophole, is drilled through soils and poorly lithified weak rocks. The depth of the tophole is typically between 500 m and 1500 m and it is drilled without pressure containment. Geotechnical problems are often encountered within well topholes during well conductor installations and drilling. These include wellbore instability and hydraulic fracture and drilling fluid loss in to the shallow formation. Drilling difficulties in the shallow section cause costly delays and may require expensive remedial measures. Uncontrolled shallow well drilling can also affect structures in proximity, as discussed in ISO19901-4 (2003), ISO 19902 (2007), HSE (1997) and by Hobbs and Senner (1998). Tophole problems can arise from a combination of difficult subsurface conditions and poor drilling practices. Despite many years of experience they still occur (Schroeder 2007), and the situation will only improve through increased collaboration between the offshore well drilling and geotechnical communities. Some loss-prevention initiatives have already been taken in BP with the geotechnical and drilling groups jointly developing internal best practices for tophole drilling and completions. At industry level, the Offshore Soil Investigation Forum (OSIF) and the SUT’s Offshore Site Investigation Group (OSIG) are currently preparing a guidance document on this subject.
Geohazardrelated text length
Foundationrelated text length
API RP2A ISO 19901-4 ISO 19902 API RP2T API RP2SK ISO 19901-7 ISO 19904-1 ISO19905-1
225p 34p 642p 142p 181p 130p 194p 288p
1p (0.5%) 2p (6%) 0p (0%) 0.2p (0.2%) 0.1p (0.05%) 0.1p (0.07%) 0.1p (0.05%) 2p (0.7%)
36p (16%) 34p (100%) 42p (6.5%) 6p (4.2%) 38p (21%) 10p (8%) 0.1p (0.05%) 46p (16%)
1%
20%
Average
validated. However, it is likely that these will still be limited to relatively simple structure-soil combinations. The lack of industry guidelines on geotechnical engineering in geohazardous areas is even starker. This is illustrated qualitatively on Table 4, Jeanjean (2010, unpublished), which shows that current API Practices and ISOs give little textual coverage to geohazards. This is clearly something that needs to be redressed but it is unlikely that any future code or standard will be able to give anything but general advice for engineering in extremely variable conditions. Engineering in such areas will therefore continue to call for geotechnical rigour. Contractual Responsibilities. The responsibility for geotechnical and geohazard risks should ideally be with those that are best able to manage them. At the present time, operators are probably in the best positions to manage geohazards since they employ the necessary specialists and have more time to investigate them and assess their impact. Similarly, experienced offshore design and installation contractors are better placed to manage geotechnical risks. However, the responsibility for geotechnical risks is not always clear, even in so-called non-geohazardous areas. For example, unproven, often bespoke, designs naturally carry greater than normal construction and operational risks for operators and their contractors. These risks increase for EPC contractors when they are required to adopt or adapt research-based design methods that are the intellectual properties of operators or niche specialists. In the short term, these additional risks are likely to be reflected in bid prices and in the numbers of contractual exclusions and claims but in time it is hoped that the new methods will be tested and calibrated by field monitoring and will evolve in to more reliable and routine solutions that attract less uncertainty and risks. In the meantime, the challenge for offshore operators, their specialist advisors and their contractors is to find ways of working together to manage the transition from evolving best practices to established design methods, while protecting the interests of all the stakeholders.
2.7 Inexperience and uncertainty (Root Cause 8) One of the biggest issues currently faced by offshore industry in general is working in ever deeper water in new geographical areas and with new technology. The geotechnical discipline is no different and is on a fast-track learning programme that is most evident at the delivery end of projects, namely in engineering procurement and installation. As described earlier, the ground conditions are a major source of uncertainty on all construction projects and are a common cause of delays and cost overruns. Although, great advances have been made in offshore construction there are still a lot of challenges associated with the engineering and installation subsea structures in deep water, particularly in geohazard-prone areas. Codes and Guidelines for Geohazards and Foundations. The relative lack of engineering experience is also evident by the shortage of guidelines, standards and recommended practices for the design of deep water structures. Jeanjean et al. (2005) suggested that the paucity of formal guidelines reflects the newness and uniqueness of many deepwater structures and the natural time lag required for field validation of new design practices before code acceptance. The implication is that best practices and new code-based procedures will emerge, as methods mature and are © 2011 by Taylor & Francis Group, LLC
Standard
Total length
15
As the offshore industry moves in to areas with increasing geological complexities and uncertainties the roles and responsibilities for geotechnical engineering and geotechnical risks need to be clear and equitable. In this respect, the industry may benefit by adapting the risk management practices developed in the underground construction industry where projects invariably have significant geotechnical exposure. The tunneling industry has over many decades developed best practices that centre on the production of Geotechnical Baseline Reports (GBRs) that define the interpreted or assumed reference geotechnical conditions and risks at the beginning of the contract (Essex 2007).The main purpose of GBRs is that they provide a set of geotechnical conditions against which the actual conditions encountered can be assessed and compared and therefore provide clarity on risk sharing. 2.8
Communication of geotechnical risks and costs
Figure 12. Benchmarking risk severity and manageability.
For the reasons discussed above, shallow subsurface risks in deepwater frontier areas, and therefore the time and costs of investigating and mitigating them, are often much greater than for traditional shallow water projects. Possibly the biggest challenge of all for shallow geospecialists is therefore to convey clear matter-of-fact messages about these new challenges to non-specialists who may lack this insight. The ultimate objective would be to provide timely fit-forpurpose geotechnical guidance that would allow the project team to understand and manage the geotechnical elements of the development efficiently and systematically, ideally as non-critical activities. Table 1 summarises the main natural and manmade controls over shallow subsurface conditions for offshore oil and gas developments. The main hazards are natural variable and/or weak soils and dormant or active geological features and processes, and the field activities that may aggravate these conditions. These subsurface conditions may pose significant threats to the integrity and safety of offshore developments during and after installation and construction. Equally, there are opportunities and strategies by which these threats can be managed and mitigated. It is important that all these negative and positive factors are communicated and discussed with the various stakeholders in the right ways and at the right times, so that they are interpreted correctly for inclusion in risk registers and are used effectively in budget planning, design and construction. Communication of geotechnical and geological risks (threats and opportunities) can be a major asset to the project when done properly but if done poorly, or at the wrong time, may be misleading and counterproductive. At BP advice on the shallow subsurface is generally tailored to meet the needs of six distinct project stages: Access: Geologic appraisal and prospect evaluation Appraise: Project feasibility Select: Choose the preferred project(s) Define: Develop the preferred project and fund Execute: Detailed design and build Operate: Production
Access. When considering acquisitions, investors and financial planners are most concerned in asset ‘prospectivity’ and value and the shallow subsurface has little influence on investment decisions. However, interest increases once access has been gained, especially with respect to the risks and costs associated with drilling exploration wells. The production of shortterm shallow drilling hazard risk assessments is the major activity for shallow geospecialists at this time. However, longer-term risks may also need to assessed even at this very early stage should exploration or appraisal wells be considered for use in production, ie so-called ‘Keeper’ wells. Appraise. Following exploration success the information of immediate use to project planners and managers is an assessment of the main threats to development, benchmarked against other projects, and the levels of investment that may be needed to manage these threats. A simple way of benchmarking levels of risk and complexity/manageability is illustrated on Figure 12 (HSE 2006). This figure also provides guidance on specific methodologies might be needed to assess risk, including Qualitative (Q), Semiquantitative and Quantified risk assessment (QRA) methods. Early Appraise is also the stage at which the data acquisition and risk assessment strategy should be discussed and agreed with project funders. Ultimately, at the end of the Appraise stage the shallow geo-specialists need to assure the project that there are no showstoppers and that there are unlikely to be major geotechnical surprises; in effect a ‘declaration of geotechnical feasibility’. Select and Define. During the Select and Define stages the main role of geotechnical engineers and the rest of the shallow subsurface team, is to build partnerships with project planners and facilities and well engineers to help develop economic and reliable geotechnical designs and pragmatic risk reduction solutions. Risk mitigation can be achieved by avoidance through optimised layout planning or
© 2011 by Taylor & Francis Group, LLC
16
Execute. Offshore engineering and construction are usually, although not always, carried out through lump sum EPC contracts. Whatever, the type of contract it is important that roles and responsibilities for geotechnical engineering and the associated risks are fully understood by the operators and main contractors. Good geotechnical communication is needed between the operator and the main contractors and Geotechnical Baseline Reports supported by fit-for-purpose factual Geotechnical Data Reports (Essex 2007) may be the key to this; especially if the main contractors’ views can be incorporated in the GBRs before contract award. Operate. The geotechnical conditions encountered during construction are often different to those assumed beforehand. It is important that these differences and any design changes associated with them are recorded and included in an updated as-built geotechnical report that could be used as a reference for future field expansion and maintenance, and for decommissioning. The preparation of as-built geotechnical reports is not an industry norm at present but would appear to be a natural extension of current best practices.
3
FACING THE CHALLENGES
3.1 Evolving practices The number of oil and gas facilities that have been built by BP and other operators in geohazardous and geotechnically difficult areas has increased significantly over the past 15 years. During this time BP has developed strategies and methods to maintain project reliability and cost-effectiveness in such environments. Many of these practices are basic engineering common sense while others have evolved to meet new challenges and opportunities. In their ISFOG 2005 keynote, Jeanjean et al. (2005) shared some of the best practices and methodologies that BP had developed for engineering in geohazardprone areas over the previous decade, with special reference to pioneering projects in the GoM. The authors championed the use of integrated multidisciplinary teams of geospecialists and phased geophysical and geotechnical surveys and supported the development of new enabling technology for these surveys. Amongst other things, they described (1) the use of shallow seismic data and the SHANSEP normalisation technique for inferring the undrained shear strengths of soils over large areas, (2) deterministic and probabilistic methods for assessing seabed stability, (3) model testing to calibrate non-code based geotechnical designs and (4) reliability-based geotechnical designs. The authors also recommended a consistent and systematic approach to geotechnical and geohazard risk assessment so that the geo-risks could be ranked alongside other project risks. No clear step changes have emerged in geotechnical and geohazard risk management in the past five
Figure 13. Risk assessment methods.
by geohazard-resistant design and economies can be made by exploring value engineering options such as risk based design, performance based design and observational engineering. Ideally, the relative costbenefits of the geohazard-optimised options would be assessed using probability based techniques that account for increased Capex and reduced Opex that may accrue compared to the ‘do nothing’ case. The level of residual risks should be tracked continually during this period and assessed using methodologies that are compatible with those used by the project for assessing risks from other sources. These may include Qualitative (Q), Semi-quantitative methods (SQ), in which frequency and severity/impact are quantified approximately within ranges and Quantified risk assessment (QRA) in which both are fully quantified. Examples of Q, SQ and QRA are shown in Figure 13. The objective at the end of Select is a technically feasible geohazard-tolerant field layout. The goal at the end of Define is an optimized reference scheme that is suitable for taking forward into Execute. © 2011 by Taylor & Francis Group, LLC
17
years and shallow subsurface conditions cotinue to pose difficulties for offshore exploration and development and many projects are still dealing with complex poorly defined conditions with no prescriptive solutions. However, BP’s experiences continue to show that an effective way of managing the risks and costs is through clear-ownership and teamwork, and through the early and continuous engagement of geospecialists with the BP project teams and with the contractors who plan, engineer and install offshore wells and facilities. In about 2004 prospects for multiple sequential projects emerged in Angola and Egypt that allowed BP to learn from GoM experiences and to take advantage of some important longer-term investment and learning opportunities for geotechnical engineering and geohazard risk management in these two areas. The next sections develop some of the geotechnical engineering best-practice themes described by Jeanjean et al. (2005) and describe how these have evolved and been adapted to meet new circumstances, lessons learnt and fresh challenges. Particular reference is made to the use of dedicated regional geoteams, regional ground modeling and possible methodologies for engineering facilities. 3.2
Figure 14. Centralised geotechnical and geohazard teams.
Figure 15. Shallow geospecialists – Project-life support.
Shallow geotechnical and geohazard assessment teams
Consultancy Group (GCG), Imperial College, URS Corp, Fugro William Lettis & Associates and AtkinsBoreas. The GAT approach is particularly cost-effective for multiple-project programmes by providing critical mass and know-how and the means to capture and transfer lessons learnt and develop standard design and risk assessment practices and engineering solutions. The co-location of GATs from different regions provides additional advantages in the form of economies of scale, cost-sharing, technology transfer and consistent and standardised corporate engineering practices.
Evans et al. (2007) describe how the industry-wide shortage of offshore shallow geohazard specialists, especially in engineering contracting, had motivated BP Exploration Technology Group (EPT) to set up dedicated multidisciplinary shallow geotechnical and geohazard assessment teams, GATs, to support a rolling programmes of offshore projects in Egypt and Angola. The GATs for these two regions are co-located with each other and close to their respective planning and engineering teams, as illustrated on Figure 14. The rationale behind this concept is that a GAT has the critical mass and competencies to provide sustained specialist support that is tailored to the project’s whole-life needs and activity levels as shown on Figure 15. The discipline bias of the GAT is geophysics and geology at the start of a project, gradually changing to geotechnical engineering as work progresses in to development planning and engineering. A key benefit of co-located GATs is that they encourage continuous dialogue between the geospecialists and BP’s project team and contractors so that they can manage the shallow subsurface engineering risks and costs together. BP’s Egypt and Angola GATs were established in 2004 and are presently supporting projects at various stages of development from Appraise to Execute. The teams are formed from a central core of shallow geophysicists, geologists, geomorphologists, sedimentologists and geotechnical engineers from within BP and from resident consultants Halcrow Group Limited and Fugro GeoConsulting. Additional external support is provided by other consultants and contractors, including the Norwegian Geotechnical Institute, D’Appolonia, Senergy, Geotechnical © 2011 by Taylor & Francis Group, LLC
3.3
Geotechnical and geohazard (G&G) mitigation strategy
The shallow geotechnical and geohazard (G&G) mitigation strategies BP is adopting for its rolling programmes of projects in Angola and Egypt are similar and in each case was developed early in the programme cycle following a desk-top geotechnical review or ‘Health Check’. The desk studies were performed using in-house or published regional and local geophysical and geotechnical (G&G) data and hydrocarbon exploration seismic quality geophysical data. Desk studies were the important first steps in the geotechnical and geohazard risk management process since they provided immediate insights in to the levels of risk, the resources and investment likely to be needed to manage these risks, and the best strategy for acquiring data to help this process. The results of the desk study were used to develop a fit-for-purpose geotechnical and geohazard risk management plan to investigate, characterize and mitigate geohazards and to provide geotechnical engineering support to the
18
Figure 17. Geophysical data acquisition equipment.
3 if necessary, to carry out supplementary geotechnical investigations to support detailed design and construction when the scheme layouts and architectures are defined better. Figure 16. Geotechnical and geohazard mitigation process.
The prospect of multiple projects allowed the Egypt and Angola Business Units to take a far-sighted view of surveys and make longer term arrangements with suppliers, as a hedge against the shortage of worldwide survey vessels. Integrated Surveys. As described earlier, deep water subsea tie-back schemes like those being developed by BP in the WND and others based on floating systems similar to those being built in Angola cover 100’s km2 of the seabed and huge volumes of soil. For cases like these, it is clearly not feasible or practical to investigate every seam, layer or stratum of soil or geological feature within the development footprints. Spatial interpolation is unavoidable and needs to be carried out systematically by experienced geospecialists to ensure the main geotechnical objectives are achieved at least cost. The integrated survey and ground model approaches favoured by BP in Angola and Egypt help this optimisation process by providing an initial regional view with gradual focus on areas or soils that ultimately have the greatest influence on project safety and costs. The geophysical data acquisition technologies used for imaging the seabed and shallow subsurface to provide the skeletal ground models in Egypt and Angola are shown in Figure 17, which is reproduced from Moore et al. (2007). The geotechnical acquisition priorities for a ground model strategy are to provide data for calibration of the shallow geophysics and characterisation of geohazards and to get representative geotechnical data for designing and installing wells and facilities. An important feature of BP’s optimisation strategy in Angola and Egypt is to achieve these goals with least effort and at lowest cost, by maximising the multi-functional site investigation activities, as shown on the Venn diagram in Figure 18. BP’s deepwater geotechnical investigation methods include the use of sampling and in situ testing in seabed and drilling modes. In the seabed mode BP frequently investigates the soils at a single location using a combination of sampling and testing tools, including
projects. The Geotechnical and Geohazard Mitigation (GGM) plans were based on the overall project plans, the inferred complexity of the shallow subsurface conditions and the perceived challenges these pose to developments. The prospect of multiple projects in both Egypt and Angola had a significant influence on the planning strategy and was instrumental in the decision to set up regional GATs and to adopt the regional ground model methodology. The key steps in the GGM process are shown in Figure 16. They comprise an integrated sequence of Data Acquisition, Ground Modeling, Risk Assessment and Layout Planning and Engineering designed to help the projects develop geohazard-tolerant and geotechnically robust reference schemes that can be taken forward in to EPC contracts for detailed design and construction. 3.4 Data acquisition Phasing. Although the actual data acquisition strategy and scope will depend on the nature and timing of the offshore development, it is good practice to take a staged approach. This involves phases of geohazard and geotechnical engineering quality geophysical surveying followed by geotechnical investigations to calibrate the geophysics and to provide geotechnical data for engineering. BP adopted acquisition campaigns based around Strategies 2 and 3 in Table 2 for subsea developments offshore Egypt and subsea and floating developments offshore Angola. The aim in each case was to acquire sufficient data to build up a development-wide predictive ground model. The main steps in this process were: 1 To carry out front-end development wide 2D ultra high resolution and Autonomous Underwater Vehicle (AUV) surveys, 2 to perform geotechnical reconnaissance investigations to fully calibrate the geophysics for preliminary design, and © 2011 by Taylor & Francis Group, LLC
19
Figure 18. Optimised geotechnical data acquisition.
Figure 19. Geotechnical investigations in seabed mode.
box cores, gravity cores, gravity piston cores, ‘T’bar testing and piezocone penetration tests (PCPTs) as illustrated on Figure 19. This combination, or ‘Set’, of tools is generally sufficient to fully characterise the shallow subsurface conditions for assessing shallow geohazards and for designing flowlines, pipelines, shallow foundations and anchors, including suction installed caissons. However, in recent years BP has also actively supported the development of other seabed in situ testing tools such as Fugro’s SMARTPIPE® and SMARTSURF™ systems (Figures 20a and 20b) to improve the understanding of soil conditions and pipe-soil interactions at or very close to the mudline (Denis & De Brier © 2011 by Taylor & Francis Group, LLC
Figure 20. New equipment.
2010). BP is also considering Benthic Geotech’s remotely drilling system Prod shown in Figure 20c, that has already been used by other operators such as Woodside, Total and Statoil (Tjelta 2010).
20
Figure 21. Ground model.
One innovation favoured by BP in Egypt is the acquisition of 3-m to 20-m long gravity piston cores exclusively for geological, sedimentological and ichnological logging and geochronological testing, as described by Thomas et al. (2010). The continuous geological cores were taken alongside geotechnical cores of similar length and proved to be a sound investment. The core logging significantly improved the WND GAT’s understanding of the types, frequencies and magnitude of historical submarine slides, and helped them optimise the geotechnical testing programmes for engineering shallow foundations and assessing the potential of future shallow slide activity. A key advantage of continuous core logging is that it can help to detect discontinuities and pre-existing shear surfaces, and possibly identify soils that may be especially susceptible to mechanical disturbance. Boreholes were used selectively in Angola and Egypt to perform in situ tests, and to provide samples for geological logging and geotechnical testing. The borehole data were used for calibrating regional seismostratigraphic models, exploring deeper geohazards, inferring in situ pore pressures and providing geotechnical data for engineering well topholes and deep foundations such as piles. The downhole work included high resolution geophysical logging, piezocone and temperature cone testing and piezorobe testing.
subsurface conditions anywhere within it, (2) to use the interpreted data for geohazard risk assessment, field layout planning and engineering at most locations and (3) limit site-specific geotechnical investigations to areas or structures that pose the biggest risks or offer the best value engineering opportunities to the project. The integrated geological modelling is discussed in ISO 19901-4 and the advantages of the approach are largely self evident. However, the success of the method is highly dependent on having sufficient data of the right quality to populate the model, a competent interpretation team and purposeful deliverables that are communicated clearly with end-users. The challenge is particularly onerous in geologically and geotechnically complex areas but the multidisciplinary GAT and phased data acquisition concepts described earlier have proven to be well suited to the task. The ground modelling methodology adopted by BP is illustrated on Figure 21. The procedure involves the integration of environmental, geophysical and geotechnical data and the detailed mapping, interpretation and calibration of these data to develop a seismostratigraphic-soil model of the area of interest, and to capture this in GIS-format for interfacing with the project team and its engineering contractors. The main outputs from the regional model are shown on Figure 21 and in an idealised flow diagram on Figure 22. These may be summarised as:
3.5
– Terrain Units. These are developed from the geomorphological interpretation of Multibeam Echo Sounder, Side Scan Sonar, 2D UHR and AUV geophysical data. Each Terrain Unit has a distinctive geological setting and depositional history with similar geological forms and processes. – Geohazards. Terrain units that are associated with past geohazard events such as submarine slides, and therefore require more detailed assessments, are generally distinguished from those that have
Ground modelling
The predictive ground model approach that BP is adopting for its deep water projects offshore Egypt and Angola is centered on the creation of a 3-D block model that captures the geomorphology, main stratigraphic units, geological features, geohazards and representative geotechnical conditions across the development footprint. The main objectives of the method are to: (1) build a model that is reliable enough to infer the shallow © 2011 by Taylor & Francis Group, LLC
21
Figure 22. Ground model outputs for geohazard risk assessment and engineering.
sediment types, one layered and the other more disturbed, are shown Figure 23(b). The hemipelagites (Set 1) have high Iv values that are at, or above, the SCL and they are relatively compressible. Conversely, Set 2, interpreted from geological evidence as mass flow deposits, have Iv’s close to the ICL and they are less compressible in the stress ranges of interest for most offshore foundations. These findings are consistent with the inferred geological stress histories. However, a second group of soils, Set 3, also interpreted as mass flow sediments, have higher Iv values closer to the SCL. This suggests that these soils may have retained some of their sedimentation structures during deformation. Alternatively the facies interpretation may be imprecise and may be need to be reconsidered and refined. Figures 24 shows similar plots for recently deposited shallow sediments from two regions in deepwater Angola, Regions 1 and 2. The data envelope shown in Figure 24(a) is for oedometer tests performed on samples of hemipelagic soils from Region 1. These soils are interpreted to be unaffected by geological processes such as salt diapirism and fluid expulsion and the test data broadly straddle the SCL. Figure 24(b) shows the results of oedometer tests on four samples of shallow soils from Region 2, in an area influenced by salt and erosion. These results suggest that the effects of uplift and erosion may include ‘destructuring’ and overconsolidation as implied in Figure 7. The results of the oedometer tests in Figures 23 and 24 clearly show the potential insights in to the stress histories and mechanical properties of marine soils that may be gained from using frameworks such as proposed by Chandler (2000). These frameworks offer the means to calibrate terrain units
more benign origins. These geohazard terrain units are subject to further detailed sedimentological, ichnological and geochronological assessment and geomechanical back analyses to estimate the frequency and magnitude of these historical events (Thomas et al. 2010). The results of this work are fed in to the future geohazard risk assessment. – Soil Provinces. Soil Provinces are linked closely to Terrain Units and are idealised areas of the seabed and shallow subsurface that have broadly similar geotechnical characteristics and properties to depths of interest for engineering. Each Terrain Unit may comprise one or more Soil Provinces and some Soil Provinces may be common to a number of Terrain Units. – Soil Units. These are the main soil layers and strata within each Soil Province that are interpreted to have broadly similar mechanical properties. Some Soil Units are common to more than one Soil Province and they are interpreted by dividing the ‘geophysical’ column into significant seismostratigraphic units separated by regional unconformities and defining the sediments that make up these units as one or more distinct soil facies. The soil facies represent different depositional and post-depositional processes and histories and they are interpreted from sedimentological, lithological, ichnological and chronological logging and analyses and geotechnical data obtained by selective sampling and in situ testing. As described earlier, it can be instructive to consider the results of incremental oedometer tests on some of the interpreted soil facies in void index – vertical effective stress space. Figure 23(a) shows envelopes of oedometer results from tests performed on WND normally consolidated hemipelagic clays and others inferred to be mass flow deposits (debrites, mudflows and slides). The contrasting soil fabrics for these two © 2011 by Taylor & Francis Group, LLC
22
Figure 23. One-dimensional consolidation tests – WND Soils.
Figure 24. One-dimensional consolidation tests – Angolan soils.
and soil units and may provide the geologicalgeomechanical link that is required for a truly integrated approach to geotechnical engineering, particularly in geologically complex areas. – Geotechnical Modifiers. Some of the Soil Units may have local depositional fabrics or structures or have undergone local post-depositional changes © 2011 by Taylor & Francis Group, LLC
similar to those discussed earlier in this paper. These features and characteristics may not necessarily be resolvable by geophysics but nevertheless may have a significant effect on soil behaviour. Examples include sediments with high void indices and soils affected by cementation, hydrocarbon
23
contamination, bioturbation and pre-existing shear surfaces – Regional Soil Parameters. Each of the main Soil Units is assigned representative mass soil parameters, such as undrained shear strengths and unit weights, which are inferred from the results of the geotechnical calibration programmes. These regional soil parameters are defined as credible ranges to reflect the levels of uncertainty with the data interpretation. The SHANSEP technique was not used to derive regional soil parameters in Angola and Egypt since the soils of greatest interest for engineering were considered to be too brittle or too variable or complex to apply the method reliably over large areas. However, the technique has been used selectively to infer operational undrained shear strengths for specific geotechnical designs or geohazard assessments where there are abrupt changes in OCR/YSR. – Operational Geotechnical Parameters.The geotechnical parameters used for engineering wells and facilities and carrying out geohazard assessments are necessarily problem-specific since they need to account for: (1) the mass properties of the soils, including the effects of geotechnical modifiers, (2) the type of structure or geohazard, (3) the imposed loads, and (4) the methods of analysis or design (code-based, reliability-based etc). For most of BP’s work in Egypt and Angola the regional soil parameters derived from the ground model databases have been used as the reference data for deriving operational design parameters. Location-specific data are used where local conditions are significantly different to the regional model and/or when geotechnical conditions may have an important bearing on safety or cost. Operational geotechnical parameters are key input to Geotechnical Baseline Reports used as reference for detailed design and construction contracts and as such, should be selected with support from the project engineering teams. The use of regional geotechnical data as the basis for design generally leads to higher factors of safety to achieve that same level of reliability as obtained using location-specific data. However, the strategy is pragmatic in largely uniform marine conditions and is broadly in line with one of the design approaches described in Eurocode 7 (2004). It also provides the flexibility to progress the geotechnical planning and design productively despite inevitable scheme changes. Nevertheless, significant effort is needed to identify atypical conditions that may control the mechanical behaviour of soil and invalidate the regional model. 3.6
Scheme definition – geohazard screening and risk assessment, and scheme layout planning
The geohazard management strategy being adopted in Angola and Egypt may be summarised as: – Assess the zones of influence of potentially active geohazards and where possible avoid them. © 2011 by Taylor & Francis Group, LLC
24
– Where, for overriding reasons, ‘geohazardous’ areas are unavoidable, assess the risks in terms of probability of occurrence and impact and consider special geohazard-resistant engineering measures to reduce the risks to tolerable levels. Geohazard Screening and Risk Assessment. The first step in this process is to screen the development area for geohazard ‘hotspots’ that denote major risk drivers.The method involves considering potential event scenarios associated each of the geohazards in the ground model database and eliminating those that are not significant because they are very unlikely to occur during the project life, would have little impact on the development or could be accommodated by routine design. The geohazard scenarios considered in this study include natural incidents similar to those in the past, reactivated geohazards, completely new events and man-made effects. The magnitudes and frequencies of the historical events, such as submarine slides, were assessed by multidisciplinary expert consensus opinion informed by the ground model terrain evaluation work, as illustrated on Figure 25. This process was supported by geomechanical back analysis and forecasting using simplified inhouse methods that are suitable for screening large areas. For example, the potential for shallow submarine slides was investigated using an undrained sliding block model (Newmark, 1965) that has been adapted to account for the effects of residual excess pore pressures, earthquake loading and soil brittleness. Similarly, the return periods for delayed failures of scarps (Potts et al. 1997, Leroueil 2001), formed in previous slides were estimated by a simple pore pressure equilibration model. Regional geotechnical parameters were generally used in the screening process but the local effects of exceptionally weak or brittle layers, or pre-existing shear surfaces (geotechnical modifiers) were also considered. Complementary engineering studies were performed to assess the potential impacts of geohazards such as submarine slides on typical seabed infrastructure (Parker et al. 2008, 2009). The risk assessment process is being facilitated by a series interactive multidisciplinary workshop sessions and the results are integrated in to the overall project risk assessment using the Qualitative, Semi-quantitative and Quantitative Risk Assessment methods shown on Figure 13. Scheme Layout Planning. The results of the geohazard screening and risk assessment are used with other criteria to optimise the scheme layouts. This is a multidisciplinary task that depends on a number of factors such as reservoir access, flow assurance and cost, as well as geohazard threats. The ideal objective for the GAT is to avoid all the buffer zones associated with the major geohazards. However, compromises are inevitable, and for technical or economic reasons it may be necessary to encroach some of the buffer zones. In such cases more detailed location- and structure-specific assessments are performed to define the geohazard risks more accurately.
Figure 25. Assessment of frequencies and magnitudes of submarine slides – Fully-developed slab slides.
Specialised geohazard-resistant engineering measures are also being considered to reduce the risks in these areas. Optimised Reference Scheme. The layout planning is an iterative process performed throughout the Appraise to Define stages of a project with the ultimate objective of developing a geohazard-tolerant reference scheme that can be taken forward in to detailed engineering and construction, generally performed through an EPC contract. 3.7
solutions, often derived from first principles, are being sought for complex structures for which there is little or no experience. Geohazardous Areas. Similar design approaches are also being applied in geohazardous areas, although additional special geohazard-resistant design measures are also considered. Although sometimes difficult to achieve, the overall objective is to develop safe and functional standard engineering solutions in both non-geohazardous and geohazardous areas and to limit the use of bespoke designs. Code-based Designs. As discussed earlier industry and international guidance for the design of shallow and deep offshore foundations has improved considerably in recent years and is beginning to address the complex behaviour of many subsea structures (Jeanjean et al. 2010). However, codes still provide solutions for relatively simple structures designed to avoid failure, with less consideration for serviceability criteria such as deflections. BP’s normal practice is to use published codes such as ISO 19901-4 (2003), ISO 19902 (2007) and API RP2A (2000) where possible but to consider alternative methods should these standard methods not be applicable, or if they are known to be excessively conservative. Value-engineering Solutions. Value engineering is a process by which design function and/or design reliability are increased at no additional cost, or cost
Engineering
The design strategy being followed by BP in their West Nile Delta and Angolan deep-water programmes is shown on Figure 26. The methods are based on the nature of the seabed and shallow subsurface conditions and the types of structures, and are driven by the scheme layout and architecture. Non-geohazardous Areas. In areas designated as non-geohazardous, simple structures with relatively predictable performance are being designed using codified methods and where such methods do not exist or are uneconomic or unsuitable, more pragmatic value engineering solutions are being adopted. The use of centrifuge tests to calibrate the design of suction anchors in layered soils for the GoM Mad Dog project described by Jeanjean et al. (2005) is an example of such a pragmatic design approach. New design © 2011 by Taylor & Francis Group, LLC
25
Figure 26. Design strategies for non-geohazardous and geohazardous areas.
longer term it is important to validate and/or refine these methods through field monitoring. The idealised flow diagrams in Figures 27 and 28 illustrate just two possible options for developing and calibrating these new designs methods. Figure 27 shows an approach that is based on the observational method. In this case a structuresoil interaction behavioural model is developed and used with project-specific soil and structures data to produce a job-specific prototype design to meet the project’s performance objectives. Provision is then made to monitor the structural response during field operations. If observed behaviour matches expectations the design may be verified. Otherwise, preplanned intervention and remedial measures may have to be implemented to assure acceptable life-of-field performance. Data from the field monitoring is also used to update and improve the behavioural model for future designs. Monitoring of flowline buckling and walking in Angola and the design and assurance process applied for the West of Shetland Clair Phase 1 Platform piles (Evans et al. 2010), are examples of the use of the observational method in BP. The method shown in Figure 28 is often used in onshore geotechnics to confirm design assumptions at sites or areas where there is little or no previous experience and is one of the ways recommended in ISO 19901-4 to manage geotechnical uncertainty Typically, it involves carrying out a field trial in advance of construction contract to validate or calibrate design assumptions for a specific site or types of soil and to update these accordingly. An example would be preliminary trials to confirm the design methodologies for predicting the axial capacities and load-deflection responses of piles which would then be used in projectspecific designs. It would be very rare for a single operator to do this in a frontier deep water area for a single project because of the high cost. However, the idea is more appealing when multiple projects are planned and trials can be performed opportunistically
is reduced without affecting function and reliability. Keaton and Eckhoff (1990) describe the approach in geologic hazard risk management terms as eliminating or reducing those aspects of a system that add cost without reducing risk. In BP the application of value engineering includes such things as challenging conventional thinking, doing things differently, design innovation, lateral thinking, adapting existing practices, extending the performance envelope and risk-based design with rigorous validation. In recent years BP has spent considerable effort investigating and developing cost-effective and safe engineering solutions based on these principles. Installation is a major cost driver and there is a big incentive to reduce the size and weight of subsea foundations and anchors. BP is tackling this challenge by developing reliability- and function-based (performance-based) designs that require the project to set performance objectives, such as failure probability, system displacements and tolerable levels of damage. These methods may also involve performance monitoring to manage the residual uncertainties, as practised in the observational engineering approach (Eurocode 7 2004). Other cost-saving initiatives presently being pursued include the development of composite foundations, the adaptation of wellconductor jetting techniques for installing piles and the use of numerical and physical modeling to challenge conventional design practices such as the use beamcolumn analyses with API RP2A p-y springs (Jeanjean 2009). New Design Methods. Many subsea structures have complex foundation behaviours and seabed interactions for which there is very little or no operational experience. In such cases it is necessary, at least in the short term, to develop design methods largely from first principles using mathematical and physical modelling and the results analogous structures or research together with expert judgement. In the © 2011 by Taylor & Francis Group, LLC
26
Figure 27. New design solutions using observational methods.
Figure 28. New design solutions using field trials.
by increasing component flexibility and/or strength. The key conclusions drawn from these studies are:
in the early projects to benefit later ones. For example in Angola BP is planning on leveraging selective deepwater pile design and installation experiences on a later project currently in Execute to extend the foundation design options on later projects. Investment in frontier trials would also more attractive when performed through joint industry projects (JIPs) sponsored by a number of operators, contractors and other interested parties with common interests. As engineering experience with new behavioural models increases and design confidence grows the plan would be to adopt them as standard company practices. In the longer term, this field experience may also provide the bases for industry best-practices, particularly if it could be pooled with similar data from other offshore operators. Geohazard-resistant Design. Preliminary studies for BP (Parker et al. 2008, 2009), have shown that the susceptibility of wells and bottom-founded facilities to geohazards such as submarine slides may be reduced © 2011 by Taylor & Francis Group, LLC
– Pipelines. The vulnerability of pipelines in landslide areas may be reduced by laying them in curves, a common practice in the GoM. – Sub-seabed soil movements. Subsea structures may be made more resistant to sub-seabed soil movements (landslides, erosive run-outs, fault movement etc) by increasing foundation capacities; driven piles being more effective for this purpose than suction-installed caissons due to the smaller area exposed to soil forces. Conversely, structural connections may be made more tolerant of this type of movement by increasing their flexibilities rather than strength. – Supra-seabed soil movements. Stronger foundations would also increase resistance to impact forces above the mudline (turbidity currents, mud volcano outflows, fluid non-erosive debris flows)
27
Procurement and Construction (EPC) contracting. In the short term, it may even be necessary for Operators to accept more responsibility than normal for the operational performance of designs based on methods that are not field-calibrated. Irrespective, EPC tender and contract documents will need to be adapted to ensure that the risks for new design methods are fully defined and roles and responsibilities for the work are clearly understood. As mentioned earlier, one possible way of defining and agreeing contractual risks for non-standard design solutions is through the use of Geotechnical Baseline Reports (GBRs) which are discussed further below. 3.8
Earlier in this paper, the communication of the shallow subsurface conditions to stakeholders was identified as a big challenge for geo-specialists. The approach taken by the Egypt and Angola GATs has been to merge the results of their work in a single document called the Geotechnical Engineering and Geohazard Mitigation (GGM) Report that is used to assist project decision making and to underpin more purposeful deliverables that support engineering and construction, as shown on Figure 30. The GGM Report evolves as more geophysical and geotechnical data are gathered and the project’s development plans emerge and represents the consensus state-of-awareness of the shallow subsurface conditions and how these affect the planned development. The GGM Report is updated whenever there is a significant change in the state-of-knowledge or project status, triggered for example by a programme decision gate, additional survey data or major scheme changes. The GGM Report embodies the ground model and amongst other things, describes the interpreted geological and geotechnical conditions and geohazards and, with input from the engineering teams, provides general guidance on geohazard avoidance, layout planning and engineering.The GGM database is captured in GIS format for ease of communication and interfacing with others in the project team. The GGM Report also provides the platform for the production of more engineering-focused and contractual documents such as the Geotechnical Basis of Design (BoD) that is specifically tailored to the scheme taken forward through FEED in to an EPC contract. The two key elements of the Geotechnical BoD are the Geotechnical Data Report (GDR) that includes all the factual data gathered to support the project and the Geotechnical Baseline Report (GBR) that contains interpreted or assumed ‘baseline, or ‘reference’ statements about the ground conditions and risks. EPC tenderers are encouraged to propose supplementary surveys and geotechnical investigations to support the BoD should they consider the data provided at the bid stage to be insufficient to support their specific design and installation plans. The GBR represents the project’s consensus assessment of the ground conditions and geohazards and how these will be taken in to account in design and for
Figure 29. Geohazard impact framework.
and piles and caissons are equally suitable for this purpose. Structural connections also need to be strengthened but this is a design challenge since it conflicts with the need for increased connector compliance to accommodate sub-seabed ground movements – ‘Weak-link’ Design. When safety and environmental considerations are accounted for, the most pragmatic geohazard-resistant design strategy may be to provide weak links that are relatively easy to repair. Equipment, structures, foundations and anchors used routinely offshore will have some inherent reserve capacities to resist geohazard-related loads However, the resistance thresholds may be increased and risks reduced by additional engineering as illustrated on Figure 29. The investment in geohazard-resistant design would be justified if it is less than the additional operating costs and production losses that would arise if the measures were not taken. Probability-based cost-benefit analyses would support this decision. Additional Considerations. The motives behind value-engineering and new design solutions are to enhance design by increasing reliability or reducing costs, or to enable developments to proceed in controlled and responsible ways in the absence of significant field experience. However, the methods are not universally applicable and therefore need to be used selectively and with care. For example, the observational method may not be suitable for safetycritical structures or where soil-structure interactions are insufficiently ductile to give warning and time to implement planned interventions. Performance-based methods may also need to be avoided if there is potential for brittle behaviour. There may also be some contractual considerations since many of the non-routine design methods are not ideally suited to traditional offshore Engineering, © 2011 by Taylor & Francis Group, LLC
Deliverables
28
Figure 30. Geotechnical documentation.
that may not be justified or practical for a single project. These include opportunities for:
construction. It also provides reference conditions for tendering or setting cost targets and for comparing with actual conditions encountered to support the interpretation of any ‘Differing or Unforeseen Site Condition’ clauses included in contracts. GBRs are most effective when they are prepared jointly by the Operator’s design team and EPC contractors and are agreed before contract award.
4
– Long term agreements with geophysical and geotechnical survey contractors. – Early geophysical and geotechnical data acquisition – The use of regional ground models to support the systematic assessment of geotechnical and geohazard risks and the geotechnical engineering for wells and facilities. – Use of observational methods and field trials to support value engineering solutions or new design methods for foundation and anchor design and for predicting complex soil-structure interactions. – Applying best-practices from onshore civil engineering projects with high geotechnical exposures, such as tunnelling, to develop strategies for managing offshore geotechnical risks within an EPC contracting framework that satisfy all parties.
CONCLUDING REMARKS
This paper describes the problems faced by offshore oil and gas companies operating in geohazardous and geotechnically-challenging areas and some of the ways that BP’s geotechnical engineers and other geospecialists are tackling them. Jeanjean et al. (2005) described some guiding principles and best-practices from the company’s pioneering deep water developments in geohazard-prone areas in the Gulf of Mexico. These included: (1) operator whole-life ownership, (2) multidisciplinary geoteams, (3) phased development-wide geophysical surveys and geotechnical site investigations and (4) the development of new engineering solutions using numerical and physical modeling and reliability-based design. This paper develops these ideas and concepts with particular reference to strategies and practices that are unfolding from deep-water development programmes in the West Nile Delta and Angola. Multiple projects are planned in both these regions which allowed BP to take a long-term holistic approach to geotechnical and geohazard risk management and to take advantage of a number of investment and leveraging opportunities © 2011 by Taylor & Francis Group, LLC
The approach that BP is taking to manage and mitigate geotechnical and geohazard risks in Egypt and Angola is clearly not the only way and may not necessarily be suitable or practical for all projects. Other, equally valid, management and mitigation strategies that embody the best practices and principles described by Jeanjean et al. (2005) and in this paper are being adopted by the company in other regions such as GoM and in the Caspian Sea. Some of the ideas and practices discussed in this paper are work-in-progress and others are still aspirational. However, to date the methodology adopted has been relatively successful. Arguably, the most significant steps were: (1) the establishment of semipermanent geotechnical and geohazard assessment
29
teams, GATs, for both regions to help provide projectlife geotechnical support and (2) to co-locate the GATs centrally, close to their respective project teams. The centralised GATs have technical and commercial advantages over more ad-hoc arrangements generally made for single projects, including: critical mass, proximity to project teams, technology transfer, shared best-practices and learning, standardisation, training opportunities and economies of scale. Perhaps the most important benefit of the GAT approach is that it has improved the interactions between the shallow geospecialists, deep geospecialists and the engineering teams and encouraged the holistic teamwork that is so often the hallmark of a successful project.
stress design, API recommended Practice RP-2A-WSD, 21st edition. Bingham, D., Drake, T., Hill A & Lott, R. 2002. The Application of Autonomous Underwater Vehicle (AUV) Technology in the Oil Industry – Vision and Experience, TS4.4, Hydrographic Surveying II, Fig XXII, International Congress, Washington DC, USA Brunsden, D. 2010. ‘A review of literature on submarine slope instability processes’. BP Egypt GAT Internal Project Report, unpublished. Bruton, D.A.S., Carr, M. & White, D.J. 2007. ‘The Influence of Pipe-Soil interaction on lateral Buckling and walking of Pipelines – The Safebuck JIP’, 6th International Conference, Offshore Site Investigation and Geotechnics, SUT, London. Burland, J.B. 1990. ‘On the compressibility and shear strength of natural clays’. 30th Rankine Lecture, Géotechnique, 40, No 3, 327–378. Chandler, R.J. 2000. Clay Sediments in Depositional Basins: the Geotechnical Cycle, Quarterly Journal of Engineering Geology and Hydrogeology, 33, 7–39. CIRIA. 1999.The Observational Method in ground engineering: principles and applications, Construction Industry Research and Information Association, Report 185, London. Clayton,C. & Power,P. 2002. Managing geotechnical risk in Deepwater., 5th International Conference, Offshore Site Investigation and Geotechnics, SUT, London. Clukey, E.C., Banon, H., & Kulhawy, F. 2000. Reliability Assessment of Deepwater Suction Caissons. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 12192. Cotecchia, F & Chandler, R.J. 2000. ‘A general framework for the mechanical behaviour of clays’. , Géotechnique, 50, No 4, 431–447. De Gennaro, V., Delage, P. & Puech, A. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea. First symposium in Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Denis, R. & De Brier, C. 2010. Deep Water Tool for Insitu Pipe-soil Interaction Measurement: Recent Developments and System Improvement. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20630. DNV. 1992. Foundations, Classification Notes No 30.4, Det Norske Veritas. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origins of near-seafloor “crust zones” in deepwater. First symposium in Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Essex, R.J. 2007. Geotechnical Baseline Reports for Construction. Technical Committee on Geotechnical Reports of the Underground Technology Research Council, American Society of Civil Engineers. Evans, T.G., Usher, N. & Moore, R. 2007. Management of Geotechnical and Geohazard Risks in the West Nile Delta, Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. Evans, T.G., Finnie, I., Little, R., Jardine R.J. & Aldridge, T.R. 2010. BP Clair Phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till, Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Eurocode 7. 2004. BS EN 1997-1:2004 Geotechnical Design. General Rules, BSI. Gens,A. 2010. Soil -environment interactions in geotechnical engineering. Géotechnique 60, No 1, pp 3–74. Gilbert, R.B, & Gambino, S.J. 1999. Reliability-based Approach for Foundation Design without Site-Specific
ACKNOWLEDGEMENTS The author is grateful to BP Exploration and Production Technology (BP EPT) for permission to publish this paper, although the views expressed are wholly his own and not necessarily those of BP. The author wishes to thank his BP geotechnical colleagues, Junius Allen, Jim Clarke, Ed Clukey, Paul Dimmock, James Hansen, Kevin Hampson, Andy Hill, Hugo Galanes-Alvarez, Philippe Jeanjean,Attasit Korchaiyapruk, Eric Liedtke and Tony Lusted for their views and support, and Bryn Austin and Mike Fiske for leading the GAT geophysical interpretation work. Special thanks to Mike Sweeney for his support and encouragement. Also, thanks to Giles Thompson and his colleagues at Senergy Survey & Geoengineering, Tore Kvalstad of NGI, Eric Parker and his colleagues at D’Appolonia and Alan Niedoroda of URS Corp for strengthening the team when it matters. Finally, this paper would not have been possible if had not been for the dedication, hard work and professionalism of a diverse team of specialists from Halcrow Group Limited and Fugro GeoConsulting Limited who have underpinned BP’s Angola and Egypt GATs over the past 6 years. The individuals are too many to mention here but I will single out three; Roger Moore and Denys Brunsden of Halcrow and Steve Thomas of Fugro, who have been here since the beginning. Many thanks. REFERENCES AGS. 2003. Response to UK Government Minister for Construction. Association of Geotechnical and Geoenvironmental Specialists. Aldridge, T.R., Carrington, T.M., Jardine R.J., Little, R., Evans, T.G. & Finnie, I. BP Clair Phase, 2010. ‘Offshore foundation design in extremely hard till’, Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Alm,T., Snell, R.O., Hampson, K.M.& Olaussen, A. 2004. Design and Installation of the Valhall piggyback structures. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 16294. API. 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – working © 2011 by Taylor & Francis Group, LLC
30
Jeanjean, P., Liedtke, E., Clukey, E.C., Hampson, K. & Evans,T. 2005. ‘An Operator’s perspective on offshore risk assessment and geotechnical design in geohazard-prone areas’, First International Symposium on Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds) Jeanjean, P., Watson, P.G. Kolk, H.J. 7 Lacasse, S. 2010. RP GEO: The new API recommended Practice for Geotechnical Engineering. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20631. Kay, S., Goedemoed, S.S. & Vermeijden, C.A. 2005. Influence of Salinity on Soil Properties. First International Symposium on Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Keaton, J.R. & Eckhoff, D.W. 1990. Value Engineering Approach to Geologic Hazard Risk Management. Transportaion Research Record 1288, Geotechnical Engineering, 1990, 168–174, Transport Research Board, Washington DC. Kvalstad, T.J. 2007. What is the Current “Best Practice” in Offshore Geohazard Investigations? A State-of-the -Art Review’, Proceedings Offshore Technology Conference, Houston, Texas. OTC 18545. Kuo, M.Y.H., Bolton, M.D., Hill, A.J. & Rattley, M. 2010. Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Ladd, C.C. & Foot, R. 1974. New Design Procedures for Stability of Soft Clays. ASCE, Journal of the Geotechnical Engineering Division, Vol. 100, No. GT7, 763–786. Leroueil, S. 2001. ‘Natural slopes and cuts: movement and failure mechanisms’. 39th Rankine Lecture, Géotechnique, 51, No 3, 197–243. Le, M-H., Nauroy, J-F., De Gennaro, Delage, P., Flavigny, E., Thanh, N., Colliat, J-L., Puech, A. & Meunier, J. 2008. Characterization of Soft Deepwater West Africa Clays: SHANSEP Testing is Not Recommended for Sensitive Structured Clays. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19193. Locat, J. 2001. ‘Instabilities along ocean margins: a geomorphological and geotechnical perspective’. Marine and Petroleum Geology, 18, 503–512. Locat, J. & Lee, H.J. 2000. Submarine Landslides: Advances and Challenges. Proceedings of the 8th International Symposium on Landslides, Cardiff, UK, June. Mithchell, J.K. 1976. Fundamentals of Soil Behaviour, New york, Wiley. MMS. 1998. US Department of the Interior Minerals Management Service. Notice to lessees and Operators of Federal Oil, Gas and Sulphur Leases in the Outer Shelf Gulf of Mexico OCS Region, Shallow Hazard Requirements, NTL 98-20, September 15, 1998. Moore, R., Usher, N. & Evans, T. 2007. ‘Integrated Multidisciplinary Seismic Geomorphology Assessment of West Nile Delta Geohazards’, Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. Newmark, N. 1965. Effects of earthquakes on dams and embankments. Géotechnique, 15, No 2, 139–160. Nimblett, J.N., Shipp, R. C. & Strijbos, F. 2005. Gas Hydrate as a Drilling Hazard: Examples from Global Deepwater Settings. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 17476. NORSOK. 2004. Marine Soil Investigations, NORSOK Standard G-001. Power PT & Clayton, C.R. 2003. ‘Managing Geotechnical Risk in Deepwater off West Africa’, 7th Annual offshore West Africa Conference and Exhibition, Windhoek, Namibia.
Soil Borings. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 10927. Griffiths, D.V., Huang, J. & Gordon, A. F. 2009. On the reliability of earth slopes in three dimensions. Proceedings of the Royal Society, A 2009 465, 3145–3164. Hadley, D., Peters, D. & Vaughan, A. 2008. Gumusut-Kaakap Project: Geohazard Characterisation and Impact on Field Development Plans, IPTC 12554, International Petroleum Technology Conference. Hight, D.W., Bond, A.J. & Legge, J.D. 1992. Characterization of the Bothkennar clay:: an overview. Géotechnique, 33, No 2, 327–340. Hight, D.W. & Lerouiel, S. 1993. Characterisation of soils for engineering purposes. Characterisation of soils for engineering purposes. Tan et al. (eds), Vol 1, 255–362. Hill, A.J. & Jacob, H. 2008. In-situ Measurement of Pipe-Soil Interaction in Deep Water. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 19528. Hill, A.J., Fiske, M., Fish, P.R. & Thomas, S. 2010a. Deepwater Angola: Geohazard Mitigation. Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia, November 2010. Hill, A.J., Evans, T.G., Mackenzie, B. & Thompson, G. 2010b. Deepwater Angola: Geotechnical Challenges. Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Hobbs, R. & Senner, D.W.F. 1998. Safety Implications for Offshore Foundations of Conductor and Shallow Well Drilling. 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. HSE. 1997. Drilling and Installation Effects on Foundations. Offshore Technology Report-OTO94 005, UK Health and Safety Executive, June 1997. HSE. 2006. Guidance on Risk assessment for Offshore Installations, Offshore Information Sheet No 3/2006, UK Health and Safety Executive. ICE. 1991. Inadequate Site Investigation. Institution of Civil Engineers, Thomas Telford, London. Imperial College Consultants/Geotechnical Consulting Group. 2007. Geotechnical Analysis of Salt Diapir Effects on Properties and Stress States of Deepwater Sediments. Report to BP Exploration, April 2007, unpublished. ISO 19901-4. 2003. Petroleum and Natural Gas Industries – Specific Requirements for offshore Structures – Part 4: Geotechnical and Foundation Design Considerations. ISO 19902. 2007. Petroleum and Natural Gas Industries – Fixed Steel Offshore Structures. ISSMGE. 2005. Geotechnical & Geophysical Investigations for Offshore and Nearshore Developments. Technical Committee 1, International Society for Soil Mechanics and Geotechnical Engineering. Swan Consultants Ltd (ed). Jardine, R., Chow ,F., Overy., R. & Standing, J. 2005. ICP Design methods for Driven Piles in Sands and Clays. Thomas Telford, London. Jeanjean, P. 2009. Re-assessment of P-Y Cirves for Soft Clays from Centrifuge Testing and Finite Element Modeling. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20158. Jeanjean. 2010. Private communication. Jeanjean, P., Hill, A. W. & Taylor, S. 2003. The Challenges of Siting Facilities along the Sigsbee Escarpment in the Southern Green Canyon Area of the Gulf of Mexico, Framework for Integrated Studies. Keynote lecture, Proceedings, Offshore Technology Conference, Houston, Texas, OTC 15156. © 2011 by Taylor & Francis Group, LLC
31
Parker, E.J., Traverso, C., Moore, R. & Evans, T. 2008. Evaluation of Landslide Impact on Deepwater Submarine Pipelines. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19459. Parker, E.J., Traverso, C., Del Giudice, T., Evans, T. & Moore, R. 2009. Evaluation of Landslide Impact on Deepwater Submarine Pipelines. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19459. Potts, D.M., Kovacevic, N. & Vaughan, P.R. 1997. Delayed collapse of cut slopes in stiff clay. Géotechnique, 47, No 5, 953–982. Randolph, M., Cassidy, M., Gourvenec, S. & Erbrich, C. 2005. Challenges for offshore geotechnical engineering. International Conference on Soil Mechanics and Geotechnical Engineering, ICSMGE. Osaka, Japan, 123–176. Schroeder, F.C., Jardine, R.J., Kovacevic, N. and Potts, D. M. 2007. Potential Effects of Well Drilling Operations on Foundation Piles in Clay. Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. SUT, Society of Underwater Technology. 2003. Guidance Notes on Geotechnical Investigations for Subsea Structures. Offshore Soil Investigation and Geotechnics Group. SUT, Society of Underwater Technology. 2004. Guidance Notes on Geotechnical Investigations for Marine Pipelines, Offshore Soil Investigation and Geotechnics Group.
© 2011 by Taylor & Francis Group, LLC
Sultan, N. 2007. Gas hydrates stability Law, Hydrate Fraction and acoustic Velocities of Gas-Hydrate Bearing Sediments: Theoretical Study and empirical Expressions, Gas Hydrate Research Project. West Africa Deepwater Operators, (WADO), Deliverable 1. Sultan, N., Adam, S., De Gennaro, V., Lakshmikantha, M.R. & Puech, A. 2009. Acoustic properties and mechanical behaviour of marine sediments partially saturated by gas. IFREMER Scientific Report Task 2, Joint Industry Funded Project. Thomas, S., Bell, L., Ticehurst, K. & Dimmock, P. S. 2010. ‘An investigation of past mass movement events in the West Nile Delta’. Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Tjelta,T. I. 2010. ‘Prod probes Statoil’s seabed soils’. Offshore Engineer, February 2010. UKOOA. 1997. UK Offshore Operators Association. ‘Guidelines for the Conduct of Mobile Drilling Rig Site Surveys’, Volume 1, March 1997 and Volume 2, April 1996. USNCTT. 1984. U.S National Committee on Tunneling Technology, Geotechnical Site Investigations for Underground Projects, National Research Council, Washington D.C. (2 Volumes) Wheeler, S.J. 1988. The undrained shear strength of soils containing large gas bubbles. Géotechnique, 38, No 3, 399–413.
32
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot University of Massachusetts Amherst, Amherst, MA, USA
T. Lunne Norwegian Geotechnical Institute, Oslo, Norway
T.I. Tjelta Statoil, Norway
ABSTRACT: The paper presents recommendations for conducting site characterisation programs to determine the geotechnical properties of cohesive offshore sediments with the primary focus being on clays. The engineering behaviour of clays is complex and characterisation of their in situ properties is magnified by additional challenges presented by offshore environments. Site investigation programs are best conducted using well calibrated in situ tests and laboratory testing of high quality undisturbed samples. Results from these test programs should be coupled with geophysical data and collectively evaluated in the context of a regional geological framework. The paper reviews clay behaviour, describes the unique conditions of offshore environments, and lists key cohesive soil parameters required for infrastructure design and geohazards analysis. Best practice recommendations founded on these fundamentals are then presented including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing and evaluation of test data. The paper concludes with an assessment of present and future challenges. 1
INTRODUCTION
regional geologic framework that encompasses the past and current geologic states. The paper discusses characterisation of cohesive offshore sediments, which can encompass clays, plastic silts and organic soils, although the main focus is on soft to medium consistency clays. This is in part due to the authors’ experience is primarily with clay deposits, but also because a large portion of offshore sites consist of soft clay deposits, especially in deeper waters. There are equally important challenges with characterisation of other offshore deposits such as sands, calcareous soils, and non-plastic silts. While some of the site characterisation methods described are to varying degrees relevant to such soils, they are not the focus of the paper. The paper begins with a review of clay behaviour since such knowledge is essential to designing and implementing an appropriate site characterisation program. The unique and complex challenges presented by offshore environments such as over pressured zones, ultradeep waters, high salinity, ice gouging, etc. are described. Specific soil parameters required for design of offshore infrastructure such as anchor systems, pipelines, conductor installation, and geohazards analysis are listed. These background sections provide the context for the main objective of the paper which is to present recommended best practice for geotechnical site characterisation programs to determine reliable
Offshore geotechnical site characterisation programs are used to determine soil stratigraphy, in situ pore water pressure, and soil parameters for foundation design of infrastructure and geohazards analysis. Cohesive soils can have highly varied geologic histories making systematic quantification of their stressstrain behaviour complex. They exhibit significant stress history effects, can have a high degree of anisotropy, and exhibit strain rate effects. They are difficult to sample and test without causing excessive and irreversible disturbance. These challenges are magnified for offshore site characterisation programs since many offshore regions have complex geologic environments and site investigations are increasingly being conducted in deeper waters. Geotechnical site characterisation programs ideally combine in situ testing, in situ instrumentation, collection of high quality samples and follow-on laboratory testing. Assessment of sample quality is essential for evaluating the accuracy of laboratorymeasured mechanical properties. Coupling laboratory data with results from well calibrated in situ tests greatly enhances the reliability of site characterisation programs. The resulting geotechnical data should be studied concurrently with geophysical data and the collective data set should be evaluated within a
33 © 2011 by Taylor & Francis Group, LLC
soil parameters. Recommendations are given for all geotechnical aspects of the site characterisation process including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing, and evaluation and selection of design profiles. This paper follows other related overview papers on offshore site characterisation, notably Randolph (2004), Kolk & Wegerif (2005), and Andersen et al. (2008). The paper stresses the importance of geophysical and geological work as part of a comprehensive site characterisation program but these topics are not within the authors’ expertise. ISSMGE (2005) reviews offshore geophysical methods.
2
FUNDAMENTAL CLAY BEHAVIOUR
An understanding of clay behaviour is necessary to formulate and conduct a successful site characterisation program. It is important for the planning and execution of the in situ testing, soil sampling, and laboratory test programs so that the most relevant tools and test methods are utilized. It is also valuable during the process of comparing and analyzing the measured data sets and ultimately in development of design profiles for key soil properties such as in situ vertical effective stress (σv0 ), preconsolidation stress (σp ), consolidation behaviour, undrained shear strength (su ), and stiffness. This section reviews the fundamentals of intact (i.e. undisturbed) clay behaviour, the influence of sample disturbance on this behaviour, and the behaviour of remoulded clays. Leroueil & Hight (2003) and Ladd & DeGroot (2003) each provide a detailed treatise on clay behaviour; Lunne & Andersen (2007) emphasize behaviour of soft, deepwater clays. The focus herein is on monotonic testing behaviour; Andersen (2004, 2009) provides details on cyclic behaviour of clays.
Figure 1. Fundamentals of 1-D consolidation behaviour: compressibility, hydraulic conductivity, coefficient of consolidation vs. vertical effective stress (after Ladd & DeGroot 2003).
Key aspects of intact clay behaviour are stress history, consolidation, strength anisotropy, non-linear stressstrain behaviour, and strain rate effects.
Figure 2 plots su versus σp data for 27 low to medium OCR clays worldwide (22 from offshore locations) and with a majority having a plasticity index (PI) between 10 to 50% (NGI 2002). The su values were determined from triaxial compression tests conducted on good to excellent quality samples that were anisotropically consolidated to estimated in situ effective stresses prior to undrained shear (CAUC). The σp values were determined from 1-D consolidation tests conducted on companion test specimens taken from the same sample. The strong link between su and σp implied in Figure 2 is often expressed in terms of normalised soil parameters such that
2.1.1 Stress history All significant aspects of clay behaviour are influenced by stress history. It is quantified through the preconsolidation stress and the corresponding overconsolidation ratio OCR = σp /σv0 . The preconsolidation stress is essentially a yield stress, which separates small, mostly elastic vertical strains (εv ) from large, mostly plastic strains, as illustrated in Figure 1. It is more appropri ately referred to as the vertical yield stress (σvy ) but the familiar σp notation is used in this paper. In terms of consolidation behaviour, Figure 1 also illustrates the significant changes in the coefficient of consolidation (cv ) as a function of stress levels relative to σp . The hydraulic conductivity (kv ) decreases with an in increase in σv with an approximate linear relationship between void ratio (e) and logkv .
with a common value of m = 0.8 ± 0.1 and S depending on the particular su measurement method/mode of shear being investigated (e.g. Ladd 1991, Ladd & DeGroot 2003). Mesri (1975) and Terzaghi et al. (1996) reason that su is directly proportional to σp (i.e. m = 1) such as for example, in the case of stability problems with su (ave) = 0.22σp , where su (ave) = the average mobilized su . Linear regression to the data set shown in Figure 2, which assumes m = 1, results in su (CAUC) = 0.28σp . Jamiolkowski et al. (1985) describe several mechanisms that cause an OCR = 1 soil deposit to become overconsolidated including vertical stress relief, desiccation, ageing/drained creep (often referred to
2.1 Intact clay behaviour
34 © 2011 by Taylor & Francis Group, LLC
Figure 4. su (CAUE) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).
Figure 2. su (CAUC) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).
Figure 5. Schematic of shear modulus versus shear strain (modified after Jardine 1992, Leroueil & Hight 2003). same database as the CAUC (σ1f = 0◦ ) data plotted in Figure 2. Linear regression to these data results in su (DSS) = 0.23σp and su (CAUE) = 0.17σp . The regression coefficients imply su anisotropy ratios of Ks = su (DSS)/su (CAUC) = 0.82 and su (CAUE)/ su (CAUC) = 0.61. While there is scatter in the data sets presented in Figures 2-4, the trends are evident and similar to that presented for terrestrial soils (e.g. Ladd 1991, Terzaghi et al. 1996) showing significant su anisotropy for these low to moderate plasticity soils.
Figure 3. su (DSS) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).
as apparent preconsolidation), and physicochemical effects (e.g. cementation). This highlights the importance of developing an understanding of the geologic history of a site so that stress history data from in situ and/or laboratory testing can be properly evaluated and understood.
2.1.3 Non-linear stress-strain behaviour The stress-strain behaviour of clays is highly nonlinear. Clay stiffness, expressed as Young’s modulus (E) or the shear modulus (G), is greatest at small strain levels (less than ∼ 0.01% for clays) and degrades with increasing strain. Jardine (1992) presented a conceptual model and examples of soils in which the stiffness-strain degradation curve is divided into several zones, as shown schematically in Figure 5. This includes linear-elastic behaviour (up to Pt. A), followed by non-linear elastic behaviour (Pts. A to
2.1.2 Anisotropy Undrained shear strength is said to be anisotropic when its magnitude depends on the orientation of the major principal stress at failure (σ1f ), with su decreasing as σ1f rotates from vertical (same direction as deposition) to horizontal. Figures 3 and 4 plot results from consolidated direct simple shear ∼ 45◦ ) and anisotropically consolidated (DSS; σ1f triaxial extension tests (CAUE, σ1f = 90◦ ) for the
35 © 2011 by Taylor & Francis Group, LLC
B), plastic straining (Pts. B to C), and large plastic straining (beyond Pt. C). The shear stiffness within the region up to Pt. A is the maximum value (nearly constant) and is commonly referred to as the small strain shear modulus Gmax (or G0 ). It is often determined using elastic shear wave measurements such that
where ρt = total density and Vs = shear wave velocity. Like su , Gmax is also anisotropic and thus it should be reported with appropriate subscripts Gij to explicitly indicate the direction of propagation (i) and polarization (j) of the shear wave. For example, downhole seismic testing results in a measurement of Vvh and a computed Gvh while cross-hole testing yields either Ghh or Ghv . Figure 6. Normalised static undrained shear strength versus rate of shear strain for clays (from Lunne & Andersen 2007).
2.1.4 Strain rate effects Clays are sensitive to the rate of strain or loading and can exhibit a significant increase in su at fast rates of undrained shearing. Figure 6 shows an example of the influence of shear strain rate on su of several clays. The data show not just an increase in su with an increase in strain rate, but furthermore that the change in su per log cycle increases as the strain rate increases. The magnitude of the effect can depend on many factors including clay type, OCR, stress path, and may depend on whether the load is strain- or stresscontrolled and whether the shearing is static or cyclic (Lunne & Andersen 2007). Nevertheless, the effect is present for all such conditions and must be considered when evaluating laboratory and in situ test data relative to field conditions, all of which typically have significant differences in the rate and duration of loading. The rate effect shown in Figure 6 is often modeled using a simple log function with most data showing an increase in su between 5 to 20% per log cycle increase in strain rate. The effect can also be modeled using an inverse hyperbolic sine function (e.g. Randolph 2004), which allows for decaying strain rate effects (as is evident in Figure 6) below a specified threshold rate.
Figure 7. Hypothetical stress path for a low OCR clay element during tube sampling, specimen preparation and undrained shear (from Ladd and DeGroot 2003).
Each stage of the sampling process, from initiation of drilling to preparation of laboratory specimens, causes potential disturbance. The most important effect of sample disturbance in low to moderate OCR clays is a destructuring of the soil, which is accompanied by a significant reduction in the sample effective stress (σs ). For example, Figure 7 shows how the reality of sampling and testing can vary unpredictably from ). This figure shows the anticithe perfect sample (σps pated stress path for a low overconsolidated clay as stresses change from the in situ stress state (Point 1) to the stress state at laboratory testing (Point 9) as a result of disturbance caused by sampling, storage and handling. There are similar impacts on the small strain shear modulus, as depicted schematically in Figure 8 in terms of the shear wave velocity. Stress relief alone results in a reduction in Vvh , which is presumably recoverable during laboratory reconsolidation, however, destructuring results in an irreversible decrease in Vvh . It is common in geotechnical engineering practice to rely on the behaviour measured from laboratory
2.2 Influence of sample disturbance Sample disturbance is the most significant issue affecting the quality and reliability of laboratory test data for clays. It causes changes in the natural soil state and structure and as a result all key design parameters such as E, G, σp and su are adversely influenced by sample disturbance. Figure 1 schematically shows this for the one-dimensional consolidation properties with sample disturbance resulting in a more rounded compression curve with greater εv at all stress levels. This tends to obscure and usually lower σp , especially for more structured soils. The only parameters not significantly affected by sample disturbance are cv well beyond σp and the e–logkv relationship, unless there is severe disturbance.
36 © 2011 by Taylor & Francis Group, LLC
Figure 8. Hypothetical reduction in sample effective stress and shear wave velocity of a low OCR clay for the perfect sample and a disturbed sample.
index strength tests (e.g. unconfined compression tests, UUC). In such tests, the specimen is at an effective stress state represented by Point 9 in Figure 7 when, in fact, the in situ behaviour under field loading starts at Point 1. Moreover, not only is the effective stress state incorrect, but destructuring that occurs during poor sampling and handling further magnifies differences between measured laboratory and in situ behaviour. As an example of these effects, Figure 9 plots data showing the influence of different samplers on the undrained shear behaviour of the Onsøy, Norway, clay. The results are from CAUC tests conducted on Sherbrooke Block, 76 mm tube and 54 mm tube samples. The Sherbrooke Block sampler generally collects high quality samples followed by decreasing sample quality for the 76 mm tube sampler and especially for the 54 mm tube sampler (e.g. Lunne et al. 1997a, 2006). There are significant differences in the strain-strain-strength behaviour as sample disturbance increases, including: decrease in su , decrease in rate of strain softening, and increase in large strain shear strength. Lunne et al. (2006) show additional examples from DSS and CAUE tests for the Onsøy clay and several other clays.
Figure 9. CAUC recompression tests conducted on samples of Onsøy, Norway, clay from 14.5 m: a) stress-stress curves, and b) effective stress paths (from DeGroot et al. 2007).
been performed using the field vane and more recently full-flow penetrometers. As an example of some of these issues, Figure 10 presents results of motorized laboratory vane (MLV) tests conducted on an intact sample of Troll clay from the Norwegian sector of the North Sea. The sample was first tested to measure the intact su and the post peak residual shear strength. It was subsequently remoulded through multiple rotations of the vane (VR) and again tested. Thereafter the sample was thoroughly remoulded by hand (HR), tested, and followed by a final remoulding through multiple vane rotations (HRVR). Different values of sur also resulted from other methods including that of the fall cone (FC). In fact, for the FC, there are different calibration factors in current use (e.g. Norwegian Standard, Swedish Standard, ISO) and different values result from these various methods. The resulting sensitivity values St = su /sur depend on which measure of sur , is used to compute the sensitivity, and in the case presented in Figure 10, ranges from 2 to 12 for the MLV. Lunne & Andersen (2007) present data from UUC tests conducted on remoulded clays at different strain rates. The results indicate that the rate effect clearly observed for intact clays (Figure 6) also occurs for remoulded clays. Preliminary tests conducted at NGI also showed sur may be anisotropic.
2.3 Remoulded clay behaviour Some design solutions (as outlined in Section 4.0) require information on the remoulded undrained shear strength (sur ) and also its strength and stiffness evolution with time after remoulding. Laboratory measurement of sur has the significant advantage over measurement of su because sample disturbance is generally not an issue. However, it remains a relatively complex parameter to evaluate because results can vary greatly depending on the degree of remoulding and the measurement method used. Some test devices require a remoulded test specimen to be prepared (e.g. fall cone) while other devices can be used to remould an intact soil (i.e. device remoulding) and also directly measure sur (e.g. laboratory or in situ vane test). In situ methods of estimating sur , which all employ device remoulding by default, have long
37 © 2011 by Taylor & Francis Group, LLC
Table 1.
Summary of key aspects of clay behaviour.
Clay behaviour
Significance
Stress history (σv0 , σp , OCR) su anisotropy
Most important factor, controls all significant aspects of clay behaviour Very significant for low to medium PI clays, su (CAUC) > su (DSS) > su (CAUE)† Significant degradation in stiffness (E or G) with increasing strain Increase in su with increase in strain rate Value depends significantly on how measured and method of remoulding. Exhibits rate effects and anisotropy Most significant issue affecting quality and reliability of laboratory test data
Non-linear stress-strain Rate effects Remoulded sur Sample disturbance
Note: † except for varved clays where su (DSS) < su (CAUE)
should always be considered when selecting equipment and test methods to conduct a geotechnical site investigation. Figure 10. Shear stress vs. angular rotation for motorized laboratory vane tests conducted on Troll clay from 4.1 m depth.
3
UNIQUE CONDITIONS IN OFFSHORE ENVIRONMENTS
The various aspects of fundamental clay behaviour presented in the previous section challenge any site investigation program. However, these challenges are magnified in the offshore environment because of numerous unique logistical issues and geologic conditions that are often present. This section presents some of these conditions and describes how they add another layer of complexity to characterising the engineering properties of offshore sediments. Many of these conditions are classified as offshore geohazards, a topic for which keynote papers were presented at ISFOG 2005 by Jeanjean et al. (2005) and at ISFOG 2010 by Evans (2010). 3.1
Figure 11. Fall cone undrained shear strength (left axis) and shear modulus (right axis) vs. time for thixotropic test conducted on remoulded Troll clay.
Remoulded clays, if left undisturbed at constant water content and temperature will typically exhibit significant thixotropic hardening. Figure 11 plots results for the same Troll clay in Figure 10 showing the significant gain in su and Gvh (as measured using bender elements) with time after thorough hand remoulding. The rate in gain in su and Gvh are similar implying a near constant rigidity index IR = Gvh /su during thixotropic hardening. 2.4
Summary
Table 1 lists a summary of the fundamental aspects of clay behaviour described in this section. It is evident that clay behaviour is complex and this complexity
38 © 2011 by Taylor & Francis Group, LLC
Non-hydrostatic pore pressures
Many offshore regions have in situ pore pressures that are in excess of hydrostatic (i.e. overpressures). These overpressure zones can be caused by rapid sedimentation, mud volcanoes, pore fluid flow, manmade activities, and other mechanisms as summarized by Dugan et al. (2010). Given that the in situ vertical effective stress profile is critical to determining soil design parameters, detection of overpressured zones and measurement of the overpressure is an essential part of a site characterisation program. Overpressures are believed to be a contributing factor to the massive Storegga submarine landslide in the Norwegian Sea (Kvalstad et al. 2005) and have caused major difficulties during drilling operations in the Caspian Sea (Allen et al. 2005). Figure 12 presents an example of overpressures for the West Azeri field in the AzeriChirag-Gunashli (ACG) development in the Caspian Sea as measured using pore pressure dissipation tests
Figure 13. Schematic of reduction in effective stress state for consolidated OCR = 1 clay subject to post deposition influx of pressurised pore water.
platform. There are many such cases in SE Asia where poor cementing results in gas charging of shallow permeable layers. 3.2 Figure 12. Pore water pressure versus depth, West Azeri, Caspian Sea (after Allen et al. 2005).
(Section 5.5). The magnitude of overpressure equals about 800 kPa at 230 m below the seabed, and over the profile is supporting 40 to 50% of the soil buoyant unit weight. Determining the genesis of overpressures is also important. If it is a result from ongoing self-weight consolidation in a rapid sedimentation environment then the deposit is considered “under-consolidated” relative to the eventual final equilibrium σv0 . In such cases the deposit will be relatively weak and assuming hydrostatic pore pressures could lead to significant errors in selecting consolidation stresses for laboratory strength testing and evaluating final su design profiles. Conversely, a deposit that has excess pore pressures that evolved after self-weight deposition pore pressures have come to equilibrium should not be categorized as under-consolidated. One mechanism that can cause this is upward or lateral fluid flow from adjacent overpressure regions. If the onset of fluid flow excess pore pressure occurs after a previous equilibrium effective stress state, then the deposit will be overconsolidated (Figure 13). In some regions excess pore pressure from both rapid sedimentation and fluid flow from adjacent overpressure zones cans occur concurrently which makes for a complex in situ state to measure and evaluate. Although Dugan et al. (2010) note that recent advances in modeling overpressures have better defined different mechanisms and spatial scales. Overpressures can also result from manmade activities. For example, Lunne et al. (1996) report on gas charging of permeable layers from deep gas leaking upwards through weak cement around conductors in the Duyong B field, Malaysia. It was discovered during geotechnical drilling for a new nearby platform when a blow-out occurred. The gas charging built up pore pressures in adjacent clay layers threatening the safety of the foundation system of the existing
3.3
Highly irregular seafloor topography
There are numerous past and present geologic events that create highly irregular seafloor topography. Examples include mud volcanism, seabed slumping, small to mega-submarine landslides, debris flows, channel erosion, faulting, salt diapirism, and ice gouging. Such complexity in the seafloor topography presents obvious challenges to locating and installing offshore infrastructure, especially for routing of pipelines. It also implies that there can be significant spatial variability in soil properties, which requires careful planning and layout of the geotechnical site exploration program. 3.4
Gas exsolution
Stress relief from sampling, especially in deep waters, can cause exsolution of even small amounts of gas dissolved in the pore water. The subsequent expansion, if undrained, can damage the soil structure which in turn will impact the quality of the measured soil behaviour
39 © 2011 by Taylor & Francis Group, LLC
Low effective stress state
For flowlines and pipelines, the largest category of offshore infrastructure, soil-structure interaction takes place in the upper few decimeter of soil where there are very low effective stress (e.g. σv0 ∼ 2 to 3 kPa at 0.5 m depth). This, along with temperature induced cyclic loading effects (from episodic flows causing pipe contraction and expansion), complicates characterisation of the in situ soil properties such as modulus and shear strength. Laboratory measurement using consolidated test specimens (e.g. CAUC) is difficult to do reliably at such low effective stresses. For in situ testing methods, such as the piezocone and full-flow penetrometers, the tip force acting on the device from hydrostatic water pressure at the sea bottom can be very large relative to the penetration resistance of the upper few decimeters of soil. This creates measurement inaccuracy issues; a problem that is especially significant for soft soils and/or as water depth increases (e.g. Lunne 2010).
and design parameters. Lunne et al. (2001) conducted a laboratory test program that simulated deepwater sampling of clay with various amounts of dissolved gas. The measured CAUC stress-strain-strength behaviour was influenced by the degree of gas saturation, with a large reduction in su with increasing gas saturation. When disturbance caused by gas exsolution was coupled with simulated tube sampling disturbance, the combined effect was significant. CAUC reconsolidation volumetric strains were larger implying greater disturbance from the combined effect. But the stressstrain curves tended to strain harden, presumably due to the large volumetric strains and reduction in void ratio prior to undrained shear. The results indicate that disturbance caused by gas exsolution depends on the state of soil structure prior to exsolution with an intact structure being better able to resist gas exsolution than an initially disturbed structure.
3.5
Surficial crust
Figure 14. Evidence of surficial crust zone based on CPTU data, offshore Nigeria (from Ehlers et al. 2005).
Ehlers et al. (2005) describe the presence of a thin crust zone near the seabed in several deepwater regions as shown for example in Figure 14. This su profile, with the relatively high strength within the upper meter, is unusual compared to the more familiar deepwater soft clay profile which starts with the lowest values at the seabed and increases near linearly with depth. It is hypothesized that these crusts are a product of bioturbation and geochemical transformation, primarily into pyrite (Ehlers et al. 2005). Detecting the presence of these thin surficial crusts and properly characterizing the strength and stiffness properties is especially important to design and installation of shallow foundation systems, flowlines and pipelines (Section 3.2).
3.6 High salinity Some offshore regions have been found to have extremely high pore fluid salinity. Figure 15 plots data for the West Azeri field in the Caspian Sea (same site as Figure 12). The pore fluid salinity reaches values as high as 250 g/kg in comparison with Caspian Sea water of 12 g/kg. This high pore fluid salinity is believed to be associated with mud volcano activity as there appears to be higher pore water salinity at locations that are close to mud volcanoes (Kay et al. 2005). The upward migration of materials within the mud volcanoes includes pore water from greater depths and may also form conduits for further fluid flow. Excess pore water pressure in the vicinity of the mud volcanoes provides a mechanism for migration of dense brine vertically and laterally. High pore water salinity needs to be taken into account for computation of in situ pore water pressure. It also impacts many aspects of laboratory testing including measurement of index properties that rely on an oven dry mass and the need to match water used for saturation and testing of consolidated specimens to the pore fluid salinity.
Figure 15. Pore fluid salinity vs. depth below seabed for West Azeri region, Caspian Sea (after van Paassen & Gareau 2004).
In the Caspian Sea ACG development, there are high plasticity clay units that were found to be slickensided (Allen et al. 2005). This indicates that the sediments were subjected to failure conditions with repeated shearing. These soil units were not exposed to subaerial conditions and thus have not been desiccated. The likely mechanisms are mechanical, such as deformation of the soil units over the crest of the underlying Apsheron Ridge, or physio-chemical. Synaeresis cracks can occur in submerged sediments due to volume changes induced by changes in pore fluid salinity (Collinson & Thompson 1982). These post-deposition salinity changes are present at several of the ACG fields where the original pore fluid has been displaced by high salinity pore fluid migrating
40 © 2011 by Taylor & Francis Group, LLC
Figure 16. Normalised preconsolidation stress versus temperature (from Leroueil & Marques 1996).
from adjacent overpressure zones. It is likely that the fluid migration is cyclic, and in combination with the high plasticity swelling clay minerals present, is a condition conducive to synaeresis. Figure 17. Preconsolidation stress for ice-gouged and non ice-gouged sites in Smith Bay, Beaufort Sea (modified after Young 1986).
3.7 Temperature changes In offshore polar regions and in deep water, the seabed temperature often approaches 0◦ C (and even slightly below) with the pore fluid freezing prevented by the salt concentration. For example, Bugge (1983) reported that on the Norwegian continental margin at latitude of 64◦ N, the sea water temperature equals 0◦ C at 850 m and reaches an equilibrium temperature of −0.9◦ C at 1300 m. In the Mexican part of the Gulf of Mexico, Vidal et al. (1994) report temperatures lower than 5◦ C in 1200 m water depth. Collecting soil samples from a 0◦ C degree environment and testing them at room temperature (∼20◦ C), which is the common practice, raises the issue of temperature effects on the measured behaviour. Another temperature effect is that during production, hot gas or oil in pipelines can cause heating of surrounding soil. Leroueil & Marques (1996) studied the influence of temperature on σp and found a nearly 1% decrease in σp per ◦ C increase for temperatures ranging from 5 to 40◦ C, as shown in Figure 16. Since su is linked to σp , as described in Section 2.1.1, a similar temperature effect on su is expected (Perkins & Sjursen 2009). 3.8
out to sea at approximately 20 m water depth. Both sites involve the same Pleistocene deposit with Site W being heavily overconsolidated with nearly constant σp suggesting mechanical precompression due to erosion. Site T is at a greater water depth and in a zone of extensive ice gouging, which was determined in part from geophysical data. The repeated reworking at Site T caused significant softening within the upper 2 meters of the initially heavily overconsolidated deposit. Below 2 meters the σp values rapidly increase with depth as the effects of ice gouging diminish and coincide with those of Site W. Another interesting feature in offshore arctic regions is the presence of subsea relict permafrost below existing unfrozen sediment. For example, in regions of the Beaufort Sea, the surficial soils in shallow waters are often soft sediments that have a temperature just below freezing, but are unfrozen due to higher pore fluid salinity. These deposits were found to be underlain by permafrost, which formed during the last ice age when the sea level was much lower than present and the then subaerial deposits were subject to extreme freezing temperatures. During subsequent sea level rise, some of the top portion of the permafrost thawed and in some regions was covered with Holocene sediments. As a result, a veneer of soft sediment exists over a semi-rigid layer of permafrost. Foundation analysis for gravity based structures in such circumstances will involve modeling squeezing of a relatively thin layer of soft sediments. Additional problems, such as settlement, can result if oil and gas production melts the permafrost. In polar regions, ice loading can impose large lateral forces on offshore infrastructure such as gravity platforms. Ice loading depends on the nature of the ice encountered (i.e. landfast, transition or polar pack) and will have both mean and cyclic components. When
Ice gouging, subsea relict permafrost, and ice loading
Offshore regions that have been, and continue to be, influenced by the presence of sea ice often have several unique geologic features. Sea ice keels that interact with the seabed can gouge the surficial sediments. Repeated ice gouging reworks the seabed and can significantly reduce σp and su . It can also result in significant variations in seafloor topography. Figure 17 plots an example for the Smith Bay region of the Beaufort Sea, offshoreAlaska.The Sites W andT are located about 17 km from each other with Site W closer to land at approximately 10 m water depth and Site W further
41 © 2011 by Taylor & Francis Group, LLC
Table 2. Some unique offshore conditions and implications for conduct of site characterisation programs.
combined with gravity forces imposed by the structure, it creates a complex stress condition within the foundation soil. In some cases it can include significant shear stress reversals that are difficult to characterise (e.g. DeGroot et al. 1996).
Condition
Site characterisation remarks (details in Section 5)
Overpressures
3.9
Essential to detect and measure, use in situ dissipation tests and/or install piezometers Low effective Use in situ tools with high sensitivity stress conditions or hydrostatically compensated CPTU, collect high quality surficial samples and test on the vessel deck (e.g. box coring) Highly irregular Select in situ testing and sampling topography borehole locations to ground truth geophysical data and check for anomalous soil properties, carefully assess soil spatial variability Gas exsolution Use water sampling probe (e.g., BAT) to measure dissolved gas, consider use of pressure sampler Surficial crust Conduct accurate shallow depth in situ testing, collect thick (∼1m) box cores (if possible) and test on vessel deck High salinity Use corrected γw for computation σv0 , correct measured soil properties for salt content, could be indicator of overpressures Temperature Research ongoing, results may lead to changes development of data correction procedure Ice gouging, Select in situ test and soil sampling subsea relict borehole locations to ground truth permafrost geophysical data, carefully assess soil spatial variability Shallow gas and Use soil sampling boreholes to ground gas hydrates truth geophysical data, consider use of pressure samplers Cyclic loading Must collect undisturbed samples and conduct laboratory cyclic test program
Shallow gas and gas hydrates
Shallow gas and gas hydrates are hazards that can be encountered during site investigation and production drilling. Shallow gas can cause dangerous blow-outs during drilling operations (e.g. Lunne et al. 1996). The presence of gas hydrates, which have been found in many deep water regions, is a potential hazard because of the large volume expansion that can occur if they melt or dissociate. Geophysical methods are typically used during a site investigation program to detect the presence of shallow gas and gas hydrates. Ideally follow-on soil sampling should be conducted to ground truth findings from the geophysical surveys. This may require special pressure sampling equipment that can maintain in situ conditions in samples during their retrieval and storage (e.g. Kolk & Wegerif 2005). 3.10 Cyclic loading Cycling loading of offshore infrastructure can result from wind, waves, ice, and earthquakes. In production facilities such as pipelines, cyclic loading can be induced by temperature changes due to episodic flow of product. In soft cohesive soils, cyclic loading causes a reduction in effective stress due to pore pressure generation and hence a reduction in su . The behavioural response during cycling loading depends on the stress path and the combination of average and cyclic shear stresses. It is another level of complexity beyond that of the monotonic shear response of cohesive soils as covered in Section 2. Andersen (2004, 2009) presents comprehensive reviews of the cyclic behaviour of soils and provides guidance on testing and design applications.
3.12
Table 2 lists the unique and demanding offshore conditions discussed in this section. It also provides some comments that are a preview of the recommendations given in Section 5 for conduct of geotechnical site characterisation programs to evaluate these conditions.
3.11 Deep to ultra deep waters Site investigations in deep (>500 m) to ultra deep (>2000 m) water depths pose a multitude of logistical and soil testing and behaviour issues. Many of the challenges noted in the previous sections are exacerbated when working in deep to ultradeep waters. Some of these include: the regions are usually remote and thus have limited support from onshore facilities, handling of equipment can take a long time due to the large water depth, large hydrostatic water pressures may limit the accuracy of some in situ tools (Section 3.2), the degree of stress relief for samples is very large and there is a higher potential for gas exsolution (Section 3.4) and dissociation of gas hydrates (Section 3.9), the soil units are typically very soft (although sometimes a thin surficial crust exists Section 3.5), and the seabed temperature is often near 0◦ C in some regions (Section 3.7).
4
REQUIRED SOIL PARAMETERS FOR DESIGN AND ANALYSIS
This section lists soil parameters required for design and analysis of offshore infrastructure and assessment of geohazards involving cohesive soils. This information when coupled with a fundamental knowledge of soil behaviour and an awareness of the unique conditions that can often exist at offshore sites should be used to formulate the required outcomes from a geotechnical site characterisation program. Table 3 provides a list of soil parameters and examples of design concepts and geohazards for which these
42 © 2011 by Taylor & Francis Group, LLC
Summary
Table 3. Summary of cohesive soil properties required for various design concepts and geohazard analyses (modified after Andersen et al. 2008). Soil property
Requirements
Soil classification Index strength (e.g. FC, MLV, UUC, torvane, etc.)
– required in all cases – not required in many cases but almost always performed because tests are quick and easy to conduct – used in empirical design methods – required in all cases /σv0 for some – need K0 = σh0 applications (e.g. allowable drilling pressure) but difficult to measure – CR for set-up analysis of some anchor systems and skirted seabed structures; analysis of pipelines and risers – CR and Cαε for settlement computations (e.g. gravity based structures) – set-up analysis of some anchor systems, skirted seabed structures, piles; analysis of pipelines and risers – settlement time rate computations (e.g. gravity based structures) – anisotropic su required for most applications (e.g. tip resistance of anchors, pile shaft friction, shallow foundation bearing capacity, jack-up spudcans, slope stability) – Gmax and anisotropic stress-strain behaviour required for slope stability, skirted seabed structures – sur required for penetration resistance of suction anchors; capacity of anchor systems and skirted seabed structures; analysis of flowlines, pipelines, risers and debris flow – thixotropy and reconsolidated sur required when calculating set-up effects (e.g. anchor systems, skirted seabed structures) – set-up analysis of some anchor systems (e.g. suction, torpedo) and skirted seabed structures
In situ stress state – σv0 – u0 – σp Compressibility – consolidation CR = ε/logσv – creep Cαε = ε/ logσv
Intact soil flow properties – kv and cv – rk = kh /kv
Undrained shear strength – su anisotropy – Gmax and anisotropic stress-strain behaviour (including strain softening) Remoulded undrained shear strength – sur or St = su /sur – thixotropy – reconsolidated sur
Remoulded compressibility and flow properties – CR, kv and cv Cyclic properties – permanent/cyclic shear strains, cyclic pore pressures Dynamic properties – strength, modulus and damping Rate effects Temperature effects
Figure 18. Map of soil tests and design applications (from Randolph et al. 2007).
parameters are required. The information presented in this table is abstracted from the comprehensive design focused papers by Randolph et al. (2005) on challenges of offshore geotechnical engineering and Andersen et al. (2008) on deepwater geotechnical engineering. Clearly all site characterisation programs will involve tests necessary to classify the various soil units encountered. In terms of specific soil parameters, the in situ stress state (σv0 , u0 , and σp ) is the most important and required in all cases. For cohesive soils, su is also required for almost all cases. The need for other parameters depends on the specific design concept and geohazard being analyzed. Some of the factors that influence clay behaviour have a compensating effect, although to variable and typically unknown degrees. It must be considered when linking results from laboratory and in situ tests with soil response conditions anticipated for a design application. Figure 18 illustrates this interplay for the undrained shear strength. It maps the relationship between the influence of remoulding (decreases su ) versus the influence of strain rate (increases su ) for lab and in situ testing relative to that of various offshore systems.
5
– required of all applications involving cyclic loading
Competent site investigation programs utilize an integrated approach that engages a multidisciplinary team of geo-specialists. This is especially important for assessment of complex geohazards as described for example by Jeanjean (2005), Kvalstad (2007), and Evans (2010). Geophysical, geological and geotechnical investigation methods are often conducted in a phased approach to take advantage of the best that each method offers and as project requirements evolve. A collective evaluation of the resulting data sets by a multidisciplinary team provides the greatest opportunity
– required for all applications involving earthquake loading – consideration required in most cases – required for all cases with soil temperature change > about 10 to 20◦ C
43 © 2011 by Taylor & Francis Group, LLC
BEST PRACTICE RECOMMENDATIONS FOR GEOTECHNICAL SITE CHARACTERISATION
detailing the requirements for the project. It is also important that the field and laboratory factual reports explicitly describe the equipment and test methods used.This documentation is especially important when multidisciplinary teams, which are often composed of members from different organizations and far removed from the work conducted, need to use the factual reports to interpret the results and develop design parameters. In terms of notation for soil parameters, examples of poor practice in site characterisation reports persist. One such example is reporting of su values. As discussed in Section 2, su is not a unique parameter and its value depends on how it is measured. It is therefore essential to explicitly indentify the measurement method used when reporting su values, be it laboratory measured, e.g. su (FC), su (CAUC), etc. or interpreted from in situ test data using N factors relative to a reference su measurement, e.g. Nkt,CAUC , NT-bar,CAUC , etc. (see Section 5.8 for definitions of these N factors). Recommendations: While publication of the proposed ISO standard on Marine Soil Investigations will not be a panacea, and it may not necessarily be thoroughly adopted worldwide, it will provide a muchneeded reference framework. Project specifications should be developed detailing required site investigation activities. Field and laboratory factual reports should thoroughly document equipment, procedures, and data processing methods used. Notation used for soil parameters should be more explicit than is often the case to avoid ambiguity.
for an effective and reliable site characterisation outcome. In the sections to follow, best practice recommendations are given for the geotechnical component of site characterisation. The focus is primarily on presentation of key recommendations with many of the details provided in the references cited. These recommendations should be tailored to the scope of the investigation (i.e. preliminary, intermediate or final), anticipated challenges (Table 2), applications being considered (Table 3), schedule, and budget. 5.1 Work scope Several non-engineering factors such as availability of equipment and personnel, time constraints, and budget, often dictate the final work scope. Nevertheless, proper planning of the investigation involves making decisions on the type, depth, and number of in situ tests and/or borings that should be planned for. Most often this is governed by the depth of stress influence of the design concept and soil spatial variability. In terms of depth, the extremes include pipelines for which only the upper few meters are of interest to analysis of large submarine landslides that may require borehole depths of 100 m or more. Recommendations: Specific recommendations on number and depth of boreholes are not given here because it is highly project dependent (i.e. application, location, soil variability, anticipated geohazards present, etc.). In principle, it is recommended that both in situ tests and high quality sampling, followed by laboratory testing, should be performed. SUT-OSIG (2000, 2004) and ISSMGE (2005) provide general guidance on developing exploration plans. 5.2
5.3 Vessels and deployment modes ISSMGE (2005) gives an overview of the type of platforms and vessels that can be used for offshore investigations. In shallow waters, the common options are anchored barges, jack-up rigs, and anchored vessels, which all must provide a stable platform. In deeper waters, options include specialized soil drilling vessels with a moon pool or geophysical survey vessels that handle geotechnical tools (i.e. soil sampling and in situ testing devices) using an A-frame, crane or winch. Use of dynamic positioning is more efficient than vessel anchoring systems in shallow to moderate water depths, and is required for deep water investigations. The draft ISO standard on Marine Soil Investigations (ISO 2010) defines deployment modes as being either “drilling mode” or “non-drilling mode” (traditionally referred to as downhole and seabed mode, respectively). In drilling mode, a borehole is advanced into the seabed using rotary drilling. At selected depths a geotechnical tool is lowered into the drill string and advanced into the seabed from the bottom of the borehole. The drilling can be performed from a vessel or stable platform (“vessel based drilling”) or from a seabed system (“seabed drilling”), which involves a remotely operated drill rig that is placed on the seabed. Non-drilling mode involves using geotechnical tools that are advanced from the seabed. This includes simple free fall devices or more sophisticated equipment
Standards, reporting and notation
There is a large body of standards that cover geotechnical site investigations with most of these developed for in situ and laboratory testing of terrestrial soils. These standards have been developed by various countries and organizations (e.g. ISO, BS, ASTM, etc.) and in some cases there is a significant difference amoung such standards for conduct of the same test. This creates opportunities for confusion and mistakes in using the resulting data. At present the only comprehensive standard that was specifically prepared for offshore investigations is NORSOK G-001 (2004). This standard was originally developed in the 1980s for projects offshore Norway but evolved over time with input from an international group of North Sea operators. In 2007, ISO formed an international committee to develop a new standard on Marine Soil Investigations (ISO 2010), using NORSOK G-001 as a basis. At the time this paper was written (May 2010) a draft of the full standard was completed and under review. In spite of standardization efforts, some standards leave considerable latitude on how to conduct a test and in some cases no standards exist. For each soil investigation, project specifications should be prepared
44 © 2011 by Taylor & Francis Group, LLC
Table 4.
that is landed on the seabed prior to commencement of testing. The important distinction here is that no borehole drilling takes place and thus the depth of penetration is limited. There is a large variety of nondrilling mode equipment being used and the quality of data obtained can vary significantly. More details and recommendations on this are given in Section 5.4 for in situ testing and Section 5.6 for sampling methods. The technology for seabed drilling systems continues to evolve with the development of increasingly versatile and sophisticated robotic units. For example, the Portable Remotely Operated Drill (PROD) as originally described by Carter et al. (1999), is currently used in commercial applications. Upon landing of the system on the seabed, rotary drilling, in situ testing, and sampling are controlled via robotic assembly of the drill string. Casing, rods, in situ tools, and samplers are stored on a carousel. In situ tools can be deployed directly from the seabed or at the bottom of a drilled borehole. Borehole depths in excess of 100 m have thus far been achieved using this equipment. In deep water, this type of seabed drilling system is more efficient than vessel drilling due to less handling of the drill string. Recommendations: Depth accuracy during drilling, in situ testing, and sampling is critical. For vessel based drilling, it is important to utilize a heave compensation system to stabilize the drill string. Good systems use the so-called “hard-tie” method that stabilizes the drill string against surface motions by using an installed seabed template as the reference point. Set-down of seabed systems and seabed templates can disturb the upper seabed and also apply a stress field to it. To minimize these problems, the set-down must be carefully controlled and the equipment should be designed to minimize imposed seabed stresses within the region of in situ testing and soil sampling (see Lunne et al. 2010).
Deployment method Vessel based drilling Seabed drilling Non-drilling mode – seabed system ROV mounted Free fall
Rod type and potential penetration depth (m) drop-in or wireline (>100 m) carousel w/ casing and drill string (75 m) or wireline (150 m) straight (40–50 m), carousel (40–50 m) or coiled (15–30 m) straight, coiled or split (all 3 m+) 15–20 (?) m
Figure 19. Schematic of recommended sequence for recording deck-to-deck reference readings for seabed in situ testing (from Lunne et al. 2010).
not necessary for coiled rod systems. Remotely operated vehicles (ROV) have the significant advantage of being able to maneuver to specific locations. The obvious disadvantage is limited thrust capacity and hence penetration depth; adding suction anchors to the ROV will increase capacity but such a system has not yet been deployed in practice. Free fall penetrometers are either expendable or retrievable with a cable system and are relatively easy to deploy. Recommendations: Long continuous push strokes are preferred for push-in tools such as the CPTU and full-flow penetrometers. It is also important when deploying these tools to monitor zero reference readings at all stages of the test as shown in Figure 19 (i.e. on deck prior to deployment, at sea bottom before and after push, back on deck). This deck-to-deck logging is considered an important quality control procedure and should always be conducted (Lunne et al. 2010).
5.4 In situ testing 5.4.1 In situ testing deployment systems Table 4 lists the various deployment systems that are available for in situ testing. Vessel based drilling with downhole in situ testing is conducted using pressurized drilling mud or a hydraulic system to advance the tool (the depth limitation is normally 500 to 600 m in the latter case). For tools such as the piezocone (CPTU), stroke lengths of up to 3 meters are possible. A disadvantage of using pressurized mud that does not use a data umbilical line is that data are stored on a memory module and cannot be viewed in real time. Seabed drilling systems that use pushing by a drill string have the advantage of being able to push an in situ tool 10 m or more from the seabed, or the bottom of the borehole. Seabed drilling systems that use wire line deployment are in development (e.g. Boggess & Robertson 2010) and should be able to test at greater depths, although the push stroke for in situ tools will likely be limited to about 3 m. Non-drilling mode seabed systems that use straight rods require the use of a constant tension winch or a tower to support the rods, which is
5.4.2 In situ tools Table 5 lists the main in situ tools available for site investigations with the CPTU being the most common tool. There is a large body of experience in using the
45 © 2011 by Taylor & Francis Group, LLC
Deployment systems for in situ testing tools.
Table 5.
In situ testing tools.
In Situ Test/ Measurements
Comments
Piezocone (CPTU) – tip resistance (qc ), sleeve friction (fs ), pore pressure (u), inclination
– most common tool – best to use differential measurement but equipment still in development
Seismic CPTU – same as CPTU + Vvh
– requires a seabed shear wave velocity source
Full-flow penetrometers T-bar, ball – intact and remoulded penetration resistance
– higher resolution than CPTU, minor overburden correction, cyclic testing can be used to measure shear strength degradation
Field vane – intact and remoulded shear resistance
– remoulded shear strength depends on number of rotations used to remould soil
Piezoprobe – in situ pore water pressure
– data often difficult to inter-pret (covered in Section 5.5) – collects pore water sample to analyze for dissolved gas
Deep water gas probe – water sample, temperature
standard CPTU it is a more sensitive device (for same capacity load cell). Full-flow probes are deployed in the same manner as the CPTU using the same penetration rate of 20 mm/s. Cyclic testing using short strokes (e.g. ± ∼0.5 m) at selected intervals during the penetration phase of the test allows for measurement of shear resistance degradation upon remoulding. The field vane test (FVT) has been used extensively in soft Gulf of Mexico soils. For terrestrial soils, it has been calibrated for use in analyzing stability problems to account for anisotropy, rate effects and insertion disturbance (Bjerrum 1973). However, in offshore practice, the Bjerrum correction factor is typically not applied (Kolk et al. 1988, Randolph 2004) and the measured results are directly used to report su (FVT). This is justified in part because the reduction in shear resistance due to insertion disturbance is partially compensated by the high strain rates applied by the test (Randolph 2004). Expendable penetrometers and recoverable free-fall penetrometers continue to be studied (e.g. Aubeny & Shi 2006, Mosher et al. 2007) with some promising results. However, more research is needed before it can be recommended as a reliable tool. Recommendations: Lunne et al. (2010) provide guidelines for use of the CPTU, full-flow penetrometers, and the FVT in deepwater soft clays. Most of these recommendations are also relevant to their use in shallow water depths. DeJong et al. (2010a) outline recommended practice for onshore full-flow penetrometer testing, much of which is also relevant for offshore practice.
CPTU and in interpreting the measured data to determine soil parameters (e.g. Lunne et al. 1997b). Saturation of the pore pressure measurement system is not a problem offshore, in contrast to terrestrial applications. However, there are accuracy problems when testing in deep waters because of the high hydrostatic pore water pressure acting on the sensors at the seabed – a problem that is exacerbated when testing soft sediments. Measuring the differential cone resistance and pore pressure would solve this problem but no such cone has yet been deployed in routine practice. Lunne (2010) points out that due to differences in design, the sleeve friction depends on the type of cone penetrometer used. This makes sleeve friction values less reliable compared to the tip resistance and pore pressure. The seismic CPTU uses geophones or accelerometers to measure shear wave velocity arrival times which can be used to estimate the small strain shear modulus via Equation 2. The shear wave source is usually deployed at the seabed. Most seismic CPTU systems use a single set of sensors and the depth specific travel time is estimated as the difference between the arrival times for successive push intervals (pseudo-interval). However, a more accurate set-up is to use two sets of sensors located 1 m apart in the cone to record the shear wave arrive at both locations simultaneously (true-interval). In recent years, full-flow penetrometers such as the T-bar and ball have started to be used in offshore practice. Because they allow the soil to “flow” around the device, the vertical stress correction (as required for full displacement devices such as the CPTU) is minimal and with an area typically 10 times that of the
– The CPTU is recommended as the main tool to be used for in situ testing. It is the best tool for soil profiling and identification of soil behaviour type. It can be used for estimating σp and su but is not reliable for estimating sur . – The seismic CPTU should be used more often to measure Vvh (ideally using true-interval equipment), which can be used to estimate Gvh . – Full-flow penetrometers have proved to be as reliable as the CPTU for estimating su of soft clays and in some circumstances are better, e.g. shallow test depths and/or very soft sediments. They are the best option for measuring soil shear degradation and estimating sur . Both penetration and extraction resistance should always be measured when conducting full-flow tests. – The FVT remains a good tool for estimating su and sur (provided the soil is remoulded through multiple fast-rotations of the vane). However, the CPTU and full-flow penetrometer tests are quicker to perform and give a continuous profile of penetration resistance. 5.5
Determining the in situ pore water pressure condition is critical to conduct of a reliable site characterisation program. As described in Section 3.1, there are a number of offshore regions where overpressured zones
46 © 2011 by Taylor & Francis Group, LLC
Measurement of in situ pore water pressure
exist and have caused significant problems. Simply assuming hydrostatic pore water pressure conditions without verification is a risk. There are two options available at present for directly measuring in situ pore water pressure: piezoprobes and piezometers. 5.5.1 CPTU and piezoprobes The CPTU can be used to measure in situ equilibrium pore water pressure (u0 ) during a pause in penetration. Excess pore pressures that are generated during penetration will eventually dissipate back to u0 with the rate depending on cv . In clays with a low cv , this can take several days for the standard diameter CPTU (d = 36 mm) which is not practical for most site investigations. Since the rate of dissipation is proportional to d2 , small diameter piezoprobes have been developed to speed up dissipation. These probes consist of a tapered section that narrows down to a diameter of about 5 to 8 mm with a filter element near the tip. However, Whittle et al. (2001) showed that the pore pressure developed by the regular diameter push rod, to which the piezoprobes are attached, eventually reaches the smaller diameter tip and influences the dissipation time (and it can take as long as with a regular diameter CPTU). Whittle et al. (2001) thus developed an interpretation method that uses data from a dual element piezoprobe which consists of two pore pressure filters, one at the small diameter tip and one at the normal location on the shoulder of the cone (Figure 20a). The simultaneously recorded dissipation data are interpreted to estimate u0 . Piezoprobes have successfully been used in site investigations but unresolved issues remain. The lengthy, small diameter extension, which ideally should be as long as possible to reduce influence of the larger diameter push rod, is vulnerable to damage during penetration. When deployed in vessel drilling mode, imperfect heave compensation of the drill string can produce erratic results that are difficult to interpret. The use of a dual-element piezoprobe and the corresponding interpretation theory of Whittle et al. (2001) are still been researched (e.g. Flemings et al. 2008) and have not yet been validated for use in commercial practice.
Figure 20. a) T2P dual element piezoprobe (from Flemings et al. 2008), and b) example of multi-piezometer string installed in borehole (from Strout & Tjelta 2007).
Figure 21. Pore water pressure measurement options for piezometers (from Strout & Tjelta 2007).
and the soil formation. This is especially important for shallow depth installations. Three methods can be used to measure the in situ pore pressure at the depth of installation of the filter element (Figure 21, Strout & Tjelta 2007): a) direct measurement of u0 at the filter; b) differential pressure measurement at the filter relative to the seabed; and c) differential pressure measurement at the seabed. Practical considerations include saturation of the connecting fluid column for differential pressure measurements and the density of this fluid column needs to match that of the in situ pore water, sensors must be protected against corrosion, and the logistics of collecting data from the instrumentation package. Recommendations: Strout & Tjelta (2007) conclude that the most reliable method for measuring in situ pore pressure in low permeability sediments is to use piezometers. The key advantage is that they provide time history data. The cost is not trivial and new
5.5.2 Piezometers Piezometers can now be installed as part of a soil investigation in a predrilled borehole (i.e. drilling mode) or via non-drilling mode using a seabed push frame. In drilling mode, the creation of a borehole allows for the installation of a single piezometer or a string of piezometers. The piezometers are grouted in place and an instrumentation package is connected to the top of the string at the seabed (Figure 20b). An important practical consideration is the effectiveness of the grout, which must create a stable hydraulic seal around the piezometers with a hydraulic conductivity that is equal to or less than that of the formation soil. For push-in piezometers, the push stroke must be straight and concentric so as to not potentially create an open void, and thus a short circuit between the piezometer rise pipe
47 © 2011 by Taylor & Francis Group, LLC
Table 6.
solutions are being developed and this effort needs to continue. Seabed sediment overpressures was the focus of a 2009 workshop (Sheahan & DeGroot 2009) that identified challenges in dealing with overpressures. For current practice, workshop recommendations were made for measurement and installation options as follows:
Soil sampling equipment.
Deployment
Sampler and diameter (mm)
Non-drilling, DWS 110 seabed mode
– for shallow depth (<∼30 m below seabed or up to depth possible to penetrate from the seabed using a seabed system) push-in piezoprobes or piezometers can be used. They should be deployed at the beginning of an investigation to immediately start collecting data. Provisions could be made to retrieve the units at the end of the investigation, or later on if possible, for reuse. – for investigations using drilling mode (i.e. deep investigations), options depend on whether a piezoprobe can be pushed without damage; if not then piezometers are the only option. If so, both options are a consideration. Piezoprobes have the potential to give immediate data (depending on soil type, ship time available for waiting during dissipation). If time history data is required, then piezometers must be used. – advanced seabed drilling platforms, such as the PROD, provide new opportunities for more efficient installation in deep water, and to significant depths below seabed, as described by Tjelta and Strout (2010). – it is probably not yet possible with complete certainty to measure u0 with an accuracy greater than 10% of σv0 .
Potential penetration m Comments 20–25
Piston fixed relative to seabed, penetration measured reliably, underpressure measured behind piston
STACOR 105
20
Piston fixed relative to seabed, penetration not measured reliably
Kullenberg 80–110
3 - >50
Piston relative to vessel, problems with recovery and sample quality
Box corer typically 500 mm square
0.5–1.0
Miniature in situ tests can be conducted in box upon retrieval, subsamples can be collected
Seabed drilling
Piston or push, 44/75 Percussion or rotary, 75
75–150
Depth depends on system casing/rod storage capacity with wireline systems thus far being able to go deepest
Vessel based drilling
Piston, push, >100 or percussion 75
Advanced by push, drive or percussion using mud pressure or hydraulics with power from an umbilical.
5.6 Soil sampling, preservation and storage 5.6.1 Sampling methods Table 6 lists the different types of soil sampling equipment presently available. Section 2.2 discussed the issue of sample disturbance and its impact on the reliability of laboratory measurement of design parameters such as σp and su . When such advanced laboratory tests are to be conducted it must be a key objective of the site characterisation program to use a sampler that can collect good quality samples. Lunne & Long (2006) outlined the key features that an offshore sampling system should have in order to have the potential to collect good quality samples. These include (Figure 22): sharp cutting shoe, stationary piston relative to the seabed, minimal inside and outside friction, small area ratio, sample diameter of 100 to 120 mm, and a retractable core retainer. In terms of deployment, the sampler should be pushed at a steady rate of ∼20 mm/sec, the push rate should be measured, and the underpressure developed behind the piston should also be measured. The Deep Water Sampler (DWS) was developed with the features displayed in Figure 22 as a goal for its design and deployment (Lunne et al. 2008). It has thus far been used on several onshore and offshore investigations, including the Troll site (Norwegian North Sea), and has shown a capability to collect good to
© 2011 by Taylor & Francis Group, LLC
Figure 22. Design features of ideal sampler for soft to medium-stiff clays (from Lunne & Long 2006).
high quality samples (Lunne at al. 2008). The depth of these investigations was less than 10 m and its good performance remains to be verified for the design full penetration depth of 20 to 25 m. The STACOR gravity sampler (Fay et al. 1985) uses a piston that is fixed relative to a small frame that sets-down on the seabed during deployment. Thus if the frame remains stationary during sampling then so
48
does the piston. Penetration is not directly measured. Numerous cases have reported collecting good quality samples using the STACOR sampler (e.g. Borel et al. 2005). The Kullenberg type of sampler also penetrates the seabed by gravity and can collect very long samples of up to 50 m or more, which can be an efficient way of collecting long sample cores. However, there are problems with this type of sampler including uncertain penetration depth and low recovery, and it has been found that sample quality tends to decrease with depth. These latter two problems are largely a result of the use of a thick walled, blunt tipped, cutting shoe and the fact that the piston is fixed with a wire to the vessel and not positioned relative to the sea bottom. Downhole sampling in vessel based drilling mode has long been shown to be able to collect good quality samples providing a fixed piston sampler is used in conjunction with a thin walled sample tube having a sharp cutting edge (e.g. 5◦ for soft clays), small area ratio and zero inside clearance ratio, and furthermore that the drilling is carefully conducted with an efficient heave compensation system. As noted in Section 5.3, several multipurpose seabed drilling systems have been developed. For the PROD system, which uses a drill string to deploy a sampler, potential sampling depth depends on the number of rods the carousel can hold (which is a function of the diameter casing being used). Box cores are good for characterizing the upper 0.5 m of the seabed. Common box cores normally collect a 0.5×0.5×0.5 m sample and can be tested immediately after retrieval to the vessel. For soft soils, miniature laboratory vane and full-flow probes can be used to test the sample while it is still in the box (e.g. Low et al. 2008). Because the effective stress of these surficial sediments is quite low, it is generally not possible to conduct reliable advanced laboratory tests on these samples; thus, the use of these miniature probes is a good, practical solution. Figure 23 shows an example of such data recorded for a soft seabed sediment. Subsamples can be collected for follow-on onshore laboratory testing. The large volume of collected soil can be saved from multiple box cores and resedimented in the laboratory for scale model testing of shallow soil structure interaction problems as pipelines and risers (e.g. 1g model tests described by Andersen et al. 2008). Recommendations: Handling of long recovery samples is difficult and the vessel must be equipped to handle such samples without disturbance. Box core testing using miniature full-flow probes and MLV is recommended for characterizing the upper (∼0.5 m) sediments for shallow depth investigations. Tube samplers are ranked as follows relative to their potential to obtain as good a quality sample and as high a recovery ratio as possible:
Figure 23. a) Penetration, extraction and cyclic resistance versus depth for miniature T-bar test conducted in box core sample of a soft seabed sediment. b) degradation plot for cycle 2.
– For non-drilling mode seabed based sampling: 1) piston sampler fixed relative to seabed, favorable geometry, steady penetration with real time measurement of penetration, and accurate measurement of recovery (e.g. Figure 22, DWS), 2) gravity sampler with piston fixed relative to seabed (e.g. STACOR), 3) gravity sampler but piston fixed relative to vessel (e.g. Kullenberg), and 4) gravity sampler without fixed piston. 5.6.2 Sample handling and storage Sample handling and storage must avoid handling disturbance and must maintain sample moisture and temperature conditions. Samples should not be subject to excessive temperature changes from warm climates, overheated wax and especially freezing temperatures. This is especially important for all samples that may be used for advanced laboratory testing (Section 5.7) Evidence suggests that soft clay samples can survive commercial transport when properly protected (and not subjected to catastrophic impact such as dropping of the sample box). For example, Figure 24 plots results of constant rate of strain consolidation (CRS) tests performed on specimens trimmed from Sherbrooke block samples of the Onsøy clay at NGI in Oslo, Norway and again at UMass Amherst, USA after commercial air transport.
– For downhole sampling in drilling mode (vessel or seabed based): 1) thin walled fixed piston sampler with favorable geometry as noted above; 2) thin walled push sampler (although sampling without a fixed piston is not recommended).
© 2011 by Taylor & Francis Group, LLC
49
Table 7.
Figure 24. Comparison of CRS tests conducted on Sherbrooke Block samples of Onsøy clay at NGI, Oslo, Norway and again after air transport to UMass Amherst, USA (from DeGroot et al. 2009).
Recommendations: Sample tubes/liners should be sealed with a mechanical seal and/or with wax and capped. The samples should be stored vertically and should not be able to move inside the sample tube or liner. Soft clays should be kept in the tube or liner for transport to the onshore laboratory. For stiffer clays, the sample may be extruded in the field and sealed in a cardboard cylinder using plastic, aluminum foil and wax. Samples should as quickly as possible be placed in a constant temperature controlled container that has been set at the estimated in situ temperature. Samples should be transported upright in containers that damp shock and vibration. Once at the onshore laboratory, samples should be stored in a container or room with humidity and temperature control.
Test Category
Test Types
Comments
Classification and Basic Index Testing
– – – – –
– simple and relatively quick to perform – necessary part of any site characterisation program – cannot provide design parameters
Express/Index Strength Testing Parameters measured: su , sur
– pocket penetrometer – fall cone – torvane – laboratory vane, – unconfined compression – unconsolidated undrained triaxial
– simple and relatively quick to perform – equipment commonly available – often gives scattered results – successful use in design requires soil/ site specific correlations
Advanced Laboratory Testing Parameters measured: CR, Cαε , σp , cv , kv , su , sur , St c , φ , Gmax
– IL and CRS Oedometer – CAUC, CAUE & DSS – ring shear – permeability – resonant column – cyclic CAUC & DSS – bender elements
– provides best control of soil state and conditions in the laboratory – relies on good quality samples – provides direct measure of design parameters – equipment more complex but automation very efficient
water content density Atterberg Limits grain size specific gravity
Advanced laboratory tests, as listed inTable 7, are an essential component of any site characterisation program requiring design parameters for compressibility, consolidation/flow, and strength behaviour. The need for such testing is highlighted in Table 3 for the various design concepts and geohazards listed. Section 5.7.3 to follow gives an overview of recommended scope and methods for conduct of advanced laboratory test programs. For some projects, geochemical an dating tests may also be required for assessing the geologic origin and history of sediments.
5.7 Laboratory testing Table 7 divides laboratory tests into three primary categories: classification and basic index testing, index strength tests, and advanced laboratory tests. Every design concept will first and foremost require classification of the various soil units encountered and such testing should always be conducted as part of any site investigation. The express/index strength tests listed in Table 7 are popular because they are generally quick, easy to perform and can readily be conducted offshore. However, it is important to note that these tests generally use fast shear rates, different modes of shear, and test small soil volumes, and the results are greatly affected by sample disturbance. Undrained shear strength profiles developed using these devices often show significant scatter. Therefore, the data from these devices represent, at best, relative strengths rather than values suitable directly for design. They should be relied upon only to indicate the general consistency of soil layers. Reliable determination of su values for design should focus on appropriate in situ tests and use of laboratory equipment that can conduct consolidated-undrained (CU) tests.
5.7.1 Field laboratory For logistical reasons, offshore laboratory testing is usually limited to soil description, classification and index strength tests. In some cases it can be advantageous to conduct CRS tests offshore in order to get an early indication of stress history, which provides a preliminary indication of the soil state and aids in planning the advanced laboratory test programme. The CRS data can also be used to evaluate sample quality as described in Section 5.7.2 to follow. If box cores are collected they should also be tested offshore as described in Section 5.6.1. 5.7.2 Evaluation of sample quality While no definitive method exists for determining sample quality, valuable information can be obtained
50 © 2011 by Taylor & Francis Group, LLC
Laboratory tests for fine grained soils.
Table 8. Evaluation of sample quality for low to medium OCR clays using Lunne et al. (2006). e/e0 at in situ stresses for sample qualities 1 to 4 OCR
1 (very good to excellent)
2 (fair to good)
3 (poor)
4 (very poor)
1 to 2 2 to 4
< 0.04 < 0.03
0.04 – 0.07 0.03 – 0.05
0.07 – 0.14 0.05 – 0.10
> 0.14 > 0.10
from making use of both qualitative and quantitative methods. Visual observations on the vessel during sample processing and testing in the offshore laboratory can provide useful qualitative information on sample quality (e.g. evidence of cracking). Qualitative (visual) assessment of sample quality is best made by examination of sample X-rays. Radiographic methods allow for non-destructive visual evaluation of sample quality, layering and the presence/quantity of stones, cobbles and other inclusions. It is also possible to obtain computerized tomography (CT) scans of soil samples that are contained in non-metallic sampling tubes or liners. CT scans provide a visual image along the length of a sample and cross-sectional images.The sample radiographs and/or CT scans should be used to select the location within a tube or liner sample of the most critical advanced laboratory tests. Quantitative assessment of sample quality for intact, low to medium OCR clays, can be done by measuring volume change during laboratory recon solidation to the estimated in situ stress state (σv0 , σh0 ). The normalised sample quality parameter e/e0 of Lunne et al. (2006) is computed as
Figure 25. Example application of e/e0 sample quality evaluation for three samplers used at Troll field (from Lunne et al. 2008).
where e = change in void ratio during reconsolidation to in situ stresses, e0 = initial void ratio, and εvol = volumetric strain (= V/V0 ) from reconsolidation to in situ stresses. These data are available from 1-D consolidation tests (IL and CRS) and the consolidation phase of CU strength tests (triaxial and DSS). The corresponding sample quality is determined using Table 8. Figure 25 plots an example application for samples collected from the Troll field using the DWS, borehole drilling with tube sampling, and a gravity core sampler. The technique of evaluating sample quality using shear wave velocity continues to be developed (e.g. Landon et al. 2007, Donohue & Long 2007). The method involves measuring Vvh of a short sample section using bender elements mounted in a jig such as the one shown in Figure 26. The advantage of this method is that by using small strain shear waves generated by the bender elements, the measurement is nondestructive and can be performed immediately after sample collection. Measurements can also be repeated on the same samples again in the onshore laboratory to potentially track any additional disturbance from sample sealing, transport and storage.
Figure 26. Schematic of device for combined sample shear wave velocity and soil suction measurements.
Figure 27 shows the proof of concept for terrestrial soils by comparing Vvh normalised by the in situ Vvh , in this case measured with downhole seismic CPTU, versus e/e0 . Ideally this method could be used to quickly and non-destructively screen samples on the vessel and again later on in the onshore laboratory prior to making decisions on selection of samples for
51 © 2011 by Taylor & Francis Group, LLC
that are important to proper and reliable conduct of these tests. Lunne & Andersen (2007) describe an approach used at NGI for large testing programs to optimize the use of in situ and laboratory testing, and when su anisotropy data are required. A brief summary is as follows: 1) select a key representative borehole and conduct sufficient number of CAUC tests on the best quality samples to establish a reliable su (CAUC) profile, 2) in the same borehole conduct some CAUE and DSS tests to establish anisotropy ratios, 3) correlate su (CAUC) values with the results of CPTU and/or full-flow tests to develop a site specific correlation for the continuous in situ data (generic correlations for these in situ tests are discussed in Section 5.8), and 4) if possible conduct a few CAUC and DSS tests in other borings. When selecting the location of a test specimen, soil from within 1 to 1.5 times the tube/liner diameter from the tube top and bottom should not be used because of greater disturbance near the sample ends (Lacasse & Berre 1988). Sample sides should be trimmed during specimen preparation to remove potentially disturbed material. Sub-sampling should not be used for preparation of final specimen dimensions. Very soft clays will need to be supported during trimming and mounting without touching the specimen by hand during preparation. The best test to measure compressibility, flow and σp is the CRS. The CRS test is preferred over incremental loading (IL) consolidation tests because of the ability for continuous measurement of deformation, vertical load, and pore pressure for direct calculation of the stress-strain curve, kv and cv . The loading strain rate should be selected such that normalised base excess pore pressure (ub /σv ) is less than about 15% in the normally consolidated stress range. Sandbækken et al. (1986) report that rates of 0.5 to 1%/hr are adequate for most clays. Experience at NGI indicates that σp from CRS tests gives a value about 10 to 15% greater than that determined from IL tests that use 24 hr load increments. The CRS tests should be conducted as early as possible. Stress history information is important for preliminary assessment of various design concepts and is also for conduct of the CAUC/E and DSS strength tests. The CRS results will also give a direct indication of sample quality via the e/e0 method described in Section 5.7.2 and Table 8. Evaluation of undrained shear strength (su ) anisotropy is often a critical aspect of many offshore design concepts as discussed in Section 2.1.2 and listed in Table 3. In the laboratory, this is best evaluated using a combination of CAUC, DSS and CAUE test. This is especially important for low to medium plasticity soft clays for which the undrained shear strength anisotropy is typically the most significant (e.g. Figures 2–4). The rate of strain for undrained shear should be selected to account for strain rate sensitivity of clays and typical field loading rates. Recommended rates include 0.5 to 2.0% per hour for triaxial tests and 5%/hr shear strain for DSS (Germaine & Ladd 1988, Lacasse & Berre 1988).
Figure 27. Normalised sample shear wave velocity versus CRS measured e/e0 for samples of three soft clays (from DeGroot et al. 2007).
advanced laboratory testing. Once samples are tested, confirmation e/e0 data can be collected during the consolidation phase. The method does require in situ Vvh , with downhole seismic CPTU being the best option (Section 5.4.2). Measurement of sample suction may be used as an additional refinement of this non-destructive sample evaluation scheme. Suction in a clay sample is the negative equivalent of the sampling effective stress (σs in Figure 7) and its value relative to σv0 is an indication of sample disturbance. Measuring both Vvh and σs could be used in the framework shown in Figure 8 or that presented by Donohue & Long (2007) to assess sample quality. Sample suction can be measured using a portable suction probe such as that described by Ridley & Burland (1993). The suction probe in Figure 26 at the base of the bender element jig is based on that of Poirier & DeGroot (2010). While these non-destructive methods of evaluating sample quality show promise, more research and development is needed before it can be recommended for regular use in offshore practice. 5.7.3 Advanced laboratory test procedures The most important advanced laboratory tests for characterisation of clays include the CRS test for measurement of compressibility, flow and stress history, and CAUC, DSS, and CAUE for measurement of stress-strain-strength behaviour and anisotropy. Other measurements listed in Table 3 such as small-strain behaviour, remoulded behaviour, and cyclic response are also important in many cases but the CRS and CAUC/E and DSS tests are critical to almost all applications. The following discusses procedure issues
52 © 2011 by Taylor & Francis Group, LLC
Ideally triaxial and DSS specimens should be anisotropically consolidated to the estimated in situ effective stress state (σv0 , σh0 ), i.e. the Recompression method (Bjerrum 1973). In the absence of measured K0 data, the Brooker & Ireland (1965) correlation amoung K0 , PI and OCR can be used to estimate K0 and compute the laboratory consolidation stress σhc ∼ σh0 . The quality of su data from Recompression tests depends on sample quality. Increasing sample disturbance can destroy structural bonds and subsequent loss in measured su ; however, large volume changes during reconsolidation can result in an increase in measured su . If sample quality is found to be poor to very poor then consideration should be given to conducting SHANSEP (Ladd 1991) tests, which involves laboratory K0 consolidation to a OCR = 1 state of stress and then to various mechanically overconsolidated stress states. Otherwise the SHANSEP testing protocol (e.g. evaluation of anisotropy, shear rates) is the same as that of the Recompression method. The advent of reliable automated triaxial stress path cell systems greatly enhances the efficiency of conducting K0 consolidation. SHANSEP triaxial and DSS test data for each major soil unit are used to evaluate the S and m parameters for Equation 1 and the best estimate OCR profile is used in Equation 1 to compute the in situ su profile. For under-consolidated and normally consolidated clays, OCR = 1 SHANSEP testing is the only viable option. Ladd (1991) and Ladd & DeGroot (2003) provide further details on the Recompression and SHANSEP methods and describe the advantages and disadvantages of both. For laboratory measurement of sur , the motorized laboratory vane (MLV) is the most versatile device. It covers a large range of shear strength, can use different rates, and can measure shear resistance of samples in various states. Several measures of the sur are possible using the MLV as discussed in Section 2.3 and it is important that the method(s) required or used be clearly documented. The fall cone (FC) is also recommended to measure sur . While not as versatile as the MLV, there is a large amount of experience with the FC, it can cover a large range of shear strength (through the use of different size cones), and uses only a relatively small volume of soil. However, calibration factors for converting FC penetration to sur differ among standards (e.g. ISO, Norwegian, and Swedish standards use different factors), and thus the calibration factor used should be documented. Measurement of sur using UUC tests is not recommended if sur is less than about 5 kPa because the tests are difficult to perform and the membrane and area corrections are significant. The UUC test is better suited for stiffer soils and has the advantages of being suitable for investigating rate effects and to investigate sur anisotropy. Recommendations: Major recommendations for conduct of advanced laboratory tests include:
– it is essential to evaluate sample quality (Table 8) and the e/e0 value should be reported for all CRS (or IL) and consolidated strength tests – 1-D compressibility, flow, and σp are best determined using CRS tests – CAUC/E and DSS tests should be used for measurement of stress-strain-strength behaviour and su anisotropy. For good quality samples, specimens should be consolidated to the estimated in situ effective stress state. – for poor to very poor quality samples the triaxial and DSS test program should involve conduct of SHANSEP tests. SHANSEP testing should be used for under-consolidated and OCR = 1 soils. – sur is best measured using the MLV, and the method of remoulding must be specified. Companion tests may be run using the FC. – Andersen et al. (2008) give recommendations for cyclic, thixotropy, creep, and reconsolidated remoulded testing and present example results. 5.8
5.8.1 Stress history recommendations profiles are developed using soil total unit In situ σv0 weight measurements and estimated in situ pore pressures. In the early stages of a site investigation, it is common to assume that the in situ pore pressure profile is hydrostatic with a sea water unit weight γw = 10.1 kN/m3 . However, measurement of in situ pore pressure (Section 5.5) and pore fluid salinity will be required in regions with non-hydrostatic pore pressures and high salinity. There are many methods of estimating σp using CRS or IL data and the K0 consolidation phase of SHANSEP tests with no method being definitively better than any other. The authors typically use two to three methods including Casagrande, constrained modulus (Janbu 1969) and strain energy (Becker et al. 1987) and report values from all methods in the factual laboratory report. The Andresen et al. (1979) su /σv0 , PI and OCR correlation is also often used. CPTU data can be used to estimate σp using the correlation (Lunne et al. 2007b).
where qnet = CPTU net penetration resistance and k is typically within the range of 0.25 to 0.35. These data should be used with the laboratory σp values and geologic history information to develop a best estimated in situ σp profile. This profile should then be used to compute the corresponding OCR profile. If the
– samples should be X-rayed to check for features such as layering and visual signs of disturbance
53 © 2011 by Taylor & Francis Group, LLC
Data evaluation and interpretation
The following presents recommendations for data evaluation and interpretation with a focus on stress history and su . The evaluation of sample quality should be used to determine if σp and su . from the advanced laboratory tests are reliable. Generally it is desired to use data from tests conducted on samples of e/e0 quality rating 1 or 2 (Table 8).
Table 9. Recommended soft clay CPTU and T-bar N-factors for evaluation of su and sur (modified after Low et al. 2010). N-factor N factor, Nrem factor Nkt,CAUC Nkt,suave Nu NT-bar,CAUC NT-bar,suave NT-bar,rem,UU NT-bar,rem,fc NT-bar,rem,MLV
Definition
Mean
Range
qnet /suCAUC qnet /suave or qnet /suDSS (u2 – u0 )/suCAUC qT-bar /suCAUC qT-bar /suave or Tbar /suDSS qT-bar,rem /sur,UU qT-bar,rem /sur,fc qT-bar,rem /sur,MLV
12.0 13.5
10.0–14.0 11.5–15.5
6.0 10.5 12.0
4.0–9.0 8.5–12.5 10.0–14.0
20.0 14.5 14.0
13.0–27.0 12.5–16.5 12.0–16.0
Notes: su N values for clays with St < 8. Use NBall = NT-bar . Table 10. Typical SHANSEP S and m values for su (ave) (modified after Ladd and DeGroot 2003). Soil Description
S
m
Sensitive cemented marine clays (PI < 30%, LI > 1.5) Homogeneous sedimentary clays of low to moderate sensitivity (PI = 20 to 80%) Sedimentary silts and organic soils (Atterberg Limits plot below A-line) and clays with shells Lacustrine varved clays
0.20
1.0
0.22
0.8
0.25
0.8
0.16
0.75
Figure 28. Undrained shear strength data for low OCR soft clay site, Haltenbanken area of the Norwegian Sea.
as those presented in Figures 2–4, are also useful in evaluating the data. For low OCR soils, reference lines of minimum expected su ratios should be plotted, i.e. assume OCR = 1 in Equation 1 and select an S value for the su of interest [e.g. Table 10 presents typical S and m values for su (ave)]. The recommendations given in this paragraph are for evaluating su data and not intended to be a substitute for in situ and laboratory testing. Figure 28 plots an example data set for a lightly overconsolidated, soft, clay site in the Haltenbanken area of the Norwegian Sea. Site and soil layer specific CPTU Nkt factors were developed in reference to the su (CAUC) tests which were conducted on good quality samples. The most notable aspect of the data is the scattered su values from the strength index tests and that many of these values are very low. This result is represents a lower a common occurrence. The 0.28σv0 bound su (CAUC) reference line assuming OCR = 1 ). (i.e. Figure 2 with σp = σv0
Note: PI = plasticity index, LI = liquidity index
σp data are considered reliable, then a site specific k coefficient should be developed. 5.8.2 Undrained shear strength recommendations The CPTU and full-flow penetrometer data can be converted to su using an appropriate N factor
N values are specific to the reference su and in situ test conducted. Low et al. (2010) and DeJong et al. (2010b) present recommended N factors for interpretation of CPTU, T-bar, and ball data for soft clays. Table 9 presents the Low et al. (2010) factors, which were developed using a database that included several offshore sites. The laboratory su data should be plotted together with that estimated from the in situ tests using the N factors in Table 9. Depending on the quality of the laboratory su data, site specific N values should be developed. The su profiles should always be evaluated in conjunction with the stress history data. As described in Section 2.1.1, su and σp are strongly linked and Equation 1 can be used to help in evaluating the consistency of the su profiles relative to the σp profile. Andresen et al. (1979) present correlations amoung su /σv0 , PI and OCR based on Equation 1. Database correlations, such
6
Many innovative systems and test methods have been developed for geotechnical characterisation of offshore sediments. Significant progress has been made in implementing improved deployment systems, in situ tools, and samplers in practice. However, challenges remain, some of which are a matter of convincing practice to implement more widely what are already known best practice methods. For example, sample quality is
54 © 2011 by Taylor & Francis Group, LLC
PRESENT AND FUTURE CHALLENGES
so critical to the reliability of laboratory data that the use of proper sampling tools must be made a priority. Another example is project specifications, which must be developed with an understanding of appropriate procedures and standards to specify. This largely involves clear communication amoung all parties and use of proper QA/QC programs. Universal adaptation of the proposed ISO standard on Marine Soil Investigations will provide a valuable reference framework. At the other end of the spectrum, some challenges are more daunting, such as for example developing geophysical techniques that yield reliable soils parameters. Other important challenges and related questions include:
the quality and reliability of in situ and laboratory test data. The paper describes these phases and gives best practice recommendations for each. Some of the most important recommendations include: accurate depth control relative to the seabed during deployment of in situ and sampling tools, CPTU and full-flow penetrometers for in situ testing, piezometers for measurement of in situ pore pressure, non-drilling seabed sampling with a piston sampler that is fixed relative to the seabed and with an appropriate geometry for sampling of cohesive sediments, evaluation of sample quality, and conduct of CRS, CAUC/E and DSS tests to measure consolidation, stress history and anisotropic stress-strain-strength behaviour.
– implementation of new seabed based drilling systems is an exciting development and the innovation needs to continue. Their use should be more widespread, especially for deep waters. – implementation of hydrostatically compensated CPTU for deep water investigations and increased use of seismic CPTU to obtain small strain shear modulus profiles. – in situ pore pressure u0 : 1) piezometers should become standard practice for projects where knowledge of u0 is essential, 2) develop inexpensive reliable drop-in piezometers, 3) develop reliable indirect indicators of overpressures, e.g. could geophysical measurements be used? – characterisation of the upper 1 to 2 m of sediment, especially in deep waters. Is box core testing good enough and how to evaluate the quality of a box core sample? How accurate are in situ tools in characterizing this surficial zone? – in situ testing of intermediate soils such as silts that can undergo partial drainage during testing. Can variable rate testing become a reliable and practical approach for offshore investigations? – do ambient pressure and temperature samplers need to be used more often or can we compensate for stress relief and temperature change in laboratory measurements? 7
ACKNOWLEDGEMENTS The authors thank their colleagues at UMass Amherst, NGI and Statoil for their contribution to the numerous research and consulting projects from which results and findings are presented in the paper, and Thomas Sheahan who reviewed the manuscript.The first author thanks NGI and the University of Western Australia for providing sabbatical opportunities to work on offshore site characterisation and acknowledges the US National Science Foundation for its support on grant OISE-0530151.
REFERENCES Allen, J.D., Hampson, K., Clausen, C.J.F., & Vermeijden, C. 2005. Well deformations at West Azeri, Caspian Sea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 999–1004. Andersen, K.H. 2004. Cyclic clay data for foundation design of structures subjected to wave loading. Proc. Int. Conf. on Cyclic Behaviour of Soils and Liquefaction Phenomena, CBS04, Bochum, Germany, 371–387. Andersen, K.H. 2009. Bearing capacity under cyclic loading – offshore, along the coast and on land. 21st Bjerrum Lecture presented in Oslo, 23 November 2007. Can. Geotechnical J. 46: 513–535. Andersen, K.H.A., Lunne,T., Kvalstad, T. & Forsberg, C.F. 2008. Deep water geotechnical engineering. Proc. XXIV Nat. Conf. of the Mexican Soc. of Soil Mechanics, Aguascalientes, 26–29 November, 2008. Andresen, A., Berre, T., Kleven, A. & Lunne, T. 1979. Procedures to obtain soil parameters for foundation engineering in the North Sea. Marine Geotechnology, 3(3): 201–66. Aubeny, C.P. & Shi, H. 2006. Interpretation of impact penetration measurements in soft clays. J. Geotech. and Geoenvironmental Eng. 132(6): 770–777. Becker, D. E., Crooks, J.H.A., Been, K. & Jefferies, M.G. 1987. Work as a criterion for determining in situ yield stresses in clays. Can. Geotechnical J. 24: 549–564. Bjerrum, L. 1973. Problems of soil mechanics and construction on soft clays. Proc. 8th Int. Conf. on Soil Mech. and Foundation Eng., Moscow, 3: 111–159. Boggess, R. & Robertson, P.K. 2010. CPT for soft sediments and deepwater investigations. Proc. 2nd Int. Symp. on Cone Penetration Testing, Los Angeles. 9–11 May 2010.
CONCLUDING REMARKS
This paper presented best practice recommendations for geotechnical characterisation of offshore cohesive sediments. The particular focus has been on clays which have a complex mechanical behaviour, are not easily characterized, and are vulnerable to irreversible disturbance when sampled. This complexity is exacerbated for offshore investigations because of the many unique and challenging conditions that are present offshore. Many tools and methods have been developed to meet these challenges. Competent geotechnical site characterisation programs should combine the best of in situ testing and laboratory testing, i.e. using well calibrated in situ tools and collection of high quality undisturbed samples for advanced laboratory testing. Each phase of the site characterisation process, from drilling to evaluation of laboratory data, can influence
55 © 2011 by Taylor & Francis Group, LLC
International Standards Organization. 2010. Marine Soil Investigations. ISO TC 67/SC 7 N. In preparation. Jamiolkowski, M., Ladd, C.C., Germaine, J.T. & Lancellotta, R. 1985. New developments in field and laboratory testing of soils. Proc., 11th Int. Conf. on Soil Mech. and Found. Eng., San Francisco, 1: 57–154. Janbu, N. 1969. The resistance concept applied to the deformation of soils. Proc. 7th Int. Conf. on Soil Mech. and Found. Eng., Mexico City. 1: 191–196. Jardine, R.J. 1992. Some observations on the kinematic nature of soil stiffness. Soils and Foundations. 32(2): 111–124. Kay, S., Goedemoed, S.S. & Vermeijden, C.A., 2005. Influence of salinity on soil properties. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 1087–1093. Kolk, H.J., Hoope, J. & Imms, B.W. 1988. Evaluation of offshore in situ vane test results. Vane Shear Strength Testing of Soils: Field and Laboratory Studies, ASTM STP 1014: 339–353. Kolk, H. & Wegeriff, J. 2005. Offshore site investigations; New frontiers. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 145–164. Jeanjean, P., Liedtke, E., Clukey, E.C., Hampson, K. & Evans, T. 2005. An operator’s perspective onoffshore risk assessment and geotechnical design in geohazard-prone areas. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 115–143. Kvalstad, T.J. 2007. What is the current “best practice” in offshore geohazard investigations? A state-of-the-art review. Proc. Offshore Technology Conf., Paper 18545-MS. Kvalstad, T.J., Andresen, L., Forsberg, C.F., Berg, K. & Bryn, P. 2005. The Storegga Slide: Evaluation of triggering sources and slide mechanisms. Marine and Petroleum Geology, 22 (1-2): 245–256. Lacasse, S. & Berre. T. 1988. State-of-the-Art: Triaxial testing methods for soils. Advanced Triaxial Testing of Soil and Rock, ASTM STP 977, 264–289. Ladd, C.C. 1991. Stability evaluation during stage construction. J. of Geotechnical Eng., 117(4): 540–615. Ladd, C.C. & DeGroot, D.J. 2003. Recommended practice for soft ground site characterization: Arthur Casagrande Lecture. Proc. 12th Panamerican Conf. on Soil Mechanics and Geotechnical Eng., MIT, 1: 3–57. Landon, M.M., DeGroot, D.J., & Sheahan, T.C. 2007. NonDestructive sample quality assessment of a soft clay using shear wave velocity. J. of Geotechnical and Geoenvironmental Eng., 133(4): 424–432. Leroueil, S. & Hight, D.W. 2003. Behaviour and properties of natural soils and soft rocks. In Tan et al. (eds.) Proc., Characterisation and Engineering Properties of Natural Soils, Balkema, 1: 29–254. Leroueil, S. & Marques, M.E.S. 1996. Importance of strain rate and temperature effects in geotechnical engineering. Measuring and ModelingTime Dependent Soil Behaviour. ASCE Geotechnical Special Publication GSP 61, 1–60. Low, H.E., Lunne,T.,Andersen, K.H., Sjursen, M.A., Li, X. & Randolph, M.F. 2010. Estimation of intact and remoulded undrained shear strength from penetration tests in soft clays. Géotechnique. doi: 10.1680/geot.9.P.017. Low, H.E., Randolph, M.F., Rutherford, C., Bernie, B.B. & Brooks, J.M. 2008. Characterisation of near seabed surface sediment. Proc. OTC., Houston, USA, Paper 19149. Lunne. T. 2010. The CPT in offshore soil investigations – a historic perspective. Proc. 2nd Int. Symp. on Cone Penetration Testing, Los Angeles. 9–11 May 2010.
Borel, D., Pench, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 921–927. Brooker, E.W. & Ireland. H.O. 1965. Earth pressures at rest related to stress history. Can. Geotechnical J. 2(1): 1–15. Bugge, T. 1983. Submarine Slides on the Norwegian Continental Margin, with special emphasis on the Storegga area. IKU, Publication 110, 152 p. Carter, J.P., Davies, P.J., & Krasnostein, P. 1999. The future of offshore site investigation – robotic drilling on the sea-bed. Australian Geomechancis. 34(3): 77–85. Collinson, J. D. & Thompson, D. B. 1982. Sedimentary Structures. George Allen & Unwin, London, U. K. DeGroot, D.J., DeJong, J.T., Yafrate, N.J., Landon, M.M. & Sheahan, T.C. 2007. Application of recent developments in terrestrial soft sediment characterization methods to offshore environments. Proc. OTC, Houston, Paper18737MS. DeGroot, D.J., Ladd, C.C. & Germaine, J.T. 1996. Undrained multidirectional direct simple shear behavior of a cohesive soil. J. of Geotechnical Eng. 122(2): 91–98. DeGroot, D.J., Landon, M.M. & Lunne, T. 2008. Synopsis of recommended practice for sampling and handling of soft clays to minimize sample disturbance. Proc. 3rd Int. Conf. on Site Characterization, Taipei, Taiwan. DeJong, J.T., Yafrate, N.J. & DeGroot, D.J. 2010b. Evaluation of undrained shear strength using full-flow penetrometers. J. of Geotechnical and Geoenvironmental Eng. In press. DeJong, J.T., Yafrate, N., DeGroot, D.J., Low, H.E., & Randolph, M. 2010a. Recommended practice for full-flow penetrometer testing and analysis. Geotechnical Testing J. 33(2): DOI: 10.1520/GTJ102468. Donohue, S. & Long, M. 2007. Rapid determination of sample quality using shear wave velocity and suction measurements. Proc. Conf. on Soil Invest. and Found. Behaviour, SUT, London, Sept. 2007, 63–72. Dugan, B. Sheahan, T.C., Thibault, J.M. & Evans, T.G., 2010. Offshore sediment overpressures: overview of mechanisms, measurement and modeling. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. In press. Ehlers, C. J., Chen, J. Roberts, H. H., & Lee, Y. C. 2005. The origin of near-seafloor “crust zones” in deepwater. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis. 927–933. Evans, T. 2010. A systematic approach to offshore engineering for multiple-project developments in geohazardous areas. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. Fay, J.B., Montargès, R., Le Tirant, P., & Brucy, F. 1985. Use of the PAM self-boring pressuremeter and the STACOR large sized fixed piston corer for deep seabed surveying. Proc. Conf. Offshore Site Invest., SUT, London, 187–199. Flemings, P.B., Long, H., Dugan, B., Germaine, J., John, C., Behrmann, J.H., Sawyer, D., & Expedition 308 Scientists, 2008. Pore pressure penetrometers document high overpressure near the seafloor where multiple submarine landslides have occurred on the continental slope, offshore Louisiana, Gulf of Mexico. Earth and Planetary Science Letters, 269(3-4): 309–32. Germaine, J.T. & Ladd C.C. 1988. State-of-the-Art: Triaxial testing of saturated cohesive soils. Advanced Triaxial Testing of Soil and Rock, ASTM STP 977, 421–459. ISSMGE 2005. Geotechnical and geophysical investigations for offshore and nearshore developments. Int. Society for Soil Mechanics & Geotechnical Engineering. Handbook.
56 © 2011 by Taylor & Francis Group, LLC
Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc., 6th OSIG, SUT, London, UK, 151–176. Lunne. T., Andersen, K.H., Low, H.E., Randolph, M. & Sjursen, M. 2010. Guidelines for offshore in situ testing and interpretation in deepwater soft clays. Can. Geotechnical J., In press. Lunne, T., Berre, T., Andersen, K.H., Strandvik, S. & Sjursen, M. 2006. Effects of sample disturbance and consolidation procedures on measured shear strength of soft marine Norwegian clays. Can. Geotechnical J., 43: 726–750. Lunne, T., Berre, T. & Strandvik, S. 1997a. Sample disturbance effects in soft low plastic Norwegian clay. Proc. Recent Developments in Soil and Pavement Mechanics. Rio de Janeiro, Brazil, 81–102. Lunne, T., T. Berre, S. Strandvik, K.H. Andersen & T.I. Tjelta 2001. Deepwater sample disturbance due to stress relief. Proc. Int. Conf. on Geotechnical, Geological and Geophysical Properties of deepwater Sediments. College Station, Texas, 64–85. Lunne, T., Isa, O. & Tan, M. 1996. Shallow gas problem at Duyong B offshore Malaysia. Proc. 11th Offshore South East Asia Conference. Singapore. Lunne,T. & Long, M. 2006. Review of long seabed samplers and criteria for new sampler design. Marine Geology. 226: 145–165. Lunne, T., Robertson, P.K., & Powell, J.J.M. 1997b. Cone Penetration Testing in Geotechnical Practice. Spon Press, London. Lunne, T., Tjelta, T.I., Walta, A. & Barwise, A. 2008. Design and testing out of deepwater seabed sampler. Proc. Offshore Technology Conf. Houston, USA, Paper 19290. Mesri, G. (1975). Discussion of ‘new design procedure for stability of soft clays. J. of Geotech. Eng., 101(GT4): 409–412. Mosher, D.C., Christian, H., Cunningham, D., MacKillop, K., Furlong, A. & Jarrett. K. 2007. The Harpoon free fall cone penetrometer for rapid offshore geotechnical assessment. Proc. 6th OSIG, SUT, London, UK, 81–90. NORSOK Standard 2004. Marine soil investigations. G-001, Rev. 2, October 2004. Norwegian Geotechnical Institute 2002. Early soil investigations for fast track projects: Assessment of soil design parameters from index measurements in clays. Summary Rep./Manual, NGI Report 521553-3, 15 Jan. 2002. Perkins, S. & Sjursen, M. 2009. Effect of cold temperature of unfrozen Troll clay. Can. Geotech. J. 46(12): 1473–1481. Poirier, S.E. & DeGroot, D.J. 2010. Development of a portable probe for field and laboratory measurement of low to medium values of soil suction. Geotech. Testing J. 33:3.
© 2011 by Taylor & Francis Group, LLC
Randolph, M.F. 2004. Characterisation of soft sediments for offshore applications. Proc. 2nd Int. Conf. on Site Characterization, Porto, Portugal, 1 209–231. Randolph, M. F., M. J. Cassidy, S. Gourvenec & C. J. Erbrich 2005. Challenges of offshore geotechnical engineering, Invited State of the Art paper. 16th Int. Conf. on Soil Mechanics and Foundation Eng., Osaka, Japan, 123–176. Randolph, M.F., Low, H.E. & Zhou, H. 2007. In situ testing for design of pipeline and anchoring systems. Keynote lecture. Proc. of the 6th OSIG, SUT, London, UK, 251–262. Ridley, A.M. & Burland, J.B. 1993. A new instrument for the measurement of soil suction. Géotechnique 43(2): 321–324. Sandbeakken, G., Berre, T., & Lacasse, S. 1986. Oedometer testing at the Norwegian Geotechnical Institute. Consolidation of Soils: Testing & Evaluation, ASTM STP 892, 329–353. Sheahan, T.C. & DeGroot, D.J. 2009. Seabed Sediment Pore Pressure: Genesis, Measurement and Implications for Design/Analysis. Final Report Int. Workshop for US Natl. Sci. Fdn. Project, www.offshoregeohazards.org, 114 pp. Strout, J.M. & Tjelta, T.I. 2007. Excess pore pressure measurements and monitoring for offshore instability problems. Proc. OTC, Houston, TX. Paper18706. SUT-OSIG 2000. Guidance Notes on Geotechnical Investigations for Subsea Structures. OSIG-Rev. 2, Prep. by Subsea Working Group of Offshore. Soil Invest. Forum. SUT-OSIG 2004. Guidance Notes on Geotechnical Investigations for Marine Pipelines. OSIG-Rev. 3, Prep. by Pipeline Working Group of Offshore. Soil Invest. Forum. Terzaghi, K., Peck, R.B., & Mesri, G. 1996. Soil Mechanics in Engineering Practice. John Wiley and Sons, New York. Tjelta, T.I. & Strout, J. 2010. Piezometer installation in deepwater Norwegian Sea. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. van Paassen, L.A. & Gareau, L.F. 2004. Effect of pore fluid salinity on compressibility and shear strength development of clayey soils. In Hack et al. (eds). Eng. Geology for Infrastructure Planning in Europe. Springer-Verlag. 327–340. Vidal, V.M.V., Vidal, F.V., Hernandez, A.F., Meza, E. & Zambrano, L. 1994. Winter water mass distributions in the western gulf of mexico affected by a colliding anticyclonic ring, J. of Oceanography, 50: 559–588. Whittle, AJ, Sutabutr, T, Germaine, JT, & Varney, 2001. A Prediction and interpretation of pore pressure dissipation for a tapered piezoprobe. Géotechnique. 51(7): 601–617. Young, G.A. 1986. The strength deformation properties of Smith Bay arctic silts. SM Thesis, MIT, Cambridge, MA.
57
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat & H. Dendani Total, Pau and Paris, France
A. Puech Fugro France, Nanterre, France
J.-F. Nauroy Institut Français du Pétrole, Rueil-Malmaison, France
ABSTRACT: The paper describes the geotechnical properties of deepwater sediments encountered on the continental slopes of the Gulf of Guinea, in water depth ranging from about 400 m to 2000 m. After more than 10 years of site investigations, a large database on the behaviour of these sediments is now available and is tentatively summarized, with illustrations from various Gulf of Guinea sites located between Nigeria and Angola. The main characteristics of the deepwater West Africa clays are addressed, comprising plasticity, carbonate and organic contents, particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. It is shown that specific laboratory and in-situ testing procedures are required for proper determination of fine particle sizes and for the measurement of the relatively high soil sensitivity. The presence and origin of a near seabed “crust” is highlighted. Its origin is questioned, and potential implications on the design of soil-pipeline interactions are emphasized. Typical results of installation of suction piles, driven piles and VLA anchors are presented, which further illustrates the clay behaviour.
1
INTRODUCTION
(CLub pour les Actions de Recherche sur les Ouvrages en Mer) and with the financial support of CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). As a key regional characteristic, the Gulf of Guinea is a “geotechnically remote” area, meaning that no soil investigation drilling vessel is permanently present in West Africa. In 1998, the first deepwater Gulf of Guinea soil investigation, at the Girassol site in 1300 m of water offshore Angola, was done in conventional drilling mode. Following this, the soil investigation strategy adopted for the Gulf of Guinea was based on the use of 30–40 m penetration seabed-based in-situ testing tools and 15–25 m long gravity piston corers. In soft deepwater sediments, such seabed-based equipments are limited by their stroke range, not thrust capacity. The requirements for geotechnical engineering purposes for suction piles and anchors (main type of foundation in deep offshore West Africa clays) are then covered. The drilling of deep soil borings (needed for e.g. driven piles for TLPs) currently still represents a small minority of cases in the Gulf of Guinea, which is a major difference with the Gulf of Mexico practice. The physical and geotechnical properties of the Gulf of Guinea deepwater sediments are described in detail in this paper. The main characteristics of these soft clays are addressed, comprising index properties
In about ten years, the deep offshore Gulf of Guinea has become a mature oil province with rapid development of numerous oilfields, like in the Gulf of Mexico or offshore Brazil. Compared to the Gulf of Mexico, the Gulf of Guinea is known to present benign environmental conditions. If subsea currents are indeed generally low and maximum storm waves limited to less than 8–10 m, strong squalls may impose severe design loads for the moorings of floaters (i.e. for temporary moorings of MODUs, or for permanent moorings of FPUs or FPSOs). The Gulf of Guinea is also known to present relatively gentle seafloor conditions, but the full list of geohazards can be encountered off Angola and Nigeria, such as faults, pockmarks, shallow gas, gas hydrates, salt and mud diapirs, seabed slope instabilities, and sub-seabed mass transport deposits. These deepwater geohazards, which may represent serious challenges in terms of geotechnical design, have been studied, principally by Total and Ifremer, since the early 1990’s, with several scientific cruises in Nigeria, Gabon, Congo and Angola (e.g. Cochonat et al. 1996, Sultan et al. 2007 & 2010). Geotechnical characterisation of the Gulf of Guinea deepwater sediments has been the topic of various collaborative R&D projects, carried out under the umbrella of CLAROM
59 © 2011 by Taylor & Francis Group, LLC
(with the noticeable exceptions of Nigeria and ultra deepwater Angola, see Sultan et al. 2007, Hill et al. 2010, or Evans 2010), a new soil investigation strategy has emerged with the use of 30–40 m stroke seabedbased in-situ testing tools and large-diameter gravity piston corers allowing the recovery of 15–25 m long samples (Borel et al. 2002 & 2005). A first application for the Kuito FPSO anchors in 430 m of water offshore Cabinda is described in Alhayari et al. (2000). Adopting this survey strategy (also applicable to other remote deepwater provinces, such as the Australasia area) may allow a reduction of mobilisation times by performing the soil investigation from vessels of opportunity. Although still rarely used, vessels of opportunity may become important in fast track development projects. Deep soil borings are of course still required, either for the design of long driven piles (e.g. for the Kizomba TLP or the Benguela-Belize CT in Angola, Labbe & Perinet 2004, Will et al. 2006), for the installation of jetted well conductors (Evans et al. 2002), or in areas where more sophisticated testing and sampling might be needed in relation to specific geohazards such as in ultra deepwater offshore Angola (Hill et al. 2010, Evans 2010). But the drilling of 100 m+ deep soil borings represents a small minority of cases in the Gulf of Guinea, which is a key difference with the Gulf of Mexico practice. Currently, three geotechnical drilling vessels are capable of drilling in up to 2000 m of water, and two more can reach 3000 m of water. The PROD (Portable Remotely Operated Drill) is an available alternative, with a seabed-based system that combines the ability to take samples (piston sampling or rotary rock coring) and perform penetration testing (CPTs BPTs). The PROD1 and PROD2 systems are capable to operate in up to 2000 m and 3000 m of water, respectively (Kelleher & Hull 2008). By saving the time for tripping the drill-string, such a seabed-based system can be more efficient at deepwater sites than conventional drilling tools deployed from a vessel deck at the sea surface (Osborne et al. 2010). In 2009, the PROD1 was applied for the first time in the Gulf of Guinea for a soil investigation in water depths of 1100–1400 m offshore Angola.
(water content and plasticity, carbonate and organic contents), particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. Having possible implications for the design of pipelines, the near seabed “crust” of stiffer clay, locally encountered at several sites in Nigeria and Angola, has been the subject of several studies. Some key results are discussed, but the paper shows that the origin of this “crust” remains largely unexplained. Where relevant, specific laboratory and in-situ testing procedures are emphasised and illustrated by case examples taken from a number of West Africa deepwater sites located between Nigeria and Angola. The examples presented in the paper, covering water depths in the range of 400 to 2200 m, come from several sites operated by Total (containing both published and proprietary data) and from data published by the other three operators actively involved in the Gulf of Guinea, i.e. BP, Chevron and ExxonMobil. Some data presented in the paper were previously published but are completed by a larger database. Other unpublished data come from recent R&D studies. Some preliminary results from ongoing work on new topics of interest, such as strain rate effects, interface friction and modelling, are also presented. The design issues related to soil-pipeline interactions are briefly discussed. Finally, typical results of installation experiences are presented, covering suction piles, driven piles and VLA plate anchors. Such installation case studies provide a further illustration of the behaviour of Gulf of Guinea deepwater sediments, in particular with respect to friction degradation (i.e. clay sensitivity) and regain of strength with time (i.e. thixotropy and set-up). 2
SOIL INVESTIGATION STRATEGY
2.1 A strategy adapted to regional requirements The first deepwater soil investigation in the Gulf of Guinea was done in 1998 for the Girassol site at a water depth of 1300 m offshore Angola. It was carried out in conventional mode, i.e. from a specialised geotechnical drilling vessel and with down-hole sampling and in-situ CPT tools operated through the drill-string. Extending the shallow water soil investigation strategy to deep water (as was done in the Gulf of Mexico too) presented operational efficiency drawbacks, like the long time required for tripping the drill-string and the use of a 1.5 m stroke CPT tool, thus making the acquisition of data in deep borings rather time consuming. With FPSO and subsea wells being the preferred option for deepwater oilfield developments in West Africa, most of the foundations and anchors to be designed and installed only require a detailed investigation of the first 20–30 m of sediments, with emphasis on the top 1–2 m for the soil-pipelines or soil-risers interaction. Coupled with relatively gentle seafloor conditions in many Gulf of Guinea development areas
2.2
Large-diameter (100 mm) piston corers have been shown to provide quality of soil cores similar to 75 mm (3 ) thin-walled pushed-in piston samplers with the conventional down-hole method from a specialised geotechnical drilling vessel (Young et al. 2000). However, gravity piston corers may by-pass some of the top 1.5–2 m of very soft seafloor sediments. This may also be the case in extremely soft seabed sediments for the STACOR® which uses a base plate and a truly fixed piston (Wong et al. 2008). In its current configuration, the STACOR® corer has a 25 m long core barrel and will generally recover 12–25 m long samples, depending on the water depth or the clay strength gradient (Borel et al. 2002, Enjaume et al. 2010). Longer gravity piston corers may be able to recover up to 30–35 m long
60 © 2011 by Taylor & Francis Group, LLC
Scope to 30–40 m for foundations or anchors
cores in specific conditions, i.e. with a heavier ballast weight and in extremely soft sediments or highly sensitive clays (Montarges et al. 1987), but the operation of such long corers is not a routine practice and can be detrimental to the sampling quality. Using accelerometers for accurate monitoring of the coring operation is a significant improvement for better understanding the corer behaviour and improving core recovery and quality (Buckley et al. 1994, Bourillet et al. 2007). In addition to the CPT, a number of in-situ testing tools can be deployed from seabed-based systems, the most commonly used being the field vane (VST), and the full-flow penetrometers (T-bar TPT and ball BPT; Randolph et al. 1998 & 2005). In soft sediments, the maximum investigation depth is not limited by the thrust capacity but by the maximum achievable stroke (generally 30–40 m). Based on years of practice, the undrained shear strength of the clay is determined by correlation with the cone resistance (Lunne & Andersen 2007). Other in-situ testing tools have the advantage of either a direct measurement (VSTs), or an increased accuracy in very soft sediments (TPTs and BPTs). Typical examples of Gulf of Guinea measurements are provided in subsequent sections of the paper. The use of various in-situ testing tools offers a superior data quality and a better definition of the soil strength intercept at the seafloor. Borel et al. (2010) propose a soil investigation strategy, where “clusters” of in-situ tests (CPT, VST and cyclic TPT or BPT) are performed at a number of selected locations, in combination with high quality sampling for advanced laboratory testing and calibration of the Nk factors. At other locations, the scope can be limited to penetrometer (CPT or full flow) testing and a reduced number of sampling for assessing the variability in soil conditions over the site. An accurate calibration of the Nk (alternatively Nt ) factor then allows a relatively precise determination of the clay strength, and the main geotechnical design parameters can be defined shortly after the soil investigation on the basis of the penetrometer test results.
Figure 1. Evidence of seabed crust from lightweight CPT testing and laboratory testing in box core (after Borel et al. 2005).
soil disturbance and frame penetration in the soil. An example of results obtained with the SEASCOUT® is shown in Figure 1. Piston gravity corers have been extensively used for pipeline soil investigations in deep waters. The analysis of the top decimetres of samples however often discloses partial loss or remoulding of the material which cannot be used for accurate measurements at very low stresses. Concerns have also arisen with giant piston corers regarding their capacity to capture subtle changes in the geotechnical parameters in the top decimetres below seabed, due to possible washing-out of the extremely soft seabed materials during the initial phase of free-fall penetration. Until recently, box corer sampling was the best option for recovering intact seabed samples. The standard box corer, typically weighing about 200–300 kg, penetrates 40–50 cm into the seabed sediments under its own weight. A trigger mechanism then releases a latch which causes the swivel base to close-off the captured sample, before the whole unit is recovered on board the vessel. The recovered samples are of high quality, but the extremely soft nature of seabed soils cannot allow any sub-sampling without disturbing the material. “In situ” testing inside the box therefore appears as the only reasonable way to obtain relevant shear strength data. This can be done with a minivane, but using a mini T-bar has the advantages of (a) obtaining continuous profiles and (b) measuring the soil sensitivity by performing quick cyclic tests. A manually operated mini T-bar (called DMS) was developed at the Centre for Offshore Foundation Systems (COFS, Perth, Australia) for strength profiling in box cores (Randolph et al. 2007). In 2008, Fugro developed the DECKSCOUT™ system (Fig. 2), in which the tool can be fitted with full flow penetrometers (T-bar or ball probes) with an actuator which can apply constant rates of penetration in the range of 0.01 to 2 cm/s (Puech et al. 2010); cyclic testing for sensitivity measurement can be automatically monitored.
2.3 Scope to 1–2 m depth for pipelines or risers Soils encountered in deep waters generally present shear strength profiles linearly increasing with depth (strength gradient between 1.0 and 1.5 kPa/m), starting at very low values at seabed (typically 1 to 2 kPa). In these very soft soils, assessing the longitudinal and lateral restraints of pipelines or flowlines laid on the sea bottom is a challenging issue for the industry. This requires both a very accurate measurement of the intact and remoulded undrained shear strengths over the first decimetres (accuracy in the order of or better than 1 kPa) and a precise soil-pipe interaction model. In the Gulf of Guinea the random presence of the so-called seabed “crust” (see details in section 4.9) increases difficulties in assessing accurate shallow shear strength profiles and impact on pipe penetration. In-situ tests (CPTU and T-bar tests) have been performed from lightweight frames in order to reduce
61 © 2011 by Taylor & Francis Group, LLC
Figure 3. SMARTPIPE® and SMARTSURF™ modules. Table 1. Country
Site
Water depth (m)
Nigeria
A K O Q∗ U M Z
350–650 1250–1500 1500–2000 1400–1500 750–800 550–1100 1300–1400 400 1300–1400 700–1100 650–700 800–1400 1400–2250 1200–1500 1350 1300–1400
Figure 2. Fugro DECKSCOUT™ system in operation on box corer.
It is now widely accepted that soil-pipe interaction laws in very soft soils should better be obtained directly on the sea bottom. The Fugro SMARTPIPE® system measures directly in-situ the penetration of an instrumented pipe section and the longitudinal and lateral restraints resulting from static or cyclic loads (Hill & Wintgens 2009, White et al. 2010). The system is equipped with a 1 m stroke mini T-bar to characterise the seabed sediments at the test location. The second version of the equipment was successfully used offshore Angola at the end of 2009. As the SMARTPIPE® will necessarily be used at a limited number of discrete locations to obtain detailed information on the soil-pipe behaviour, a companion and complementary tool called SMARTSURF™ has been designed (Puech et al. 2010). It is aimed at providing accurate information on seabed soil properties required at a large number of locations regularly distributed all along the pipeline routes to (a) assess whether the chosen test locations give representative results and (b) extrapolate data with confidence to the entire pipeline network. The SMARTSURF™ is equipped with a 3 m stroke standard T-bar or CPTU, a 1 m stroke mini-T-bar and a pushed-in stationary piston sampler specifically designed to recover 2 m long cores of soft undisturbed soil. The SMARTPIPE® and SMARTSURF™ can be operated with the same launch and recovery system (LARS), allowing safe handling from any vessel of opportunity fitted with a convenient A-frame (Fig. 3). The SMARTPIPE® and the SMARTSURF™ can also be deployed from a specialist drilling vessel. The equipment is rated for up to 3500 m water depth. A site investigation strategy to obtain fast and reliable design parameters for soil-pipe interaction assessment entirely based on in-situ testing is proposed by Borel et al. (2010).
Congo Angola
∗
3
B∗∗ D F G I L P R S
Ehlers et al. (2005) ∗∗ Dutt & Ehlers (2009)
PHYSICAL PROPERTIES
Physical and geotechnical properties described in sections 3 and 4 were derived from a review of industrial or R&D data from sixteen sites. Details are given in Table 1. 3.1 Water content and submerged unit weight Deepwater sediments in the Gulf of Guinea are characterised by very high water contents (Fig. 4). At the seafloor, w is typically comprised between 150 and 250%, and decreases to 100–200% over the very first metres of penetration (typically 6–8 m). Over these upper 6–8 m the water contents are close to or over the liquid limit (giving 1 < LI < 1.2). Beyond that depth the rate of decrease becomes very slow over the depth of interest for the engineering of most deepwater structures (i.e. 20–30 m below seabed). Most of the sites exhibit water contents between 80 and 150%. The highest values are mainly (but not exclusively) representative of the deepest sites. A few deep penetration boreholes are available. They confirm a decreasing trend (Fig. 5), but water
62 © 2011 by Taylor & Francis Group, LLC
Gulf of Guinea sites considered in paper.
Figure 4. Water contents versus penetration in top 20 m (from STACOR® samples).
Figure 5. Water contents versus penetration depth (deep boreholes).
contents still present relatively high values (80–120% at about 120 m of penetration in the deepest boreholes). The site B borehole appears as an exception but it is worth noting that site B is in only 400 m water depth (Dutt & Ehlers 2009). There is a relatively large scatter in the values of water contents at a particular depth. This observation is not only valid for the global data set but also for a particular site or field, and even for a given borehole as highlighted in Figure 4. The bulk unit weights are very low, starting at 12–13 kN/m3 at seabed and reaching 13–15 kN/m3 below 6–8 m. Submerged unit weights of Gulf of Guinea soils are significantly lower than in the Gulf of Mexico where submerged unit weights become rapidly higher than 6 kN/m3 (e.g. Quiros & Little 2003). 3.2 Carbonate content Carbonate contents are generally lower than 20% of the total soil weight (Fig. 6). Variations with depth are mostly erratic even along the same profile (see for example site M). Scanning electron microscopy (SEM) observations indicate that carbonates are mainly shell debris or foraminifers randomly distributed in the clay matrix. The proportion of calcite detected in the finer fractions (<80 µm) is generally small. Infrared spectrometry measurements were performed on a sample of site F which presents a carbonate content of about 20% (Fig. 7). The absorbance peak corresponding to carbonates at a wave number of 1430 cm−1 is strongly reduced after sieving at 80 µm.
Figure 6. Carbonate contents versus penetration depth.
Calcite cannot be considered as a cementing agent of the soil matrix. Therefore, the compressibility and shear strength of Gulf of Guinea clays are not driven by the carbonate content, as is explained in sections 4.2 and 4.3. 3.3
Particular care should be taken in the measurement of the organic content of Gulf of Guinea sediments. Past experience has shown that the organic content of
63 © 2011 by Taylor & Francis Group, LLC
Organic content
Figure 7. Infrared spectrometry measurements for site F.
these soils is dramatically overestimated when using the traditional loss on ignition (LOI) method, which provides organic matter content in the range 10–20%. This overestimation is mainly due to the high fraction of kaolinite in the Gulf of Guinea clays. Kaolinite decomposes into metakaolin at 450˚ C, thereby loosing up to 13% of its weight in the form of water, which is interpreted as a loss of organic matter. A more appropriate method consists in deriving the organic matter content from the total organic carbon (TOC). The TOC represents the percentage in weight of atomic carbon involved in organic bonds, and can be quantified as the sum of the pyrolysable carbon (PC) and the residual carbon (RC) as measured by e.g. Rockeval or Poluteval techniques (Thomas et al. 2007). Several samples tested at IFP provided TOC percentages of 1–2%, in good agreement with values given by Holtvoeth (2004). The total organic carbon is not easily related to the bulk organic matter content since other atoms than carbon are also included in the organic matter: oxygen, hydrogen, sulphur, nitrogen. The organic content (percentage in weight of the organic matter) can be derived by multiplying theTOC by a conversion factor aimed at taking into account the percentage of carbon in the bulk organic matter. For this purpose, Broadbent (1953) recommended the use of a factor ranging from 1.9 to 2.5. For the Gulf of Guinea deepwater sediments, the conversion factor was assessed at IFP with the help of the Kerogenatron method, designed for pure kerogen isolation with no loss nor alteration. Analysis of specimens from various sites yielded organic matter contents in 100◦ C dried samples ranging from 3 to 6%. Although this method is rigorous, it is not suitable for routine applications due to its cost and complexity. In the light of the Kerogenatron results, a conversion factor of 2.5 appears to be reasonable, suggesting that the organic contents of Gulf of Guinea deepwater soils generally do not exceed 6–8% (Fig. 8). The dichromate oxidation method, as recommended by e.g. the British Standard BS 1377, yields organic content values typically between 3 and 10%, with the exception of site P.Although no direct comparison with
Figure 8. Organic contents versus penetration depth.
the TOC method is available, the dichromate oxidation method then appears as a reasonable alternative. 3.4
Particle size distributions (PSD) of Gulf of Guinea sediments are below 400 µm and very often below 100 µm. It has been observed that PSDs obtained from standard hydrometer testing after dispersion by sodium hexametaphosphate (HMP) are poorly representative of the actual particle sizes. SEM micrographs show an open flocculated structure consisting of series of linked clay aggregates (see later Figs. 18–19). The size of the unit particles appears to be small (a few micrometres) whereas crushing of the dry sample results in aggregates bigger than 5–10 µm (Fig. 9). These aggregates are difficult to de-flocculate by polyphosphate dispersive agents. This results in a general shift of the PSD towards coarser particles with stepping distribution curves where classes of particle sizes are missing. Several dispersing agents were systematically tested at the INPL/LEM laboratory (Nancy, France) in a Malvern Mastersize laser granulometer. Cation exchangers in the form of polysulfonate resin beads (100–500 µm in diameter) have been found to be the most efficient (Fig. 10). This technique offers the advantage to release Na+ and to adsorb the multivalent cations, which are removed from the dispersed suspension instead of remaining as soluble complexes. Ultrasounds applied to an untreated sample produce also strong dispersion. Combining ultra sounds and resin is the ideal solution when the laser granulometer is used.
64 © 2011 by Taylor & Francis Group, LLC
Particle size distribution
Figure 9. SEM photograph of a typical Gulf of Guinea clay. Figure 12. X-ray powder diffraction patterns of a typical Gulf of Guinea clay. M/I mica or illite, K kaolinite, Ph phyllosilicates, Q quartz, H halite, C carbonate, Py pyrite (from Thomas et al. 2007).
It is recognised that (a) hydrometer sedimentation overestimates the content of the finest particles, especially when these are dominant (which is the case in the present study), and (b) laser diffraction overestimates the largest particles because the intensity of the diffracted light varies with the second power of the volume, i.e. with the sixth power of the size, which means that a tenfold larger particle scatters one million stronger light. For Gulf of Guinea sediments the main effect originates from the platy shape of the analyzed particles (Fig. 9). A correct agreement between the sedimentation curve and the laser diffraction curve can be achieved using a modified Stokes law with simple disc model. Comparing the results obtained by sedimentation and by laser diffraction further yields additional information about the average aspect ratio of the particles.
Figure 10. PSD curves obtained with various dispersive agents (after Thomas et al. 2005).
3.5
Figure 11. Densitometer versus laser granulometer PSDs (after Thomas et al. 2007).
As was discussed by Thomas et al. (2007), differences in PSD are associated with different measurement techniques. Interpreting PSD curves must be done in relation to the method used because standard hydrometer (densitometer) and laser granulometer yield dramatic different results. As shown in Figure 11, the clay content, defined as the amount of particles passing the 2 µm sieve, is 20–30% with the granulometer and 60–80% with the hydrometer.
65 © 2011 by Taylor & Francis Group, LLC
Mineralogy
Figure 12 exemplifies typical X-ray diffraction powder patterns which can be routinely obtained on sediment specimens of the Gulf of Guinea. The detected crystalline phases are quartz, calcite, halite and phyllosilicates. Among the latter minerals, kaolinite appears as the only one significantly detected (2θ = 14.5◦ /7Á). No clear modulation around 7◦ can suggest the possible presence of smectites. The different clay types are better identified on oriented preparations (Fig. 13). The 2:1 clays (smectite or micas) are clearly identified around 7–8◦ / 14–16Á, illite at 10.5◦ /12Á and its first harmonic at 21◦ /6Á; kaolinite at 14.5◦ /7Á and its first harmonic at 29◦ /3.5Á. Adsorption of ethylen glycol reveals the presence of smectite (swelling clay) by the shift of the 7–8◦ /14Á peak towards 6◦ /18–20Á. Calcination at 550◦ C eliminates kaolinites and smectites and reveals the part of illite at 10–11◦ /10.5Á. The specific surface areas measured by absorption of methylene blue from aqueous solutions are between 180 and 260 m2 /g, which confirms the significant presence of smectite (Le, 2008).
Figure 13. X-ray diffraction patterns on oriented preparation sample after separation by sedimentation. LN: oriented preparation, EG: ethylene glycol treatment, 550◦ C: calcinated at 550◦ C.
Whereas properly applied X-ray diffraction allows a clear qualitative identification of the different clay species, their quantification remains a challenging task. Using the X-ray diffraction pattern is not relevant, because the height and area of a diffraction peak are determined not only by the abundance of species but also by the structure factor. Sophisticated and arduous techniques have shown the extreme difficulty in investigating the morphological and textural complexity of the clay phases, and in distinguishing between mixed illite-smectite phases, mixed kaoliniteillite phases and aggregated species. Transmission electron microscopy (TEM) showed that kaolinite particles occur in small size and strong anisotropy. The poor crystalline quality of kaolinite was confirmed by infrared spectroscopy (Thomas et al. 2007). As a semiquantitative assessment, one can assume that kaolinite is the dominant species (probably about 40–50% of the clay minerals) and that smectites are in significant proportion (>20% and up to 40%).
Figure 14. Atterberg limits versus penetration depth.
3.6 Plasticity The plasticity index (PI) of Gulf of Guinea deepwater clays is typically between 70 and 130 but can reach values as high as 150 near the seabed. These values are much higher than those frequently encountered in the “highly plastic” Gulf of Mexico clays with PI typically in the range 30–70. The plastic limit (WP) is quasi constant with depth, in the range of 50 ± 10% whereas the liquid limit (WL) typically decreases from 150–200% at the seafloor to 130–170% below 6–10 m of penetration (Fig. 14). When represented in the Casagrande diagram, the soils plot close to the A-line and classify as highly plastic clays (CH) to highly plastic silts (MH), as shown in Figure 15. 4
Figure 15. Gulf of Guinea clays in Casagrande diagram.
to define the stress history of the Gulf of Guinea soils, since no implicit assumption regarding the past overburden pressure is introduced. The YSR is determined from the ratio of the vertical yield stress σvy to the actual overburden pressure σv0 :
The vertical yield pressures derived from a number of standard oedometer tests performed on several typical deepwater sites are plotted versus depth in Figure 16 and are used to compute the yield stress ratio. is higher than the vertical effective The yield stress σvy stress σv0 . The distance between the profiles expressed − σv0 ) is quasi constant by the difference (σ = σvy with depth, typically in the range of 15–40 kPa for most sediments. The corresponding YSR values decrease
GEOTECHNICAL PROPERTIES
4.1 Yield stress ratio The term yield stress ratio (YSR) is preferred to the commonly used term over-consolidation ratio (OCR)
66 © 2011 by Taylor & Francis Group, LLC
Figure 17. Sensitivity framework for site Z clay. Figure 16. Vertical yield pressure and YSR versus penetration depth for typical Gulf of Guinea sites.
The properties of a natural clay differ basically from its intrinsic properties due to the influence of the soil structure (fabric and/or bonding). The structure of a natural clay depends on many factors, such as depositional conditions, ageing, cementation and leaching. The evolution of the void ratio with penetration depth, or with the effective vertical pressure σvo in a natural clay deposit, is called the sedimentation compression curve (SCC). The majority of normally consolidated natural clays have sedimentation compression curves which (after normalisation) lie in a narrow band above the ICL. The regression line through this band has been called the sedimentation compression line (SCL) by Burland. The SCL is not a fundamental line but represents the trend followed by a majority of natural sedimentary clays. Temporal variations may result in “saw-toothed” curves instead of smooth regular curves. The fact that the SCL lies to the right of the ICL implies that, for a given void ratio, a natural clay is capable of carrying an effective overburden pressure higher than the corresponding reconstituted clay. Cotecchia & Chandler (2000) have generalised Burland’s approach to sensitive clays. They have shown that the SCL is not unique (as implicitly assumed in Burland’s work) but depends on the soil sensitivity. Typically, for site Z clay, the sensitivity framework approach is illustrated in Figure 17 by plotting the following data:
from about 3 at 2–3 m of penetration to 1.1–1.3 at depth. Higher values of YSR can be found within the first 2 m below seabed, depending on the presence of a so-called “crust” (see sections 2.3 and 4.9), but this particular point is not discussed here. Yield stress ratios in excess of 1 are not the result of over-consolidation in the geological sense (i.e. no past overloading of the material). De Gennaro et al. (2005) have shown that these soils exhibit a significant structural effect, and that the (σvy − σv0 ) difference is a quantitative measurement of the “extra-strength” due to the material structure. 4.2 Compressibility De Gennaro et al. (2005) and Le et al. (2008) have highlighted the interest of interpreting the compressibility data of Gulf of Guinea deepwater sediments in the light of the conceptual framework introduced by Burland (1990) for comparing and interpreting the compressibility of natural sedimentary clays and reconstituted clays. A normalising parameter called void index Iv is introduced to aid in correlating the compression characteristics of various clays:
– the ICL, as proposed by Burland (1990); – the ICL obtained for Gulf of Guinea deepwater sediments (Le 2008). For effective vertical stresses in excess of 10 kPa (i.e. soils below 3 m penetration), a very good fit is observed; – the SCL proposed by Burland, which corresponds to a sensitivity St of about 4 to 5; – the SCL proposed by Cotecchia & Chandler (2000) for a sensitivity of 6; – the envelope and mean value of the SCCs obtained from the water content measurements on the site Z samples.
where e∗100 and e∗1000 = intrinsic void ratios corresponding to an effective vertical stress σv of 100 kPa and 1000 kPa respectively (determined by oedometer tests on reconstituted sample); and C∗c = intrinsic compression index. The properties of reconstituted clays are called “intrinsic” properties since they are inherent to the soil and independent of its natural state. The compressibility of reconstituted clays can be described by a reasonably unique line called the intrinsic compression line (ICL) passing by the points (Iv = 0, σv = 100 kPa and Iv = −1, σv = 1000 kPa). For a particular clay, e∗100 and C∗c are related to the value of the void ratio at the liquid limit eL . © 2011 by Taylor & Francis Group, LLC
Based on the trend of the sedimentation compression curves, this analysis suggests that the clay has a marked
67
structure and a sensitivity in the order of 4 to 6. It is worth mentioning that the sensitivity measured from in situ cyclic T-bar testing was found close to 6. According to the sensitivity framework, a normally consolidated soil with a post-sedimentation structure is characterised by compressibility curves that cross the SCL at their in-situ stress (σv = σvo ) before show ing an abrupt increase in compressibility at σv = σvy (yield stress). This is exactly what is observed in Fig ure 17 for site Z. At stresses above the yield stress σvy , the compression curves plunge and reach the experimental ICL at higher stresses (about 200–300 kPa). The compressibility is maximal after the yield point and values of compression index Cc as high as 2 can be observed. In summary, the Gulf of Guinea deepwater clays (a) are normally consolidated, (b) have a marked post sedimentation structure, and (c) have a sensitivity in the range of 3 to 6, and this sensitivity can be interpreted as a quantitative measurement of the structural effect. 4.3
Figure 18. SEM micrographs of site D clay samples from 0.5 m and 14.0 m penetration.
Microstructure
The microstructure of intact samples has been investigated at ENPC/CERMES (Le 2008) by using scanning electron microscopy (SEM) and mercury intrusion porosimetry (MIP) in parallel, as suggested by Delage & Lefebvre (1984).To preserve the microstructure during dehydration, samples were frozen as quickly as possible so as to get a crypto-crystalline structure of ice with no volumetric expansion due to freezing according to the methodology proposed by Delage et al. (2006). Figure 18 shows two SEM pictures of surface (0.5 m) and deeper (14 m) samples. At both depths, a clear evidence of large voids can be observed, with diameters equal or larger than 1 µm, typical of soft soils. Voids appear to be as frequent as the solid phase, corresponding to an approximate porosity of 50%. Connecting clayey bridges are well observed, giving to the assemblage the aspect of a honeycomb microstructure. Clay minerals are easily observable with an average thickness smaller than 0.1 µm and an average diameter close to 0.5 µm. In Figure 19, the SEM photos of the 0.5 m sample submitted to vertical stresses of 50, 200, 800 and 1600 kPa emphasise the changes in microstructure during compression.The progressive collapse of the material structure, principally due to a decrease of the pore sizes, is responsible for the high compressibility of the clay. Mercury intrusion porosity (MIP) is based on the principle that a non-wetting fluid (here mercury) cannot enter a porous medium unless a pressure is applied. Assimilating pores to cylindrical capillaries, the pressure p can be related to an equivalent entrance pore radius r by Laplace’s law:
Figure 19. SEM micrographs of site D clay samples (0.5 m penetration) after compression to 50–1600 kPa (from Le 2008).
As the pressure p is increased, pores of smaller and smaller radius are filled with the intruding liquid. Results are plotted under the form of a cumulative curve giving the pore size distribution (eHg ) of the porous medium. In Figure 20, the MIP pore size distribution curves of intact samples from various penetration depths define a particular pore entrance value of about 0.2 µm. This radius is compatible with the intra-aggregate pore sizes observed through SEM. Below 0.2 µm, the curves are superimposed and this domain is known to be typical of the inter-aggregate porosity and to be poorly sensitive to macroscopic changes in void ratio. 4.4
4.4.1 CPT cone resistance and T-bar profiles Typical cone and T-bar resistance profiles are presented in Figure 21. They are representative of the large number of CPTs performed on West Africa continental slopes. Below 2 m of penetration and down to 30–40 m, the penetration resistance increases quasilinearly with depth. The gradient in net cone resistance lies between 10 and 30 kPa/m with a general tendency
where σ = surface tension of the intruding liquid; and θ = contact angle.
68 © 2011 by Taylor & Francis Group, LLC
Undrained shear strength
Figure 22. Gradient of net cone resistance versus water depth.
Figure 20. MIP pore size distribution curves for site D clay (from Le 2008).
Figure 23. T-bar and VST test results with Ntv = 11.5 (from Borel et al. 2005).
4.4.2 Undrained shear strength and sensitivity Vane shear tests (VST) provide a measurement of the in-situ peak shear strength Suv at typically 1 m or 1.5 m intervals. The net cone resistance and the vane shear strength can be related by a cone factor Nkv such as qn = Nkv . Suv . The Nkv factor is in the range of 10–15. More detailed analyses performed on a small number of sites where 30 to 40 m profiles of VST and CPT tests could be closely correlated suggest an Nkv factor increasing with depth from about 11 near the seabed to 14–15 at depth (Puech et al. 2005). The Ntv factor applicable for T-bar data has been found close to 11.5 either for in situ standard T-bar testing (Fig. 23) or for mini T-bar testing (Puech et al. 2010). For water depth in excess of 700–800 m, the average gradient in shear strength ranges from 1.0 to 1.5 kPa/m. Slightly higher values can be locally encountered at shallower sites. Ultra deep water sites investigated so far (in a water depth of about 2000 m) have gradients on the lower bound. However, gradients of the same order have been found at sites in only 1400 m of water are depth. Normalised shear strength ratios SuVST /σvc high, with values in excess of 1 over the very first
Figure 21. CPT net cone resistance and T-bar profiles (site K).
to decrease with increasing water depth as shown in Figure 22. In areas unaffected by the presence of geohazards, resistance profiles are highly repetitive (additional examples of CPT profiles are given in Colliat & Colliard 2010). Over the top 2 m of penetration two different types of profiles may be encountered i.e. profiles linearly increasing with depth or profiles presenting a peak in resistance corresponding to the seabed crust. Comparisons are presented later in Figure 33. The origin and consequences of such peaks in geotechnical engineering are discussed in sections 4.9 and 5.
69 © 2011 by Taylor & Francis Group, LLC
Figure 24. Composite plot showing intact shear strength profiles (CPT and VST) and residual shear strength profiles (VST) compared to CPT sleeve friction and cyclic T-bar tests (cycles 10–30 shown). Figure 25. Design remoulded shear strength profiles (assuming St = 3 to 4) compared to CPT sleeve friction, VST tests (remoulded strength after 3 rotations) and cyclic BPT test results (cycles 1–10 shown).
1–2 m of penetration, decreasing to about 0.4–0.5 below 10 m. The latter is compatible with a ratio = 0.36 mentioned by Andersen & Jostad SuDSS /σvc (2004), assuming SuDSS = 0.75 SuVST (Fig. 27). The sensitivity St is expressed as the ratio of the intact undrained shear strength Su,i to the remoulded undrained shear strength Su,r, measured using the same instrument. The measurement of clay sensitivity is a topic that attracted specific attention, in particular since installation experiences suggested that it tended to be under-estimated by the measurements provided by both laboratory tests (UU) and in situ tests (see section 6.1). Farrar et al. (2008, cited in Robertson 2009) claim that the CPT sleeve friction is often close to the remoulded undrained shear strength measured by the VST in soft NC clays. But Lunne & Andersen (2007) have shown that the CPT sleeve friction measurements do not agree with other remoulded shear strength measurements. Figures 24 and 25 present a comparison of data obtained with the VST, the CPT and full flow penetrometers (Tbar and ball) at two different sites in Angola. In Figure 24 are shown:
Figure 25 present results from another site, showing: – the CPT sleeve friction; – the remoulded shear strength as obtained with the VST but after only 3 rotations of the blades instead of 10 as per ASTM recommendations (limitations due to time constraints); – the remoulded shear strength derived from the ball tests (Nb = 12) after 10 cycles; – the design remoulded shear strength profiles based on labvane tests and past installation experiences in the area (assuming a clay sensitivity of 3 to 4). These data illustrate typical results obtained in the normally consolidated but structured clays of the Gulf of Guinea, i.e.: • The CPT sleeve friction clearly compares with the
residual (not remoulded) VST shear strength values; • The VST remoulded shear strength measurements
(when the number of blade rotations is limited to 2 or 3 as is often the case in offshore practice) plot below the residual shear strengths but provide underestimated values of the clay sensitivity (i.e. in the range of 2 to 3); • The cyclic T-bar or ball tests provide higher values of sensitivity, generally consistent and less scattered
– the cone resistance (divided by a cone factor of 13) and the peak VST shear strength; – the CPT sleeve friction and the residual shear strength obtained with the VST; – the remoulded shear strength as derived from cyclic T-bar with Nt = 11.5 (cycles 10 to 30 are shown). © 2011 by Taylor & Francis Group, LLC
70
Figure 27. Correlation between VST and DSS shear strengths.
value representative of normally consolidated conditions (K0nc ) measured in the post-yield domain is close to 0.5 (in the range of 0.45–0.55 for the small number of specimens tested). Higher values are obtained in the pre-yield domain. 4.5.3 Shear strength anisotropy Series of direct simple shear (DSS) tests and anisotropically consolidated (CAUc and CAUe ) triaxial tests were performed. The shear strengths obtained are respectively noted SuDSS , Suc and Sue . The ratio Suc /SuDSS has been found close to 1.2. The values obtained for the Sue /SuDSS ratio are more scattered, due to the difficulty in controlling extension tests on very soft soils and at very low confining pressures, but a value of about 0.8 seems appropriate on the basis of the most reliable sets of data. A fair correlation has been found between the shear strengths obtained from VST and DSS tests, of the type SuDSS # 0.75 SuVST (Fig. 27). This result may be interesting for fast track engineering when only insitu data are available. It is noted that the data imply relatively high conversion factors between penetrometer resistances and average laboratory strengths (e.g. NtDSS # 15). This may be explained by both the high rate sensitivity of Gulf of Guinea clays and the robustness of their fabric (reduced softening with first pass of penetrometer).
Figure 26. Typical results from cyclic T-bar tests in site Z clay.
than labvane measurements. Sensitivities are often in the range of 3 to 4 but can reach 6 for site Z as shown in Figure 26. These values are in good agreement with the interpretation of compressibility data in the sensitivity framework (section 4.2) and were confirmed by back-analysing suction pile installation data at the respective sites (see section 6.1). Recent (unpublished) data obtained with the mini T-bar indicate that sensitivity values in excess of 10 may be obtained within the “crust”. It is generally admitted (e.g. Randolph et al. 2005) that 10 cycles are sufficient to reach stabilisation and derive a sensitivity value from full flow penetrometers. In the structured clays of the Gulf of Guinea, Borel et al. (2010) and Puech et al. (2010) emphasise the need to perform at least 25–30 cycles to reach stabilisation (Fig. 26).
4.6 Thixotropy 4.5 Advanced laboratory testing
Thixotropy is measured as the regain in undrained shear strength with time of remoulded samples maintained at constant water content and sheared at increasing time intervals. All available results (21 tests from 8 different sites) were compiled to provide the envelope curves presented in Figure 28. These results are consistent with those presented by Andersen & Jostad (2002). They are also in agreement with the set-up behaviour of suction piles at various Gulf of Guinea sites described by Colliat & Colliard (2010). The results obtained for site Z may suggest that the higher the sensitivity the faster the undrained
4.5.1 Sample quality A large majority of samples recovered in the deep waters of the Gulf of Guinea were taken using the STACOR® giant piston corer. All results referred to in this paper were derived from high quality STACOR® samples with the Lunne et al. (1998) quality index generally rating fair to excellent. 4.5.2 Coefficient of lateral earth pressure at rest K0 The coefficient of lateral pressure K0 can be obtained in drained triaxial tests under zero radial strain. The
71 © 2011 by Taylor & Francis Group, LLC
Figure 28. Results from thixotropy tests.
shear strength is regained, but data are too scarce to draw definite conclusions. 4.7
Consolidation properties
The coefficient of vertical consolidation Cv of Gulf of Guinea clays measured in incremental oedometer tests is generally found between 1 and 10 m2 /year in the virgin compression zone. The coefficient of consolidation Cv,r corresponding to recompression in the stress range of 0.5 σvo to σvo (often used to evaluate reconsolidation around suction emplaced piles) lies in the same order of magnitude. These values apply in the range 2 to 20 m with a slightly decreasing trend with depth. An interesting point is that similar results are obtained on oedometer tests carried out on horizontally or vertically trimmed specimen, which is consistent with SEM observations showing identical microstructure in both directions. Unfortunately the authors are not aware of experimental field data (settlement of structures, pore pressure dissipation around instrumented pile walls) which could be used to confirm the validity of these values at large (foundation) scale. Thorel et al. (2010) have performed centrifuge tests on reconstituted Gulf of Guinea clays. The Cv values calculated from the observed settlements by the Asaoka method are of about 30 m2 /year, i.e. 5 to 6 times higher than Cv values measured from oedometer tests on in situ samples of the same clay. Of particular interest is the value of Cv at mudline for pipeline applications. Laboratory testing on the extremely soft surface material is highly challenging and poorly reliable. Hill & Wintgens (2009) and White et al. (2010) report consolidation measurements performed with the SMARTPIPE® system (Fig. 3) where an instrumented pipe section was allowed to settle on the seabed. The pore pressures dissipate rapidly showing that lay-induced pore pressures under a real pipeline would dissipate within days, not weeks. A satisfactory match between field observations and Finite Element Analyses is obtained for Cv values of about 75–100 m2 /year.
Figure 29. Results of triaxial shearing tests on site M clay (from Le et al. 2008).
4.8
72 © 2011 by Taylor & Francis Group, LLC
Effective parameters from triaxial testing
The shearing behaviour is largely explained by the structure of the clays. Le (2008) and Le et al. (2008) performed two types of tests on a natural clay: (a) triaxial compression and extension tests, in which the samples were consolidated to their estimated in-situ effective stress state, then sheared in undrained condition, and (b) SHANSEP type tests, in which the samples were consolidated isotropically (or anisotropically) beyond their in-situ states of stress before being sheared in undrained condition. Typical results are presented in Figure 29. The TN4 (compression) and TN7 (extension) tests performed at in-situ stress state present a non-plastic response in the q − p space (p nearly constant before yield). Failure is characterised by high effective friction angles of 40◦ or even more (M > 1.6), and such results are typical of all sites investigated so far (Fig. 30). The non-plastic pre-yield response is related to the strong structure of the clay. The high effective friction angles may be explained by a “sand-like” behaviour of the material where the aggregates of fine particles play the role of grain-sized elements. Friction angles in excess of 40◦ have been found in other clays, e.g. Mexico City clays (Diaz-Rodriguez et al. 1992).
Figure 31. Interface shear test results: peak shear strengths from ring shear tests on remoulded clay compared to Cam-shear tests on undisturbed clay.
Figure 30. Composite plot of triaxial test results in the q − p space.
smooth interfaces simulating different pipeline coatings (Kuo et al. 2010). To simulate interface conditions at shallow pipeline embedment, the stress level was very low (<10 kPa), and rates of shearing ranging from 0.3 to 30 mm/min were used (where the slowest rate was expected to allow the determination of a drained behaviour, which was apparently not achieved). The apparently quasi linear increase of the peak shear strength with applied normal stress may appear satisfactory, but both sets of results have a number of apparent inconsistencies in common:
The TN16 (compression) and TN17 (extension) tests were performed at a stress state four times higher than the in-situ stress state. The response is different and failure is characterised by smaller effective friction angles (about 30◦ with 1.15 < M < 1.25). CIU tests performed on some specimens at even higher stress states have yielded effective friction angles as low as 20–25◦ (not shown here). This sharp decrease in the mechanical properties of the clay under increasing stress level is attributed to the progressive destructuration of the material. A practical consequence is that particular care should be taken in choosing a testing methodology (i.e. “in-situ stress” versus SHANSEP test procedure), depending on the geotechnical problem considered. It also confirms the original Ladd & Foott (1974) statement that clays with a high degree of structure do not exhibit a normalised behaviour because the structure is significantly altered during consolidation to high stresses.
• Data are highly scattered in the same test series; • No clear difference can be observed between slow
(drained) and fast (undrained) tests; • Rough interfaces may give a lower strength than
smooth interfaces.
Several aspects of the behaviour of Gulf of Guinea clays are still poorly understood and deserve further investigations. Work is presently ongoing in relation with the topics briefly described below.
Kuo et al. (2010) invoke the presence of faecal pellets as a source of pore water, inducing “hydroplaning” during interface shearing. This could explain the high variability observed for both peak and residual shear strengths. However, no pellets were observed in the samples tested at IFP. Obviously, more work is needed to clarify the soil-pipeline interface friction, in particular the potential for extremely low friction values when samples are sheared on a rough pipeline coating. Further Cam-shear tests on intact Gulf of Guinea clays are in progress as part of an ongoing CITEPH project.
4.9.1 Interface friction Peak and residual frictions that can be mobilised at soil-structure interfaces are important parameters for designing anchors, gravity bases or assessing the stability of pipelines or risers. Figure 31 presents results of ring shear tests (i.e. remoulded material) performed at Institut Français du Pétrole (IFP) on rough versus smooth steel interfaces and at moderate stress levels. The rates of shearing were respectively of 0.017 and 44 mm/min. On the same graph are plotted results of Cam-shear tests on undisturbed clay samples, performed with rough and
4.9.2 Rate effects Rate effects in the highly plastic Gulf of Guinea clays are significant. Figure 32 presents the results of a CIU triaxial test where the rate of shearing was varied in a ratio of 1 to 100. The response is clearly of the isotach type, meaning that the effect of strain rate changes is permanent and that separate constant rate of shearing curves can be generated. The increase in shear strength is about 18% per log cycle of shearing rate. The yield envelope is affected but not the critical state line. Rattley et al. (2010) carried out a series of constant rate of strain DSS tests on deepwater Angola
4.9
Further work
73 © 2011 by Taylor & Francis Group, LLC
2005). Work is ongoing within CITEPH, and first modelling attempts are promising. In parallel, the RASTRA model (Pereira & De Gennaro 2009, De Gennaro & Pereira 2010) reproduces correctly the rate effects, as shown in Figure 32. Efforts are now deployed to couple S3-SKH and RASTRA into a single model. 4.9.4 Effect of gas Gas may be occasionally encountered in West Africa deepwater sediments. It may be of thermogenic origin or associated to the dissociation of hydrates. A Joint Industry Project (JIP) on the behaviour of Gulf of Guinea clays containing gas was recently completed. The work, supported by BP, ExxonMobil, Total, Fugro, Ifremer and ENPC-CERMES, had the following objectives: – to investigate mechanisms of gas generation in marine sediments; – to identify and quantify the amount of free gas by correlating free gas content and acoustic properties of marine sediments; – to develop specific laboratory procedures for testing gassy soils and quantify the effects of various gas contents on the shear strength and compressibility; – to build a database of high quality tests on three different sediments to serve as reference for further analyses and interpretation by operators and academic organisations.
Figure 32. Strain rate effects on site D clay.
Whereas the results remain confidential, the JIP opens two areas for further R&D work. Firstly, the amount of free gas and the resulting damage are well detected by variations in the acoustic properties of the sediment. The availability of a tool capable of measuring in-situ and simultaneously the compressive and shear velocity of the sediments (VP and VS ) is essential to identify and characterise gassy soils. Developing such equipment is the aim of the “GEOPS” R&D project jointly conducted by Ifremer and Fugro and supported by Total. Secondly, limitations were highlighted in existing models dedicated to the description of the behaviour of gassy sediments (e.g. Sobkowicz & Morgenstern 1984, Wheeler 1988, or Grozic et al. 2005), and possible improvements were identified. Work is under way in the framework of an ongoing CITEPH project.
clay samples, with strain rates ranging between 0.003 and 300% per hour (the latter being 100 times faster than the reference DSS testing rate). A clear trend for increasing shear strength with increasing strain rate was obtained, with ratios of 16% to 19% per log cycle. According to Rattley et al. (2010), “the viscous mechanisms controlling this rate dependency appear to be consistent for both strain rate and creep behaviour, as described during one-dimensional compression, and provide evidence of the applicability of a unique stressstrain rate relation for the offshore Angola clay which is likely to be related to the ratio of its mineralogical components”. A practical consequence is that, based on specific testing procedures, optimised foundation design could be considered by taking full advantage of the increase in available soil strength under rapid loading (e.g. squalls) presented by Gulf of Guinea clays.
4.10 About the near seabed “crust” 4.10.1 Characterisation of the crust The undrained shear strength profile of deep ocean sediments is generally characterised by very low values at seabed (typically in the range of 1–2 kPa) and a quasi-linear increase with depth (gradient lying between 1.0 and 1.5 kPa/m). In some areas around the world, where there is evidence that the sediment is historically normally consolidated (NC), the shear strengths within the upper 2 m (exceptionally 3 m) of penetration are found higher than normally expected. Values of up to 10–20 kPa can be observed. The shear
4.9.3 Modelling Based on our present knowledge of the behaviour of Gulf of Guinea clays, it can be inferred that a constitutive model should be capable to reproduce effects of non-linearity at small strains, stress history, structure (initial sensitivity and rate of destructuration) and strain rate. The Baudet & Stallebrass (2004) S3-SKH model, including effects of structure and destructuration in its formulation, offers attractive features (Baudet & Ho
© 2011 by Taylor & Francis Group, LLC
74
strengths then decrease to reach progressively the “normal” linearly increasing profile typical of NC clays. The seabed zone with a “higher than expected” undrained shear strength is often called “crust”. This term will be used hereafter, but it is emphasised that the consistency of the sediment in this so-called “crust” layer remains very soft to soft, with Su in the range of 7–15 kPa. This phenomenon can be very pronounced in the Gulf of Guinea, with a clear peak in shear strength between 0.3 m and 1.2 m below seabed and a sharp decrease below that depth to reach the “normal” linear profile around 2 m of penetration (Fig. 33). It has been suggested that this peak might be an artifact due to the weight of seabed-based modules influencing the actual in-situ shear strength, but similar “crust” layers have also been highlighted by lightweight CPT modules (Peuchen 2000). Furthermore, there is now full evidence of their existence, e.g. from direct shear strength measurements in box corers (Fig. 1). Le (2008) has shown that the peak zone was associated with local decreases in water contents. 4.10.2 Origin of the crust Crusts have been found in the seabed sediments of various oceans, without any apparent correlation with physical parameters such as water depth or seabed temperature. At a particular field or site, a crust can be locally present or absent. No evidence of crustal zones has been found so far at depths in excess of 2–3 m, suggesting that either strengthening mechanisms are recent (Holocene period) or that the crust vanishes with time or sedimentation process. It is known from geological evidence that these sediments have never seen overburden stresses in excess of the present state, meaning that the relatively high shear strengths of the crust are not the result of mechanical over-consolidation. The apparent over-consolidation ratio (AOCR) is typically of the order of 3 to 5 within the crust zone. From a geotechnical point of view, it is certainly preferable to use the term yield stress ratio (YSR) which does not refer to the past effective vertical stress (section 4.1). Based on a study done by Brausse (2001), three families of mechanisms might be considered as possible drivers of apparent over-consolidation in marine sediments: (a) strong interparticle physiochemical bonding and ion exchange which may be combined with effects of cementation and cause higher attraction between particles (through ionic and van der Waals forces), (b) inorganic cementation involving cementing agents such as silicates, carbonates, iron oxides, or alumina, and (c) bioturbation and organic cementation. No unique or convincing explanation has been proposed so far. Physico-chemical effects are favoured by Sultan et al. (2001) who propose an interpretation of the apparent over-consolidation using a microscopic model based on the theory of the diffuse double layer (Gouy 1910, Chapman 1913) which relates the mechanical behaviour of the soil to the variations in ionic concentration. More recently, Ehlers et al. (2005)
© 2011 by Taylor & Francis Group, LLC
Figure 33. CPT profiles in seabed “crust” measured at various deepwater sites offshore Angola and Nigeria compared to site F with no “crust”.
have presented X-ray radiographs of surficial cores from offshore Nigeria showing a correlation between the most intensively burrowed sections and the highest soil shear strength, thus favouring the role of biological activity (bioturbation). Research work on this fascinating issue is ongoing at Cambridge University under the direction of Prof. M. Bolton. Kuo & Bolton (2009) and Kuo et al. (2010) consider the presence of faecal pellets in the studied samples as the cause of higher shear strengths. However crusts have been observed on other sites without such presence of pellets in the samples. Obviously further effort is needed to understand the origin of the crust, the conditions for its development and the reasons for its demise. The low rate of sedimentation prevailing in the Gulf of Guinea may play a particular role. 5
DESIGN ISSUES RELATED TO SOIL-PIPELINE INTERACTIONS
The purpose of this section is to describe the geotechnical considerations related to the design of flowlines (transporting an unprocessed product), pipelines (generally for a single phase fluid after processing) and risers. For the purpose of this section, the generic term “pipelines” is used. Pipelines in deep offshore areas are generally laid on the seabed under their own weight, without burial in a trench. They may experience lateral buckling and axial “walking” due to expansion caused by their operating temperatures and pressures. Analysis
75
Figure 35. Principle of the NGI model test rig.
Figure 34. Model test rig used for 1-g physical modeling tests carried out at NGI (from Dendani & Jaeck 2007).
Significant advances in the understanding of soilpipeline interactions in soft soils have come from extensive physical modelling performed in the framework of R&D projects (e.g. the SAFEBUCK JIP, Bruton et al. 2006) or sponsored by oil companies (e.g. Dendani & Jaeck 2007). Figures 34 and 35 show an example of model test rig used for 1-g soilpipeline interaction tests in soft deepwater clays from the Gulf of Guinea. Dynamic soil-pipeline stiffness tests were carried at 3S-R, Grenoble on mixtures of kaolin and bentonite aimed at approaching the mineralogical properties of GoG clays (Orozco et al. 2007). A limitation of these experiments is the incapacity to properly reproduce the natural soil structure (see section 4). It is recognised that:
of pipe-soil interaction is required to assess these phenomena. As mentioned by e.g. Cathie et al. (2005) and White & Cathie (2010), valuable experiences and progress have been gained by the geotechnical and pipeline engineering and construction communities toward a better understanding and modelling of the soil-pipeline interactions. These experiences are related to the following aspects: – site characterisation and geotechnical investigation (see section 2.3); – modelling of soil-pipeline interactions; – analysis of pipeline inspection data to improve and calibrate the interaction models. 5.1
• Initial pipeline penetration largely conditions the
axial and lateral responses and should be accurately predicted. This penetration is not only driven by the submerged weight of the pipeline but also by installation conditions (i.e. vertical overstressing, dynamic motions). • Pipeline “walking” is controlled by the longitudinal soil-pipeline friction; axial transfer curves should account for both peak shear strength and residual shear strength at large displacements. • The lateral pipeline behaviour is complex, and the transfer curves should capture first break-out resistance, post-peak resistance, additional resistance due to berm formation, and subsequent evolutions during loading/unloading cycles.
Modelling soil-pipeline interactions
Pipelines in deep water are submitted to production (start-up and shut-down) cycles. Internal variations in temperature and pressure cause successive expansions and contractions of the line which induce both axial and lateral displacements. Large cumulated displacements can be experienced, which are commonly designated by “walking” (in longitudinal direction) or “snaking” (in lateral direction). These movements generate stresses in the pipeline itself, at the connection spools and on subsea structures. This may result in fatigue damage or lateral buckling. Extensive structural analyses are performed, using both analytical and finite element analyses, in order to demonstrate that pipeline movements cannot lead to excessive stresses or that mitigation procedures are required to control the displacements. Interaction conditions between pipeline and seabed are fundamental for the integrity evaluation of the pipelines. It is now well established that simple friction coefficients are not sufficient to capture the complexity of the longitudinal and lateral behaviour of pipelines on soft soils. Interaction curves are generally incorporated into the structural models, using an approach similar to the t-z and p-y load transfer methods used for analysing the response of foundation piles. © 2011 by Taylor & Francis Group, LLC
5.2
Pipeline penetration
The assessment of pipeline embedment is fundamental for the evaluation of the axial and lateral resistances. The initial pipeline embedment depends directly on the shear strength profile in the upper half-metre of soil, and the presence of a seabed crust (see section 4.9) has a significant impact on the resulting penetration. The consequences regarding pipeline stability can be high, and heavy expenditures might be required if artificial stabilisation is needed. The embedment under self-weight may be estimated from conventional bearing capacity equations modified for the curved shape of a pipeline
76
Figure 36. Axial soil resistance model (from Dendani & Jaeck 2007). Figure 37. Lateral soil resistance model (from Dendani & Jaeck 2007).
(Randolph & White 2008, Bruton et al. 2009). However, field observations show that the as-laid pipeline embedment is generally greater than the penetration calculated from the static pipeline submerged weight alone. During laying operations, the pipeline is submitted to dynamic motions in the touch-down zone, resulting in the force applied by the pipeline being much greater than its submerged weight. A force amplification factor Ka for dynamic motion shall then be considered to reproduce the observed penetration. For a 12 pipeline on soft clay, Cathie et al. (2005) suggest a load concentration factor equal to 2.0. For the soft deepwater Gulf of Guinea sediments, a back analysis of the observed penetrations suggests an amplification factor Ka in the range of 1.5 to 2.0. Alternatively, the pipeline penetration can be assessed from model in-situ pipe tests and measurements with the SMARTPIPE® and SMARTSURF™ tools (see section 2.3).
against model tests. The break-out resistance can be divided into a frictional component, and a passive resistance component depending on pipeline embedment and undrained shear strength:
where W = submerged pipeline weight (kN/m); Clat = lateral bearing factor; Suav = average undrained shear strength (kPa) over depth z; and z = pipeline bottom penetration (m). Dendani & Jaeck (2007) reported that Clat = 2.3 gives a fairly good match with pipeline model test results performed on Gulf of Guinea clays. The lateral behaviour of a partially penetrated pipeline is usually brittle and involves high break-out resistance followed by lower residual resistance during sweeping. During each lateral sweep, a soil berm develops ahead of the pipeline and causes a rise in lateral resistance. A schematic lateral transfer curve used for pipeline design in deepwater Gulf of Guinea soils is given in Figure 37.
5.3 Axial resistance The axial (longitudinal) resistance for a pipeline on soft clay may be estimated using an approach similar to the total stress method (α method) used for pile capacity:
5.5
Pipeline inspection data in deep offshore Gulf of Guinea obtained in several fields operated by Total show the following main results:
where α = adhesion factor; L = arc length of pipeline embedded in soil; and su = intact undrained shear strength. The intact value of undrained shear strength should be used because set-up will take place between pipeline installation and operation. Dendani & Jaeck (2007) reported that the axial resistances recorded during the model tests are best predicted by taking α = 0.7 and α = 0.35 for the peak resistance and residual resistance, respectively. An example of axial resistance curve is given in Figure 36.
• Pipeline penetration is generally in the range 30–60
% of the pipeline diameter; • Soil berms are identified (see Fig. 38); • No excessive lateral or axial movements are
observed, suggesting that the soil-pipeline interaction models used for the design are generally acceptable. 6
5.4 Lateral resistance
DESIGN ISSUES FROM INSTALLATION EXPERIENCES
The Nkossa production barge, in a water depth of 200 m offshore Congo, marked the first step of deepwater developments in the Gulf of Guinea. It was also
The break-out lateral (horizontal) resistance is conventionally assessed using empirical methods calibrated
© 2011 by Taylor & Francis Group, LLC
Inspection data
77
Table 2. Characteristics of foundations for typical Gulf of Guinea deepwater structures. Water depth (m)
Figure 38. Pipeline on seabed from a video survey in deep offshore Gulf of Guinea.
the first production FPU anchored with suction piles in soft deepwater sediments (Colliat et al. 1997). Similar to other major offshore oil provinces, deepwater Gulf of Guinea oilfields have subsequently been developed rapidly, and currently cover water depths between 400 m and about 2000 m, in particular offshore Angola and Nigeria. The main types of foundations are suction piles and driven piles, applied for the full spectrum of production structures, i.e. FPU, FPSO, OLT, TLP and CT. Because of relatively mild environmental conditions, the Gulf of Guinea has also seen the installation of two compliant towers in Cabinda (the CT concept originally described by Des Deserts in 1992 and Reusswig & Nair in 1994 was used in Cabinda in 2005, see Will et al. 2006), and the very first FPDSO for the Azurite field in 1400 m of water offshore Congo. Within existing deepwater production structures, the deep draught SPAR and semi-submersible floaters are the only two which have not yet been utilised in the Gulf of Guinea. Table 2 summarises the main characteristics of the foundations and anchors that are used for illustration in this section. All case studies given in Table 2 are from offshore Angola (or the Cabinda enclave for BenguelaBelize), with the exception of the Akpo field, located offshore Nigeria. 6.1
1300
Girassol RTA
1300
Rosa RTA
1350
Dalia FPSO
1300
Akpo FPSO
1300
Kizomba A TLP
1200
Benguela-Belize compliant tower Kizomba MODU moorings Akpo standby moorings
400
∗
1200 1300
Suction piles 4.5 m OD Gravity piles 8.0 m OD Gravity pile 8.0 m OD Suction piles 4.9 m OD Suction piles 5.0 m OD Driven piles 2.14 m (84 ) Driven piles 2.74 m (108 ) SEPLA 28 m2 plate VLA 15 m2 plate
Penetr. Depth (m) 16.5 19.5 23 20.5 24 123 150 25∗ 30
Installed with a suction follower pile
design principles for calculating the friction resistance along the outside and inside walls of suction piles in different types of cohesive sediments, where specific requirements for high plasticity clays apply to deepwater West Africa sediments. The environmental conditions in the Gulf of Guinea also have some design implications, some of which are common to other deepwater areas and others being more specific, in particular the cyclic undrained shear strength of the clay and the reverse end bearing capacity of suction piles in relation to the design mooring load case. 6.1.1 Cyclic undrained shear strength No systematic study of the cyclic response of Gulf of Guinea clays has been performed so far. However a compilation of existing cyclic DSS or triaxial test data allows “average” contour diagrams to be proposed. An example is given for DSS loading in Figure 39, where the response of the Drammen clay is also shown for comparison. Gulf of Guinea clays are characterised by a strong rate effect and a slow degradation process. This good resistance to cyclic loading combined with the West Africa storm composition, which for anchoring systems involves a significant static loading component and relatively low peak dynamic loads, explains why the undrained cyclic shear strength is typically found about 5–10% larger than the static reference strength. Hence, designs are de facto based on static shear strengths.
Suction piles in taut leg moorings
Subsea field development with oil storage in an FPSO on site is the preferred solution offshore West Africa. As per 2010, a total of 20 FPSO or FPU vessels and OLT buoys are operated in the Gulf of Guinea, representing the installation of about 500 suction piles in less than 10 years. To the Authors’ knowledge, all existing FPSO vessels and OLT buoys, in water depths between 200 m and 1500 m, are anchored with suction piles in semi-taut to taut leg configuration, i.e. with the mooring line making an angle of 20◦ to 45◦ to the horizontal or seafloor and a significant vertical loading component at the anchor. Andersen & Jostad (2002 & 2004) and Andersen et al. (2005) have described the
6.1.2 Reverse end bearing capacity In TLMs with a significant vertical loading component at the anchor point, reverse end bearing (REB) is an important contribution to the ultimate capacity of a
78 © 2011 by Taylor & Francis Group, LLC
Girassol FPSO
Foundation type & size
Figure 39. Normalised diagram showing number of cycles to failure for an “average” Gulf of Guinea clay and the Drammen clay (DSS mode, OCR = 1).
suction pile in clay. In early cases (e.g. the Nkossa FPU or the Girassol FPSO, Colliat et al. 1997 & 2007), REB was not considered in the pile capacity. With the increased industry confidence and perceived reliability, REB is now taken into account in full (as is usually the case in the Gulf of Mexico) or a proportion of it only (typically about 30% in the Gulf of Guinea, on the basis of the ratio of the double amplitude cyclic load to the total mooring load, as introduced by Bureau Veritas). A regional issue that has driven a certifying authority to conservatively limit REB capacity is the lack of long term monitoring of sea states and storm events, and the difference in dynamic loads between 100-year storm and squall conditions. Specifically, the design load case for the mooring of an FPSO vessel offshore Nigeria will be governed by squall wind loading and it is considered that the peak squall load will last longer than a peak storm wave load. However, both rapid and slow pull-out load tests of model piles in the centrifuge do show that high excess pore pressures develop at the pile tip, corresponding to a significant REB capacity (Colliat et al. 2010).
The same result is obtained for the FPSO piles of the D and K sites, where the predicted self-weight penetration ranged from 8.3–11.5 m to 9.6–13.5 m when the observed values were respectively equal to 12–13 m and 15–16 m (Fig. 40). From these larger self-weight penetrations, a sensitivity of 4 to 5 is back-calculated, whereas laboratory and in-situ test results (VSTs with only 2 revolutions) available at time of design suggested a sensitivity ranging between 2.5 and 4. An even lower sensitivity of about 2 would have been obtained had the remoulded shear strength been taken directly equal to the CPT sleeve friction (see discussion in section 4.4). Therefore, it is recommended to put more emphasis on tests that better mimic large strains at the soil-pile interface, such as cyclic T-bar or ball penetration tests with sufficient number of cycles (N > 25).
6.1.3 Sensitivity estimated from self-weight penetrations During penetration, either by self-weight or when suction is applied, the friction resistance along the outside and inside pile wall is assumed to be governed by the remoulded clay strength at the soil-pile interface, i.e. by the soil sensitivity. From the back-analysis of the self-weight penetration of a number of suction piles at various deepwater West Africa sites, Colliard & Wallerand (2008) suggest a clay sensitivity larger than expected.
6.1.4 Installation behaviour Figure 40 summarises the installation behaviour of three sets of FPSO piles, with a self-weight penetration equal to 50–67% of the final embedment depth, which is typical of suction piles in soft deepwater sediments. At final penetration, the suction ranges from 70–95 kPa (at 16.5 m for site S) to 100–135 kPa (at 20.5 m and 24.0 m for site D and site K, respectively). Required suction reduces with increased pile diameter (for a given penetration) since it acts over an increased internal pile cross section. With the suction piles from
Figure 40. Typical suction versus depth curves for installation of FPSO suction piles.
79 © 2011 by Taylor & Francis Group, LLC
site D and site K having similar diameter and submerged weight, together with similar undrained shear strength profiles at both sites, the lower suction measured for the K piles suggests a higher sensitivity of the deepwater clays from this site. The suction curves given in Figure 40 are typical of the penetration resistance of suction piles with a diameter in the order of 5 m and a height-to-diameter ratio of about 4 to 5 (see Table 2). Colliat (2006) showed that similar installation behaviour is obtained for suction piles in the Gulf of Mexico when the gradient in undrained shear strength is similar (i.e. of about 1.0 to 1.5 kPa/m). However, it should be noted that the suction curves shown in Figure 40 were not always measured directly inside the pile and have not been corrected for venturi effects or hydraulic losses in the pumping system; this may be the reason for the step increase of suction immediately after self-weight penetration for the D and K piles. 6.1.5 Effect of ring stiffeners The lack of heavy lift vessels in the Gulf of Guinea initially promoted fabrication solutions in favour of reducing the suction pile weight by using internal ring stiffeners. This was the case for the Nkossa FPU or the Girassol FPSO piles (Colliat et al. 1997 & 2007). When such ring stiffeners are relatively wide and close to the pile tip, it is now well known that they have a direct influence on the soil-pile interface friction. The inside friction is reduced, in relation to an increased remoulding effect and down-dragging of a mixture of seafloor soil and water, whereas the outside friction may be increased by displacing a larger volume of the inside soil plug out of the pile during penetration. For the latter effect, Andersen & Jostad (2004) describe how the normal stresses along the external pile wall are initially taken by the pore pressures, then give increased effective stresses and friction resistance as pore pressures dissipate. Pull-out tests of model piles with and without stiffeners in the centrifuge show that the net pull-out capacity of stiffened piles is 10–35% higher than for piles without ring stiffeners (see Fig. 43 and Colliat et al. 2010).
Figure 41. Examples of installation and retrieval behaviour.
Gulf of Guinea sites is described in Colliat & Colliard (2010), with set-up periods ranging between 1 day and 3½ years. The results obtained suggest a rapid increase in friction resistance, from 35–45% in one week to about 70% in one month, but little gain between one month and 3½ years. This is in relatively good agreement with (a) the results of thixotropy tests in high plasticity clays (Andersen & Jostad 2002, Dendani 2003, Fig. 28), and (b) similar data obtained at deepwater Gulf of Mexico sites (Jeanjean 2006). The interpreted interface friction factor ranges from 0.25 to 0.35 at time of installation, and increases to 0.40–0.56 in the longer term. Despite some inevitable uncertainties related to too scarce a database and other operational issues (in particular, when the measurements of the retrieval over-pressure and submerged pile weight are inaccurate), obtaining a long term friction factor below the generally accepted design value of 0.65 for high plasticity clays is a concern (where the friction factor is equal to 1.95/St with St = 3, from Andersen & Jostad 2002). A thorough confirmation or verification would require a larger set-up database, or (preferably) the performance of specific field tests.
6.1.6 Set-up behaviour Suction piles might have to be retrieved when they are installed out of tolerances (generally within one or two days after installation), or in the event of other operational constraints such as the replacement of a mooring line (i.e. several months or years after installation). In both cases, retrieval of the suction pile by over-pressurising allows a direct measurement of the increase in friction resistance with consolidation time (set-up). One example of installation and retrieval behaviour of two FPSO piles is given in Figure 41, showing an installation suction of about 90 kPa at 16.5 m depth, and an over-pressure of 180 kPa and 165 kPa at time of retrieval one and two days later, respectively. Although limited to a database of six retrieval cases, the set-up behaviour of suction piles at three different
6.2
The case of the Girassol RTA piles experiencing an abnormally low friction resistance due to the painted outside wall (contractor’s choice for protection against corrosion) has been published in detail (Dendani & Colliat 2002, Dendani 2003, Colliat et al. 2007) and is not repeated here. But, as a direct detrimental
80 © 2011 by Taylor & Francis Group, LLC
Suction piles for riser towers
Figure 42. Installation predictions and results for large RTA piles (from Colliat et al. 2007).
consequence, the design of following RTA piles was more conservative and did not rely on passive suction. For the Girassol RTA piles, the design shear strength (determined from DSS tests) was multiplied by 0.70 for the effect of constant tension loading on the undrained clay strength (thus increasing the material factor from 1.30 to 1.86), and passive suction (REB) was considered only for the short term loading case. After installation, the reduced friction capacity of the painted piles was balanced by doubling the RTA submerged weight with ballast, thus making it a kind of “gravity friction pile”. On the other hand, the RTA pile for site R was originally designed as a gravity friction pile, with the submerged weight of the RTA equal to 1.25 times the service tension load case, in order to have the RTA working under compression loading. A similar type of foundation with counter-weights is also mentioned by Zimmermann et al. (2009) for the Greater Plutonio riser tower in Angola. Figure 42 shows a comparison of the installation results and predictions for both the site S and site R RTA piles. Being the first suction piles ever installed in soft deepwater Gulf of Guinea sediments, a rather conservative prediction was done for the site S RTA piles, but the much lower penetration resistance is mainly related to the paint effect (Colliat et al. 2007). The good installation prediction for the site R RTA pile confirms that the design model is correct. However, more field experience will be required before the industry regains confidence in REB capacity for RTA piles under tension service loading.
Figure 43. Centrifuge pull-out testing of suction piles with and without stiffeners (from Colliat et al. 2010).
This RTA experience was also studied by model pull-out tests in the centrifuge, specifically performed with similar deepwater clay samples reconstituted in the centrifuge, and with emphasis on the effect of ring stiffeners on both penetration and pull-out resistance (Thorel et al. 2010). It was observed that significant passive suction is mobilised in rapid pull-out loading of both the stiffened (S) and un-stiffened (U) piles, although it is necessary to displace the pile in excess of one metre to develop 100% of the REB capacity (Fig. 43). Similar results were obtained by Raines et al. (2005) with centrifuge tests carried out with Speswhite clay samples. 6.3
With a penetration of 150 m, the foundation piles of the Benguela-Belize compliant tower (together with those of the Tombua-Landana CT in the same area) are within the longest piles ever driven in soft deepwater sediments. The 2.74 m (108 ) diameter CT piles were driven with a Menck MHU 2100 hammer, and relatively low driving blow-counts (i.e. 22–24 blows/25 cm at final depth) are reported by Will et al. (2006). On another hand, with a penetration of 123 m, the 2.14 m (84 ) diameter piles of the Kizomba A TLP are within the range of pile diameters and penetrations for equivalent Gulf of Mexico TLPs (Doyle 1999).
81 © 2011 by Taylor & Francis Group, LLC
Driven piles
Despite a slight difference in pile diameter (i.e. 96 for Kizomba, versus 72 –96 in the Gulf of Mexico), a fair comparison of the Kizomba pile driving behaviour is possible since the same type of hydraulic underwater hammer (Menck MHU 500T) is used in the Gulf of Mexico. For the Kizomba TLP piles, Labbe & Perinet (2004) report a lower SRD than expected and lower driving blow-counts at final depth (i.e. 28 blows/25 cm instead of a predicted value of 45 blows/25 cm). In the Authors’ opinion, this lower than expected SRD is due to the use of the Stevens et al. (1982) method, which assumes implicitly a clay sensitivity of 2 and will then provide conservative drivability predictions in soft clays with a larger sensitivity. For the Kizomba TLP piles, a better drivability prediction would have been obtained by using an alternative SRD calculation method with a friction degradation model (e.g. Puech et al. 1990, Colliat et al. 1993, or Dutt et al. 1995). The largest difference between the Kizomba and Gulf of Mexico sites is in the self-weight penetrations of the TLP piles, with 51–55 m for Kizomba (for a prediction ranging between 34 m and 52 m), as compared to a typical range of 30–40 m for similar pile sizes in the Gulf of Mexico (Doyle 1999). At Kizomba, Labbe & Perinet (2004) also report the performance of a re-drive test, showing a rapid gain of pile shaft friction resistance with time. Interestingly, similar rapid set-up of large diameter driven piles is described by Dutt & Ehlers (2009), also comparing two case studies from two deepwater sites in West Africa and in the Gulf of Mexico. These pile driving results from various Gulf of Guinea sites suggest a sensitivity of the clay higher than considered in standard driveability prediction methods and a rapid regain of resistance with time. This is in good agreement with the installation behaviour of suction piles described above.
deepwater clay, the required UHC is typically in the order of 4–5 MN for temporary MODU moorings, as compared to about 8–9 MN for semi-permanent or permanent moorings (Colliat 2006). In a recent application for three stand-by moorings for supply boats, 15 m2 Stevmanta anchors were installed in 1300 m of water offshore Nigeria, reaching an embedment depth of 28–32 m. Published Stevmanta or Dennla application cases (Ruinen & Degenkamp 2001, Murff et al. 2005, Magne 2008) mention embedment depths of 15–25 m for 10–15 m2 VLA anchors (with the smallest values of penetration and plate area corresponding to VLAs installed into stiffer clays offshore Brazil). In comparison, the result obtained with this Nigeria example suggests a lower penetration resistance in the Gulf of Guinea sediments, in a similar way to suction piles or driven piles. 7
Originally considered as an area with relatively gentle soil conditions, the deep offshore Gulf of Guinea has proven to be unique in certain aspects, highlighted by a number of differences with the Gulf of Mexico characteristics. After more than 10 years of investigations in the Gulf of Guinea, a large database on the behaviour of these deepwater sediments is now available, and has been tentatively summarised. A soil investigation strategy adapted to the regional requirements of this remote area is proposed, with emphasis on the importance of in-situ testing. Geotechnical characterisation is based on the use of conventional CPTs and VSTs, with the need for performing cyclic full-flow penetrometer tests (T-bar or ball tests) with at least 25–30 cycles for the determination of the soil sensitivity. Specific laboratory testing procedures were developed to address some key physical and geotechnical properties of these unusually high plastic clays, e.g. for the measurement of organic content and the determination of fine particle sizes. Since the Gulf of Guinea deepwater sediments have a high degree of structure, triaxial testing of high quality samples under in-situ stresses should be preferred to the SHANSEP test procedure usually applied for Gulf of Mexico clays. This high degree of structure is also the cause for measuring an apparent overconsolidation ratio of about 1.5–2 in the top 10–20 m of sediments, but the Gulf of Guinea deepwater sediments are generally normally consolidated clays (away from topographic highs created by diapirism). Two important properties of the Gulf of Guinea deepwater sediments are confirmed by the results of installation experiences, i.e. a relatively high sensitivity (St = 4 to 6), giving a lower than expected resistance to penetration of suction piles or driven piles from friction degradation, and a rapid and important regain in friction resistance with time from thixotropy and set-up effects. Some geotechnical issues still require further studies, in particular: (a) origin and characterisation of the near seabed “crust”, (b) interface soil-pipeline friction
6.4 Vertically loaded plate anchors In 1995, conventional drag anchors were used for the moorings of the Nkossa oil FSO in a water depth of 125 m offshore Congo (Colliat et al. 1997). To the Authors’knowledge, the first application of VLA plate anchors in TLM moorings in the Gulf of Guinea was for the semi permanent mooring systems of MODUs, for the drilling of production wells at the Kizomba field offshore Angola. SEPLA anchors were installed by means of a suction follower, and the same type of SEPLA anchors has then been used at the TombuaLandana field, offshore Cabinda, for the same kind of application. On another hand, both the Vryhof Stevmanta and Bruce Dennla VLAs have been used in a number of pre-installed temporary moorings for exploration drillling MODUs. For permanent moorings, VLA plates of about 25–30 m2 in surface area are generally used, whereas smaller anchors, with areas of the order of 12–15 m2 , are used in the latter case of temporary moorings. The plate area is directly related to the required anchor UHC (Murff et al. 2005). For an estimated embedment depth of about 25 m in soft
82 © 2011 by Taylor & Francis Group, LLC
CONCLUSIONS
Prof. P. Delage (ENPC-CERMES), Prof. F. Thomas (INPL/LEM), N. Sultan and J. Meunier (Ifremer) and H. Manh Le (Fugro).
resistance, (c) effect of gas and gas hydrates in the sediments, and (d) definition of a single constitutive model to reproduce the effects of stress history, structure and strain rate. These issues are targeted by ongoing R&D studies.
REFERENCES 8 ABBREVIATIONS AOCR BPT CT CPT DSS FPDSO FPSO FPU FSO ICL MIP MODU NC OCR OLT REB RTA SCL SCC SEM SEPLA SHANSEP SRD TOC TPT TLM TLP UHC VLA VST YSR
Alhayari, S., Chen, J.H. & Maerten, B. 2000. Kuito project: Successful suction pile design and installation based on innovative soil investigation. Proc. Offshore West Africa Conference, OWA, Abidjan. Andersen, K.H. & Jostad, H.P. 2002. Shear strength along outside wall of suction anchors in clay after installation. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Kyushu. Andersen, K.H. & Jostad, H.P. 2004. Shear strength along inside of suction anchor skirt wall in clay. Proc. Offshore Technology Conference, OTC paper 16844, Houston. Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E.C., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P. & Supachawarote, C. 2005. Suction anchors for deepwater applications. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet, B.A. & Ho, E.W.L. 2005. Modelling the effects of structure in deep-ocean sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet B.A. & Stallebrass S.E. (2004). A constitutive model for structured clays. Géotechnique 54 (4), 269–278. Borel, D., Puech,A., Dendani, H. & De Ruijter, M. 2002. High quality sampling for deep water geotechnical engineering: the STACOR experience. Proc. Int. Conference on Ultra Deep Engineering Technology, UDET, Brest. Borel, D., Puech, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Borel, D., Puech, A. & Po, S. 2010. A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Bourillet, J.F., Damy, G., Dussud, D., Sultan, N. & Woerther, P. 2007. Behaviour of a piston corer from accelerometers and new insights on quality of the recovery. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Broadbent, F.E. 1953. The soil organic fraction. Adv. Argon, 5, 153–183. Brausse, M. 2001. Underconsolidation and apparent overconsolidation in marine sediments: a study of stress states in the northwestern Gulf of Mexico. Civil Engineering MsC Thesis, University of Rhode Island. Bruton, D., White, D., Cheuk, C., Bolton, M. & Carr, M. 2006. Pipe-soil interaction behavior during lateral buckling, including large amplitude cyclic displacement test by the Safebuck JIP. Proc. Offshore Technology Conference, OTC paper 17944, Houston. Bruton, D., White, D.J., Langford, T.L. & Hill, A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments. Proc. Offshore Technology Conference, OTC paper 20096, Houston. Buckley, D.E., MacKinnon, W.G., Cranston, R.E. & Christian, H.A. 1994. Problems with piston core sampling: Mechanical and geomechanical diagnosis. Marine Geology, 117, 95–106. Burland, J.B. 1990. On the compressibility and shear strength of natural clays. Géotechnique, 40 (3), 329–378.
Apparent Over-Consolidation Ratio Ball Penetration Test Compliant Tower Cone Penetration Test Direct Simple Shear Floating Production Storage Drilling & Off-loading Floating Production Storage & Off-loading Floating Production Unit Floating Storage & Off-loading Intrinsic Compression Line Mercury Intrusion Porosimetry Mobile Offshore Drilling Unit Normally Consolidated Over-Consolidation Ratio Oil off-Loading Terminal Reverse End Bearing Riser Tower Anchor Sedimentation Compression Line Sedimentation Compression Curve Scanning Electron Microscope Suction Embedded Plate Anchor Stress History And Normalized Soil Engineering Properties Soil Resistance to Driving Total Organic Carbon T-bar Penetration Test Taut Leg Mooring Tension Leg Platform Ultimate Holding Capacity Vertically Loaded Anchor Vane Shear Test Yield Stress Ratio
ACKNOWLEDGEMENTS The authors thank Total EP, Fugro France and IFP for the permission to publish this paper. Parts of the work published come from various R&D programmes conducted in the framework of CLAROM (CLub pour les Actions de Recherche sur les Ouvrages en Mer) with financial support from the private programme for the oil & gas E&P industry CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). Projects were carried out with the joint participation of Acergy, Doris Engineering, Ifremer, Saipem, Technip and Total, and in collaboration with several University laboratories. The contribution of colleagues from these organizations is acknowledged. The Authors are particularly indebted to V. De Gennaro and
83 © 2011 by Taylor & Francis Group, LLC
Cathie, D.N., Jaeck, C., Ballard, J-C. & Wintgens, J-F. 2005. Pipeline geotechnics – state-of-the-art. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Chapman, D.L. 1913. A contribution to the theory of electrocapillarity. Philosophical Magazine, 25 (6), 475–481. Cochonat, P., Bourillet, J.F., Quemeneur, P. & Colliat, J.L. 1996. Deepwater sedimentary processes in the Gulf of Guinea. Proc. Int. Conference on Deepwater Site Investigation, SUT, London. Colliard, D. & Wallerand, R. 2008. Design and installation of suction piles in West Africa deepwaters. Proc. 2nd BGA Int. Conference on Foundations, ICOF 2008. Colliat, J.L. 2006. Evaluation of suction piles and plate anchors from current deepwater mooring applications. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, San Francisco. Colliat, J.L., Boisard, P., Gramet, J.C. & Sparrevik, P. 1997. Geotechnical design and installation behaviour of suction anchor piles – Comparison with drag anchors. Proc. Int. Conference on Behaviour of Off-Shore Structures, Delft. Colliat, J.L. & Colliard, D. 2010. Set-up of suction piles in deepwater Gulf of Guinea clays. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L., Dendani, H., Jostad, H.P., Andersen, K.H., Thorel, L., Garnier, J. & Rault, G. 2010. Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L., Dendani, H. & Schroeder, K. 2007. Installation of suction piles at deepwater sites in Angola. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Colliat, J.L., Vergobbi, P. & Puech, A. 1993. Friction degradation and set-up effects in hard clays offshore Congo and Angola. Proc. Offshore Technology Conference, OTC paper 7192, Houston. Cotecchia, F. & Chandler, R.J. 2000. A general framework for the mechanical behaviour of clays. Géotechnique, 50 (4), 431–447. De Gennaro, V. & Pereira, J-M. 2010. A viscoplastic constitutive model for unsaturated geomaterials. Computers and Geotechnics (under review). De Gennaro, V., Puech, A. & Delage, P. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Delage, P. & Lefebvre, G. (1984). Study of the structure of a sensitive Champlain clay and of its evolution during consolidation. Canadian Geotechnical Journal, 21, 21–35. Delage, P., Marcial, D., Cui, Y.J. & Ruiz, X. (2006). Ageing effects in a compacted bentonite: a microstructure approach. Géotechnique, 56 (5), 291–304. Dendani, H. 2003. Suction anchors: Some critical aspects for their design and installation in clayey soils. Proc. Offshore Technology Conference, OTC paper 15376, Houston. Dendani, H. & Colliat, J.L. 2002. Girassol: Design analysis and installation of the suction anchors. Proc. Offshore Technology Conference, OTC paper 14209, Houston. Dendani, H. & Jaeck, C. 2007. Pipe-soil interaction in highly plastic clays. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Des Deserts, L. 1992. Architecture and application of various configurations of compliant towers. Proc. ASME Int. Conference on Offshore Mechanics and Arctic Engineering, OMAE, Houston.
Diaz-Rodriguez, J.A., Leroueil, S. & Aleman, J.D. 1992. Yielding of Mexico city clay and other natural clays. Journal of Geotechnical Engineering, ASCE, 118 (7), 981–995. Doyle, E.H. 1999. Pile installation performance for four TLP’s in the Gulf of Mexico. Proc. Offshore Technology Conference, OTC paper 10826, Houston. Dutt, R.N., Doyle, E.H., Collins, J.T. & Ganguly, P. 1995. A simple model to predict soil resistance to driving for long piles in deepwater normally consolidated clays. Proc. Offshore Technology Conference, OTC paper 7668, Houston. Dutt, R.N. & Ehlers, C.J. 2009. Set-up of large diameter driven pipe piles in deepwater normally consolidated high plasticity clays. Proc. ASME Int. Conference on Ocean, Offshore and Arctic Engineering, OMAE2009-79012, Honolulu. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origin of near-seafloor “crust zones” in deepwater. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Enjaume, P., Hamon, M., Epalanga, K. & Mackenzie, B. 2010. Angolan deepwater soil conditions: GIS technology development for sediment characterization. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Evans,T.G., Feyereisen, S. & Rheaume, G. 2002.Axial capacities of jetted well conductors in Angola. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Evans, T.G. 2010. A systematic approach to offshore engineering for multiple-project developments in geohazardous areas. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Gouy, G. 1910. Sur la constitution de la charge électrique à la surface d’un électrolyte. Journal de Physique Théorique et Appliquée, 9, 457–468. Grozic, J.L.H., Nadim, F. & Kvalstad, T.J. 2005. On the undrained shear strength of gassy clays. Computers and Geotechnics, 32 (7), 483–490. Hill, A.J., Southgate, J.G., Fish, P.R. & Thomas, S. 2010. Deepwater Angola: Geohazard mitigation. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Hill, A.J. & Wintgens, J-F. 2009. In-situ measurement of pipesoil interaction in deep water – Results of a successful offshore campaign. Proc. SUT Annual Conference, Perth. Holtvoeth, J. 2004. Terrigenous organic matter in sediments of the eastern equatorial Atlantic – Distribution, reactivity and relation to Late Quartenary climate. PhD thesis, Bremen University. Jeanjean, P. 2006. Set-up characteristics of suction anchors for soft Gulf of Mexico clays: Experience from field installation and retrieval. Proc. Offshore Technology Conference, OTC paper 18005, Houston. Kelleher, P. & Hull, T 2008. Quality assessment of marine sediments recovered with a hydraulically tethered piston corer. Proc. Offshore Technology Conference, OTC paper 19687, Houston. Kuo, M.Y-H. & Bolton, M.D. 2009. Soil characterization of deep water African clays: Is biology a source of mechanical strength?. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Osaka. Kuo, M.Y-H., Bolton, M.D., Hill, A.J. & Rattley, M. 2010. New evidence for the origin and behaviour of deep ocean ‘crusts’. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth.
84 © 2011 by Taylor & Francis Group, LLC
Labbe, J.P. & Perinet, D. 2004. The deepwater installation of the Kizomba-A SWHP anchor piles and related constraints for ultra deepwater lifting. Proc. Int. Conference on Deep Offshore Technology, DOT, New Orleans. Ladd, C.C. & Foott., R.1974. New design procedure for stability of soft clays. Proc. Journal of the Geotechnical Engineering Division, ASCE, 100 (7), 767–787. Le, M.H. 2008. Caractérisation physique et mécanique des sols marins d’offshore profond. PhD thesis, ENPCCERMES. Le, M.H., Nauroy, J.F., De Gennaro, V., Delage, P., Thanh, N., Flavigny, E., Colliat, J.L., Puech, A. & Meunier, J. 2008. Characterization of soft deepwater West Africa clays: SHANSEP testing is not recommended for sensitive structured clays. Proc. Offshore Technology Conference, OTC paper 19193, Houston. Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Lunne, T., Berre, T.V. & Strandvik, S. 1998. Sample disturbance effects in deep water soil investigations. Proc. Int. Conference on New Frontiers in Offshore Site Investigation and Foundation Behaviour, SUT, London. Magne, E. 2008. Pre-installed mooring systems for MODUs. Offshore Asia Conference, Kuala Lumpur. Montarges, R., Fay, J.B. & Le Tirant, P. 1987. Soil reconnaissance at great water depth. Proc. Int. Conference on Deep Offshore Technology, DOT, Monte Carlo. Murff, J.D., Randolph, M.F., Elkhatib, S., Kolk, H.J., Ruinen, R., Strom, P.J. & Thorne, C. 2005. Vertically loaded plate anchors for deepwater applications. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Orozco, M., Foray, P. & Nauroy, J.F. 2007, Pipe-soil horizontal dynamic stiffness in soft soils. Proc. 17th International Offshore Polar and Engineering Conference, ISOPE, Lisbon, Portugal. Osborne, J.J., Yetginer, A.G., Halliday, T. & Tjelta, T.I. 2010. The future of deepwater site investigation: seabed drilling technology? Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Pereira, J.M. & De Gennaro, V. 2009. Time dependent behaviour of fluids filled geomaterials: application to reservoir formation. Proc. 4th Biot Conference on Poromechanics, Poromechanics IV, Columbia University, New York. Peuchen, J. 2000. Deepwater cone penetration tests. Proc. Offshore Technology Conference, OTC paper 12094, Houston. Puech, A., Colliat, J.L., Nauroy, J.F. & Meunier, J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Puech, A., Orozco-Calderon, M. & Foray, P. 2010. Mini T-bar testing at shallow penetration. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Puech, A., Poulet, D. & Boisard, P. 1990. A procedure to evaluate pile drivability in the difficult soil conditions of the southern part of the Gulf of Guinea. Proc. Offshore Technology Conference, OTC paper 6237, Houston. Quiros, G.W. & Little, R 2003. Deepwater soil properties and their impact on the geotechnical program. Proc. Offshore Technology Conference, OTC paper 15262, Houston. Raines, R.D., Ugaz, O. & Garnier, J. 2005. Centrifuge modelling of suction piles in clay. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth.
Randolph, M., Cassidy, M., Gourvenec, S. & Erbrich, C.H. 2005. Challenges of offshore geotechnical engineering. Proc. Int. Conference on Soil Mechanics and Geotechnical Engineering, ICSMGE, Osaka. Randolph, M.F., Hefer, P.A., Watson, P.G. & Geise, J.M. 1998. Improved seabed strength profiling using T-bar penetrometer. Proc. Int. Conference on New Frontiers in Offshore Site Investigation and Foundation Behaviour, SUT, London. Randolph, M.F., Low, H.E. & Zhou, H. 2007. In situ testing for design of pipelines and anchoring systems. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Randolph, M.F. & White, D.J. 2008. Pipeline embedment in deep water: processes and quantitative assessment. Proc. OffshoreTechnology Conference, OTC paper 19128, Houston. Rattley, M.J., Hill, A.J., Thomas, S. & Sampurno, B. 2010. Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Robertson, P.K. 2009. Interpretation of cone penetration tests – A unified approach. Canadian Geotechnical Journal, 46, 1337–1355. Ruinen, R. & Degenkamp, G. 2001. Anchor selection and installation for shallow and deepwater mooring systems. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Stavanger. Reusswig, G.H. & Nair, N.N. 1994. A compliant tower design for mild environments. Proc. ASME Int. Conference on Offshore Mechanics and Arctic Engineering, OMAE, Houston. Sobkowicz, J.C. & Morgenstern, N.R.1984. The undrained equilibrium of gassy sediments. Canadian Geotechnical Journal, 21, 439–446. Stallebrass, S.E.. & Taylor, R.N. 1997. The development and evaluation of a constitutive model for the prediction of ground movements in overconsolidated clay. Géotechnique 47 (2), 235–253. Stevens, R.S., Wiltsie, E.A. & Turton, T.H. 1982. Evaluating pile drivability for hard clay, very dense sand, and rock. Proc. Offshore Technology Conference, OTC paper 4205, Houston. Sultan, N., Cochonat, P., Cauquil, E. & Colliat, J.L. 2001. Apparent overconsolidation and failure mechanisms in marine sediments. Proc. OTRC Int. Conference Honoring Prof. W. Dunlap, Houston. Sultan, N., Marsset, B., Ker, S., Marsset, T., Voisset, M., Vernant, A.M., Bayon, B., Cauquil, E., Adamy, J., Colliat, J.L. & Drapeau, D.H. 2010. Hydrate dissolution as a potential mechanism for pockmark formation in the Niger delta. Proc. Journal of Geophysical Research, Solid Earth (in press). Sultan, N., Voisset, M., Marsset, T., Vernant, A.M., Cauquil, E., Colliat, J.L. & Curinier, V. 2007. Detection of free gas and gas hydrate based on 3D seismic data and cone penetration testing: An example from the Nigerian continental slope. Marine Geology, 240, 235–255. Thomas, F., Nauroy, J.F., Rebours, B & Meunier, J. (2005). Mineralogical characteristics of the Gulf of Guinea deep water sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Thomas, F., Puech, A., Nauroy, J.F., Palix, E. & Meunier, J. 2007. Specific identification test procedures for deepwater sediments of Gulf of Guinea. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London.
85 © 2011 by Taylor & Francis Group, LLC
Thorel, L., Dendani, H., Garnier, J., Colliat, J.L. & Rault, G. 2010. Installation process of suction anchors with and without stiffeners in Gulf of Guinea clay: centrifuge modelling. Proc. Int. Conference on Physical Modelling in Geotechnics, ICPMG, Zurich. Wheeler, S.J. 1988. A conceptual model for soils containing gas bubbles. Geotechnique, 38, 389–397. White, D.J. & Cathie, D.N. 2010. Geotechnics for subsea pipelines. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. White, D.J., Hill, A.J., Westgate, Z. & Ballard, J-C. 2010. Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Will, S.A., Mascorro, E., Robertson, W, Hussain, K. & Paulson, S. 2006. Benguela-Belize compliant tower:Tower
© 2011 by Taylor & Francis Group, LLC
design. Proc. Offshore Technology Conference, OTC paper 18068, Houston. Wong, P.C., Taylor, B.B., & Audibert, J.M.E. 2008. Differences in shear strength between Jumbo piston core and conventional rotary core samples. Proc. Offshore Technology Conference, OTC paper 19683, Houston. Young, A.G., Honganen, C.D., Silva, A.J. & Bryant, W.R. 2000. Comparison of geotechnical properties from largediameter long cores and borings in deep water Gulf of Mexico. Proc. Offshore Technology Conference, OTC paper 12089, Houston. Zimmermann, C.A., Layrisse, G., De La Cruz, D. & Gordonnat, J. 2009. Greater Plutonio riser tower installation – Studies and lessons learnt. Proc. ASME Int. Conference on Ocean, Offshore and Arctic Engineering, paper OMAE2009-79028, Honolulu.
86
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Geotechnics for subsea pipelines D.J. White Centre for Offshore Foundation Systems, University of Western Australia, Perth
D.N. Cathie Cathie Associates, Brussels, Belgium
ABSTRACT: The geotechnical analysis performed for subsea pipeline design involves challenges that are not common in conventional foundation engineering. This paper reviews recent research in pipeline geotechnics and shows examples of how this research is being applied in practice. A general theme running through this paper is the twin challenges of the changes in seabed topography and the changes in soil properties that occur through the installation and operating life of a pipeline. Results from in situ and element testing of soils that replicate the loading and disturbance imposed by pipelines are used to show that significant changes in strength are induced. Soil generally weakens during the episodes of remoulding that accompany pipeline laying, buckling, walking and storm loading, and during ploughing and trenching. The soil strength recovers during subsequent episodes of reconsolidation between storms, and between startup and shutdown events. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A unifying theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths through cyclic episodes, and the importance of recognising the widelyvarying rates of shearing involved in pipe-soil processes. Pipeline geotechnics can involve drained behaviour in fine-grained clayey soils – for example, during slow axial expansion of pipelines – and undrained behaviour in coarse-grained soils – for example during ploughing. Concepts from critical state soil mechanics often provide a simple framework for clarifying this behaviour.
1
INTRODUCTION
soft fine-grained soils, where the management of thermal and pressure-induced expansion is a critical design issue. The second is the stability of light large-diameter pipelines in shallow water, where primary and secondary stabilization measures represent a significant capital expenditure, and where geotechnical analysis techniques are not well established. The first scenario is relevant to pipeline design in almost all deepwater frontiers globally.The second scenario is particular relevant off the coast of Australia, where many hundreds of kilometers of gas trunkline are currently planned, to bring gas from deepwater fields to onshore LNG facilities. Section 1.2 provides a brief overview of some key pipeline design considerations that have a strong geotechnical influence. This overview is intended to provide a brief introduction to some of the most novel aspects of pipeline geotechnics, for those who are unfamiliar with this area of geotechnics. In Section 1.3 comparisons are made between pipeline geotechnics and conventional foundation engineering. Some of the most relevant aspects of soil behaviour are highlighted in Section 2, using recent experimental observations, including some from unusual forms of penetration testing. These observations provide a backdrop to the mechanisms and analysis techniques that are outlined in Sections 3, 4 and 5 for assessing
1.1 Scope of paper The purpose of this keynote paper is to set out the challenges of pipeline geotechnics and to highlight some recent developments in this area that the authors have been involved with. This is not an exhaustive treatment of the subject, but is intended to – provide an overview of areas in which intensive research has recently been published. – highlight certain novel analysis techniques for design that this research is beginning to permit. More complete introductions to pipeline geotechnics are provided by the relevant chapters of books on offshore geotechnical engineering by Randolph and Gourvenec (2010) and Dean (2010). Cathie et al. (2005) presented a more exhaustive review of research across the whole of pipeline geotechnics. The present paper is intended to be of value to pipeline engineers as well as geotechnical specialists, and includes some basic geotechnical content where we consider this useful. The topics in this paper are generally relevant to one of two design scenarios. The first is high pressure high temperature pipelines laid in deepwater on
87 © 2011 by Taylor & Francis Group, LLC
Figure 1. Some geotechnical aspects of pipeline design.
pipeline embedment, lateral pipe-soil resistance and axial pipe-soil resistance respectively. Sections 6 and 7 are focussed on the geotechnics of pipeline trench construction by ploughing and jetting respectively. Space limitations preclude discussion of in-trench pipeline stability, and the associated upheaval, backfill liquefaction and flotation issues. The paper finishes with brief conclusions. 1.2
Pipe-soil interaction processes
Many of the areas of pipeline design that have geotechnical aspects are illustrated in Figure 1. Offshore pipelines are often left on the seabed, unburied, if this does not lead to unacceptable instability under hydrodynamic loading. The interaction between the pipeline and the seabed feeds into many aspects of the pipeline design. If the pipeline must be buried, for stability or to avoid fishing gear, the shielding of the pipeline via the construction of a trench (possibly backfilled) requires geotechnical design. On-bottom pipelines are increasingly being designed to allow movement during their operation, either under hydrodynamic loading or under thermal and pressureinduced expansion. Steel catenary risers, which are extensions of pipelines that connect to surface facilities, inevitably move where they touchdown on the seabed, in response to oscillation of the floating facility. Throughout the lay process and during subsequent operating cycles, the pipeline is subjected to geotechnical forces where it is in contact with the
88 © 2011 by Taylor & Francis Group, LLC
seabed. A pipeline is a forgiving structure, being able to tolerate significant deformation and gross movements across the seabed, except at points of fixity such as end terminations. Such instabilities are unacceptable for the platforms and foundations that conventional geotechnical engineering is equipped to design. Indeed, if a pipeline was not permitted to move across the seabed under thermal loading, this would often induce unacceptable thermally-induced stresses. The forgiving flexibility of a pipeline does not, therefore, alleviate the need to quantify the pipesoil resistance forces to a sufficient accuracy that the robustness of the design is demonstrated. One of the most difficult aspects of pipeline design, which is an increasing challenge as operating temperatures and pressures rise, is the management of thermal and pressure-induced loading. Controlled on-bottom lateral buckling is an attractive design solution but one which requires the pipe-soil responses to be bracketed: both high and low geotechnical resistance can hamper a design (Bruton et al. 2007; AtkinsBoreas 2008). A second and related behaviour that arises from the thermal and pressure-induced loading is the tendency for pipes to ‘walk’ axially over cycles of startup and shutdown (Tornes et al. 2000; Carr et al. 2006). This phenomenon can be driven by the asymmetry of the heat-up and cool-down processes or by the presence of a seabed slope or end-of-line tension (which creates an asymmetry in the mobilized axial pipe-soil resistance).Accurate assessment of the axial pipe-soil resistance forces is required for robust modelling of this process.
Table 1.
Comparison of pipeline geotechnics and conventional foundation engineering (after White and Gaudin 2008). Foundation
On-bottom pipeline
Problem geometry
Known, controlled.
Design criteria for in-service behaviour
To remain fixed, movement uD.
Surrounding soil conditions
Similar to in situ state. Relatively unaffected by installation.
Soil-structure interaction
Usually minimal. Imposed loads are not strongly affected by foundation displacements. Scour and wave-induced liquefaction may require mitigation Usually available. Can assume lowest credible geotechnical capacity
Uncertain. Embedment affected by lay process and metocean conditions. Subsequent pipeline movements disturb seabed topography. May be required to displace significantly, u D, through hundreds of cycles of operation or hydrodynamic loading. Soft soil is significantly affected by installation. Remoulding, heave and reconsolidation affect the local strength. Often significant. Local pipe-soil load-displacement relationship affects overall pipeline response.
Soil-ocean interaction Single conservative design approach
Another significant design issue that is particularly relevant in the shallow waters offshore Australia, is pipeline stability under hydrodynamic loading from storm-induced currents and waves. In this situation a conservative approach is to adopt a low value of soil resistance. However, the cost of stabilization measures such as concrete coating is huge, and there is a strong incentive to refine the geotechnical analysis to remove any unnecessary conservatism in the design seabed resistance. In shallow water a pipeline may require additional – ‘secondary’ – stabilisation for hydrodynamic stability. Secondary stabilisation solutions revolve around reducing the hydrodynamic loading and increasing the available lateral resistance. An open trench provides partial shielding from hydrodynamic load. Burial of the pipe eliminates direct hydrodynamic loading (although soil liquefaction under hydrodynamic loading can destabilise a buried pipe). Geotechnical assessments must be made of the trenching process – which may be by ploughing, cutting, jetting, dredging or a combination. Other secondary stabilisation techniques include continuous rockdumping, or engineered solutions to provide local anchoring at intervals along the pipe. These solutions include flexible concrete mattresses, anchor blocks or saddles placed over the pipeline, or small piles on either side of the pipeline. The stability of these objects must also be assessed in design, taking account of the additional cyclic loading transferred to them by the unstable pipeline.
foundations and piles. This is partly because it is only recently that design codes have permitted gross pipeline movement, and so designers have not needed to explicitly assess the interaction forces as pipelines sweep across the seabed. Also, it is only recently that some of the complexities of the underlying soil behaviour have been recognized. A generally accepted framework for routine analysis has not emerged. The contrasts between pipeline geotechnics and conventional foundation engineering are summarised in Table 1 and illustrated in Figure 2. The designer’s task in the geotechnical design of a pipeline is aided by the structure’s tolerance of movements and mild deformation, but is hampered by the difficulty of assessing the geometry of the scenario and the operative soil properties. The laying of a pipeline and any subsequent lateral or axial movements disturb the topography of the seabed. The changed geometry and the altered soil properties need to be captured in calculations of the available pipe-soil resistance. Even the intact soil properties are difficult to establish at the shallow embedments relevant to pipeline geotechnics. Undisturbed sampling of soft near-surface soils is difficult and penetrometer tests at shallow embedment require particular interpretation techniques (Puech and Foray 2002, White et al. 2010a). Some soil properties and parameters such as friction angles and undrained strength ratios tend to be different at the very low stress levels relevant to pipeline geotechnics. A further complication is that interaction between the ocean and the seabed – leading to scour and liquefaction – can be significant in shallow water. The result is a tripartite interaction between the ocean, the pipeline and the seabed, which is illustrated in Figure 3. Cross-disciplinary design approaches for
1.3 Comparison with foundation engineering Geotechnical design procedures for pipelines and risers are relatively undeveloped compared to
89 © 2011 by Taylor & Francis Group, LLC
Scour and wave-induced liquefaction can dominate behaviour Often unavailable. Both upper and lower bound geotechnical capacity may adversely effect structural response
Figure 3. Tripartite interaction between the seabed, the ocean and a pipeline in shallow water during storms.
buckles to sweep laterally across the seabed – and storms – which create high hydrodynamic loading. Over the operating life of a pipeline, the surrounding soil may therefore be subjected to a large number of episodes of disturbance followed by recovery and reconsolidation. The soil strength will generally fall and rise with each episode, and the net effect may be an overall increase or decrease in the strength of the soil. Coupled with the associated changes in seabed topography, this may lead to a rise or a fall in the geotechnical restraint on the pipeline. Figure 4a illustrates the case of a buckling pipeline on a fine-grained soil (in deepwater, where hydrodynamic loading is negligible). Each startup or shutdown of the buckle causes gross monotonic remoulding of the surrounding soil, comparable to the high disturbance created during passage of a penetrometer. The startup and shutdown episodes are separated by a period of time which may or may not be sufficient for full dissipation of the excess pore pressures generated during the previous disturbance – depending, obviously, on the consolidation characteristics of the soil relative to the frequency of the startups and shutdowns. Typical pipeline designs require several hundred or even one thousand shutdown and startup events to be considered. Figure 4b illustrates the case of a pipeline on a sandy or silty soil in shallow water, where storms create high hydrodynamic loading, which in turn causes the pipeline to exert cyclic loads on the surrounding soil. A storm loading event is perhaps best considered as a pre-failure cyclic disturbance as distinct from the gross (undrained) monotonic remoulding of the previous example. In this illustration the cyclic loading leads to a weakening, associated with pore pressure buildup, which is subsequently compensated for by reconsolidation and an associated densification of the soil. Storms occur at a frequency such that full dissipation occurs between events, for the soil considered here. In design, it is necessary to consider the effect
Figure 2. Comparison of pipeline geotechnics and conventional foundation engineering (images from Jayson et al. 2008 and Fisher and Cathie 2003).
this interaction are in their infancy. There is not yet a routine basis for assessing pipeline stability under hydrodynamic action that incorporates all three interactions concurrently (Damgaard and Palmer 2001, Cheng et al. 2010).
2 2.1
RELEVANT SOIL MECHANICS Illustrations of soil behaviour near pipelines
The soil close to a pipeline is grossly disturbed as the pipe is laid. If that disturbance happens rapidly enough for excess pore pressure to be generated then the subsequent reconsolidation process generally leads to an increase in the strength and density of the soil. Subsequent events may disturb the pipeline and the surrounding soil further. These events include the startup and shutdown of the pipeline – which cause
90 © 2011 by Taylor & Francis Group, LLC
Figure 5. Changes in the undrained penetration resistance of fine-grained soils during cyclic penetrometer tests (after Gaudin and White 2009).
remoulded state, with this ratio being termed the sensitivity. Much higher sensitivities can sometimes be found, particularly in carbonate soils. In the analysis of problems involving significant disturbance, it is necessary to identify the relevant soil strength, which may lie somewhere between the intact and remoulded values. Cyclic T-bar or ball penetrometer tests allow this behaviour to be quantified. The progressive reduction in net bearing resistance through cycles of disturbance is shown for various different soils in Figure 5. The strength degrades exponentially with the number of cycles of disturbance, which allows the response to be characterized by two parameters – the sensitivity, St (or its inverse, δrem ) and a parameter related to the ductility, N95 (which is the number of cycles of disturbance after which the resistance has decayed by 95% of the difference between the intact and remoulded values). It is evident that both of these parameters vary significantly between soil types (particularly St ), but the form of the decay is similar. Analysis techniques for design can capture the decaying soil strength by adopting a value that represents the relevant level of disturbance. The general approach is to firstly convert the penetrometer ductility parameter, N95 , to an equivalent strain level (e.g. Zhou and Randolph 2009). The relevant strain level for the problem being considered is then used to deduce the operative undrained strength. Such techniques have been proposed for spudcan penetration (Erbrich 2005, Hossain and Randolph 2009). These methods utilise the type of strength degradation curves shown in Figure 5 to link the strains and operative strength around a spudcan to those around a T-bar. T-bar and spudcan penetration involves a comparable level of disturbance to monotonic pipe embedment. However, the dynamic motions that accompany pipe laying mean that a greater level of disturbance and hence a lower operative soil strength is applicable, compared to that mobilised during initial T-bar penetration. The dramatic reductions in strength evident in Figure 5 are partly due to the generation of positive
Figure 4. Illustrative histories of soil element behaviour near unstable pipelines: episodes of disturbance and recovery.
of these disturbances on the available lateral pipe-soil resistance – which controls the stability. This stability is affected by changes in the pipeline embedment as well as the changes in the strength of the soil around the pipe during disturbance. The same history of disturbance and reconsolidation shown in Figure 4b is applicable to the backfill above a trenched pipeline. These illustrations of the history of soil behaviour are clearly very idealized, and it is rare that the related changes in soil strength are tracked explicitly within a design analysis. However, it is important to recognize these effects, since they have a significant influence on the geotechnical restraint on a pipeline. These illustrations provide a convenient background to the following three examples of soil element behaviour. These examples show the changes in soil strength that can accompany episodes of disturbance and recovery, and also showcase novel testing techniques that may in the future be utilized to quantify this behaviour for design. The examples are: – episodes of undrained disturbance and reconsolidation seen by a T-bar penetrometer; – episodes of undrained disturbance and reconsolidation seen by a vertical rod penetrometer; – drained and undrained failure of different soils; – low stress friction response of fine-grained soils.
2.2 Disturbance and recovery: T-bar tests The strength of soft fine-grained seabed soils typically reduces by a factor of 2–5 from the intact to the fully
© 2011 by Taylor & Francis Group, LLC
91
Figure 6. Undrained strength through episodes of remoulding and reconsolidation (test in lightly overconsolidated kaolin in the UWA beam centrifuge) (White and Hodder 2010).
Figure 7. Critical state interpretation of episodes of remoulding and reconsolidation (White and Hodder 2010).
the attraction of setting this behaviour within an effective stress framework, but it does rely on a very crude simplification of the overall behaviour, in which the response of all the elements of soil around a penetrometer is lumped into a single representative effective stress level and specific volume. More refined models will allow this behaviour to be more accurately quantified and numerical simulations will test the validity of this simplification. The key point, however, is that the rises in soil strength during episodes of reconsolidation can eclipse the reductions in soil strength during the preceding episodes of remoulding.
excess pore pressure in these contractile materials during undrained shearing. As this positive pore pressure dissipates and the effective stress rises back to the geostatic state the material densifies and the subsequent undrained shear strength may be higher. Cyclic T-bar penetrometer tests with periods of reconsolidation between episodes of cycling show this regain in strength. Figure 6 summarises the results of a cyclic T-bar test in kaolin clay reported by White and Hodder (2010), expressing the T-bar strength at a particular depth – 2.25 m – during each cycle. After just three episodes of full remoulding and reconsolidation, the current remoulded strength was comparable to the original intact strength. These results quantify the contrasting effects of disturbance and recovery shown in Figure 4a for this soil and the particular disturbance pattern imposed by a T-bar. This behaviour is easily understood within a critical state-type framework, since this provides an explicit link between moisture content (which reduces as positive pore pressures dissipate) and undrained strength. This interpretation can be extended to a quantitative treatment, expressed in terms of the operative strength averaged over all of the soil near the penetrometer (rather than of a single soil element). An accurate back-analysis of the results shown in Figure 6 can be achieved by defining two failure lines in stressvolume space, which represent the intact and fully remoulded strengths of the soil, as proposed by White and Hodder (2010). As shown in Figure 7, this back-analysis of the T-bar resistance at a depth of 2.25 m (which corresponds to an in situ effective vertical stress of σvo = 12 kPa) is based on the intact strength line (ISL) being reached during the initial T-bar stroke of a episode, and the effective stress point migrating towards the remoulded strength line (RSL) according to an exponential trend (i.e. the reduction in effective stress per T-bar stroke is proportional to the difference between the current effective stress and the effective stress at the remoulded state for the current specific volume). This analysis has
2.3
The example above involves only 3 episodes of reconsolidation. A significantly larger number of episodes of reconsolidation are involved in the second example. A novel vertical rod penetrometer has been used on a recent centrifuge project at UWA, with the aim of quantifying the resistance and strength of surficial material, as the soil is forced to flow past the penetrometer. This device is a cylindrical bar, oriented vertically, 4 mm in diameter. The device is embedded until the tip is typically 5 – 10 diameters below the soil surface. The bar is equipped with multiple levels of strain gauging located above the soil surface, which allow the magnitude and distribution of the pressure on the bar to be derived. In one test the bar was embedded in soft kaolin to a depth of 45 mm then cycled laterally by a distance of 20 mm at a rate of 0.3 mm/s. This rate corresponds to a dimensionless velocity of vD/cv ∼ 10 which is almost fully undrained, based on the limits demonstrated by Finnie and Randolph (1994) (albeit for a different geometry of problem). The elapsed time between the bar passing the mid-point of each lateral stroke was 100 seconds, which corresponds to a dimensionless consolidation time of T = cv t/D2 = 0.5. This value is indicative of significant (∼50%) pore pressure dissipation, based on limits provided by Randolph (2003) (again for a slightly different geometry of problem).
92 © 2011 by Taylor & Francis Group, LLC
Disturbance and recovery: vertical rod tests
Figure 8. Lateral resistance on a vertical bar penetrometer through episodes of disturbance and reconsolidation.
Figure 9. Drained and undrained resistance during penetration.
The resulting variation in the average lateral resistance on the bar penetrometer is shown in Figure 8. Initially the resistance reduces, as the remoulding damage exceeds the recovery from reconsolidation. However, a steady rise in resistance is soon evident, with the strength after several episodes exceeding the initial (intact) value. After this test the level of the soil surface around the penetrometer had lowered. This is evidence that the soil surrounding the device had densified through the episodes of reconsolidation, which is consistent with the changing strength. This example highlights again the changes in soil strength that can occur through the episodes of disturbance and recovery that can be imposed on the soil surrounding a seabed pipeline. As shown in Figure 8, the strength continued to increase through more than 25 cycles of heavy remoulding, but ultimately should reach a limiting value, when the soil has densified sufficient that there is no longer a tendency to generate positive excess pore pressure when disturbed.
plough share (Peng and Bransby 2010). In all cases the velocity is normalized by the coefficient of consolidation of the seabed and an appropriate drainage distance – generally the size of the foundation. There is some variation in the dimensionless velocities at which fully drained and fully undrained behaviour occurs, although this may be partly due to the difficulties in establishing appropriate values of the coefficient of consolidation. There are more significant differences between the relative magnitude of the drained and undrained resistance. These arise from the state of the soil – and hence its tendency to generate positive or negative pore pressure in undrained conditions – and also the particular boundary value problem. For example, for a soil with a particular undrained strength and angles of friction and dilation, the relative magnitude of the drained and undrained bearing capacity depends on both the applicable bearing factors (i.e. Nq and Nγ for drained conditions and Nc for undrained conditions) as well as the soil strength properties. The relative magnitude of the drained and undrained strengths of an interface is not affected by the geometry of the problem, so provides a more simple differentiation between the drained and undrained behaviour of a particular soil (albeit in combination with a particular interface). Figure 10 shows the steady residual resistance measured during monotonic shearing of normally consolidated kaolin clay over a rough steel surface at different velocities. The kaolin was normally consolidated to a stress of 2.5 kPa prior to shearing. These tests used a direct shear box at UWA that has been modified to operate at the low stress levels relevant to pipeline geotechnics. The four tests show a trend of increasing resistance with reducing velocity. This trend is consistent with a hyperbolic backbone curve of the same form as used in Figure 9, drawn between the fully drained and fully undrained limits. These limits correspond to an interface friction angle of 30◦ and an undrained = 0.25, both of which are constrength ratio of su /σvc sistent with other published results for this stress level
2.4 Drained and undrained soil responses The shear strength of a given soil in a particular state depends on whether drained or undrained conditions are imposed, as well as the mode of shearing. In conditions in which drainage is permitted, the shearing can be imposed at rates that span from fully drained (i.e. in which no excess pore pressure builds up) to fully undrained (i.e. in which effectively no pore pressure dissipation occurs, despite drainage being permitted). Various authors have recently explored the continuous variation in mobilized soil strength and geotechnical resistance between drained and undrained conditions. Results summarized in Figure 9 show the variation in penetration resistance of circular surface foundations (Finnie 1993), and cone penetrometers in soft clay (Randolph and Hope 2004) and dense silt (Silva 2005). Similar relationships have been derived for the uplift resistance of buried pipelines (Bransby and Ireland 2009) and the sliding resistance of a pipeline
93 © 2011 by Taylor & Francis Group, LLC
Figure 11. High effective stress friction and a non-linear failure envelope during low stress interface tests and axial pipe-soil movement (data from Bruton et al. 2009, White et al. 2010b).
Figure 10. Interface shear resistance at varying rates, from drained to undrained.
(Bolton and Barefoot 1997; Pedersen et al. 2003; Bolton et al. 2009). The movements of an on-bottom pipeline in a given soil may span the ranges of velocities and therefore drainage conditions shown in Figure 9 and Figure 10. Thermal expansions and the associated lateral movements span from zero at ‘virtual anchor points’ to millimetres per second of axial movement near ends and buckles, and even metres per second of lateral movement during initiation of lateral buckles. A further range of velocities arises from pipeline movements driven by hydrodynamic action – which include oscillations in the touchdown zone during laying that are created by vessel motion, or oscillations in response to direct hydrodynamic action on the pipe during storms. A high velocity applies during the ploughing of a pipeline trench. In combination with the large size of a ploughshare, this leads to undrained conditions even in sands. As a consequence of the varying velocities involved in these processes, it is common for the geotechnical analysis for pipeline design in a fine-grained soil to require an assessment of the drained response. Conversely, in coarse-grained soils an assessment of the undrained behaviour can be required. Also, there are often occasions when the actual response involves partial drainage, and it is necessary to tie together drained and undrained assessments in order to predict the most likely behaviour, and the potential range of responses.
2.5
The same trend appears in low stress soil-soil and soil-interface shearing of fine-grained soils (Pedersen et al. 2003, White and Randolph 2007, Hill and Jacob 2008, Bruton et al. 2009, White et al. 2010b). Nonlinear failure envelopes that express the friction angle or limiting stress ratio as a function of effective stress can capture this variation (e.g. Figure 11). It is important to recognise that the friction angles measured at conventional geotechnical stresses may not be appropriate for the assessment of drained pipe-soil resistance. The high friction angles are also reflected in the higher undrained strength ratios found at low stresses – a link highlighted in the approximate expression for normally consolidated undrained strength ratio su /σvc = φ/100 derived from the analysis of Wroth (1984), where φ is the friction angle in is the consolidation stress. Changes in degrees and σvc friction angle affect both the drained and undrained strengths of soil, since the underlying behaviour is principally frictional. 2.6
As well as the conventional aspects of soil behaviour that are considered in the analysis of foundations, pipeline geotechnics is also often concerned with a greater degree of soil disturbance, and intervening periods of recovery and reconsolidation. These processes also take place at lower stress levels compared to conventional geotechnics, and can lead to significant changes in the state and therefore the strength of a soil through the operating life of a pipeline. Also, due to the relevant drainage distances and rates of movement, it is can be necessary to focus on the drained response of fine-grained soils and the undrained response of coarse-grained soils.
Low effective stress friction
A final feature of soil behaviour that is particularly relevant to pipeline geotechnics is the variation in friction angle with stress level. At the low stresses relevant to pipelines, higher friction angles are found compared to more usual geotechnical stress levels. The peak friction angle of sands increases with reducing stress level (Bolton 1986). Results from experiments performed on Earth (Fannin et al. 2005) and onboard the space shuttle (Sture et al. 1998) also show that the critical state or constant volume friction angle is higher at very low stresses.
3 3.1
PIPELINE EMBEDMENT Pipelaying mechanics
The as-laid embedment of a pipeline affects the subsequent pipe-soil resistance, as well as the thermal
94 © 2011 by Taylor & Francis Group, LLC
Summary of soil behaviour
Figure 12. Pipeline laying notation (Randolph and White 2008a).
Figure 13. Maximum stress concentration factors for pipe laying on an elastic seabed (Randolph and White 2008a).
insulation. The installation process involves soilstructure interaction, since the maximum vertical pipe-soil force during the lay process will exceed the submerged pipe weight, W , by an amount that depends on the seabed stiffness and the geometry of the catenary created by the S-lay or J-lay arrangement. The configuration of a pipeline during laying is shown in Figure 12. A key parameter is the horizontal component of tension, T0 , which is constant through the suspended part of the pipeline and can be assessed from the pipe weight, water depth and hang-off angle. The maximum contact force (per unit length) with the seabed, Vmax , and hence the local force concentration factor, flay = Vmax /W , is a function of the seabed stiffness, k (defined as the secant ratio of force per unit length, V, to embedment, w) in addition to the bending rigidity, EI, and T0 . The force concentration factor reduces with increasing water depth and decreasing seabed stiffness. A characteristic length, which relates to the length over which the bending stiffness moderates the catenary behaviour, is given by λ = (EI/T0 )0.5 . Parametric solutions for the static lay conditions have been presented by Randolph and White (2008a), who showed that for horizontal tension of T0 > 3λW (which holds for most pipelines), results from analytical solutions (Lenci and Callegari 2005) and numerical analysis using OrcaFlex (Orcina 2008) all converge to unique design lines. The value of flay may be expressed approximately as (Figure 13):
Section 3.4, but firstly the significant influence of dynamic pipe movements during the laying process is highlighted. During J-lay or S-lay installation, dynamic movement of the pipe occurs within the touchdown zone, driven by the vessel motion and hydrodynamic loading of the hanging pipe. These loads induce a combination of vertical and horizontal motion of the pipeline at the seabed (Lund 2000, Cathie et al. 2005). In addition to vessel motion due to swell and waves at the sea surface, cyclic changes in pipeline tension may occur (depending on the accuracy of the tensioning system) if the offloading of the pipe is not smoothly coincident with the vessel advancement. This dynamic movement, although often of very small amplitude, leads to local softening of the seabed sediments and can push soil away to either side of the pipe alignment, creating a narrow trench in which the pipe becomes embedded. An illustration of the significant additional embedment that can occur simply due to small amplitude cyclic motions is shown in Figure 14. These results are from a centrifuge model test on lightly overconsolidated kaolin clay (Cheuk and White 2010a). A model pipe was penetrated to a normalised embedment of w/D = 0.1 (point A), when the vertical pipe-soil load was fixed constant (the normalised vertical load, V/su D reduced with pipe embedment, due to the increasing su with depth). A series of packets of horizontal oscillations were then imposed, increasing in amplitude (Figure 14a). The adopted amplitudes of motion reflect ROV observations during laying, although these are dependent on the lay geometry and metocean conditions (Westgate et al. 2010a). The aim in this experiment was to represent dynamic lay motions in an idealised manner. The pipe initially settled at a rapid rate, by an amount that far exceeds that due to the combined vertical and horizontal loading alone. The lateral soil resistance mobilised during the first two cycles, when the embedment doubles (to point B), corresponds to an equivalent friction factor of H/V < 0.25. As the embedment increases, a greater soil resistance is mobilised for a given amplitude of
3.2 Seabed disturbance during pipelaying Equation 1 is derived based on a single value of secant seabed stiffness, V/w, which would be applicable to a purely elastic seabed. The actual seabed response is generally non-linear during vertical penetration, and is stiffer during unloading since the seabed has been plastically deformed. As a result, the actual operative secant stiffness varies along the touchdown zone. Theoretical solutions for monotonic vertical pipe penetration into undrained soil are described in
© 2011 by Taylor & Francis Group, LLC
95
motion, reflecting the increasing constraint on the pipeline. The softening of the surrounding soil is evident in the reductions in lateral resistance at points C and D, with increasing disturbance. Complementary assessments of the level of soil remoulding during pipeline laying can be made using large deformation finite element analysis. A study reported by Wang et al. (2009) replicated the first stage of the model test shown in Figure 14 (with horizontal motions of +/−0.05D) using continuum finite element analysis. The soil constitutive model included softening through a reduction of the undrained strength with accumulated plastic strain, in a manner consistent with the behaviour shown in Figure 5. The resulting patterns of lateral resistance and embedment and the local soil remoulding are shown in Figure 15. Even small lateral motions of just +/−0.05D lead to a remoulded zone that extends by almost one pipe diameter to each side, and the pipe itself rests on fully remoulded soil. These results highlight the importance of assessing pipeline embedment using an appropriately degraded value of soil strength, as well as accounting for the catenary overstress via Equation 1 or some comparable approach. 3.3 As-laid pipeline survey observations Similar conclusions can be drawn from the results of ROV surveys following pipe laying. The variation in embedment along a pipeline varies for a variety of reasons including local variations in the seabed strength. However, consistent variations have also been identified due to wave height, lay rate, downtime events, and changes in lay angle. The first three of these effects influence the level of dynamic motion that the pipe is subjected to in the touchdown zone and the latter effect influences the vertical stress concentration (Westgate et al. 2010a). The resulting range of embedment can be expressed as a statistical variation, and this range can be compared with calculations performed using theoretical solutions for the catenary overstress and the seabed penetration resistance, using both intact and remoulded soil strengths. Figure 16 shows the distribution of embedment from survey measurements taken at 1 m intervals along a 13 km long pipeline (excluding short lengths of pipeline that had a far greater embedment due to downtime events). The pipeline was laid on almost uniform deepwater soil conditions, in a water depth of 1215– 1450 m (Westgate et al. 2010b). The lay process took several days, during which the sea state varied with a significant wave height of between 0.6 m and 1.7 m. The calculated pipeline embedment based on the static catenary overstress (Equation 1) coupled with intact and fully remoulded strengths are highlighted, along with estimates based on a dynamic analysis of the vertical stress concentration performed using Orcaflex (Orcina 2008) (Figure 16). The fully remoulded strength coupled with the static overstress
Figure 14. Effect of lateral motions on pipeline embedment (after Cheuk and White 2010a).
matches well with the most frequent embedment, although the agreement was not as good for some other pipelines at the same site. This result and other comparisons from a limited range of post-installation surveys, suggest that the fully remoulded soil strength leads to reasonable estimates of the average pipe embedment for average lay conditions. However, there remains significant scatter between different pipelines in the same conditions and variations in embedment along a single pipeline (Westgate et al. 2010a, Westgate et al. 2010b). An additional effect appears to be that lighter pipelines are more susceptible to dynamic lay effects – probably due to their reduced inertia – and so a greater degradation in soil strength applies.
96 © 2011 by Taylor & Francis Group, LLC
Figure 15. LDFE simulation of soil strength after dynamic pipe laying (after Wang et al. 2009).
breakout resistance since it is not immediately obvious what combination of input parameters will lead to upper and lower bound outcomes. It is also attractive to set the geotechnical components of pipeline analysis within a full probabilistic framework. This is consistent with the probabilistic structural reliability analyses that are increasingly performed during the assessment of pipeline on-bottom stability or lateral buckling. 3.4 Solutions for vertical pipe penetration Using an appropriate operative soil strength, su , the shallow embedment of a pipeline in undrained conditions can be calculated via a bearing capacity expression that reflects the appropriate geometry of failure mechanism. Experimental results presented by Dingle et al. (2008) revealed the internal soil failure patterns during pipe penetration (Figure 17a). These detailed observations, coupled with the vertical force-displacement response from other tests, have been used to validate numerical simulations (e.g. Figure 17b, Merifield et al. 2008a) and plasticity limit analyses (Randolph and White 2008b). The resulting bearing capacity, Vult , can be calculated as the superposition of components related to the soil strength at the pipe invert (via a bearing factor, Nc , that varies with embedment, pipe roughness and strength heterogeneity) and the soil buoyancy – enhanced by heave (Randolph and White 2008a, Merifield et al. 2009):
Figure 16. Distribution of as-laid embedment for a single pipeline (Westgate et al. 2010b).
The use of the remoulded soil strength over-predicts the embedment in the case of minimal pipeline motions (for example in calm weather or during lay down of the final catenary section of pipe) and under-predicts embedment during severe weather or downtime events, again based on limited field data (Westgate et al. 2009; Westgate et al. 2010b). Due to this inevitable scatter, including variability due to effects that cannot be predicted in advance of the pipelaying, assessments of pipeline embedment are subject to uncertainty. If only an upper or lower bound embedment is required for design then this uncertainty can be circumvented using conservative assumptions. However, both upper and lower bound embedments are usually critical for different design considerations. It is therefore necessary to perform lower and upper bound assessments using opposite extremes of the input parameters. These can be extended to a full probabilistic assessment, using statistical variations in input parameters, as illustrated in Section 4.4. Such an approach is desirable when assessing the subsequent
The nominally embedded cross-sectional area is denoted A , the soil submerged unit weight is γ and
97 © 2011 by Taylor & Francis Group, LLC
Figure 19. Undrained failure envelopes for lateral breakout in uniform soil.
4
LATERAL PIPE-SOIL INTERACTION
4.1 Theoretical failure envelopes The same combination of experimental, numerical and analytical techniques has been used to derive failure envelopes in vertical-horizontal load space, which allow the lateral breakout resistance in undrained conditions to be assessed (Dingle et al. 2008; Merifield et al. 2008a, Randolph and White 2008b). Different forms of failure mechanism apply, depending on whether tension can be sustained at the rear of the pipeline (Cheuk et al. 2008; Merifield et al. 2008b) as shown in Figure 19. For the no-tension case (and w/D < 0.5), the failure envelopes pass through the V-H origin, and there is a ‘frictional’ cut-off corresponding to a failure mechanism involving the pipe riding up at an angle shown as θ in Figure 19, with no soil deformation occurring. In drained conditions, experimental results show that the failure envelope has a similar shape to the no-tension undrained case (Zhang et al. 2002a). Centrifuge modelling results from tests on soft clay show that the breakout resistance is well predicted by the unbonded failure envelopes if the pipe is installed with some level of dynamic movement during laying (such as the simulation in Figure 14) (Cheuk and White 2010b). In this case, for very soft clay, the appropriate undrained strength appeared to be the intact value. In this particular case the effects of remoulding and reconsolidation approximately cancel out. Without dynamic movement the breakout resistance is often higher, but is highly brittle, reflecting the opening of a crack behind the pipe prior to mobilisation of a full two-sided mechanism (Dingle et al. 2008). It is thought that any dynamic movement leaves a skin of weak remoulded soil at the pipe surface. This can fail in a local mechanism behind the pipe, even if some tension is sustained, providing negligible extra resistance above the unbonded case. Modified versions of these failure envelopes can be created for other boundary conditions, such as a pipe running parallel to a slope (e.g. Morrow and Bransby
Figure 17. Failure mechanisms during vertical pipe penetration (Dingle et al. 2008, Merifield et al. 2008a).
Figure 18. Bearing factors during vertical penetration into heterogeneous and uniform soil (Chatterjee et al. 2010b).
the factor fb captures the enhancement of soil buoyancy beyond that given by Archimedes’ principle. A value of fb = 1.5 is typical. Solutions for the bearing factor, Nc , in this expression have been provided for uniform and linear soil strength profiles (Aubeny et al. 2005, Merifield et al 2008a), and with modifications to account for heave (Merifield et al. 2009, Chatterjee et al. 2010a, 2010b). A pair of bearing factor profiles derived from large deformation finite element analyses are shown in Figure 18, for two different profiles of soil strength (Chatterjee et al. 2010b).
98 © 2011 by Taylor & Francis Group, LLC
Figure 21. Effect of load path on equivalent breakout friction.
load path reaches the unbonded failure point marked A, with the equivalent friction factor marked µeq . The load path marked B represents hydrodynamic loading, when the pipe is subjected concurrently to both horizontal drag and upwards lift. In this case the equivalent friction factor at failure is higher (although the constant V path marked A provides a conservative assessment). The theoretical failure envelopes of the form shown in Figure 19 can be re-expressed as equivalent friction factors, varying with load path and load level, W /Vult . For illustration, results have been derived from the unbonded envelopes corresponding to a smooth pipe interface, uniform soil strength and embedments of w/D = 0.25 and 0.45. Load paths varying from dV/dH = 0 to dV/dH =−2 (i.e. a lift force of twice the drag) have been used (Figure 20). This form of presentation clarifies how much the seabed response differs from the simple Coulomb friction that is often assumed. As the pipe weight reduces, the equivalent friction factor rises, particular at low load levels. The effect of load path is shown in Figure 21 which compares µeq for two ratios of dV/dH, using the pure drag case of dV/dH = 0 for normalisation. The load path has a significant influence on the equivalent friction factor at breakout. For the load path with an uplift of dV/dH = −2, the equivalent friction at failure is typically 1.5-2 times higher than pure drag at w/D = 0.25. The difference rises to >2.5 for w/D = 0.45. At low loads, however, the load path does not affect the equivalent breakout friction because the same tangential mechanism applies. This limiting friction factor at low loads that is caused by the tangential mechanism is not captured in conventional ‘friction + passive’ models for breakout resistance. The component of passive resistance in these models is uninfluenced by the vertical load level, leading to a finite breakout resistance under a vertical load of zero. This is incorrect and unconservative for stability design. The tangential failure mechanism leads to zero breakout resistance under zero vertical load for any embedment of w/D < 0.5, if tension cannot be sustained at the pipe-soil interface (regardless of the pipe roughness).
Figure 20. Equivalent friction factors from failure envelopes.
2009). For this case the failure envelope is rotated by the slope angle and changes shape slightly, depending on the ratio of soil strength to weight. Due to the curved shape of the failure envelopes, the equivalent friction factor for lateral failure in a downslope direction is not simply the value for level ground, adjusted by the slope angle according to Coulomb friction (unless failure is via the tangential mechanism). 4.2 Failure envelopes as friction factors Failure envelopes depict the limiting pipe-soil loads in a different manner to the empirical expressions that are commonly used in practice (e.g. DNV 2007a, Verley and Sotberg 1994, Verley and Lund 1995, Bruton et al. 2006). These approaches divide the soil resistance into frictional and passive resistance components that are superposed. It is then common for the resulting resistance to be expressed as an equivalent friction factor, by dividing by the vertical pipe-soil load. In this way, the geotechnical resistance is linked only to the pipe weight (and lift), and any structural analysis software need not include any soil parameters, nor incorporate explicit calculations of the pipe embedment. The theoretical failure envelopes can also be manipulated to express the pipe breakout resistance as an equivalent friction factor, µeq . For a given initial state, the equivalent friction at failure, µeq = H/V depends on the load path in H-V space. The initial state is typically well within the failure envelope (W /Vult << 1) due to the unloading beyond the maximum overstress of the catenary, dynamic lay effects and also from any gain in soil strength after laying. A general parameter to describe the current level of unloading is the overloading ratio, R = Vult /V, where V is the current vertical load and Vult is the current vertical bearing capacity (Zhang et al. 2002a; Zhang and Erbrich 2005). In lateral buckling analysis, the load path is generally assumed to involve horizontal loading, whilst the vertical load remains constant and equal to the pipe submerged weight (i.e. dV/dH = 0). In Figure 19 this
99 © 2011 by Taylor & Francis Group, LLC
4.3
Table 2.
Coupling of geotechnical analysis and structural and hydrodynamic responses
To properly capture this strong influence of load path in an assessment of pipeline stability it is necessary to couple the failure envelopes with the imposed hydrodynamic loads. As a minimum, the load path of the critical wave condition can be combined with the failure envelopes to derive a single value of equivalent friction factor that is then used in an absolute stability check. The conservative approach of assuming a load path of dV/dH = 0 will provide a safe under-prediction of the available breakout resistance (Figure 19). Adoption of the actual load path when assessing µeq will reduce this conservatism. To provide information beyond simply the breakout resistance, the failure envelopes can be incorporated within a plasticity macroelement model, which allows the full load-displacement response to be simulated (e.g. Schotman and Stork 1987, Zhang et al. 2002b, Cocchetti et al. 2009). In essence, these models allow the pipe embedment and therefore the size and shape of the failure envelope to be updated whenever the load reaches the current failure envelope and displaces. If the load remains within the failure envelope the pipe is stable. The most simple use of this type of model in design is to track the response of a single pipe element whilst it is subjected to the most severe hydrodynamic conditions, unsupported by any neighbouring pipe (which is a conservative simplification). A more sophisticated approach is to attach multiple pipe-soil elements to a full structural model of the pipeline, allowing longitudinal load-shedding to be incorporated. This approach is described by Tian and Cassidy (2008) who incorporated the Zhang et al (2002b) model for drained pipe-soil interaction into a structural analysis of a pipeline subjected to hydrodynamic loading. This allowed full wave-pipe-soil interaction to be simulated using a more sophisticated pipe-soil model than had been previously used (an inevitable limitation being that the wave-seabed interaction depicted in Figure 3 is not captured). 4.4 Quantifying uncertainty in lateral breakout resistance: Monte Carlo analysis As discussed in Section 3.3, the as-laid embedment of a pipeline is strongly influenced by the dynamic effects in the touchdown zone during laying and the consequent reduction in soil strength. These processes depend on the laying procedures and the metocean conditions, which are not known during design. Similarly, the operative soil strength during breakout is subject to uncertainty, partly from natural variability and partly due to potential reconsolidation of the soil near the pipe after lay-induced disturbance. Despite these difficulties it is necessary to provide some assessment of the pipeline embedment and the consequent breakout resistance during design. This assessment can be performed within a probabilistic framework, with the uncertainty in each input
Parameter
Value
Pipeline diameter Submerged pipe weight, W Specific gravity, SG Touchdown lay factor, klay Soil submerged unit weight, γ Buoyancy factor, fb
0.5 m 1.5 kN/m 1.38 1.25 8 kN/m3 1.5
Table 3.
Probabilistic parameters for breakout assessment. Statistical values
Operative soil strength profile during laying Operative soil strength profile during breakout
Parameter
P5
P95
Gradient, ksu (kPa/m) Dimensionless gradient, κ = ksu D/sum Mudline value, sum (kPa) Dimensionless gradient, κ = ksu D/sum
1.5
3 20
2
4 0
parameter being given a statistical description. Such an approach is consistent with the reliability-based approaches being used for the structural aspects of pipeline design (e.g. Carr et al. 2004, DNV 2007a, AtkinsBoreas 2008, Rathbone et al. 2008). The recently-developed solutions for pipeline embedment and combined V-H capacity provide a robust theoretical framework within which a probabilistic assessment of the pipeline response can be made.This is illustrated by an example of a 0.5 m diameter smooth pipeline with the deterministic parameters shown in Table 2. The soil strength profiles relevant to the embedment and the breakout events are given a probabilistic description, as set out in Table 3. For simplicity, only the soil strength profiles are treated probabilistically, although it is straightforward to include other parameters that have natural variability, such as the soil unit weight. The adopted values are typical for a flowline being laid on a soft fine-grained soil in deep water. The operative soil strength profile, representing the disturbed state during laying, increases with depth and has a small mudline intercept. The operative value during breakout is uniform and higher. The upper bound (UB) and lower bound (LB) parameters for each strength profile differ by a factor of two and are assigned 5% and 95% exceedence values (i.e. the LB P5 value has a 5% probability of exceeding the actual soil strength). This ratio between UB and LB values is broadly typical of practice, but it is strongly affected by the level of natural variability and the spatial intensity and quality of site investigation data. The probabilistic soil strength parameters might be derived from in situ and ex situ testing on samples distributed along the entire pipeline length, which will then capture scatter related to the natural variability
100 © 2011 by Taylor & Francis Group, LLC
Deterministic parameters for breakout assessment.
Figure 22. Strength distributions in probabilistic analysis case.
(as well as scatter from the tests themselves). It is necessary also to include within these soil strengths some uncertainty to capture the lay process, and any reconsolidation. Having chosen P5 (LB) and P95 (UB) values of soil strength (or any other pair with a specified likelihood, representing upper and lower bounds) a probabilistic distribution must be fitted. In this case a gamma distribution has been chosen because it does not feature negative values (since negative values of strength are not possible). A gamma distribution is also free of the long positive tail that is present in a lognormal distribution (which is sometimes adopted for non-negative variables). The resulting cumulative distribution functions for each soil strength parameter are shown in Figure 22. Using these distributions coupled with the embedment and failure envelope analyses described earlier, a Monte Carlo analysis has been performed using a software program called MCWHIPLASH, developed by Advanced Geomechanics, to illustrate the influence of soil strength uncertainty. The analysis involved 10,000 realisations, which is sufficient to generate smooth distributions of the outputs. The first step of the calculation involves estimation of the as-laid embedment. The probabilistic operative strength leads to the variation in embedment shown in Figure 23. The as-laid embedment varies monotonically with the single variable – the operative strength during laying – in a non-linear manner. Although the P95 and P5 strength parameters differ by a factor of 2, the ratio between the P95 and P5 embedments is only 1.38 (Figure 23b). This reflects the non-linearity of the bearing capacity expression (Equation 2). A halving of the soil strength, for example, does not lead to a doubling of the embedment because the bearing capacity factor rises with depth. Also, the strength of the soil is not the only mechanism supporting the pipeline – the soil buoyancy has an influence and this is not given a probabilistic variation in this example. In practice, the soil unit weight has a narrow range of variability, so the soil buoyancy term generally attenuates the variability arising from the soil strength. Overall, there is a narrowing of the
Figure 23. Variation in as-laid embedment for example probabilistic analysis.
outputs relative to the inputs due to the particular form of the mechanisms that govern pipe embedment. The next step in the analysis is the calculation of breakout resistance, which is expressed here as an equivalent breakout friction factor, Hbrk /V. In this example, the load path to breakout has been assumed to be purely horizontal loading under a maintained vertical load corresponding to the submerged pipe weight (V = W ), which is the usual assumption in a lateral buckling analysis. The breakout resistance is affected by both the aslaid embedment and also the (probabilistic) strength at the time of breakout, which are assumed to be uncorrelated in this example, for simplicity. The variation in breakout friction factor with as-laid embedment is shown in Figure 24a and the variation with soil strength at breakout is shown in Figure 24b. The resulting distribution of friction factor is shown in Figure 24c. The overloading ratio, R = Vult /V, varied between 2.3 and 9.5 at breakout in this example, with a mean value of 4.9. Consequently, none of the Monte Carlo realisations involved failure via the tangential cut-off mechanism (indicated by the cap on Hbrk /V for w/D = 0.25 in Figure 20). A trend line through the centre of the clouds of results in Figure 24a and Figure 24b indicates that the breakout friction factor increases approximately linearly with both soil strength and embedment (although this is not the case for other examples that could have been adopted).
101 © 2011 by Taylor & Francis Group, LLC
Figure 25. Normalised variability of input and output parameters in example probabilistic analysis.
Figure 24. Variation in breakout friction factor for example probabilistic analysis.
For this example, the P5 and P95 values of Hbrk /V are 0.94 and 1.45 respectively, which have a ratio of
© 2011 by Taylor & Francis Group, LLC
102
1.54.As for the embedment, this variability is narrower than the range of input strengths. The cumulative distributions for the strengths and the resulting embedment and breakout friction factor are compared in Figure 25, with each variable being normalised by its P50 value. This form of presentation highlights the relative uncertainty in each parameter. As noted previously, the analysis leads to a narrower output range than input range: a given uncertainty in soil strength leads to a lower uncertainty in breakout friction factor. This is a useful outcome, given the difficulty in assessing accurately the soil strength close to the seabed. This effect can be contrasted with the response of a surface foundation (with or without short skirts). The lateral capacity of the foundation varies in proportion to the surface soil strength, so the soil strength and the foundation capacity have the same level of variability. The deterministic results based on combinations of the P5 (LB) and P95 (UB) values are also marked on Figure 24a and Figure 24b. These results highlight the value of a Monte Carlo analysis. The friction factors found using the extreme values of soil strength in combination (i.e. LB/LB and UB/UB) lie within the centre of the distributions (at the P34 and P55 percentiles respectively). These do not provide a useful indication of the potential variability. The friction factors calculated using the more unlikely assumptions of the UB/LB and LB/UB soil strength combinations give breakout values that have percentiles of P1 and P99 . The (un)likelihood of the results found using these assumptions is also strongly dependent on the particular combination of pipe and soil parameters being used, due to the non-linearity of the governing mechanisms. There is therefore no simple method for linking the probability of the inputs to the probability of the outputs, without a Monte Carlo analysis or a similar statistical technique. If it is assumed that the operative strengths during laying and breakout are correlated, it might be assumed that the LB/LB and UB/UB cases would then provide a true indication of the potential variability in Hbrk /V.
However, this is not the case. In fact, for the example shown here, with fully correlated strengths, the highest value of Hbrk /V is found for an analysis based on the P31 strength during laying and breakout. The maximum breakout resistance does not arise from either the weakest or strongest soil, but an intermediate case. Why does this surprising observation arise? Soft soil gives high embedment, and therefore a high wall of passive soil on breakout. However, the beneficial embedment is countered by the weakness of the passively loaded soil. Due to these counteracting effects coupled with the non-linear shape of the V-H failure envelope (and its non-linear growth with embedment), the extreme values of breakout resistance are often not associated with the extreme values of soil strength. This same interaction underlies the narrower range of breakout friction factors compared to the input range of soil strengths. The form of Monte Carlo analysis shown here provides a solution to this difficulty, capturing the uncertainties associated with soil strength and other parameters that influence pipe embedment and breakout. The simple example shown here includes only the variability associated with soil strength. In practice the other geotechnical inputs such as the soil unit weight can also be treated probabilistically. It is also possible, using MCWHIPLASH, to incorporate other parameters such as the pipeline bottom tension during laying, via the solution given in Equation 1. The uncertainty associated with the geotechnical parameters arises from the site variability and the quantity and quality of site characterisation data. To provide a consistent approach throughout the entire analysis process – from interpretation of the geotechnical SI data through to the pipeline structural analysis – a probabilistic approach can also be used to assess the soil strength parameters (e.g. Lumb 1966, DNV 2007b, Lacasse et al. 2007). However, these methods do require a sufficient quantity of site characterisation data, which is often not the case. The uncertainty related to the changes in soil strength during laying and recovery can be reduced through model tests that simulate these processes (White and Gaudin 2008). 4.5 Large amplitude monotonic lateral response The engineered buckles that are often used to relieve thermal and pressure-induced expansions typically involve lateral movements of several pipe diameters, as is evident in the picture within Table 1. This example is a buckle that was initiated over a sleeper, which provides a vertical upset and therefore reduces the buckling load locally, providing a reliable buckle initiation point. Only the central part of the buckle is lifted away from the seabed, and a significant length of grounded pipeline sweeps back and forth across the seabed as the buckle expands and contracts during operating cycles. Buckles are sometimes initiated on-bottom at lay route curves, and designers always need to consider the likelihood of a ‘rogue’ buckle being initiated anywhere along the route, where the
Figure 26. Deformation mechanisms during large amplitude lateral sweeping of a pipe on soft clay.
pipe rests on the seabed. It is therefore necessary to assess the lateral pipe-soil resistance over long distances of lateral movement across the seabed. Flowlines are typically designed for several hundred startup and shutdown events throughout the field life. During the early years of operation these will be more frequent. In later years, as the operating temperature and pressure reduce, the feed-in to the buckle may reduce. However, if new wells are tied-in, or if the field recovery is enhanced, then the length of feed-in into a buckle, and the consequent amplitude of movement, may increase in later life. As discussed previously, the overloading ratio of a pipeline during operation is typically greater than 2, which puts the load point on the ‘dry’ side of the unbonded breakout failure envelopes (V/Vult < ∼0.45). In this case the pipe moves upwards at failure, leading to a reduction in resistance after breakout as the pipe rises towards the ground surface. Model tests using PIV image analysis have shown that in soft clay the steady failure mechanism involves basal sliding of a berm of soil ahead of the pipe, with a thin layer of additional material being ploughed from the seabed (Figure 26a). The same mechanism has been replicated in large deformation finite element analysis (Figure 26b). The subsequent steady ‘residual’ resistance depends on the size of the berm and
103 © 2011 by Taylor & Francis Group, LLC
the strength of the soil within it (as well as the seabed strength). The size of the berm depends on the initial embedment of the pipe and its trajectory, whilst the strength within the berm depends on the remoulding it has undergone, and the sensitivity of the soil. At a given point in a large amplitude lateral sweep, these two effects can be quantified by defining an ‘effective embedment’, w /D, which amalgamates the embedment below the original soil surface, w/D, and an additional component of embedment arising from the berm, hberm /D (Figure 26c). The volume of the berm is calculated from the area swept by the pipe invert. By assuming that the berm has a particular aspect ratio, η = length/height, its height can be found from the area. This height is discounted by a sensitivity, St,berm , which represents the reduction in berm strength due to remoulding, to give the effective additional embedment, hberm /D:
This normalisation approach provided excellent agreement between a set of centrifuge model tests on soft clay, simulating pipes of different weights and initial embedments (White and Dingle 2010). The tests all showed differing values of steady residual lateral resistance, ranging from Hres /V = 0.32 to 0.65 (Figure 27a, b). There was no clear influence of simulated pipe weight, V, but a scattered trend with initial embedment, (w/D)init . However, the profiles of normalised lateral resistance, H/su D, when plotted against effective embedment, match closely regardless of the vertical load (Figure 27c). All of the responses lie close to a power law relationship between H/su D and w /D. The same behaviour is seen in large deformation finite element analysis (Wang et al. 2010). This effective embedment approach alone does not provide a predictive tool, only a robust normalisation of the response. A method for calculating the pipe trajectory is required, in order that the evolving size of the soil berm and the local embedment can be assessed. This trajectory is affected by the pipe weight, even though the resistance at a given local embedment appears not to be. This missing link has been filled by project-specific empirical models, and is currently being tackled within the SAFEBUCK JIP, via the development of a plasticity macroelement model. This model will operate in the usual macroelement manner of combining yield envelopes, a hardening law and a flow rule to describe the general force-displacement response of an element of pipe (following Zhang et al. 2002b). However, it will incorporate the influence of the changing soil strength and berm geometry on the shape of the failure envelope. This will allow the macroelement approach to be extended to large amplitude movements. In the meantime, various empirical expressions have been proposed to estimate the residual lateral resistance with reference only to the as-laid state (not the post-breakout trajectory) (e.g. Bruton et al. 2006,
104 © 2011 by Taylor & Francis Group, LLC
Figure 27. Centrifuge modelling observations of residual lateral resistance on soft clay (White and Dingle 2010).
Cardoso and Silviera 2010). Based on the six tests on soft kaolin clay shown in Figure 27, White and Dingle (2010) proposed the a simple relationship between initial embedment, (w/D)init , overloading ratio, R, and residual lateral friction factor, Hres /V:
with suggested parameters of µ = 0.3 and k = 2 (Figure 27d). The form of Equation 4 encapsulates some of the factors that influence the residual resistance: a high (w/D)init leads to a large initial berm size and a low overloading ratio, R, leads to a deeper pipe trajectory; both of which increase the passive horizontal resistance. An obvious limitation of this tentative expression is that it does not include any soil parameters, so does not explicitly capture the varying strength as soil is remoulded ahead of the pipe, or the variation in soil strength with depth. Surprisingly, this empirical expression has a different form to another one derived by a recent confidential study for the SAFEBUCK JIP (White and Cheuk 2009), and also from that published by Cardoso and Silviera (2010). In all cases these expressions are based on a particular collection of model tests, rather than any theoretical mechanism. Such contrasts provide a warning that these empirical expressions may not be as accurate when applied beyond the conditions in which they were calibrated.
Figure 28. Typical full scale modelling observations of cyclic lateral response on soft clay (White and Cheuk 2008).
4.6 Large amplitude cyclic lateral response During large amplitude cyclic lateral movements, the pipe-soil resistance remains dependent on the berm material being swept back and forth. Each time the pipe changes direction the berm is left behind, and is remobilised if the pipe approaches that point during a subsequent cycle. These aspects of behaviour are illustrated by a typical model test involving cycles of large amplitude lateral movement, between fixed displacement limits (Figure 28, White and Cheuk 2008). This test involved cycles of fixed lateral amplitude under constant simulated pipe weight. The general form of the response involves initial breakout of the pipe (point A in Figure 28) followed by a gentle increase in resistance associated with the growth of a small ‘active’ berm ahead of the pipe (B). On reversal of the sweeping direction, this response is repeated (C) and the berm generated during the previous sweep is left behind, becoming ‘dormant’. When the pipe again approaches this point during a later sweep, an increase in resistance is experienced as the dormant berm is collected (D). With repeated cycles of movement, the berms at the limits of the pipe movement grow, causing a corresponding increase in resistance. The first sweep encounters slightly higher steady resistance than later sweeps due to the larger active berm arising from the initial embedment of the
105 © 2011 by Taylor & Francis Group, LLC
Figure 29. Soil berms during large amplitude lateral movement.
pipe. An image of berms created by this type of large amplitude pipe motion are shown in Figure 29a. In the design of lateral buckles, it is important to model the constraint imposed by the soil berms, in order to provide an adequate assessment of fatigue. If the pipe-soil resistance is assumed to be constant,
with the berms ignored, then a buckle will progressively lengthen through cycles of expansion and contraction, which reduces the peak bending stresses near the crown (Cardoso et al. 2006, Bruton et al. 2007). Soil berms inhibit this lengthening, causing the high stresses generated during buckle initiation to be locked-in. Although the restraint provided by the berms also attenuates the amplitude of the pipe motion within each cycle – and therefore the cyclic stresses – the overall effect on fatigue is usually harmful, due to the higher mean stresses that are locked-in (Bruton et al. 2007). Models for the cyclic large-amplitude lateral behaviour can be based on the accumulation and deposition of berm material, as shown schematically in Figure 29b. The current berm size can be used as a hardening parameter that governs the passive resistance, rather like in Equation 4. For undrained conditions, the rate that the berm grows with lateral pipe movement is equal to the depth of soil scraped away by the pipe, from conservation of volume (White and Cheuk 2008). Re-consolidation of the soil that has been remoulded and transported ahead of the pipe will increase the berm resistance. The cyclic resistance is not only affected by the changing geometry. Depending on the soil type, pore pressure dissipation may occur during lateral sweeping, and is also likely to occur between startup and shutdown events. This leads to reconsolidation of the disturbed soil within the berm, and also swelling of the unloaded seabed that is exposed by the scraping action of the pipe. For design, the variation in lateral resistance can be estimated from a geotechnical analysis that considers in detail the mechanisms described above. Then, to incorporate the results into the structural analysis of the pipeline, they can be converted into more simple relationships. For example, the results of a geotechnical analysis can be converted into equivalent friction factors (H/W ) for the residual and berm resistance as a function of cycle number (Bruton et al. 2009). Over the design life of a lateral buckle, the depth of the trench created by the sweeping action of the pipe can be significant. As the trench deepens, soil debris may collect in the base, raising the residual resistance. Soil may even flow over the crown of the pipe. The seabed topography near the crown of an on-bottom buckle that has been operating for several years is shown in Figure 30 (Cardoso and Silviera 2010). The pyramidal berms on each side of the pipe are similar in shape to the model test observations in Figure 29. The pipe itself is almost hidden from view, being partly buried under softer soil that has accumulated within the trench. The situation evident in Figure 30 is analogous to the cyclic rod penetrometer results shown in Figure 8. Repeated remoulding and reconsolidation of the seabed will tend to increase the soil strength and consequently increase the constraint on the pipe. The accumulation of pipe embedment during a set of full-scale and centrifuge model tests is shown
Figure 30. Topography of a soft clay seabed after several years of operation of a lateral buckle (Cardoso and Silviera 2010).
Figure 31. Accumulation of pipeline embedment during cyclic large-amplitude sweeping: SAFEBUCK model test database (White and Cheuk 2009).
in Figure 31. These results are from a database of large-amplitude cyclic lateral tests collated by the SAFEBUCK JIP. These tests involved soft finegrained model seabeds. Short elements of pipe were modelled, exerting a constant bearing pressure in the range V/D = 0.5– 8 kPa. Fixed-amplitude lateral cycles of 1–10 pipe diameters were imposed. The pipes embedded at rates
106 © 2011 by Taylor & Francis Group, LLC
of 1–50% of the pipe diameter per cycle and there was no tendency for a stable embedment to be reached. It is worth noting that these tests involved up to only 60 cycles of movement, whereas most pipeline systems are designed for an order of magnitude more startup and shutdown events. An effect not accounted for in these tests was the longitudinal flexural stiffness of the pipe. In all tests the simulated pipe weight was held constant. In practice, any vertical movement of the pipe down into the trench can lead to a reduction in the vertical pipe-soil contact force, with load being transferred longitudinally to adjacent sections of pipe. This in turn will attenuate the rate at which the pipe embeds. This soil-structure interaction is difficult to capture in design without a pipe-soil interaction model that simulates the trajectory of the pipe. The macroelement plasticity models (e.g. Zhang et al. 2002b; Tian and Cassidy 2008) have this capability, but have not yet been extended to capture large deformation effects related to soil berms and remoulding. In cases where the pipe settlement has been considered important to capture, some recent projects have used a cycle-by-cycle approach in which the trench geometry is updated after each startup and shutdown, based on a geotechnical analysis of the pipe trajectory that is performed independent of the structural analysis (Bruton et al. 2009). This approach allows the changing vertical pipe-soil contact force to be captured through a structural analysis of many operating cycles, albeit in a manner that requires significant user intervention. Given the settlement implied by extrapolation of Figure 31 to the typical design number of operating cycles, it is likely that some longitudinal load shedding will occur during the design life of most on-bottom lateral buckles. Centrifuge and large-scale model testing is currently a widely-used technique to aid the assessment of cyclic large-amplitude lateral pipe-soil resistance (White and Gaudin 2008; Langford et al. 2007). In the long term, it is envisaged that analysis techniques for this behaviour will become more well-established, and model testing will not be required as often. However, given the complexity of the mechanisms shown here, and the difficulty of predicting the behaviour based on conventional soil parameters alone, it is currently common to perform project-specific model tests to assess appropriate cyclic pipe-soil model parameters. 5 AXIAL PIPE-SOIL INTERACTION 5.1 Effective stress and total stress models The axial resistance between an on-bottom pipeline and the seabed affects the feed-in response towards lateral buckles and also controls the end expansions. In addition, it influences the ‘walking’ behaviour through cycles of startup and shutdown (Carr et al. 2006). Low axial friction can be particularly problematic, and the relationship between pipeline walking rate and axial pipe-soil resistance is generally non-linear. The rate of
walking per cycle can increase rapidly as the available axial resistance decreases (Bruton et al. 2007). Compared to the lateral behaviour, the axial pipesoil response involves a more simple geometry, with failure being constrained to occur at, or close to, the pipe-soil surface. By analogy with pile design, both α (total stress) and β (effective stress) approaches can be considered. The β-method expression for the ultimate axial resistance per unit length, T, is:
where µ is the pipe-soil friction coefficient, which can be alternatively expressed in terms of a pipe-soil friction angle, δ, where µ = tan δ. The parameter ζ is a factor that accounts for the enhancement of the normal pipe-soil contact force due to a ‘wedging’ action (White and Randolph 2007). The α-method expression is
where αsu is the shear stress acting on the pipe surface at failure and DθD is the pipe-soil contact length around the pipe perimeter, with θD being the angle subtended by radii to the limits of the pipe-soil contact. The adhesion factor, α, captures the roughness of the surface (i.e. the relative strength of pipe-soil and soil-soil shearing). Also, if su is taken as the in situ soil strength then α also encompasses any differences between that strength and the strength of the soil at the pipe surface, due to downdrag, and the processes of remoulding and reconsolidation since the pipe was laid. The argument in support of β-methods is rooted – perhaps idealistically – in the fundamental concept that soil strength is controlled by effective stress friction. In addition, pipeline movements can often be so slow that drained conditions prevail even in fine-grained soils. However, excess pore pressures are generated at even modest speeds in some clays. For this situation, the β-method can be adjusted by using an excess pore pressure ratio, ru = u/σn,av , based on the average excess pore pressure (u) and total stress (σn,av ) around the pipe surface: Equation 5 is then multiplied by (1 − ru ). If the response is fully undrained it is instead tempting to use an analysis based on the relevant undrained shear strength. Estimation of the reconsolidated su beneath the pipe might be thought easier than assessing ru , although the two types of calculation are essentially interchangeable and based on the same principles. As a pipe penetrates the seabed, the surficial soil is dragged downwards. As a consequence, the pipe is bearing not on soil from below the surface, but on soil that was previously at or very close to the surface – as illustrated by the experimental data (Dingle et al. 2008) and numerical results (Zhou et al. 2008) shown in Figure 32. Also, the excess pore pressure generated during the undrained penetration process will take time to dissipate. Solutions for the dissipation of excess pore pressure around a pipe based on an elastic soil
107 © 2011 by Taylor & Francis Group, LLC
Figure 32. Seabed distortion after vertical pipe penetration (Dingle et al. 2008 (left) vs. Zhou et al. 2008 (right)).
Figure 34. Interface shearing: kaolin clay and Storæbelt clay till.
Figure 33. Post-laying pore pressure dissipation around pipe perimeter (Gourvenec and White 2010).
model are presented by Krost et al. (2010) and Gourvenec and White (2010). Figure 33 shows the variation in normalised excess pore pressure, averaged around pipe, with dimensionless time. These dissipation curves allow the level of consolidation at the pipe-soil interface prior to any in-service movements to be assessed – which is analogous to the ‘set-up’ of a driven pile. The pipe weight often applies a bearing pressure that exceeds the apparent pre-consolidation pressure of the seabed, so the soil strength will generally increase during dissipation of the excess pore pressure created by the laying of the pipe. A final complication is that if the response is undrained, then the state of the soil beneath the pipe will inevitably change through the hundreds of operating cycles, and therefore hundreds of episodes of failure followed by reconsolidation. The tendency for excess pore pressure to be generated during axial movement will be reduced, as described below. 5.2
Interface shearing and drainage
The response of a fine-grained soil during interface shear varies between fully-drained and fully undrained across a range of velocities that typically spans two or three orders of magnitude. It is useful to consider firstly the response in tests where the apparatus forces the failure to occur on a particular plane, rather than the case of a curved pipe surface on the seabed. The interface direct shear box tests from Figure 10 are shown again in Figure 34, and supplemented by results from shearing of a concrete – clay till interface at varying rates, reported by Steenfelt (1993). This study
related to the sliding capacity of shallow foundations supporting the Storebælt Crossing in Denmark, but the results provide useful insights relevant to axial pipe-soil interaction. The responses of both the kaolin clay and also the Storebælt clay till are very similar, in terms of both the limiting drained and undrained resistances and also the velocity range over which this transition occurs. A hyperbolic relationship provides a convenient fit to the response. However, these results are not generally applicable to all soils. Other proprietary tests show the drained-undrained transition at a velocity that is two orders of magnitude different, reflecting the particular characteristics of the soil. It is not immediately obvious how to normalise the drained-undrained transition for shearing on a plane. The soil permeability, k, is a relevant parameter, since it controls the rate of pore water flow and excess pore pressure dissipation, but it can be argued that the stiffness may not be: the volume of pore water to be expelled may depend principally on the shear zone thickness (and its potential to dilate or contract), rather than the bulk stiffness of the soil. This argument is pursued by Palmer (1999) in his analysis of ploughing and is discussed in Section 6. If the soil stiffness is not relevant then the coefficient of consolidation is not required within the normalisation, unlike the cases shown in Figure 9. It remains to be established whether the appropriate dimensionless group to normalise the velocities in Figure 34 is vL/cv (with L being some relevant drainage distance) or v/k. In practice, it can be more efficient to perform multiple interface shear tests at varying velocities (and at the low stress relevant to pipelines) in order to identify this response directly, rather than assessing it indirectly from a normalised theoretical curve via measurements of cv and k. The form of response shown in Figure 34 can be used to assess the relevant axial resistance for the initial axial movement of a pipeline. Over a longer period, and after a series of movements, this relationship no longer applies. The apparent friction coefficient during undrained and partially drained sliding is lower than the drained case due to the generation of positive
108 © 2011 by Taylor & Francis Group, LLC
pore pressure when shearing is initiated. Subsequent dissipation of this pore pressure leads to a rise in effective stress. This is accompanied by contraction of the interface zone, which can be observed as settlement of the platen in a direct shear test. An example of this behaviour is shown in Figure 35. These are results from shearing of normally consolidated kaolin clay on a steel interface.The shearing rate was close to the fully undrained limit and a sequence of 2 cycles of +/−5 mm is shown. No consolidation period was permitted between cycles. The initial apparent friction (based on total stresses) is τ/σn = 0.28. However, over the following two cycles the resistance rises (Figure 35a) whilst the sample settles (Figure 35b). The rate of settlement is approximately constant through the two cycles, with a slight tendency for more rapid settlement following each reversal. Meanwhile, the absolute shearing resistance, when plotted against the cumulative horizontal movement, shows a continuous rise, interspersed with reductions to zero at each reversal point (Figure 35c). After only 35 mm of movement, or an elapsed time of approximately 3 hours, the apparent friction coefficient has risen from 0.28 to 0.46 (Figure 35d). This response can be illustrated schematically in stress – specific volume space (Figure 36a). For the type of test shown in Figure 35, the stress path initially heads to the critical state line in an undrained manner (path OU). If sliding continues, then the state moves along the critical state line (CSL) following the increase in effective stress that accompanies the dissipation of excess pore pressure (UP). Ultimately the test path and the strength (or apparent friction) reaches the same state as a drained test would (point D) (Figure 36b). The drained test follows the path OD. At point D the effective stress equals the applied total stress. The idealised strength response in Figure 36b ignores the possible influence of a changing mobilised friction angle, and we are also not differentiating between a critical state line approached after monotonic shearing, and one at a denser state approached through cyclic shearing. 5.3 Episodic interface shearing and consolidation The three test paths shown in Figure 36 cover three possibilities for an initial episode of sliding. For the fully undrained case, after sliding halts, the consolidation will be accompanied by contraction following an unload-reload (κ) line (U-R, Figure 36a) – exactly analogous to the reconsolidation stages shown in Figure 7. Following this reconsolidation, a smaller excess pore pressure and hence a higher undrained strength will be mobilised in a subsequent fast sliding event – just as shown in Figure 6. These idealised responses of a single soil element through episodes of undrained sliding and reconsolidation can be used to estimate the long-term cyclic axial pipe-soil resistance during movements at undrained rates. Eventually, the process of undrained shearing
109 © 2011 by Taylor & Francis Group, LLC
Figure 35. Settlement and apparent friction behaviour during interface shearing of kaolin clay.
followed by reconsolidation will take the soil state to the point on the critical state line at the applied total normal stress. In this case there is no longer a tendency for excess pore pressure to be created during shearing, and the apparent friction will be the drained value regardless of the shearing rate.
Figure 36. Critical state interpretation of drained, undrained an partially-drained interface shear tests.
5.4 Application to axial pipeline resistance These observations from shear box testing are not directly applicable to pipeline behaviour due to the potential for failure to occur on multiple planes at different radii from the pipe surface. At first sight, Figure 35 and Figure 36 suggest that once a drained failure condition has been reached then the apparent sliding friction will always be the drained value irrespective of the rate of movement, since the soil state at the interface has reached the critical state line, and has no tendency to contract or dilate. However, in a shear box test the failure plane is prescribed by the boundary conditions of the apparatus, so only this single plane needs to be hardened. In contrast, failure can occur at different radii close to the surface of a pipeline. If the pipe surface is rough then both pipe-soil shearing and soil-soil shearing have comparable resistance, and failure will occur at whichever circumferential surface offers the least resistance. We illustrate how this affects the hardening behaviour by considering a pipe that undergoes intermittent fully undrained movements, with intervening consolidation periods. For simplicity we are setting aside the effect of partial drainage during shearing. The shear stresses created within the soil by the axial traction at the pipe-soil interface decay approximately with the inverse of the radial position. Given that a shear plane is typically a fraction of a millimetre in thickness in fine-grained soils, there are many potential shear planes carrying essentially the same shear
© 2011 by Taylor & Francis Group, LLC
Figure 37. Schematic illustration of successive undrained failures on adjacent surfaces during axial pipeline movement.
stress. If the soil is assumed to be uniform, then an initial axial movement, fast enough to be undrained, will cause failure within the shear zone immediately adjacent to the pipe (zone A in Figure 37a, b). The mobilised axial resistance will be controlled by the effective stress during undrained failure of this element, which will follow the stress path OA in Figure 38. After a subsequent period of pore pressure dissipation, shear zone A will have contracted (Figure 37c) and hardened to point A , under the stress of σn imposed by the pipe (Figure 38). However, although this element is now able to offer increased shearing resistance during a subsequent cycle, zone B has not been hardened, and will form the ‘weak link’ during the next undrained axial movement (Figure 37d). This process can continue with each successive cycle leading to hardening of another zone of soil – or, as the shear zones progressively harden, to the remobilisation of a previously-sheared zone. Ultimately all zones will reach the steady state (SS) point on the CSL
110
interpret them within the framework described here, to support assessments of axial pipe-soil resistance. 6 TRENCH CONSTRUCTION BY PLOUGHING 6.1
Figure 38. Critical state interpretation of failures on adjacent planes during axial pipeline movement (see Figure 37).
(Figure 38). Hereafter, the mobilised soil strength and hence the axial pipe-soil resistance will be the drained value, regardless of the rate of movement. However, this process of hardening the multiple potential failure zones means that a considerably greater number of cycles and episodes of consolidation are required, compared to in a shear box test. Further research involving model testing, supported by episodic interface shear tests, will shed light on these mechanisms. What we have presented here is clearly idealised, but it captures the most important features of cyclic axial pipe-soil behaviour in finegrained soils, namely: 1. The significant difference between the undrained and drained sliding resistance of ‘virgin’interfaces; 2. The potential for partial drainage within the shear zone during a long slow sliding event; 3. The consolidation and hardening that follows undrained sliding episodes in soft soils (on the wet side of the CSL); 4. The effect of the multiple potential failure surfaces at, or close to, a pipe-soil interface. The critical state framework provides a convenient basis for interpreting and analysing these events. Ultimately, given sufficient cycles of movement and reconsolidation, any shear box or pipe-soil interface should reach a state where there is no further tendency for the shear zone to contract and excess pore pressure to be generated. At this point a sliding resistance equal to the high drained value would be reliably achieved for all rates of movement. However, it remains unclear how many cycles (or what cumulative distance of shearing) is required to reach this state in a given soil. The form of interface shear box test used to illustrate this behaviour in an individual soil element is rarely conducted on fine-grained soils, precisely because of the ill-defined drainage boundary conditions, and the possibility of partially-drained behaviour occurring. In most normal geotechnical applications such behaviour would be unwanted. However, here it illustrates the influence on interface friction of partial drainage from a shear band concurrent with sliding. Given the insights shown here, it can be worthwhile to conduct project-specific interface shear tests and
Pipeline ploughs are used to excavate soil from a trench to permit a pipeline to be lowered below the seabed (Palmer et al. 1979; Reece and Grinsted, 1986; Cathie and Wintgens, 2001) and for burying cables (Jordon and Cathie, 2004). The soil mechanics of ploughing involves large soil deformations, mobilization of peak and residual shear strengths simultaneously in different zones around the share, partially drained failure, strain rate effects on soil strength, and interface shearing. In this section, some of these themes are developed with knowledge gained recently. Plough-soil mechanics is complex and field soil conditions are rarely uniform.The operational environment towing a plough with a long catenary is analogous to a stick-slip resistance on the end of a large spring hanging from a heaving vessel. Trying to keep the plough on track is difficult, let alone making clear field observations and measurements of performance. These conditions can lead trenching contractors to be satisfied with approximate correlations of plough performance in terms of trench depth, tow force applied and progress rate. However, this is unsatisfactory, at least scientifically, and it is certain that improvements can be made to the current public domain methods for pipeline plough assessment, which are summarised by Cathie and Wintgens (2001). A large pipeline plough and the key loads acting on it are shown in Figure 39. The plough advances due to a tow force (T) being applied which overcomes the soil resistance acting on the engaged parts. In normal situations these are the base of the skid and the base and upper surface of the share. Figure 40 is a close-up of a typical share of a pipeline plough for producing a V-shaped trench. Referring to Figure 39, in steady-state ploughing a force and moment balance exists between all the forces acting on the plough. In actual operations the line of action of the base share forces (Shv , Shh ) in particular is continuously changing. The line of action moves towards the tip as the plough pitches forward – for example if the operators raise the skids by rotating the skid linkage counter clockwise in Figure 39 – and towards the rear if the reverse happens. Forward or aft pitch also occurs if the downward component of the soil resistance on the top of the share changes, or if the normal component of soil resistance on the base of the share changes (e.g. if harder or softer soils are encountered). Ploughs appear to move forward in a series of accelerations and decelerations, or start-stops. This is due to the tow wire catenary behaviour, and the time dependent soil shear strength mechanisms (brittle failure, partial drainage, and thixotropy or setup).
111 © 2011 by Taylor & Francis Group, LLC
Mechanics of ploughing
Figure 39. General arrangement of a large pipeline plough showing actions and resistances.
Figure 40. Pipeline plough share (mouldboards that displace the excavated soil to the side of the trench are retracted).
The equations of motion and the catenary response can be coupled with a soil resistance model quite easily in a finite difference algorithm but the unknowns are largely associated with the soil resistance models and the simulation is of academic value except in very deep water where the catenary is particularly long and flexible.
6.2
Existing ploughing resistance model for sand
A ploughing resistance model for sand was described by Cathie and Wintgens (2001) based on some basic theory and calibrated to a limited data set of full size ploughing:
where F is the horizontal component of the ploughing resistance, W is the plough weight (increased by the roller loads i.e. the weight of pipe supported), Cw is a friction coefficient, γ is the soil submerged unit weight, and Cs is a coefficient, similar to a passive pressure coefficient. The first term is, of course, the frictional resistance of the plough and is assumed to be independent of speed. The dynamic resistance component of Equation 7 (Fdynamic = Cd vD2 ) principally arises from the potential of the soil to dilate during shear, and the change in pore pressure (normally suction) and effective stress that this induces, offset by any drainage that may occur. Palmer (1999) has shown that, for a simplified representation of a triangular plough share, the velocity-dependent component of force, Fdynamic , is a function of the forward velocity, v, the soil dilation potential, S and the soil permeability, k:
The function f is linear until pore water cavitation begins to occur in some area around the share. Similar relationships have been developed by van Os and van Leussen (1987), and van Rhee and Steeghs (1991). The former validated the theory with plane strain tests in 2D while the latter performed tests with a plough geometry (i.e. a triangular trench) and performed multipasses to simulate several plough shares at different depth settings. The soil dilation potential, S, or volumetric strain to critical state (Van Leussen and Nieuwenhuis, 1984) may be defined as
112 © 2011 by Taylor & Francis Group, LLC
where ec is the critical void ratio and e is the in situ void ratio. Considering that the value of S can only vary up to about 0.4, while the permeability k can vary by at least an order of magnitude in sands of broadly similar grain size, it is clear that the dynamic resistance to ploughing will be dominated by the soil permeability and variations in this parameter. It is likely that much of the scatter that is observed in progress rates when ploughing in sand is due to small local variations in grain size distribution and therefore permeability. Therefore, Cd , the dynamic force coefficient, will be a function of (S/k), increasing gently with density and increasing strongly with reducing permeability. The term in D3 in Equation 8 arises from a term in D for the magnitude of the suction pore pressure and D2 arising from the cross-sectional area of the trench to be cut. This ignores the actual three-dimensional shape of the share and takes no account of the actual deformation mechanism experienced by the soil around the share. The share can be considered as a horizontally advancing ‘penetrometer’increasing in size up to a certain point. When the share tip first influences the soil, the cavity formed may be considered “deep”, i.e. unaffected by the presence of the soil surface. At this point, the suction pressures will be related to the dimensions of the tip and not the depth below seabed. As the share advances the cavity increases in size and breaks out on the seabed so a depth effect (“shallow” failure) is correct. Considerable forward movement (and therefore time – several seconds) is associated with transitioning from the “deep” to the “shallow” failure at any cross-section, during which time drainage can occur. These 3D aspects of the share geometry are one reason why Cathie and Wintgens considered that the dynamic resistance actually increases with something nearer to D2 (Equation 7) rather than D3 (Equation 8). 6.3 Components of ploughing resistance in sand Further insights into the ploughing resistance formulation are provided by more recent research discussed below. 6.3.1 Static ploughing resistance Model testing in dry sand at 1g performed by Lauder (2010) has provided some useful understanding of drained ploughing resistance and the soil flow mechanism around the share and mouldboards (Figure 41). There can be a large accumulation of spoil above the share. This adds weight to the share (thus increasing share friction), adds weight as a surcharge to the soil being sheared by the share, and adds resistance due to the pressure of the spoil on the mouldboards. The flow mechanisms seen in model testing are confirmed by the pattern of scour marks on full-size plough mouldboards. Figure 42 provides a front view of ploughing showing schematically the area of undisturbed soil in the trench, the disturbed soil (for sand,
Figure 41. Model plough showing accumulation of soil above the share (Lauder, 2010).
this material is likely to be close to the critical state) and the spoil heaps. The effect of any build-up of spoil in the body of the plough was taken account of implicitly in calibrating the lumped Cs and Cw terms of the model of Cathie and Wintgens (2001) (Equation 7). However, one could argue that the drained resistance is made up of several separate components: 1. Share base friction or adhesion due to the downward component of soil resistance on the share face; 2. Shearing resistance of the undisturbed soil ahead of the share (passive pressure) augmented by the weight of the surcharge; 3. Shearing resistance of the disturbed soil acting on the mouldboards. The passive resistance component of Equation 7 (Fpassive = Cs γ D3 ) is the basic component of share resistance and the vertical component also creates additional friction on the base of a share (leading to the first term in Equation 7). This can be assessed approximately using 3D finite element analysis. Results are presented in Figure 43 for a range of peak friction angles representing loose to very dense sand. A Drucker-Prager soil model with a cap was used in ABAQUS. It can be seen that the passive resistance does not increase with D3 and is more correlated with D2 or the trench cross-sectional area, A. The effect of surcharge is not included in this analysis and the soilsteel friction angle was taken as φ – 5◦ which may be rather high for the denser soils. Peng and Bransby (2010) performed 2D FEA modeling and concluded that the static resistance parameter Cs was linearly related to Kp tanφ or K2p for the 2D model, varying with depth as D2 . This study did not include the effect of surcharge. Bransby et al (2005) and Lauder et al (2008) report 1 g model test results that appear to broadly confirm the static resistances suggested by Cathie and Wintgens (2001) but with a margin of uncertainty due to the lower effective stress level and therefore possibly higher operational friction angle in the 1g tests. The formulation of the drained ploughing resistance therefore still needs to be resolved finally. The dimensionless parameter group would be expected to be Fpassive /γ D3 but this does not appear to be the case.
113 © 2011 by Taylor & Francis Group, LLC
Figure 42. View of soil moving operation with a pipeline plough.
Figure 43. Drained resistance of share only as a function of depth and angle of friction.
The effect of surcharge in a two-dimensional idealisation has been considered by Hettiaratchi and Reece (1974, 1975) and is a function of surcharge pressure (q) and trench depth (D). For a 3D diamond shaped share, the surcharge pressure is (as a first approximation) a function of the trench cross-sectional area divided by the trench width (see Figure 42). Therefore, the surcharge pressure is a function of depth (q = f(γ’D)) and the effect on soil passive resistance a function of D2 . Surcharge pressure also increases the vertical load on the share and therefore increases friction. This component would be related to the volume of soil retained above the share and so both D2 and the length of the share, i.e. D3 for a diamond-shaped share. In respect of the component of resistance arising from mouldboard pressure, this should be related to the surcharge depth as discussed above, and therefore the resistance would be a function of D2 . Finally, work must be done to lift the soil from the trench to the spoil heap level. For an increment of forward movement, the volume of soil to be raised is a function of trench area. Assuming that the average distance it must be raised is related to D the contribution to resistance is unsurprisingly related to γ D3 . Clearly, a combination of these terms will control experimental or field data. It is essential to get an improved understanding of these components of
Figure 44. Suction pressures during soil cutting (He et al. 2005) (low and negative to hydrostatic: blue, green, yellow, red).
static resistance and their relative magnitudes before measured data can be properly interpreted to establish the dynamic resistance components. 6.3.2 Dynamic ploughing resistance in sand As noted in Section 6.2, rate-dependent effects in submerged sands are largely due to the suction created by dilation potential during undrained shearing. This subject has been studied extensively in respect of cutting sands during dredging (Van Leussen and Nieuwenhuis, 1984; van Os and van Leussen,1987; Miedema, 1987; Miedema, 2005). Experimental and numerical analysis (e.g. van Os and van Leussen, 1987; He et al, 2005) demonstrates that the normalized cutting speed v/k is a key variable controlling this resistance (Figure 44). All of the research into the cutting of sand for dredging, which is centred on groups based in Delft, has used v/k for normalisation rather than vD/cv , which is more commonly considered for normalizing penetrometer and foundation speed effects in geotechnical engineering. The answer may lie in the localisation during shearing of sand, limiting the dilation to a narrow plane. The resulting pore pressure response is essentially a seepage problem, rather than one of consolidation. Interestingly, there are exceptions within
114 © 2011 by Taylor & Francis Group, LLC
the interface properties: sud = f(S,v/k, φ, δ). Dynamic resistance would then be related to D2 considering the effect of this operative undrained strength on the passive resistance (Hettiaratchi and Reece, 1974) as suggested by Cathie and Wintgens (2001). It is possible that a large component of the dynamic resistance arises from the increased normal stresses on the base of the share due to suction. The use of an operative undrained strength would facilitate an integrated interpretation for ploughing in sands and clays and enable the resistance to be evaluated in terms of a normalised ploughing speed (which would be consistent with the framework shown in Figure 9, which is used elsewhere in pipeline geotechnics). 6.4 Figure 45. Pore water pressures around share during ploughing (low and negative to hydrostatic: black, blue, green, yellow, red; the upper figure is at the share tip, the lower figure is further back along share).
geotechnical engineering: Elsworth and Lee (2005) analyse excess pore pressure generation during cone penetration using only v/k, treating the problem as seepage not consolidation. Cathie Associates has performed 3D finite element modeling of the plough share problem to investigate the dynamic component of ploughing resistance. A Mohr-Coulomb cap soil model with limited dilation was used within ABAQUS. Differing from the He et al. (2005) work where cutting rates are faster, our FEA suggests that maximum suction pressures are relatively localized around the share, although of course the zone would grow for higher speeds or lower permeability. Ahead of the share, the suction pressures are lower due to the increased mean stress resulting from the advancing share. Figure 45 also shows clearly the effect of the free surface higher up the share where suction pressures drain much more quickly. Partial dissipation of excess pore water pressures is limited near the tip but quite substantial further back along the share, particularly near the top. The length of the share is a factor affecting the dynamic ploughing resistance just as it is for static resistance. A further conclusion arising from the 3D FEA work was that the average plastic volumetric strain around the share (a measure of the dilation potential) was not very sensitive to the soil density (or dilation angle) even if very localized high dilations occurred. Finally, the dynamic ploughing resistance also contains a contribution from the dynamic shearing of the accumulated spoil against the mouldboards. In the view of the authors, a promising approach to assess dynamic resistance is to consider a operative undrained strength, sud , that would be a function of the average pore pressures generated around the share (capped in dilatant conditions by cavitation) and
115 © 2011 by Taylor & Francis Group, LLC
Ploughing resistance in clay
Cathie and Wintgens (2001) also proposed an empirical model for ploughing resistance in clays:
where Fw is the adhesion of the underside of the skids and share to the soil during ploughing, Cc is a coefficient similar to a bearing capacity factor, su is the soil undrained shear strength, D is the trench depth, v is the plough speed and Cd is a coefficient relating the strength of the soil at normal shear strain testing rates (su ) to the strength of the soil at ploughing rates of strain. Recent numerical studies by Cathie Associates have highlighted the sensitivity of Cc to the working interface resistance between the share and the soil. Near the tip the resistance is probably close to the intact su . Further along the share the strength almost certainly drops to a residual value, possibly corresponding to the remoulded shear strength. The working interface resistance has a strong impact on the values of Cc computed. Probably insufficient attention has been given to experience and research on tillage tools in the agricultural sector. For example, Karmakar (2005) investigated the behaviour of a tool shown in Figure 46 which has some similarities to a pipeline plough share. He found that brittle failure dominated the response of the soil above the share. While his work was in unsaturated materials, ploughing experience confirms that, in stiff clays, the soil is excavated as lumps and our FEA work also demonstrated the importance of brittle tensile failure for all but soft clays. The value of Cc indicated in Cathie and Wintgens (2001) did not take account of brittle failure and considered that it was largely independent of su . This appears to result in over-predictions of the ploughing resistance in very stiff to hard clays. We assume this is because the clays actually shear in a brittle manner with the operational strength being somewhere between the intact, residual or (submerged) tensile strengths. While for clays there is still plenty of scope for theoretical work on ploughing resistance, it receives
Figure 46. Crack propagation in brittle unsaturated clays (Karmakar, 2005).
Figure 48. Examples of jet trenchers. Figure 47. Schematic of pipeline lowering using jet trenching (Vanden Berghe et al. 2008).
less attention in industry because ploughing most clays is much easier than ploughing sands.
7 TRENCH CONSTRUCTION BY JETTING 7.1
Mechanics of jet trenching
7.2
1. Sediment entrainment (erosion and fluidization of coarse-grained soils) or soil cutting (fine-grained soils); 2. Sediment transport; 3. Trench collapse; 4. Sediment deposition.
Jet trenching involves the use of water jets to erode and fluidise sands, or to cut clays, in order to permit lowering and burial of cables (Jordon and Cathie, 2004) and pipelines as shown in Figure 47 (Vanden Berghe et al, 2008). Jetting systems utilize the power of water jets to erode or cut the soil, and to transport or fluidise it to allow the cable or pipeline to be lowered into the trench (Figure 47). Jetting systems range in power from about 250 kW to 2 MW and jet swords generally comprise quite large numbers of nozzles facing forwards, downwards and rearwards depending on the manufacturer’s design. Jetting pressures are typically in the range 2–10 bar with flow rates between 100–3000 m3 /hr.
116 © 2011 by Taylor & Francis Group, LLC
Jetting in sands
The main physical processes involved in jetting of coarse-grained soils are:
The water jets erode coarse-grained soil due to their high energy and carry the suspended particles away from the forward face, transporting them rearward. Sediment erosion is a well understood and for the typical flow rates used by jet trenchers, erosion and fluidization is normally achieved. However, if there is insufficient flow compared to the trench volume and progress rate of the trencher, the fluidization will be incomplete and the jetting swords will experience resistance from the soil still in place. A nominal measure of the soil volume to be excavated is defined by the excavation rate ER :
Figure 49. Relationship between fluidization ratio and trencher speed.
where v is the trencher progress rate (m/hour), DSw is the embedded sword depth and WSw is the sword width. The ratio of the water pumped into the soil and the volume to be excavated at a given speed can be called the fluidization ratio (i.e. jetting system flow rate/ excavation rate). This is also a measure of the resulting water/solids ratio (or the solids concentration, as is considered in the materials transport industry). It is useful for any trencher to develop charts of fluidization ratio for various speeds, sword width and sword depth to provide a rapid method of assessing likely progress rates that could be anticipated. An example is shown Figure 49. Fluidisation ratio is one indicator of the likely maximum rate of progress of a jet trencher and must be sufficient to enable the soil to be transported at the flow rates available. Fluidisation ratios above 5 appear to be sufficient but this also depends on the soil grain size and the configuration of the jets. Following erosion and fluidization of the sand grains, the sand must be transported behind the trencher before it is deposited in the lower flow regime behind the swords. The process is depicted in Figure 50 based on experimental work at the University of Taiwan on behalf of CTC Marine Projects. Both the power and orientation of the forward jets, and the power of any rear facing jets affects the sediment transport and to some extent defines the length of the turbulent region through which the pipeline or cable is lowered. Lateral inflow of soil is also observed in model tests, which merges with the turbulent flow behind the swords to increase the solids content and reduce the energy in the turbulent region. Inflow is less in fine, dense sands which can support a near vertical wall for longer than in coarse, loose sands. Eventually, sediment is deposited in a hindered settling regime at a rate that is largely dependent on the mean grain size. Fine sands settle much more slowly
than coarse sands and gravels. Therefore, the fluidized zone is longer and burial often better in fine soils (see Figure 50 and Figure 51). A mathematical model of the jet trenching process has been developed by Vanden Berghe et al (2008) and Peng and Capart (2008) under contract to CTC Marine Projects. The model is based on the fundamental physical processes that occur and has been calibrated by a series of 1g model tests. Both the tests and the mathematical modelling confirm the dependence of the progress rate on jetting power and sand density, and capture the main phenomena of sidewall collapse, sedimentation rate and overspill. Overspill (or loss of sediment from the trench) occurs due to dispersion of the sediment in the turbulent flow. The effect of cross-currents, which may be quite severe in practice, has not yet been considered. The model has also been used to optimize sword design and to assess the value of rearward facing jets to lengthen the fluidized region aft of the trencher. Note that the burial depth of the cable or pipeline also depends also on the weight, stiffness and residual tension in the cable or pipeline. Lower weight, higher stiffness or higher tension all lead to a longer span length and therefore reduced burial depth (Figure 51). A quantitative treatment of this interaction is provided by Vanden Berghe et al (2010). The unit weight (or specific gravity) of the pipeline or cable during installation must be greater than that of the fluidized soil. Moreover, there must be sufficient weight margin to ensure that the lowering of the product is not affected by the turbulence and upward flow of water behind the trencher. A minimum specific gravity is believed to be about 1.8 (but without conclusive experimental support). 7.3
Jetting of fine-grained soils is essentially a cutting process rather than an erosion process. The jets must cut and break up the material ahead of the sword. After that the essential mechanisms are similar except the clay is not fully disaggregated but remains in lumps which must be transported and maintained in suspension if a pipeline or cable is to be lowered. Conventional wisdom is that clay lumps should be removed (educted) from the trench by a dredging tool located on the trencher if good lowering is to be achieved. A water jet from a static nozzle impinging on a bed of clay creates a circular depression or cylindrical cavity in the clay often up to 3 times the nozzle diameter (Machin and Allan 2010). The depth of the cavity will depend on the jet pressure, the undrained shear strength and other properties of the soil, the offset between the bed and the nozzle (stand-off distance), and the time the jet is allowed to act. Machin et al (2001) show that the depth of cut consists of a quasi-instantaneous cut depth formed in a time span of 0.1–0.5 ms. After this, time-dependent erosion of the already-formed cut takes place. Under some circumstances, and particularly in fissured clays or clays
117 © 2011 by Taylor & Francis Group, LLC
Jetting in clay
Figure 50. Mechanisms observed during jet trenching (Vanden Berghe et al. 2008).
Figure 52. Pressure distribution of a submerged water jet acting on a target (Kondo et al. 1974). Figure 51. Lowering of a pipeline or cable in fluidized zone behind a jet trencher.
with silt or sand seams, hydraulic fracture mechanisms also operate. A basic model for clay cutting using jets requires an understanding of the mechanics of a submerged jet. If the center-line velocity of a submerged jet at the nozzle is v0 , the fluid velocity decays with distance due to spreading of the jet and entrainment of the surrounding water. The velocity, v, at the centreline of the flow at a stand-off distance x from the nozzle can be estimated by:
with the deformation of the first liquid to strike the solid; 2. A lower, quasi-steady pressure as the jet begins to flow outwards; this is associated with the pressure required to deflect the liquid sideways over the surface; 3. Shear stresses caused by liquid moving over the surface at high speed from the centre of impact.
where C is a dimensionless hydrodynamic drag coefficient (C ≈ 6.2), and d is the jet diameter. Within the initial region, up to about 6d, there is no loss in velocity. When a jet impinges on a surface, the necessary deflection of the jet fluid results in a pressure applied to the surface as illustrated in Figure 52 (Kondo et al. 1974). A solid target impacted by a liquid jet is subjected to the following pressures: 1. A strong initial compression pulse with a duration on the order of a microsecond; this is associated
118 © 2011 by Taylor & Francis Group, LLC
Considering the normal impact of a steady jet of inviscid, incompressible fluid against a rigid surface, and assuming the stand-off distance is small enough, the pressure distribution may be considered as uniform in the core area of the jet. The pressure applied to the surface is known as the stagnation pressure, p:
where ρw is the mass density of the water jet. A simple model for assessing the required jet pressure to cut clay soil can be developed by considering that the normal pressure of the jet, p, acting on the soil (Figure 52) creates a bearing failure. The reality is probably much more complex with both normal and shear stresses applied to the soil, local spalling occurring, localized drainage of the surface, erosion and variations in water pressure (hammering).
Bearing capacity failure will occur if:
where Nc is a bearing capacity factor (typically Nc = 6), or more simply the condition that p > qc , the cone tip resistance, can be used. Equations 11-13 can be combined to relate the pressure at the nozzle (p0 ) required to cut a soil of a given shear strength as a function of stand-off distance, considering that Equation 12 also applies to the pressure and velocity at the nozzle (following Bernoulli). As shown also by Machin and Allan (2010), the resulting relationship between standoff distance and the nozzle pressure to cause bearing failure is:
As a first approximation, x may also be considered the depth of cut that can be made for a jet located at the surface of the soil. The relationship is shown on Figure 53. With a jetting pressure of 20 bar from a 20 mm OD nozzle, in a soil of 150 kPa shear strength, a cut depth of about 0.2 m can be anticipated from each jet excluding any stand-off distance. From a practical perspective, considering a jetting sword of the type shown in Figure 48 with multiple jets, it is unlikely that zero stand-off distance can ever be achieved for all jets simultaneously. A minimum average stand-off might be 0.1 m if the trencher is being driven hard into the trench face. Assuming a minimum acceptable cut depth for each jet (of 20 mm OD) of 0.2 m gives x/d of 15. A typical 5 bar low pressure trenching system could be expected to operate acceptably in shear strengths of less than about 20 kPa. This agrees with general experience in the trenching industry. Low pressure (say 5 bar) systems which work effectively in sand can also create a trench in very soft to soft clays. As the bearing capacity failure occurs and the hole is deepened, the actual failure mechanism becomes more complex. The bearing capacity increases with depth, and the flow is no longer as shown on Figure 52 but is constrained inside the hole. The mechanism of entrainment of water leading to the reduction in jet velocity is quite different. However, Equation 14 is considered to provide a good indication of the ability of a static jet to cut the soil and experimental work does seem to confirm that this approach provides a conservative assessment (Machin and Allan 2010). Jet trenching involves the jets traversing rather than remaining static. Atmatzidis and Ferrin (1983) investigated the influence of time and traversing speed in the laboratory with a 1 mm diameter nozzle in clean sand, silty sand, silt and clay. The depth of jet penetration into the soil target and jet effectiveness was measured as a function of exposure time, the degree of saturation of the soil, the dry density of the soil and the traversing velocity of the jet over the soil target. Their research showed that the time required for the jet to approach maximum penetration in the soil target was about 15–20 seconds for a driving pressure
Figure 53. Required jet pressure for cutting.
Figure 54. Cutting depth as a function of jet trenching speed.
of 1000 psi (69 bar), being less for coarse-grained soil than fine-grained. It was found that the jet penetration depth, x, in a given soil is related exponentially to the corresponding time from initial impact. Machin and Allan (2010) indicate that it takes several seconds to reach the full cavity depth in clays. Atmatsidis and Ferrin (1983) also show that the penetration depth, x, could be related exponentially to the jet’s traversing velocity, v:
119 © 2011 by Taylor & Francis Group, LLC
where xmax is the limiting penetration for a stationary jet after infinite time, and ζ is an empirical constant (with units of velocity) depending on jet and material properties. Combining the static and traversing equations provides an indication of the likely depth of cut that a specific jetting system could provide, if the value of ζ were known. Machin and Allan (2010) suggest that a quasiinstantaneous cavity depth is achieved for translation speeds exceeding 0.1–0.5 m/s (360–1800 m/hr) but slower speeds are required for maximum penetration. Figure 54 shows a typical result for cutting stiff clay (su = 100 kPa) considering that the quasiinstantaneous cavity depth is achieved at a speed of 500 m/hr (assumed).
soil deformations to be observed. Other insights have emerged through finite element analysis – with notable advances being the use of coupled methods in sand to capture undrained and partially-drained ploughing behaviour, and the use of large deformation techniques to capture gross remoulding of fine-grained soils during large lateral movements.
Successful lowering of a cable or pipeline also depends on the orientation of the jets so that the soil is actually cut into blocks which can be educted or transported away from the zone where the product must sink. As far as the authors are aware, the first discussion of this issue in the public domain is given in Machin and Allan (2010). 8
CONCLUSIONS ACKNOWLEDGEMENTS
This paper has reviewed various aspects of pipeline geotechnics, by reference to recent and emerging research activity from both academia and industry. The design challenges in pipeline geotechnics differ somewhat from conventional foundation engineering. Two particular challenges are the changes in both the seabed topography and also the soil properties that can occur throughout the installation and operating life of a pipeline: these are themes that run throughout this paper. Results from novel forms of repetitive in situ testing have been used to illustrate the response of soil to the forms of loading and disturbance that are induced by a pipeline. In soft fine-grained soils, these tests illustrate the balance between the reduction in strength from remoulding and the recovery that accompanies subsequent reconsolidation. Concepts from critical state soil mechanics provide a useful framework for capturing this behaviour. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A common theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths, and the importance of recognising the widely-varying rates of shearing involved in pipe-soil processes.The slow rates at which pipes move under thermal loading and the high rates at which trenching machines are driven mean that pipeline geotechnics often involves a drained response in fine-grained soils and undrained behaviour in sands. The conventional laboratory tests used to characterise these soils often require modification in order to extract the properties that are relevant for pipe-soil interaction. Some of the concepts discussed in this paper represent merely a snapshot of an evolving understanding, which will no doubt advance in the coming few years. For example, the changes in soil strength through episodes of disturbance and recovery, and through cycles of partially-drained interface shearing, have only recently been observed in experiments. The underlying mechanisms are not fully established and quantitative calculation methods for design use are in their infancy. New modelling technologies have been recently applied to pipeline geotechnics. The mechanisms of pipe-soil penetration and breakout behaviour have been quantified through sophisticated centrifuge model tests, which allow complex load sequences to be imposed on a model pipe and detailed internal
The work described here forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established at the University of Western Australia in 1997 under the Australian Research Council’s Special Research Centres Program. COFS is now supported by Centre of Excellence funding from the State Government of Western Australia. The first author is supported by an Australian Research Council Future Fellowship (grant FT0991816) The assistance from a number of colleagues at UWA and Cathie Associates during the preparation of this paper is acknowledged. In particular, Fauzan Sahdi kindly provided the data in Section 2.3 from his PhD studies. The interface shear box tests in Section 5.2 were performed by Nat McNab assisted by Binaya Bhattarai. The MCWHIPLASH software (Section 4.4) was written by the first author with David Bonjean of Advanced Geomechanics, Perth. Helpful comments provided by George Zhang, also of Advanced Geomechanics, who reviewed a draft of this paper are also acknowledged. Some of the research described in this report has been guided by the SAFEBUCK Joint Industry Project, which is coordinated by David Bruton of AtkinsBoreas. The support of the SAFEBUCK participants is gratefully acknowledged. REFERENCES AtkinsBoreas. 2008. SAFEBUCK JIP: Safe design of pipelines with lateral buckling; design guideline. Report BR02050/C, AtkinsBoreas 252pp. Atmatzidis, D.K. and Ferrin, F.R. 1983. Laboratory investigation of soil cutting with a water jet, 2nd US Water Jet Conference, Rollo, Missouri, University of Missouri, 101–110. Aubeny, C.P., Shi, H., and Murff, J.D. 2005. Collapse loads for a cylinder embedded in trench in cohesive soil. ASCE Int. J. Geomechanics 5(4):320–325. Bolton, M.D. 1986. The strength and dilatancy of sands. Géotechnique 36(1):65–78. Bolton, M.D. and Barefoot, A.J. 1997. The variation of critical pipeline trench back-fill properties. Proc. of Conference on Risk-Based and Limit State Design and Operation of Pipelines, Aberdeen. Bolton, M.D., Ganesan, S.A. and White, D.J. 2009. SAFEBUCK Phase II: Axial pipe-soil resistance: summary report. Cambridge University Technical Services, Report for Boreas Consultants (SAFEBUCK JIP), ref. SC-CUTS-0705-R01. 54pp. Bransby, M.F. Yun, G.J. Morrow, D.R. and Brunning, P. 2005. The performance of pipeline ploughs in layered
120 © 2011 by Taylor & Francis Group, LLC
soils, International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, September. Bransby, M.F. and Ireland, J. 2009. Rate effects during pipeline upheaval buckling in sand. Proc. ICE Geotechnical Engineering 162: 247–256. Bruton, D.A.S., White, D.J., Cheuk, C.Y., Bolton, M.D. and Carr, M.C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Proc. Offshore Technology Conference, Houston. OTC17944. Bruton, D., Carr, M. and White, D.J. 2007. The influence of pipe-soil interaction on lateral buckling and walking of pipelines: the SAFEBUCK JIP. Proc. 6th Int. Conf. on Offshore Site Investigation and Geotechnics, London. 133–150. Bruton, D., White D.J., Langford, T.L. and Hill, A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments Proc. Offshore Technology Conference, Houston, USA. Paper OTC20096. Cardoso, C.O., da Costa, A.M. and Solano, R.F. 2006. HP-HT pipeline cyclic behaviour considering soil berms effect. Proc. 25th Int. Conf. on Offshore Mechanics and Arctic Engineering, Hamburg, Germany. Cardoso, C.O. and Silviera, R.M.S. 2010. Pipe-soil interaction behavior for pipelines under large displacements on clay soils – a model for lateral residual friction factor. Offshore Technology Conference, Houston. Paper OTC20767. Carr, M., Matheson, I., Peek, R., Saunders, P. and George, N. 2004. Penguins flowline lateral buckle formation analysis and verification. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, OMAE2004-51202. Carr, M.C., Sinclair, F., and Bruton, D.A.S. 2006. Pipeline walking – understanding the field layout challenges, and analytical solutions developed for the SAFEBUCK JIP. Proc. Offshore Technology Conf., Houston, Paper OTC17945. Cathie, D.N. and Wintgens, J.F. 2001. Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conf., Houston. Paper OTC13145 Cathie, D.N. Jaeck, C.. Ballard, J.C and Wintgens J.F. 2005. Pipeline Geotechnics: State of the Art. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG) Perth, Australia. 95–114. Chatterjee, S., Randolph, M.F., White, D.J. and Wang, D. 2010a. Large deformation finite element analysis of vertical penetration of pipelines into the seabed. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth. Chatterjee, S., Randolph, M.F. and White, D.J. 2010b. The effects of strain rate and strain softening on the vertical penetration resistance of pipelines on an undrained seabed. Submitted for publication. Cheng, L., White, D.J., Palmer, A.C., Jas, E., Czajko, A, Fogliani, N, Fricke, R. and An, H. 2010. A new facility for research into the stability of pipelines on unstable seabeds. Offshore Pipeline Technology Conference, Amsterdam. Cheuk, C.Y., White, D.J. and Dingle, H.R.C. 2008. Upper bound plasticity analysis of a partially-embedded pipe under combined vertical and horizontal loading. Soils and Foundations 48(1):137–148. Cheuk, C.Y. and White, D.J. 2010a. Modelling the dynamic embedment of seabed pipelines. Géotechnique. Accepted August 2009, in press. Cheuk, C.Y. and White, D.J. 2010b. The influence of breakout conditions on the lateral resistance of on-bottom pipelines. Canadian Geotechnical Journal, in review. Cocchetti, G., di Prisco, C., Galli, A. and Nova, R. 2009. Soil-pipeline interaction along unstable slopes: a coupled
three-dimensional approach. Part 1: Theoretical formulation. Canadian Geotechnical Journal 46:1289–1304. Damgaard, J.S. and Palmer, A.C. 2001. Pipeline stability on a mobile and liquefied seabed: a discussion of magnitudes and engineering implications. Proc., 20th Int. Conf. on Offshore Mechanics and Arctic Engineering, Rio de Janeiro. Dean, E.T.R. 2010. Offshore Geotechnical Engineering. Thomas Telford. Dingle, H.R.C., White, D.J., and Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45(5):636–652. DNV 2007a. Recommended Practice DNV RP-F109, Onbottom Stability Design of Submarine Pipelines. (with amendments, April 2009). Det Norske Veritas. DNV. 2007b. Recommended Practice DNV-RP-C207, Statistical representation of soil data. Det Norske Veritas. Elsworth, D. and Lee, D.S. 2005. Permeability determination from on-the-fly piezocone sounding.ASCE J. Geotech and Geoenvironmental Engng., 131(5):643–653. Erbrich, C.T. 2005. Australian frontiers – spudcans on the edge. Proc. Int. Conf. on Frontiers in Offshore Geotechnics, Perth. 49–74. Fannin, R.J., Eliadorani,A. and Wilkinson, J.M.T. 2005. Shear strength of cohesionless soil at low stresses. Géotechnique 55, No. 6, 467–478. Finnie, I.M.S. 1993. The performance of shallow foundations in calcareous soil, PhD thesis, University of Western Australia. Finnie, I.M.S., and Randolph, M.F. 1994. Punch-through and liquefaction induced failure of shallow foundations on calcareous sediments. Proc., Int. Conf. on Behaviour of Offshore Structures, BOSS’94, Boston, 217–230. Fisher, R. and Cathie, D. 2003. Optimisation of gravity based design for subsea applications. Proc. International Conference on Foundations, ICOF, Dundee. 283–296. Gaudin, C. and White, D.J. 2009. New centrifuge modelling techniques for investigating seabed pipeline behaviour. Proc. XVIIth Int. Conf. on Soil Mechanics and Geotechnical Engineering. Alexandria. Gourvenec, S.M. and White, D.J. 2010. Elastic solutions for consolidation around seabed pipelines. Proc. Offshore Technology Conference, Houston. Paper 20554. He, J. Vasblom, W.J. Miedema, S.A. 2005. FEM analyses of cutting of anisotropic densely compacted and saturated sand, Western Dredging Association Conference, WEDAXXV and TAMU37, New Orleans, USA, June and www.dredgingengineering.com/dredging Hettiaratchi, D.R.P. and Reece, A.R. 1974. The calculation of passive soil resistance, Géotechnique, 24(3):289–310. Hettiaratchi, D.R.P. and Reece, A.R. 1975. Boundary wedges in two-dimensional passive soil failure, Géotechnique, 25(2):197–220. Hill, A.J. and Jacob, H. (2008) In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC 19528. Hossain, M.S. and Randolph, M.F. 2009. Effect of strain rate and strain softening on the penetration resistance of spudcan foundations on clay. ASCE Int. J. Geomech. 9(3):122–132. Jayson, D., Delaporte, P., Albert, J-P, Prevost, M.E., Bruton, D. and Sinclair, F. 2008. Greater Plutonio project – Subsea flowline design and performance. Offshore Pipeline Technology Conf., Amsterdam. Jordon P, and Cathie, D.N. 2004. Developments in cable protection by burial. SubOptic 2004. Karmakar , S. 2005. Numerical modelling of soil flow and pressure distribution on a simple tillage tool using
121 © 2011 by Taylor & Francis Group, LLC
computational fluid dynamics, PhD thesis, University of Saskatchewan. Kondo, M. Fujii, K. and Syoji, H. 1974. On the destruction of mortar specimens by submerged water jets, 2nd International Symposium on Jet Cutting Technology, Cambridge, paper B5, 69–88. Krost K., Gourvenec, S.M. and White, D.J. 2010. Consolidation around partially-embedded submarine pipelines. Géotechnique, Accepted December 2008, in press. Lacasse, S., Gutteromsen, T., Nadim, F., Rahim, A., Lunne, T. 2007. Use of statistical methods for selecting design soil parameters. Proc. 6th Int. Offshore Site Investigation and Geotechnics Conference, London, UK, 449–460. Langford, T.E., Dyvik, R. and Cleave, R. 2007. Offshore pipeline and riser geotechnical model testing: practice and interpretation. Proc. Conf. on Offshore Mech. and Arctic Eng., San Diego. Paper OMAE2007-29458. Lauder, K. 2010. Predicting pipeline plough performance by scale model testing, Presentation to CTC, April, University of Dundee. Lauder, K. Bransby, F. Brown, M. Cathie, D. Morgan, N, Pyrah, J. and Steward, J. 2008. Experimental testing of the performance of pipeline ploughs, 18th Int. Offshore and Polar Engineering Conf. ISOPE-2008, Vancouver, Canada, 212–217. Lenci, S. and Callegari., M. 2005. Simple analytical models for the J-lay problem. Acta Mechanica, 178:23–39. Lumb, P. 1966. The variability of natural soils. Canadian Geotechnical Journal. 3:74–97. Lund, K.M. 2000. Effect of increase in pipeline soil penetration from installation. Proc. of ETCE/OMAE 2000 Joint Conference; Energy of the New Millennium OMAE2000/ PIPE-5047. Machin, J.B. and Allan, P.J.A. 2010. State-of-the-art jet trenching analysis in stiff clays, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Machin J.B. Messina, F.D. Mangal, J.K. Girard, J. and Finch, M. 2001. Recent research on stiff clay jetting, Offshore Technology Conference,Houston, Paper OTC 13139. Merifield, R., White, D.J. and Randolph, M.F. 2008a.Analysis of the undrained breakout resistance of partially embedded pipelines. Géotechnique, 58(6)461–470. Merifield, R.S, White, D.J. and Randolph, M.F. 2008b. The effect of pipe-soil interface conditions on undrained breakout resistance of partially-embedded pipelines. Proc. Int. Conf. on Advances in Computer Meth. and Analysis in Geomech. Goa, India. Merifield, R., White, D.J. and Randolph, M.F. 2009. The effect of surface heave on the response of partiallyembedded pipelines on clay. ASCE J. Geotechnical and Geoenvironmental Engineering, 135(6):819–829. Miedema, S.A. 1987. Calculation of the cutting forces when cutting water saturated sand, PhD dissertation, Delft University of Technology. Miedema, S.A. 2005. The cutting of water saturated sand, the final solution, WEDAXXV and TAMU37, New Orleans, USA, June, and www.dredgingengineering.com/dredging. Morrow, D.R. and Bransby, M.F. 2009. The influence of slope on the stability of pipelines subjected to horizontal and vertical loading on clay seabeds. Proc. Conf. on Offshore Mechanics and Arctic Engineering. OMAE2009-79050. Orcina. 2008. OrcaFlex software, Orcina Ltd, Ulverston, UK Palmer, A.C., Kenny, J.P., Perera, M.R. and Reece, A.R. 1979. Design and operation of an underwater pipeline trenching plough. Géotechnique 29(3):305–322. Palmer, A.C. 1999. Speed effects in cutting and ploughing. Géotechnique 49(3):285–294.
Pedersen, R.C., Olsen, R.E. and Rausch, A.F. 2003. Shear and interface strength of clay at very low effective stresses. ASTM Geotechnical Testing J., 26(1):71–783. Peng W, and Bransby M.F. 2010. Numerical modelling of soil around offshore pipeline plough shares, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Peng A.T.H and Capart H. 2008 Underwater sand bed erosion and internal jump formation by travelling plane jets, Journal of Fluid Mechanics, 595:1–43. Puech, A. and Foray, P. 2002. Refined model for interpreting shallow penetration CPTs in sands. Proc. Offshore Technol. Conf., Houston, Paper OTC14275. Randolph, M.F. 2003. Science and empiricism in pile foundation design. Géotechnique. 53(10): 847–875. Randolph, M.F. and Hope, S.N. 2004. Effect of cone velocity on cone resistance and excess pore pressure. Proc. Conf. on Engineering Practice and Performance of Soft Deposits, Osaka, 147–152. Randolph, M.F. and White, D.J. 2008a. Pipeline embedment in deep water: processes and quantitative assessment. Proc. Offshore Technology Conference, Houston, USA. Paper OTC19128-PP. Randolph, M.F. and White, D.J. 2008b. Upper bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4):297–301. Randolph M.F. and Gourvenec S.M. 2010. Offshore Geotechnical Engineering. Taylor and Francis. Rathbone, A. Hakim, M. A. Cumming, G. and Tørnes, K., 2008. Reliability of lateral buckling formation from planned and unplanned buckle sites. Proc. 27th International Conference on Offshore Mechanics and Arctic Engineering, OMAE2008-57300, Estoril, Portugal. Reece, A.R. and Grinsted, T.W. 1986. Soil mechanics of submarine ploughs, Offshore Technology Conference, OTC 5341. Schotman, G.J.M. and Stork, F.G. 1987. Pipe-soil interaction: a model for laterally loaded pipelines in clay. Proc. Offshore Technology Conference, Houston, OTC5588. Silva, M.F. 2005. Numerical and physical models of rate effects in soil penetration. PhD thesis, University of Cambridge. Steenfelt, J.S. 1993. Sliding resistance for foundations on clay till. Proc. Wroth Memorial Conference, Predictive Soil Mechanics. Thomas Telford. 664–684. Sture, S., Costes, N. C., Batiste, S.N., Lankton, M.R., AlShibli, K.A., Jeremic, B., Swanson, R.A. and Frank, M. 1998. Mechanics of granular materials at low effective stresses. ASCE J. Aerospace Engng 11(3): 67–72. Tian,Y., and Cassidy, M.J. 2008. Explicit and implicit integration algorithms for an elastoplastic pipe-soil interaction macroelement model. Proc. 27th International Conference on Offshore Mechanics and Arctic Engineering, OMAE2008-57237, Estoril, Portugal. Tornes, K., Jury, J. and Ose, B. 2000. Axial creeping of high temperature flowlines caused by soil racheting. Proc. Conf. on Offshore Mechanics and Arctic Engineering, OMAE-PIPE5055. Verley, R. and Lund, K.M. 1995. A soil resistance model for pipelines placed on clay soils. Proc. International Conference on Offshore Mechanics and Arctic Engineering (OMAE), Copenhagen, Denmark, 5: 225–232. Verley, R.L.P. and Sotberg, T. 1994. A soil resistance model for pipelines placed on sandy soils. ASME J. Offshore Mechanics and Arctic Engineering, 116(3):145–153. Van Leussen, W. and Nieuwenhuis, J.D. 1984. Soil mechanics aspects of dredging, Geotechnique 34(3): 359–381.
122 © 2011 by Taylor & Francis Group, LLC
Van Os, A.G. and van Leussen, W. 1987. Basic research on cutting forces in saturated sand, Journal of Geotechnical Engineering, 113 (12):1501–1516. Van Rhee, C. and Steeghs, H.J.M.G. (1991) Multi-blade ploughs in saturated sand; model cutting tests. Dredging and Port Constructions, June, 37–39. Vanden Berghe, J.-F. Pyrah, J. Gooding, S. and Capart, H. 2010. Development of a jet trenching model in sand, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Vanden Berghe, J.-F. Capart, H. and Su, J.C.C. 2008. Jet induced trenching operations: mechanisms involved. Offshore Technology Conference, Houston. Paper OTC 19441. Wang, D., White, D.J. and Randolph, M.F. 2009. Numerical simulations of dynamic embedment during pipelaying on soft clay. Proc. Conf. on Offshore Mechanics and Arctic Engineering, Honolulu. Paper OMAE2009-79199. Wang, D., White, D.J. and Randolph, M.F. 2010. Large deformation finite element analysis of pipe penetration and large-amplitude lateral displacement. Canadian Geotechnical Journal. Accepted April 2009, in press. Westgate, Z., White, D.J. and Randolph, M.F. 2009. Video observations of dynamic embedment during pipelaying on soft clay. Proc. Conf. on Offshore Mechanics and Arctic Engineering, Honolulu. Paper OMAE2009-79814. Westgate, Z.J., Randolph, M.F., White, D.J. and Li, S. 2010a. The influence of seastate on as-laid pipeline embedment: a case study. Applied Ocean Research, Available online, 4 February 2010. doi:10.1016/j.apor.2009.12.004. Westgate, Z.W., White, D.J., Randolph, M.F. and Brunning, P. 2010b. Pipeline laying and embedment in soft fine-grained soils: Field observations and numerical simulations. Proc. OffshoreTechnology Conference, Houston. Paper 20407. White, D.J. and Randolph, M.F. 2007. Seabed characterisation and models for pipeline-soil interaction Proc. 17th Int. Offshore and Polar Engng. Conference, Lisbon, Portugal. CD 11pp. White, D.J. and Gaudin, C. 2008. Simulation of seabed pipe-soil interaction using geotechnical centrifuge modelling. Proc. 1st Asia-Pacific Deep Offshore Technology Conference, Perth, Dec 2008. White, D.J. and Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement. Marine Structures 21(1):59–79.
White, D.J. and Cheuk, C.Y. 2009. SAFEBUCK JIP: Pipe-soil interaction models for lateral buckling design: Phase IIA data review. Report to Boreas Consultants (SAFEBUCK JIP), UWA report GEO 09497r1. 185pp. White, D.J. and Dingle, H.R.C. 2010. The mechanism of steady ‘friction’ between seabed pipelines and clay soils. Géotechnique. Accepted April 2009, in press. White, D.J. and Hodder, M. 2010. A simple model for the effect on soil strength of remoulding and reconsolidation. Canadian Geotechnical Journal. Accepted November 2009, in press. White, D.J., Gaudin, C., Boylan, N and Zhou, H. 2010a. Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils. Canadian Geotechnical Journal, 47(2):218–229. White, D.J., Hill, A.J., Westgate, Z. and Ballard, J-C 2010b. Observations of pipe-soil response from the first deepwater deployment of the SMARTPIPE. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth. Wroth, C.P. 1984. The interpretation of in-situ tests. 24th. Rankine Lecture, Géotechnique, 34(4):449–489. Zhang, J., Stewart, D. P., and Randolph, M. F. 2002a. Modelling of shallowly embedded offshore pipelines in calcareous sand. ASCE Journal of Geotechnical and Geoenvironmental Engineering, 128(5):363–371. Zhang, J., Stewart, D. P., and Randolph, M. F. 2002b. Kinematic hardening model for pipeline-soil interaction under various loading conditions. ASCE International Journal of Geomechanics, 2(4):419–446. Zhang, J., and Erbrich, C.T. 2005. Stability design of untrenched pipelines – geotechnical aspects. Proc. International Symposium on Frontiers in Offshore Geo-technics (ISFOG) Perth, Australia. 623–628. Zhou, H., White, D.J. and Randolph, M.F. 2008. Physical and numerical simulation of shallow penetration of a cylindrical object into soft clay. Proc. GeoCongress 2008, New Orleans,ASCE Special Geotechnical Publication No. 179, 108–117. Zhou, H. and Randolph, M. F. 2009. Penetration resistance ofcylindrical and spherical penetrometers in rate-dependent and strain-softening clay. Géotechnique. 59(2)79–86.
123 © 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph Advanced Geomechanics, Perth, Australia
ABSTRACT: Design of piles in carbonate soils remains one of the most challenging geotechnical problems, fraught with traps for the unwary. Whilst carbonate soils generally behave in accordance with the same underlying rules of soil mechanics that are applicable to all soils, they often exhibit characteristics at the extremes of this continuum of behaviour. Carbonate soils themselves straddle a wide range of different material types, varying from soft uncemented silts and muds to dense well cemented calcarenites and limestones. We also find dense uncemented carbonate soils and loose and compressible cemented soft rocks (calcarenites, etc.). All combinations of these are encountered offshore Australia. This paper presents new work that addresses two key aspects of pile design in this array of carbonate material types: axial design of drilled and grouted piles in compressible cemented carbonate soils and lateral design of piles (drilled and grouted or driven) in uncemented carbonate soils, varying from soft muds to dense sands. The new work on axial pile design builds on the extensive work undertaken in the 1980s following the problems with the North Rankin A platform, and which was subsequently put to good use in the design of the nearby Goodwyn A platform. However, more recent developments have identified a need to develop a more generalised approach that can be applied at sites with a variety of characteristics. In addition, more detailed review of some of the data collected during the original work has revealed important new mechanisms that are of considerable significance but have hitherto been neglected. The new work on lateral pile design follows a different tack from any of the existing methodologies used for assessing lateral pile response in such soils, which are essentially of a purely empirical form, with minimal theoretical basis. A new method with a robust theoretical basis is presented for assessing the lateral pile response in carbonate soils, which particularly accounts for their well known strong susceptibility to degradation (liquefaction) when subject to cyclic loading. 1
INTRODUCTION
The axial and lateral response of piles is analysed almost universally in the offshore industry using load transfer approaches. Interaction between the pile and the soil is treated through non-linear springs distributed down the pile shaft, with each horizontal layer of soil treated independently. The load transfer curve thus represents the integrated response of the soil layer extending outwards from the pile. The early development of load transfer curves followed an empirical approach, derived from the results of instrumented pile load tests, with limited guidance on how to scale the curves for different pile diameters or soil stiffness. More rational frameworks have since been established, aided by modern numerical analysis, but there is still an underlying need for experimental data for aspects such as the effects of cyclic loading. The key ingredients of load transfer curves may be summarised as follows:
loading) or net lateral pressure (force per unit projected area, for lateral load transfer). 3. Post-peak softening response, particularly for axial load transfer but also for lateral response in cemented soils. 4. Allowance for the effects of cyclic loading.
1. A pre-failure portion, with initial gradient linked to the shear modulus of the soil and a non-linear transition to a limiting load mobilised at an appropriate local displacement. 2. Limiting loads under monotonic loading, generally expressed as an equivalent shaft friction (axial
125 © 2011 by Taylor & Francis Group, LLC
As in much offshore design, the last aspect above provides the greatest challenge. The most common approach has been implicit, reducing the limiting load transfer for cyclic conditions and assessing the pile response by means of an equivalent monotonic analysis, rather than attempt to carry out cycle by cycle analyses where degradation of the load transfer response is modelled explicitly. In many soil types, post-peak softening under monotonic loading is relatively minor (the API guidelines suggest a maximum reduction of 30% under axial loading) and the implicit approach for cyclic loading has proved sufficient. However, post-peak softening for piles grouted into cemented carbonate soils can be dramatic, and the effect on the overall pile capacity must be addressed in detail. This is also true in respect of lateral loading of piles driven or grouted into cemented sediments, where cracks in the surrounding sediments can extend to the free surface. In addition, pile tests at laboratory and field scale in carbonate sediments have shown that the
response under axial cyclic loading is complex, and with potential for accumulating damage to lead to progressive failure of the entire pile. A brief background is provided here with respect to the load transfer algorithms that have been developed over the last two or three decades for pile response, and in particular their application to carbonate sediments. 1.1 Axial load transfer Load transfer analysis of the axial response of piles stems from the 1960s (Coyle and Reese 1966). The initial stiffness of the pre-failure load transfer response may be linked directly to the shear modulus of the soil (Randolph and Wroth 1978), and the approach can be extended to allow for non-linear soil response, by integrating the stress-strain response radially outwards from the pile (Kraft et al. 1981). For a typical hyperbolic stress-strain response, the resulting load transfer curve turns out to match closely an inverted parabola, with an initial gradient that is about double the secant gradient at full mobilisation of the ultimate friction. This form of curve is also very similar to the generic shape recommended in the API guidelines (API RP2A 2000). In siliceous clays and sands, ratios of shear modulus to limiting shaft friction result in a load transfer curve where full mobilisation occurs at a local pile displacement of 0.25 to 2% of the pile diameter, with a typical value of around 1% (Jeanjean et al. 2010). In cemented material, it is appropriate to adopt a somewhat stiffer load transfer response, partly because of higher ratios of shear modulus to limiting shaft friction, but also because this is conservative from the point of view of increased stress concentration in the upper part of the pile, hence greater damage during cyclic loading. The load transfer software, RATZ, was developed originally in the mid 1980s (Randolph 1986), and at that stage was calibrated to match the response of laboratory rod shear and constant normal stiffness (CNS) tests, and subsequently field grouted pile tests (Randolph 1988, Randolph & Jewell 1989, Randolph et al. 1996). The program uses an explicit approach, similar to the commercial finite difference code, FLAC, to follow the non-linear response of the pile under either monotonic or cyclic loading. The current version of the software is based on Excel for convenience of data input and viewing results, although the numerical computations are carried out in a Fortran-based DLL (Randolph 2003). Details of the load transfer algorithm are described later, in relation to the CYCLOPS software, which was later developed from RATZ. The load transfer algorithm includes a non-linear pre-failure response, post-failure strain-softening, cyclic degradation calculated through a damage algorithm based on accumulated irreversible relative pile-soil displacement, and a low cyclic residual friction within the current range of plastic pile-soil displacement followed by (partial) recovery for displacements outside that zone (see Figure 1). As discussed later, the low cyclic residual
Figure 1. Schematic of load transfer algorithm in RATZ.
(CFG in Figure 1) is a key feature observed in grouted pile and CNS test responses. An important distinguishing feature of the load transfer analysis in RATZ is the provision for cycle-bycycle analysis of a complete storm loading sequence acting on the pile. This allows simulation of gradual ratcheting, and transfer of load progressively down the pile as the more heavily loaded soil in the upper part of the compressing pile accumulates displacement and softens. 1.2
126 © 2011 by Taylor & Francis Group, LLC
Lateral load transfer
Forms of load transfer curves for lateral pile response are recommended for siliceous clays and sands in the API and ISO guidelines (API 2000, ISO 2007). These are mostly based on approaches developed in the 1970s by Reese, Matlock and others. At the time of the strut strengthening developed for first generation platforms in Australia’s Bass Strait (Wiltsie et al. 1988), Exxon undertook an extensive suite of centrifuge model and small field tests in order to develop lateral load transfer curves specifically for carbonate sands (Wesselink et al. 1988). The above work led to a generic load transfer curve for monotonic loading, where the lateral load per unit length, P (or net pressure, p = P/D), was expressed as a power law function of the cone resistance, qc , and the normalised displacement, y/d. A normalised version of the load transfer formulation was later proposed, following a more extensive series of centrifuge model tests (Dyson and Randolph 2001), expressed as:
where γ is the effective unit weight of the soil. Best fit values of the parameters, R, n and m, were established as 2.7, 0.72 and 0.58 respectively. A similar expression was proposed by Novello (1999), given by
The inclusion of a depth term, z/D, in this relationship means that, in general, the ratio, p/qc , increases more rapidly with depth than using Eq. (1). For example, if the cone resistance were to vary proportionally with depth, Eq. (2) would result in the lateral load for a given displacement, y, also being proportional to depth, rather than varying with depth to the power of 0.72 using Eq. (1). In other respects though, the two formulations are quite similar. Carbonate sediments tend to have high friction angles, in the region of 40◦ . However, comparing the above relationships with the API load transfer formulations for siliceous sand, the resulting response tends to be bracketed by the API curves for 20◦ and 35◦ , at least for displacements less than 0.1D. At larger displacements, the carbonate soil formulations give lateral resistance in excess of that for the 35◦ siliceous sand (Wesselink et al. 1988). The forms of load transfer curve given by Equations (1) and (2) are not bounded as y increases, and also have an infinite initial gradient (although a finite value of around 4G0 may be adopted for numerical implementation, where G0 is the small strain shear modulus). For practical values of deflection, with y << D, the net lateral pressure on the pile remains as a small fraction of the cone resistance. Thus for qc = 10 MPa, and γ D = 20 kPa, Eq. (1) gives a net pressure, p, of 1.86 MPa for y = 0.2D, and remains below 0.5qc even for y ∼ D. The effects of cyclic loading were explicitly considered by Novello (1999), through the introduction of a multiplier on the computed monotonic resistance:
where U* is the pore pressure ratio, varying from zero to 1.0 for a fully liquefied soil, and K0 is the coefficient of earth pressure at rest. For cemented carbonate soils, commonly encountered at shallow depth in the form of cap rock, attention must be paid to the brittle nature of the stress-strain response, and the potential for cracked fragments of cemented material to fail towards the free surface. A strain-softening p-y model to simulate this was proposed by Abbs (1983). However, a more rational approach was later developed by Erbrich (2004) for application in the carbonate deposits on the North-West Shelf of Australia. The basis of the Erbrich model is the concept of wedges of ‘chipped’ material forming near the surface. This concept was combined with the kinematic mechanism of Muff and Hamilton (1993) in order to evaluate the net resistance once a chip had occurred, and to develop a criterion for the maximum depth to which chips, as opposed to deformation of intact rock (which offers significantly higher capacity than chipped material), would occur. The ‘chipper’ model proposed by Erbrich (2004) was calibrated by means of 3D finite element analyses, and also through comparisons with centrifuge
model tests. It was found to yield greater lateral pile capacity than the Abbs (1983) model, but provided a sound physical basis on which many anchor piles have been designed offshore Australia and in other regions where cemented carbonate sediments occur. 2 2.1
Introduction
After experiencing problems with the driven piles at North Rankin A, which largely free-fell to their design penetration with minimal driving, consideration once again turned to drilled and grouted piles as the preferred foundation system in carbonate soils and soft rocks (calcarenites). However, the experience with the free-falling North Rankin piles, the results of various small scale model tests and the results of direct shear CNS tests suggested that a very stiff and brittle response might be expected for a grouted insert in these calcarenites. A major project was therefore commenced which was intended to measure the performance of grouted pile sections of varying sizes (up to 2 m diameter) in presumed similar soils to those at North Rankin A (Randolph, et al. 1996). A site was identified at Overland Corner (OC), South Australia where cemented carbonate material with large void ratios (i.e. apparently similar characteristics to the North West Shelf material) was found. The results from this test program formed the basis of design for the drilled and grouted piles adopted for the nearby Goodwyn A platform which is located in very similar soils to those found at NRA. This remains the largest structure built to date in Australian waters with drilled and grouted piles. The RATZ model was the principal design tool used in the GWA design, with the key parameters tuned to give the best fit to the OC grouted section tests (GSTs). CNS direct shear tests were undertaken for both the GWA and OC soils, but these were treated only as qualitative information and were not used directly in the design of GWA or the back-analysis of the OC grouted sections. More recently it has been necessary to design large diameter drilled and grouted piles at a number of other sites offshore Australia. For these cases, CNS direct shear tests suggested potentially quite different behaviour to that observed at GWA and OC for both monotonic and cyclic loading. This seemed conceptually in keeping with the developing understanding of the underlying soil mechanics of these soils (i.e. the effect of differing density, cementation and stress level on the foundation performance), but raised a challenge as to how to best deal with this variability in the design process. It was felt that simple reliance on the calibrated GWA/OC design parameters could not be justified. Instead the axial pile response was developed by directly matching the modelled response to the CNS test response. It was considered that this approach should be robust and inherently conservative, since it was thought that scale effects associated with shearing an interface in a small scale CNS test would (usually)
127 © 2011 by Taylor & Francis Group, LLC
PART 1 – AXIAL PILE DESIGN
may vary along the length. Interaction with the soil is characterised by discreet non-linear ‘t-z’ load transfer springs. The t-z curve is a function that relates the local shaft friction at the pile-soil interface to the current local pile displacement, accounting for the previous displacement history. The load transfer function takes full account of cyclic loading, and allows for accumulation of displacement and degradation of shaft friction. 2.2.1 Form of t-z curve The general form of the t-z load transfer curve adopted in CYCLOPS is presented schematically on Figure 3. The parameters defining the shape of the t-z curve include: Figure 2. Comparison between CNS test and RATZ.
lead to a more brittle response than obtained at the grout-soil interface for a drilled and grouted pile in the field; the original OC calibration also appeared to lend support to this notion. Once this CNS test calibration process was started in earnest, it soon became apparent that a number of the characteristic features commonly observed in a CNS test (and indeed also apparent in the GSTs) could not be reproduced with the existing RATZ model. An example illustrating these limitations is presented on Figure 2, which presents both the raw CNS test data and the ‘best fit’ achieved with the RATZ model. A key problem is that the RATZ model exhibits a ‘binary’ response during cyclic loading; while cycling within the limit of previously applied cyclic displacement amplitudes only a very low ‘cyclic residual’ friction is mobilised, but once the applied cyclic displacement is increased to and beyond the previous limit it instantaneously recovers to the full post-cyclic strength. This is clearly very different from the actual test data and seems to suggest both potentially conservative and unconservative characteristics. The former because the size of the zone where only the cyclic residual can be mobilised is overestimated, which may lead to increased redistribution of the axial pile load down a pile during cyclic loading, but the latter because a much larger pile displacement will be required to mobilise the post-cyclic strength than predicted by RATZ. It was far from obvious whether these effects would cancel out or whether or not an overall conservative or unconservative bias would be found. It was therefore considered essential to incorporate these effects into any future work. The CYCLOPS model was thus born, which is an enhancement of the RATZ model to enable much improved simulation of the observed behaviour in CNS tests and GSTs. The CYCLOPS model will be described in the next section. 2.2
Model description (CYCLOPS)
CYCLOPS was developed from the computer program RATZ (Randolph, 2003). Both RATZ and CYCLOPS treat the pile as an elastic bar with properties that
– The elastic threshold (ξ). – The peak shaft friction (τpeak ). – The displacement to τpeak (wpeak ), defined in part by the maximum shear modulus (Gmax ). – The initial strain softening and strain softening parameters (i-η and η respectively). – The residual shaft friction (τres ). – The displacement from τpeak to τres (wres ). – The cyclic residual shaft friction (τcyc- res ). – The bias parameter (b), defined in turn by the initial and final bias parameters (bi and bf respectively). – The bias parameter exponent (m). – The cyclic transition width (wcyc ). The parameters ξ, Gmax , τpeak , i-η, η, τres and wres define the shape of the ‘backbone’ monotonic failure curve, indicated by the light solid line on Figure 3. A demonstration of the influence of wres on the monotonic response is presented on Figure 4, which shows monotonic post-τpeak failure curves for a uniform set of parameters and for various wres values ranging between 0.1 m and 1.5 m. Similar sets of curves are presented on Figure 5 for various η values ranging between 0.1 and 0.9, and on Figure 6 for various τres /τpeak values ranging between 0.1 and 0.9. A point worth noting is that a specific post-τpeak monotonic response described by given values of wres , η and τres /τpeak can be approximately matched by various alternative combinations of these parameters over a limited displacement range. This point is demonstrated on Figure 7, which shows the postτpeak monotonic response for a variety of cases, all of which give a similar response over the initial 0.2 m of displacement. It can be seen that this similitude of response is obtained where low values of wres are combined with high values of τres /τpeak at one extreme, and high values of wres are combined with low values of τres /τpeak at the other extreme. The parameters bi , bf , m and to a lesser extent τcyc-res define the relative shape of the maximum bias stress (τwmax ) curve, indicated by the dashed line on Figure 3. For any given total plastic displacement, the maximum bias stress curve essentially ‘mirrors’ the monotonic failure curve according to the following relationship:
128 © 2011 by Taylor & Francis Group, LLC
Figure 3. Details of CYCLOPS t-z load transfer curve.
Figure 4. CYCLOPS – Displacement to residual.
Figure 5. CYCLOPS – Strain softening parameter.
This is demonstrated on Figure 3, which shows a nominal load path that would be followed by a two-way displacement controlled cyclic CNS test. This example test comprises 3 cycles of loading to ±wtest (where wtest > wpeak ) followed by ‘positive’ monotonic loading. During the first cycle the load path follows the monotonic curve through τpeak down to τfail-1 . The sample is then displaced −2wtest and then back to the origin. An important aspect to note at this point is development of the cyclic ‘gap zone’ which, following completion of the first cycle, exists between −wtest and +wtest and represents the maximum extent of the plastic pile-soil relative displacement. The second cycle commences and the sample is loaded back to +wtest . Relative to the previous
cycle when the sample was at this displacement, the total plastic displacement that the sample has been subjected to is approximately 4wtest , which is equivalent to a positive displacement along the ‘backbone’ monotonic curve of (approximately) 4wtest . Hence, the strength on the ‘backbone’ monotonic curve has reduced from τfail-1 to τfail-2 . At the edge of the gap zone the shear stress mobilised is therefore now equal to τwmax-2 in accordance with Equation (4). After another full cycle down to a total negative displacement of −wtest followed by reloading back to +wtest the maximum bias stress reduces to τwmax-3 , and similarly after another full cycle the maximum bias stress at the end of the gap zone is τwmax-4 . As the sample is then monotonically loaded past +wtest
129 © 2011 by Taylor & Francis Group, LLC
Figure 6. CYCLOPS – Residual shaft friction. Figure 8. CNS test comparison – RATZ and CYCLOPS.
2.2.2 Enhancements specific to CYCLOPS The CYCLOPS t-z algorithm incorporates a number of enhancements compared to the RATZ t-z model, namely: – Inclusion of a variable bias parameter (defined by bi and bf ). – Inclusion of an initial strain softening parameter (i-η). – Modification of the shape of the t-z curve within and immediately outside of the ‘gap zone’ formed during cyclic loading. The following sections discuss in detail the reasons for and implementation of the enhancements outlined above. Figure 7. CYCLOPS – Matching of various parameter sets.
and beyond the gap zone, the shear stress path gradually transitions from the maximum bias stress curve (i.e. τwmax-4 ) to the ‘backbone’monotonic failure curve (i.e. τfail-4 ) over a displacement equal to the cyclic transition width (wcyc ). The displacement controlled cyclic CNS test example described above illustrates an important aspect of the CYCLOPS t-z model, in that the amount of post-τpeak plastic displacement that occurs within the gap zone during cycling corresponds to an equivalent amount of plastic displacement along the monotonic τfail curve, and therefore essentially defines the magnitude of τfail at any point during cycling. It should be noted, however, that in the example test description above, it was assumed that all of the post-τpeak displacement was plastic. CYCLOPS actually assumes that a small component of the displacement is elastic (as determined by the maximum shear modulus Gmax ), and this elastic component is not included in the ‘total’ accumulation of plastic displacement that is used to calculate τfail .
2.2.3 Variable bias parameter As briefly described in Section 2.2.1, the bias parameter (b) defines the maximum bias shear stress (τwmax ) at the edge of the gap zone during cycling, primarily as a function of the monotonic failure stress (τfail ). The importance of the bias parameter can be demonstrated by examining a typical displacement controlled cyclic CNS test response. Such a response is shown plotted as shear stress (τ) versus horizontal displacement on Figure 2 and as the maximum shear stress attained during each cycle (τmax ) versus cycle number on Figure 8. The test comprised 25 cycles of displacement controlled loading at ±5 mm followed by monotonic loading to +12 mm, then to −5 mm and then back to the origin. The two points of interest to this discussion are:
130 © 2011 by Taylor & Francis Group, LLC
– The rate at which τmax reduces with increasing cycle number. – The recovery of τ during the post-cyclic monotonic loading phase (i.e. Cycle 26 on Figure 8) as the sample is displaced beyond the ±5 mm gap zone. Figure 8 also shows the calibrated CNS test response as determined using RATZ. In RATZ, b has an implied value of 1.0, such that the maximum bias
Figure 10. CNS test and CYCLOPS comparison – Example 1. Figure 9. Variable bias parameter definition.
shear stress curve illustrated on Figure 3 is identical to the corresponding ‘backbone’ monotonic failure curve.As indicated on Figure 8, there is no increase in τ as the sample is displaced beyond the ±5 mm gap zone at cycle 26. In attempting to fit RATZ to this test, one can decide to set τres equal to the fully recovered postcyclic test value as adopted for the response shown on Figure 8, but the drawback of this is that the τmax values obtained in each cycle are significantly (and non-conservatively) greater than the actual test values. Alternatively, one could decide to set τres equal to the value of τmax at the end of cycling. This would result in a better fit to the τmax versus cycle number response, but would fail to capture the post-cyclic recovery in τ to the final τres . Hence, for these reasons it was considered essential to introduce a user-specified bias b into the CYCLOPS model. However, further detailed evaluation of various CNS tests results demonstrated that b not only varied from sample to sample but also often decreased in value over the course of each test. Therefore, CYCLOPS incorporates a non-constant b, which is specified in terms of initial and final values (bi and bf respectively). The manner in which CYCLOPS calculates the value of b at any given stage of cycling can be expressed as:
The parameter Ngap is the number of load cycles where the axial displacement is sufficient to reach either side of the edge of the gap zone, while m is a simple exponent. Figure 9 shows the relationship between b (normalised with respect to bi and bf ) and Ngap for various values of m. The benefits of adopting a non-uniform b are illustrated by two CYCLOPS calibrated responses to the previously considered CNS test. Results from the first CYCLOPS fit (Example 1; uniform b) are presented on Figure 10 (τ versus horizontal displacement) and
Figure 11. CNS test and CYCLOPS comparison – Example 2.
Figure 8 (τmax versus cycle number). Identical parameters to those adopted for the RATZ fit were used, with the exception of i-η (which is not included in RATZ), wcyc (implied value of 0 mm in RATZ) and bi = bf which were both set to 0.21. The resulting CYCLOPS response shows significantly improved agreement with the observed test behaviour compared to RATZ, particularly with regards to τmax during the latter stages of cycling and the post-cyclic recovery to τres . However, it can be seen that during early stages of cycling CYCLOPS still underpredicts τmax , which is somewhat overconservative, and implies that a value of b greater than 0.21 should be adopted for the first few cycles. The second CYCLOPS fit presented on Figure 11 and Figure 8 (Example 2; variable b) adopted identical t-z parameters to Example 1 but with bi and bf set to 0.68 and 0.21 respectively. The resulting predicted response shows generally excellent agreement with the test response, highlighting the benefit of adopting a non-uniform b. 2.2.4 Initial strain softening parameter The ‘standard’ strain softening parameter (η) is an exponent that controls the rate at which the postτpeak monotonic failure stress (τfail ) decreases with
131 © 2011 by Taylor & Francis Group, LLC
Figure 12. Initial post-peak strain softening parameter.
Figure 13. Shear stress within gap zone.
increasing displacement, as demonstrated on Figure 5. Calcarenites typical of those encountered on the North West Shelf generally exhibit a fairly brittle post-τpeak response, with η values usually less than 0.5. For these relatively low η values, the negative gradient of the τfail versus displacement curve immediately following τpeak can be very large (in absolute terms), and it has been found previously that this large gradient can result in a slight overprediction of the degradation in shaft friction resulting from cyclic loading. In order to avoid this problem an additional parameter, termed the initial strain softening parameter (i-η), was incorporated into the CYCLOPS monotonic τfail algorithm which enables the immediate post-τpeak ductility of the τfail response curve to be increased slightly. The influence of i-η on τfail is illustrated on Figure 12, which plots τ/τpeak versus the post-τpeak displacement normalised by wres , assuming η = 0.4 and for various values of i-η. It should be noted that for a relatively large i-η value the CYCLOPS response matches the ‘baseline’ RATZ response (i.e. as shown by the iη = 1 × 106 curve on Figure 12). As i-η decreases, the post-τpeak displacement at which the CYCLOPS response curve rejoins the baseline (i-η = 1 × 106 ) response curve increases. This is demonstrated on Figure 12, where for i-η values of 1000 and 100 the CYCLOPS response rejoins the baseline response at approximately 0.8% and 7% of the displacement to residual (wres ) respectively. 2.2.5 Shear stress within gap zone As noted earlier the RATZ model gives a poor fit to the stress-displacement response within the cyclic ‘gap’ zone. A new formulation was therefore implemented in CYCLOPS to address this limitation. This modification is illustrated on Figure 13 which shows curves representing τ/τfail versus w/wcyc , where w is the displacement increment relative to the edge of the gap zone. Note that a negative normalised displacement represents behaviour inside the gap zone (i.e. a w/wcyc of zero defines the gap zone edge).
Example curves are presented for b values ranging between 0.05 and 0.95. 2.2.6 Shear stress outside gap zone As discussed earlier RATZ assumes a hardwired value of b equal to 1.0, implying that the maximum bias shear stress (τw max ) developed at the edge of the gap zone is equal to the current monotonic failure stress (τfail ). However, the available CNS test and GST data do not support this assumption since it is generally found that the value of τ at the gap zone edge is less than τfail . A certain displacement must be applied beyond the edge of the gap zone for τ to recover from the value at the gap zone edge to τfail . To address this limitation the CYCLOPS model allows the user to specify the displacement over which τ recovers from τw max (as defined by Equation 4) at the edge of the gap zone to τfail . This transfer displacement is termed the cyclic transition width (wcyc ). Further examination of the available CNS test and GST data indicate that the magnitude of wcyc is correlated with the size of the gap zone; an increase in size of the gap zone is matched by an increase in wcyc but at a decreasing rate, such that wcyc eventually appears to approach a limiting maximum value. Hence the CYCLOPS model incorporates a user specified maximum cyclic transition width (wcyc-max ). The value of wcyc at any stage is then determined according to the size of the gap zone (wgap ) and the exponential relationship presented on Figure 14, which plots wgap against wcyc , all normalised by wcyc-max . The implication of this approach is that wcyc /wgap decreases from close to unity (as implied by the dashed line on Figure 14) for wgap /wcyc-max near zero to approximately 0.2 for wgap /wcyc-max = 5. The shape of the shear stress-displacement curve assumed in the CYCLOPS model immediately outside the gap zone (i.e. w/wcyc = 0 to 1.0) is illustrated on Figure 15 for various values of b ranging between 0.05 and 0.95.
132 © 2011 by Taylor & Francis Group, LLC
Figure 16. OC GST 400S2 – CYCLOPS comparison.
Figure 14. Cyclic transition width.
Figure 17. OC GST 400L – CYCLOPS comparison.
Figure 15. Shear stress within and outside gap zone.
2.3 CYCLOPS model applied to GSTs The discussion presented to date has focused on the observed response in CNS tests. However, the same characteristic features are also clearly evident from GST data. Typical CYCLOPS fits to some of the OC GST results (Randolph et al, 1996) are presented on Figure 16 and Figure 17. Similar response characteristics have also been observed elsewhere, with the recent full scale pile test results from Haberfield et al. (2010) being a notable example. Despite the ability of CYCLOPS to provide a very good match to the OC GSTs, problems arose when attempting to find a common set of parameters that could realistically model both the OC CNS tests and the monotonic and cyclic GSTs. These problems were of two basic forms, namely: – Use of the optimal set of parameters derived from the monotonic GST responses resulted in premature
133 © 2011 by Taylor & Francis Group, LLC
failure when the grouted sections were subject to cyclic loading. – The optimal monotonic and cyclic parameters derived from the CNS tests were unable to predict the actual capacity of the grouted sections, giving conservatively low results. This originates from the fact that the observed CNS test τpeak values were low compared to the GST results. Although it is possible that this represents a systematic difference between the τpeak that can be achieved in a CNS test and that which can be achieved for a GST, we believe that this is unlikely to be the case. It is suspected that there may have been some problems in the sampling or testing of the OC cores which have contributed to the observed difference, but this is subject to confirmation. In summary, despite the much more realistic modelling that is possible with CYCLOPS compared to RATZ, the available correlation with GST data still gave a conservatively low prediction of measured GST response.This is of course safe for design purposes and hence is a robust position to occupy into the future. 2.4
Scale effects
In order to try to assess whether the underprediction of the GST response using CYCLOPS might be
due to some kind of scale effect between small scale CNS tests and full scale piles, or due to some other cause, a programme of numerical analyses was conducted using FLAC (Itasca, 2005). In these analyses an axisymmetric continuum model for the pile and soil was adopted, rather than the t-z approach used in CYCLOPS. These analyses focused on comparing responses under monotonic loading for CNS tests and GSTs and specifically addressed the potential role of borehole roughness on the GST response. A particular focus of this study was whether or not a full scale pile might exhibit a thicker interface failure plane (ie. shear band) than obtained in a CNS test, which could in turn lead to a ‘stretching’ of the required displacement to mobilise the residual friction. 2.4.1 Constitutive model The standard ‘Strain Softening Plasticity Model’ provided within FLAC was used for these analyses. Prior to the onset of plastic yielding the soil response was modelled as linear elastic. The soil cementation was modelled through a cohesive strength, which degraded progressively to zero with increasing plastic strain. Following breakdown of the cementation, the soil strength reverted to a frictional material with the strength defined by a friction angle; the friction angle was also variable starting from zero prior to any plastic yielding and then increasing to a maximum value as the true cohesion degraded to zero; this ‘peak friction’ angle reflected the fact that whilst the true cementation might be gone, the soil sample was still likely to be compact and interlocked at this stage. Further plastic shearing led to a gradual reduction of the friction angle until a ‘critical state’ was achieved at larger strain, and thereafter a constant value was assumed. Another key feature of these materials is contraction when subject to plastic shearing. This was incorporated into the soil model through the adoption of a negative dilation angle; this varied from a high initial value, corresponding to the maximum rate of contraction, and thereafter gradually increased (ie. became less negative) at increasing plastic shear strains. At the ‘critical state’ a zero dilation angle was imposed. A ‘cap’ yield surface was not included since the axial pile friction in these soil types is governed by normal stresses well below the cap yield surface and hence it is the shear induced contraction, combined with the effects of cementation degradation, that should be captured for a realistic model. Whilst cap models can simulate the shear induced contraction well, most are inadequate for dealing with the brittle cementation component. 2.4.2 CNS test simulation The shear stress-displacement response predicted in the FLAC CNS test simulation, assuming a normal stiffness applicable to a 2 m diameter pile, is presented on Figure 18. The initial brittle post-peak response reflects the rapid breakdown in cementation while the gradual reduction thereafter is due to the shearing induced compaction.
134 © 2011 by Taylor & Francis Group, LLC
Figure 18. CNS test simulation – 2 m diameter pile.
2.4.3 Mesh details for GST simulation Following a program of initial analyses, the key components identified for capturing realistic GST behaviour were identified as: – Meshes needed to be vertically aligned to ensure that the dominant and known shear band orientation (ie. vertical) could be captured accurately and within a single element thickness. Borehole roughness was introduced by assigning stiff and strong ‘pile’ properties to some of the elements that were adjacent to the pile but which would ordinarily be assigned soil properties (for a ‘smooth’ pile). At certain locations on the grout-soil interface, where the shear band is forced to propagate from one radial distance to another (invariably where a distinct change in grouted section diameter occurs), a bearing type failure is initiated at the ‘lip’ of the grouted section outstand, and this would, by definition, be smeared over several soil elements in front of the ‘lip’ and would involve imposition of high normal stresses. Strain softening and localisation into a shear band is not a critical feature of such a failure mode and hence the vertical alignment of the mesh is not believed to have a significant adverse influence on these bearing type failure modes. – Very thin soil elements were used immediately adjacent to the grouted section interface since this was essential in order to capture an appropriate degree of borehole roughness. This also offered the advantage that several elements would always be involved in any local bearing type failures that developed around individual hard-points. The minimum element thickness adopted in our analysis was around 4.5 mm at the nominal grout-soil interface, increasing steadily to the radial boundary of the mesh. For the five elements in closest proximity to the groutsoil interface the average element thickness was around 6 mm to 6.5 mm, which was very comparable to the element thickness used in the CNS test simulation (5.8 mm). This similarity of thickness is vital in order to avoid spurious mesh scale effects,
Figure 20. Load displacement response – ‘rough’ and ‘smooth’ 2 m GST.
imposed on the top of the mesh to represent a burial depth of around 45 m. – The grouted section itself was modelled as an elastic solid with a resultant EA that proved to be around double that of the actual OC test sections. A Poisson’s ratio of 0.3 was also assumed.
Figure 19. Grout-soil interface details – 2 m diameter GST analyses.
since the minimum shear band thickness is limited by the element thickness which ultimately controls the displacement rate from peak to residual friction. Analyses were undertaken with a variety of different roughness assumptions – details of two of the considered cases for the 2 m diameter GST are presented on Figure 19. The meshes used for these analyses had the following key features: – The modelled GST length to diameter ratio varied from about 1.5 to 6, which covers the range for most of the actual OC GSTs. No end bearing was included below the grouted section. – The outer radial boundary of the mesh, which was fixed against radial movement, was defined at a radius of 60 times the grouted section radius. – Soil was included above and below the grouted section in order to avoid local effects, such as unrealistic stress concentrations at the base of the grouted section and under-confined mesh distortion at the top of the grouted section. – The in situ effective horizontal stress at the mid depth of the mesh was the same as assumed in the CNS test simulation. A vertical surcharge was
2.4.4 GST analysis results The shear stress-displacement responses predicted in the 2 m diameter GST simulations are presented on Figure 20 for a variety of different roughness conditions. It may be noted that for all cases the peak frictions obtained from the rough GSTs are significantly less than the simulated CNS test peak. Some of the variability in peak friction from the GSTs is due to the slightly different diameters for the differing roughness, but the enhancement for the ‘rougher’ piles was generally significantly greater than the change in diameter would imply. The reason for the lower average peak friction in the GSTs compared to that obtained in the CNS test simulation was found to be due to progressive failure; at peak mobilisation of the GST a sufficiently high strain had developed over a short section at the top and a large section at the bottom of the grouted section such that the initial cohesion had degraded to zero. A purely frictional response therefore applied over these zones, which gave a significantly smaller interface friction than implied by the initial soil cohesion. The residual frictions exhibited a different trend with higher values obtained in all of the GST cases compared to the CNS test simulation. An evaluation of the various results revealed that this was due to several reasons:
135 © 2011 by Taylor & Francis Group, LLC
– Even at low roughness, the horizontal effective stress averaged along the GST was found to be higher than the normal stress acting at the residual condition in the CNS test simulation.
When considering results from this kind of analysis due caution should be exercised since such analyses are notoriously difficult to perform reliably. Our examination of the GST simulation results did not reveal any obviously faulty features, except (in a few cases) a slightly enhanced residual friction at large head displacements resulting from some over constraint in the system. Overall, we believe that these analyses do provide some evidence for an enhanced residual friction on a full scale grouted section compared to a CNS test, although we recommend that this conclusion should be treated as provisional in the absence of additional confirmatory data. In addition it may only be applicable in certain cases such as for soils subject to shear induced contractancy, as modelled here. Other questions may be asked as to the modelling assumptions made for the shear band. An alternative strategy, adopted in some cases (usually clay soils), is to include some strain rate dependency of the soil strength, which in turn controls the minimum shear band thickness that develops. However, for the cemented soil under consideration here, it is not readily apparent why any significant viscous strain rate effects might arise. Within a continuum framework, it is not apparent that any other assumptions could be sensibly made, in the absence of empirical data. However, it is possible that a more fundamental assessment using a discrete element method (DEM) approach might potentially reveal some different types of failure mechanism (at the micro-mechanical scale) in these kinds of soil, which could suggest a scalability of shear band thickness that is not resolvable in the current work. In any event, in order to further advance our understanding of this complex matter either additional testing at the OC test locations or more full scale GST or pile load tests combined with comprehensive programs of CNS tests from suitable sites elsewhere, will be required.
– Calculation of the ‘equivalent’ CNS spring acting along the GSTs revealed that while this was generally in accordance with the CNS test simulation, strong end effects were apparent which gave rise to the enhanced average horizontal effective stress. – For the rougher interface cases, highly concentrated pressure bulbs were mobilised just above many of the significant outstands on the grout-soil interface. Further detailed inspection revealed that each pressure bulb was associated with a transition of the sliding failure plane from a larger to a smaller diameter (i.e. to become closer to the nominal diameter of the pile). These pressure bulbs were therefore associated with local bearing failures around the outstand rather than pure frictional sliding failures. At these pressure bulbs, highly concentrated horizontal effective stresses were mobilised, which offset the contraction induced reduction in horizontal effective stress that occurred in the CNS test simulation, in turn enhancing the frictional resistance. As can be seen by comparing Figure 18 and Figure 20, the initial stiffness of the simulated CNS test response was found to be very much greater than obtained from the GST simulations. This is readily understood when considering the different configurations of the CNS and GST test responses; the former only required shearing of a single 5.8 mm thick element while the applied shear stresses at the grout-soil interface of a GST propagate throughout the mesh and hence elastic deformations accumulate over a wide area. From the same figures, it can also be seen that none of the GST analyses revealed any difference in the displacement required to transition from the peak to residual friction between a CNS and full scale GST. Hence despite the imposed interface roughness in the GST simulations, the shear band thickness in these analyses was controlled by the minimum element thickness in all cases; the increasing roughness cases did nothing to enhance the shear band thickness, and hence there was no mechanism to enhance the displacement to residual. 2.4.5 Conclusion from FLAC analyses Unfortunately the results of the FLAC analyses did not reveal a definitive explanation as to why the response in a CNS test might differ significantly from that of a full scale GST. In most cases the response of the simulated CNS tests were found to be a close proxy of the numerical GST analysis. However, the analyses did hint at the potential for some differences, both due to end effects (although these might not be important for an actual pile) and due to the influence of a ‘roughened’ interface. Specifically, increasing roughness appeared to enhance the residual friction compared to an equivalent CNS test. None of the analysis results suggested that the displacement required to transition from the peak to residual friction should be any different between a CNS test and a full scale GST.
2.5
The usual process by which the design storm history is incorporated into the cyclic axial capacity analyses for drilled and grouted piles offshore Australia is as follows: 1. A structural model of the platform incorporating all dead, live and environmental loads is used to determine the total pile head axial load occurring at the peak of the maximum design wave (i.e. maximum overturning moment) and at the trough of the same wave (i.e. minimum overturning moment) for the given design condition. 2. The structural model of the jacket is then subjected to waves of various height (H) and period (T) corresponding to the design lifecycle and storm wave H-T histograms. Processing of the model output results in a pile head cyclic axial load ‘transfer function’ which relates H and T to the pile head axial load at each wave peak and trough, thereby defining the loads applicable to each cycle. 3. The pile head cyclic axial load transfer functions, together with the number of cycles of each wave
136 © 2011 by Taylor & Francis Group, LLC
Influence of storm order
type as specified in the lifecycle and storm histograms, are input into the RATZ or CYCLOPS software in some order. Historically, the lifecycle and storm wave H-T histograms were generally only defined as occurrence matrices rather than specific time histories. Therefore, the lifecycle and storm occurrence matrices were arranged to produce various ‘ideal’ time histories, including ‘increasing’ (i.e. smallest wave to largest wave), ‘decreasing’(i.e. largest wave to smallest wave) and ‘increasing-decreasing’ (i.e. smallest waves at the beginning and end, with the largest wave occurring at the storm midpoint). Generally it was found that the ‘decreasing’ arrangement would result in the highest reduction in axial pile capacity as a result of cyclic storm loading. 2.5.1 ‘Actual’ storm load time history Recently ‘actual’ wave-by-wave storm time histories have been developed for pile capacity analyses. An example of an actual storm profile, in terms of pile head axial load at the wave peak and wave trough, is presented on Figure 21. Surprisingly, it was found that the cyclic degradation inflicted on a pile subjected to an actual storm was greater than that for a pile subjected to any of the idealised storm arrangements that had been used in the past. The reasons for this were not immediately apparent but on further investigation it was found that this was related to: – The arrangement of the larger waves within the profile. – The degree to which each of the larger waves can cause zones along the pile to go post-τpeak . – The degree to which packets of smaller waves, which occur after each of the larger waves, are able to cause further damage to the post-peak zones. Fundamentally, it is a basic characteristic of both the RATZ and CYCLOPS models that the degradation process is governed by the amount of plastic degradation that occurs for soil elements that are loaded at some point during the storm such that their monotonic (peak) strength is exceeded. Cycling at loads that never reach τpeak but which exceed the elastic threshold shaft friction ξτpeak can also cause degradation, but at a much slower rate than for any element that goes post-τpeak at some stage. As an example of the shaft friction degradation process, first consider a pile subjected to an idealised decreasing storm loading profile, where the largest load is applied first and the smallest last. The process is demonstrated schematically in a step-by-step fashion on Figure 22 and described as follows: – At the start of the storm the largest wave causes the soil along the upper portion of the pile to reach its peak shaft friction and to commence along the strain-softening part of the t-z curve (i.e. the post-τpeak zone on Figure 22a). This is due to the flexibility of the pile which results in much larger pile-soil interface movements at the top of the pile
137 © 2011 by Taylor & Francis Group, LLC
Figure 21. Pile head axial load profile – actual storm.
–
–
–
–
compared to the bottom. Hence, from this point on, these soil elements operate in the post-τpeak domain, as illustrated by the resulting failure shaft friction (τfail ) profile on Figure 22b. Subsequent large waves of decreasing height will cause some additional soil elements to transfer from pre-τpeak to post-τpeak behaviour, which enables the post-τpeak zone to propagate further down the pile (Figure 22c). In addition to this, further damage is inflicted on the initial post-τpeak zone, leading to a steady reduction in the available shaft friction (i.e. τfail ) over this region (Figure 22d). It is important to note at this point that since all the large cycles are lumped together at the beginning of the storm, the total amount of post-τpeak degradation that occurs in the first few cycles is limited. Hence, as the storm progresses and the modelled wave height reduces, the ability of subsequent waves to extend the post-τpeak zone rapidly decreases. Following the first few large cycles, the subsequent very large number of medium to small cycles are only able to continue to degrade the elements that were forced post-τpeak in the first few cycles (Figure 22e). Once full degradation of the post-τpeak zone to the residual shaft friction (τres ) has occurred, which is generally after only a relatively small proportion of the total number of smaller waves in a storm, the large number of remaining small waves are unable to cause any further damage (Figure 22f). Hence, at the end of the storm, the post-τpeak zone has generally transformed almost completely into a zone where τfail = τres . However, the initial postτpeak zone caused during the first large cycle is only extended to a limited degree by the subsequent cycles.
Now consider the same pile subjected to the same storm loads but with the load profile ordered such that the larger waves are spread out over the storm’s duration and separated by packets of smaller waves. An example of such an ‘arranged’ profile is presented on
Figure 23. Pile head axial load profile – arranged storm.
–
–
Figure 22. Pile axial capacity degradation – ‘decreasing’ storm.
–
Figure 23, which has been developed from the ‘actual’ storm history presented on Figure 21. The arranged profile comprises five ‘load packets’ with each packet comprising one large wave followed by a series of smaller waves ordered according to decreasing wave peak load. The process is demonstrated schematically in a stepby-step fashion on Figure 24 and described as follows:
–
– After application of the first large wave (Figure 24a), a post-τpeak zone will form along the upper portion of the pile (Figure 24b). This is the same as for the first cycle in the decreasing storm. – A subsequent packet of smaller waves will then work on the post-τpeak zone (Figure 24c). This
138 © 2011 by Taylor & Francis Group, LLC
post-τpeak zone is not extended, but most or all of it will be transformed into a τfail = τres zone (Figure 24d). Note that once a soil element goes post-τpeak , even very small subsequent loads may propagate the degradation, because any stress that exceeds the (very small) cyclic residual friction (τcyc-res ) leads to plastic strain and hence an accumulation of plastic damage. The small waves are used much more effectively (in terms of causing damage) than in the descending storm analysis, since one large wave followed by a moderate packet of small waves may lead to a τres zone only a little smaller than caused for the entire descending storm. When the large wave from the second load packet is applied (Figure 24e), the post-τpeak zone caused by the first large cycle is unable to offer much resistance to this load, since it has been fully degraded to the τres value. Hence, this load is distributed much further down the pile than would have occurred if this were only the second wave in a descending storm history. This therefore maximises the size of the new post-τpeak zone caused by this second large cycle (Figure 24f). A subsequent package of small waves then works on this new post-τpeak zone (Figure 24g) until it has degraded to τres (Figure 24h). With subsequent ‘packets’ of a relatively large damage triggering wave followed by a set of smaller waves the process continually repeats itself, with each larger wave propagating the post-τpeak zone further down the pile and each packet of smaller waves degrading the strength in this zone towards τres .
In summary, it can be seen that by separating the larger damaging waves by sets of smaller waves, cyclic damage is maximised. In the descending storm only a relatively few of the small to medium cycles make any contribution, whereas in an ‘optimally’designed storm (for maximising damage) all the small to medium cycles will contribute to the damage process.
this model for practical design purposes a significant number of CNS tests need to be performed. For each CNS test a best fit calibration is developed, which requires considerable skill and care, since as mentioned in Section 2.2.1, superficially similar monotonic responses over a limited strain range can result in a very different cyclic response. A variety of stress and displacement controlled CNS tests are required at each tested location in order to properly anchor the calibrated CYCLOPS response. Any excessive brittleness observed in the monotonic CNS test stress strain response also needs to be treated with extreme caution since this may never be apparent in a field situation. A well calibrated view on sensible ranges for the peak skin friction is required to ensure that an unconservative parameter selection is not made. Once the CNS test calibrations are complete the values of the various parameters adopted in each CYCLOPS fit are then plotted as a function of depth in order to ascertain appropriate trends. Design lines can be fitted through the calibrated data as desired. Our experience with this procedure so far has been good, with generally sensible ranges of parameter variability found from these calibration exercises. As more experience is obtained using the CYCLOPS model over time it is hoped that some of the required parameters may generally be found to lie within tight bounds, which may reduce the number of detailed calibrations that are required at each site. Notwithstanding, it is likely that design of drilled and grouted piles under complex cyclic loading conditions will remain a challenging task for many years to come. 2.7
Figure 24. Pile axial capacity degradation – ‘arranged’ storm.
2.6 Assessment of parameters for design The CYCLOPS model outlined in this paper can accurately model measured CNS test and GST responses, but this ability comes at a cost in that many different parameters need to be defined. In order to use
Results of an example analysis are presented here to demonstrate the application of CYCLOPS to a typical full size pile, and are compared to the equivalent results obtained using the standard RATZ algorithm. A steel pile has been adopted for this example with a 2.2 m outer diameter, a constant 90 mm wall thickness and it is assumed to be grouted in a 2.5 m diameter hole. The soil parameters used in the analysis are presented in Table 1. The assumed storm load history has been applied in various ways to illustrate how important this can be: the full actual storm (Figure 21), a ‘5 packet’ storm (Figure 23), a standard decreasing storm and an increasing storm. The pile head load-displacement response predicted using CYCLOPS for the ‘actual’ storm is presented on Figure 25, along with the undegraded monotonic response. It should be noted that the somewhat spikey appearance of the cyclic response, particularly for isolated large cycles, is due to the fact that only the peaks (minimum and maximum) of each cycle are saved in the analysis for later post processing. The t-z curves extracted from the actual storm analysis at a couple of nominal depths (9.8 m and 39.8 m) are presented on Figure 26. The extent of cyclic degradation caused by the storm is readily apparent from
139 © 2011 by Taylor & Francis Group, LLC
Example analysis
Table 1. analysis.
CYCLOPS & RATZ parameters for example
Parameter
CYCLOPS
RATZ
τpeak τres /τpeak Gmax ξ wres η i-η τcyc /τpeak bi bf m wcyc-max
440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 300 0.01 0.5 0.2 5 0.1 m
440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 N/A 0.01 N/A N/A N/A N/A
Figure 25. CYCLOPS pile head response – actual storm.
these figures, leading to an 18.4% reduction in the post-storm monotonic capacity. From the t-z curves it is apparent that the response is affected by cyclic degradation, even at 39.8 m below the mudline. This is further illustrated on Figure 27, which presents the peak mobilisable friction at the end of the storm for each of the different representations of the storm. It can be seen that the actual storm causes the greatest degradation, and is slightly worse than the ‘5 packet’ storm. The increasing storm causes little damage (0.8%), but it can still be observed from Figure 27 that the peak mobilisable friction is less than the prescribed peak friction over the upper 20 m. However, this reduction is principally due to monotonic progressive failure not cyclic loading. The standard decreasing storm that has generally been used for design purposes in the past is shown to be significantly less damaging (10.2%) than either the actual storm or the ‘5 packet’ storm. The key feature evident from Figure 27 is the propagation of damage to increasing depth with increasing disorder in the storm, as anticipated by the process described in Section 2.5. The pile head load-displacement response for the equivalent RATZ analysis of the actual storm is presented on Figure 28. The cyclic degradation (15.0%) is only a little less than obtained with CYCLOPS for this particular case. The RATZ generated t-z curves from the actual storm are shown on Figure 29. By comparison with Figure 26, the more stylised t-z response obtained from RATZ at a depth of 9.8 m is readily apparent. Also it can be seen that cyclic degradation did not propagate as far as 39.8 m in the RATZ model although this did occur with CYCLOPS. The peak mobilisable friction at the end of the storm is presented for the various RATZ analyses on Figure 30. Interestingly, the cyclic degradation obtained with RATZ for an increasing storm was slightly higher, albeit still very small (1.4%) compared to CYCLOPS. However, for all the other storm representations, CYCLOPS gave slightly higher cyclic degradation than RATZ. In summary it can be seen that for the example presented here CYCLOPS gave slightly lower
Figure 26. CYCLOPS t-z response – actual storm.
post-storm capacities than RATZ. However, we have found much larger differences in other practical cases, particularly where complex stratigraphy is encountered. The general rule to date is that CYCLOPS invariably yields lower post-cyclic capacities than obtained with RATZ, all other things being equal. The importance of modelling actual storm histories rather than idealised representations is also clearly demonstrated by both the RATZ and CYCLOPS analysis results. 3 PART 2 – LATERAL PILE DESIGN 3.1
A new model for determining the lateral response of piles in uncemented carbonate soils when subject to undrained cyclic loading is presented herein. The new
140 © 2011 by Taylor & Francis Group, LLC
Introduction
Figure 29. RATZ t-z response – actual storm. Figure 27. CYCLOPS post-cyclic failure shaft friction profiles.
Figure 28. RATZ pile head response – actual storm.
method retains the basic p-y curve format used as standard for design of laterally loaded piles in the offshore industry but builds on the design approaches used for shallow foundations and suction piles. Design of such structures is now generally performed using a ‘cyclic strength approach’. For these kinds of structures the standard practice is to define a ‘cyclic strength’, which is some reduced proportion of the monotonic strength, and is a function of the number of applied cycles, the degree of cyclic loading (2-way, 1-way, etc.) and the magnitude of strain that is deemed acceptable for the soil to sustain.Appropriate factors of safety are applied to these cyclic strengths and the combined bearing and sliding capacity is then determined in order to define
141 © 2011 by Taylor & Francis Group, LLC
Figure 30. RATZ post-cyclic failure shaft friction profiles.
the maximum loads that may be safely applied to the foundation, with only limited in-service movements. Unfortunately, modelling of laterally loaded pile behaviour is more complex than design of shallow foundations or suction piles since the ultimate geotechnical capacity is generally not the governing design criteria. For a typical long and flexible pile, an increasing applied pile load will simply mobilise the soil over a greater depth and hence it is the pile bending
moments and pile head displacements that are usually the critical design parameters. To assess these it is therefore necessary to determine the appropriate load displacement response at each depth level in the soil. In addition, whilst shallow foundations and suction piles are usually pure load controlled systems (i.e. applied loads do not generally redistribute in a global sense) the upper part of a laterally loaded pile when subject to cyclic loads will exhibit a partially displacement controlled response, leading to the potential to redistribute soil reaction pressures down the pile as cycling progresses. In order to capture these features we believe that it is first necessary to define the appropriate monotonic p-y response and then to perform a cycle-by cycle analysis to capture the progression of the response over time. 3.2
Model development
Development of the new p-y model comprised four basic stages: – Stage 1: Finite element type analyses were performed to establish the p-y response of a pilesoil slice element for a range of generic stress strain curves representative of uncemented carbonate soils.These were developed for various numbers of applied load cycles. Analyses were also performed to assess the modification required to allow for near-surface effects. – Stage 2: The results from Stage 1 were used to develop an appropriate p-y curve format that can be used for any specific soil strength assumption and required number of cycles. – Stage 3: Using the results from Stage 2 along with an examination of the characteristic response of raw simple shear test data, an algorithm was developed to allow a cycle by cycle analysis to be advanced. – Stage 4: The p-y curve format and the cycle-bycycle algorithm were implemented into a new Excel program, which has been named pCyCOS. To aid in comprehension of the new model, the basic algorithm for the new p-y model is first presented followed by the detailed analyses that were performed to calibrate the various required parameters. 3.3 Envelope p-y response The envelope p-y response is the first basic building block of the p-y model and defines the p-y response of a horizontal slice through the pile-soil system when subjected to load controlled cycling for various numbers of cycles. For any element of soil with a given normalised monotonic undrained shear strength (su /σvo , where su is the monotonic undrained shear strength and σvo is the initial vertical effective stress), an S-N curve may be defined, which is a representation of the normalised cyclic stress ratio (S = τcyc /σvo ) and the number of cycles (N) to mobilise a nominal shear strain level (γ).
142 © 2011 by Taylor & Francis Group, LLC
Figure 31. Envelope p-y response – basic concepts.
By drawing vertical slices through the S-N curve it is possible to define stress strain curves for any given value of N. This is illustrated on Figure 31a and Figure 31b. By implementing these stress strain curves into a finite element type model of a lateral pile element we can then develop the appropriate p-y curve that applies for the specified number of cycles, N, as illustrated on Figure 31c. It may be noted that a normalised pressure
response is defined from this assessment (p/σvo where p is the soil pressure applied to the pile element). the process described For the specified su /σvo above is repeated for different values of N and the results compiled, as shown on Figure 31d. This latter plot essentially defines the response of a lateral soil element when subject to load controlled cycling. For example, considering a pile element that imposes a constant peak cyclic pressure p1 on the soil, the lateral displacement for that pile element would be equal to y1 after the first cycle, y2 after the second cycle, y3 after the third cycle and so on (Figure 31d). The entire process outlined above is repeated for various values of su /σvo , as required, in order to cover the range of monotonic strengths expected in the field. We have conservatively adopted pure 2-way cyclic loading conditions in the model to date, and hence 2-way S-N data has been used to define the necessary stress strain curves for developing the envelope p-y curves. The model may be readily changed for any other degree of cycling, but would require additional assessment of the various input parameters.
3.4 Cycle by cycle p-y response The envelope p-y curves that define the pile-soil behaviour under pure load controlled cyclic conditions have been defined above. If the problem under consideration were purely load controlled, then the final model would require no further sophistication than to add together all of the load controlled element responses for any given number of cycles, N, in order to give the total response for the entire system after N cycles. However, as noted earlier, the (partially) displacement controlled nature of the pile-soil system means that pile-soil elements near the top will shed load to elements further down, as the upper elements, which are subject to larger cyclic displacement amplitudes, ‘soften’ to a much greater degree than the lower elements. A cycle by cycle model is therefore required to capture this behaviour. 3.4.1 First cycle The lateral soil response used for the first cycle is simply the p-y envelope response for N = 1 as discussed in Section 3.3. By definition this describes the monotonic loading response of a pile-soil element. Monotonic (first cycle loading) is a purely load controlled process and therefore for any given pressure, p1 , applied during the first cycle, the lateral displacement for this element will be y1 (Figure 32a). 3.4.2 Second cycle The p-y response assumed for the second cycle is defined from the envelope p-y response obtained for N = 2, but is not equal to the N = 2 envelope curve. For any given pressure, p1 , applied during the first cycle the lateral displacement obtained during the second
143 © 2011 by Taylor & Francis Group, LLC
Figure 32. p-y curve format for first and second cycles.
cycle, assuming that the same pressure p1 was still applied, would be equal to y2 (Figure 32b). However, in reloading to the pressure p1 during the second cycle, the pile-soil element is assumed to follow a different
pressure–displacement curve. This reloading curve is formed from 3 parts: Part 1 – Up to an applied pressure of 25% of p1 , the pile-soil element follows a linear path starting from the origin and giving a displacement of y0.25 at P0.25 (=25% of p1 ). (Evaluation of y0.25 is considered later.) Part 2 – For a pressure exceeding 25 % of p1 , the pile-soil element follows a second linear path that passes through the point (p0.25 , y0.25 ) and through the point (p1 , y2 ). Part 3 – Where the pressure defined from Part 2 of the p-y curve exceeds that which would be obtained using the N = 1 (monotonic) envelope curve, then the monotonic envelope curve is adopted. It can be seen that a key feature of this second cycle p-y response is an initial ‘soft’ zone followed by a hardening response. Such behaviour is encountered in all simple shear tests on uncemented carbonate soils where a reversal of the cyclic shear stress is imposed. The selection of an inflection point at 25% of p1 has been made based on inspection of the results from many simple shear tests. It may also be noted that the monotonic response may (eventually) be recovered, partly due to the dilatory soil response as large shear strains are applied following partial liquefaction. This is consistent with the philosophy that the generation of excess pore pressure during cyclic loading mainly affects the soil stiffness rather than its ultimate strength. However, it should be also appreciated that when a load cycle imposes a soil pressure that is close to the ultimate limit defined for cycle N, the degree of softening that occurs for cycle N + 1 may be such that there is a complete loss of stiffness under any further cyclic loads (Figure 32c). The pile will therefore be free to move through the softened soil, but would eventually encounter fresh soil, which would constrain the ultimate displacement somewhat. Large displacement hardening of this type has not been included in the model, but in practical applications it is unlikely to be an important limitation, due to the partially displacement controlled nature of lateral pile loading, since this also serves to limit the increase in displacement from one cycle to the next. 3.4.3 Subsequent cycles The form of the p-y curve used for all subsequent cycles (N = i) depends on whether or not the peak pressure applied to the pile-soil element in cycle N = i − 1 is the maximum pressure that has ever been applied to that element (pmax ), or is less than pmax . Where the peak pressure applied in cycle N = i − 1 (i.e. pi-1 ) is equal to pmax , the p-y curve for the next cycle is defined using the same procedure as used for the second cycle (Section 3.4.2), as illustrated on Figure 33a. This approach includes a degree of oversoftening in the p-y response, since inherently it is assumed that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to
Figure 33. p-y curve format for subsequent cycles.
a lower pressure during the earlier cycles, which would be expected to lead to a lower degree of softening for cycle N = i. Where the pressure applied in cycle N = i − 1 is less than pmax , two alternative criteria are used to assess the p-y response for cycle N = i, with the criterion that gives the greatest degree of softening adopted for the final p-y response. The first criterion is identical to that presented on Figure 33a: the displacement for cycle N = i is calculated assuming that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to a greater pressure during the earlier cycles. Inherently this means that the predicted degree of softening will underpredict that which would occur in practice, and indeed if pi-1 is significantly less than the maximum pressure ever applied, it can lead to a p-y curve for cycle N = i that is stiffer than that used for cycle N = i − 1, which is clearly a physically unreasonable outcome. Hence the second criterion is to first determine the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , (i.e. as per Figure 33a), and then apply a scale back factor to the predicted degree of
144 © 2011 by Taylor & Francis Group, LLC
softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure that is less than pmax . The scale back factor used in this calculation has been derived by consideration of the available S-N data and will be discussed later. This process is illustrated on Figure 33b. 3.4.4 Modifications for surface effects The p-y curves discussed above are defined based on a horizontal slice through the pile and soil. This therefore represents the ‘deep’ failure mode where the soil can flow around the pile but with no vertical movement of the soil. However, near the mudline, vertical movements can occur in the soil giving rise to wedge type failures. To account for this effect a scaling factor (‘p-multiplier’) is applied to the pressures calculated using the ‘deep’ failure p-y curves. The scaling factor is determined as the ratio between the ultimate lateral resistance calculated accounting for the actual embedment depth, to the ultimate lateral resistance that would be obtained for the ‘deep’ failure mode. The ultimate peak lateral resistance has been defined at any depth based on the plasticity model developed by Murff and Hamilton (1993). The validity of this assumption was confirmed through a program of 3D finite element (FE) analyses, using the models discussed in the following section. These analyses showed that the simple ‘p-multiplier’ approach proposed here generally gave results within 10% of those obtained from the more rigorous FE analyses, which is considered acceptable for engineering design purposes. 3.5 Analysis of envelope p-y response The envelope p-y curves required for the new p-y algorithm have been developed from FE analyses performed with ABAQUS (ABAQUS Inc., 2006) and finite difference (FD) analyses performed using FLAC (Itasca, 2005). We used two programs for this work, since they have different strengths and weaknesses. In addition, a few cases were analysed using both programs to enable internal verification of the results obtained. Irrespective of the specific program used, the meshes generated for developing the envelope p-y curves are based on the same basic assumptions: – A plane strain ‘slice’ through the pile and soil was modelled. – Infinite elastic elements were included to model the far field behaviour; these were found to be essential during the equivalent work performed to determine the p-y curves for the CHIPPER model (Erbrich, 2004). – The pile was modelled as having a ‘smooth’ (i.e. frictionless) interface with the soil. – It was assumed that no separation would occur between the pile and the soil, due to the presumed undrained nature of the applied loading. – A small strain formulation was generally adopted (although we also performed some confirmatory analyses using a large strain formulation).
3.5.1 Constitutive model All of the numerical analyses were performed using an undrained total stress approach. For all cases we selected the Mohr-Coulomb soil model, which degenerates to a simple Tresca model when a friction angle of zero is specified, as required for the assumed undrained conditions. To account for the very high degree of non-linearity in the stress-strain response, the cohesive strength was defined with a hardening response as a function of the plastic shear strain. The elastic part of the response was modelled using a simple linear elasticity model, with a Poisson’s ratio of approximately 0.5 to ensure constant volume conditions. 3.5.2 Stress strain response Stress-strain curves characteristic of uncemented carbonate soils were established based on the available test data. From these tests, a variety of stress strain curves were developed for use in the FE/ FD analyses which reflected the full range of different normalised monotonic strengths, su /σvo . For each case, we used the procedure presented in Section 3.3 in order to develop the shear stress versus shear strain curves required as input for the numerical analysis. These curves were defined for pure monotonic loading and for pure 2-way cyclic loading at various numbers of cycles. Typical stress-strain curves for a) soft/loose = 0.25 and b) dense soil with soil with a low su /σvo a high su /σvo = 10 are presented on Figure 34a and Figure 34b respectively. We performed verification analyses in both FLAC and ABAQUS under simple shear loading conditions in order to confirm that the intended stress-strain curves were indeed modelled accurately. Figure 35 presents one such comparison, which confirms that the model operated as expected. 3.5.3 Envelope p-y analysis – results Example analyses showing the resulting p-y curves obtained from the FE/ FD analyses using the input stress-strain curves shown on Figure 34 for a) soft/loose soil with a low su /σvo = 0.25 and b) dense soil with a high su /σvo = 10 are given on Figure 36a and Figure 36b respectively. These p-y curves are presented for a number of different applied load cycles. Similar curves were developed for a range of differ ent su /σvo values, encompassing the relevant range. From the range of results obtained, it was found to be possible to interpolate linearly between the values obtained in order to determine the appropriate p-y curves for any other desired combination of su /σvo and N.
145 © 2011 by Taylor & Francis Group, LLC
3.6
Parameters for cyclic p-y response
In addition to the envelope p-y response, the two other essential components of the new p-y model are the definition of the inflection point and the stiffness degradation function to account for reducing loads from one cycle to the next.
Figure 34. Typical envelope stress-strain responses. Figure 36. Computed envelope p-y curves.
(i.e. ‘stiffening’) load displacement path. The basic form for this p-y curve was obtained from an evaluation of the stress strain curves observed in cyclic simple shear tests performed on uncemented carbonate soil samples, where N > 1. Typical responses from these tests indicate that the convex nature of the stressstrain response in the first cycle contrasts with the concave response for all subsequent cycles (Figure 37). Furthermore, the test data implies a clear trend for γ0.25 (the shear strain at 25% of the maximum shear stress) to increase with increasing γcyc (the maximum cyclic shear strain in each cycle) and a conservative best fit to the data may be expressed as:
Figure 35. Comparison of stress-strain response.
3.6.1 Inflection point As discussed in Section 3.4.2 all load cycles after the first load cycle have a three part p-y format, with the first two parts describing a linear or concave upwards
Having obtained a relationship between γ0.25 and γcyc , it was necessary to define an equivalent relationship between y0.25 and yapp , where yapp is any given lateral movement of the pile. To assess this, some postprocessing was performed of the various horizontal slice FE/ FD analyses discussed earlier. This postprocessing took the form of evaluating the shear strain
146 © 2011 by Taylor & Francis Group, LLC
Figure 37. Detailed stress-strain responses.
(i.e. γcyc ) obtained at specific load levels throughout the FE/FD meshes used for these analyses, assessing γ0.25 at each element from Equation (6), and then determining appropriate weighted average values for γcyc and γ0.25 . It was then assumed that y0.25 can be calculated as a function of yapp as:
Figure 38. p-y inflection point data.
Various methods were tried in order to assess appropriate weighted values for (γ0.25 /γcyc )ave and it was found that the best approach, in terms of overall consistency and minimising the dependence of the final result on the size of the mesh used in the numerical analysis, was as follows:
where: Ne = the number of elements in the FE/FD mesh. dA = the area of element i in the FE/FD mesh. γ0.25 and γcyc are the relevant strains for element i in the FE mesh. This method gives greater weight to the values of γ0.25 /γrmcyc obtained in parts of the mesh that experience higher cyclic shear strains compared to those regions that are more distant from the pile and therefore experience lower cyclic shear strains. The weighting by γcyc dA was selected intuitively, but it would also be considered theoretically robust if a linear trend could be demonstrated between yapp and the integral of γcyc dA over the whole mesh. This was there fore checked for a variety of cases (su /σvo = 0.35 to 1.5 and N = 1 to 30) and it was found that an approximately linear response was obtained in all cases, with no more than an 8% (and generally a lot less) deviation from a linear relationship, which is acceptable for practical engineering purposes. The minimal mesh dependency obtained with this approach is illustrated on Figure 38 which presents y0.25 /D versus yapp /D for four test cases, where D is the pile diameter. These results were obtained from four horizontal slice analyses, all of which have meshes that contain an equal number of elements but with the outer boundary steadily increased (i.e. increasing outer
Figure 39. Model tests; gap width vs. cyclic displacement.
diameter (OD) to inner diameter (ID) ratio). Very similar results were obtained irrespective of the magnitude or N, which allowed for a single trendline to of su /σvo be fitted and incorporated into the p-y model, as also shown on Figure 38. It is interesting to compare the results presented on Figure 38 with the results of a series of laterally loaded model pile tests in carbonate soil undertaken at the University of Sydney (Randolph et al., 1988). Under 2-way loading these tests revealed a pronounced ‘S’ shaped hysteresis curve, with low lateral resistance obtained in the central region, which was assumed to be associated with a gap zone or softened soil. The size of this ‘gap zone’ (ug ) was measured and, as shown on Figure 39, was plotted in an identical format to that for y0.25 on Figure 38. Comparing these two figures it may be seen that the design line for y0.25 follows the same trend as the various lines presented for ug , but with a slightly steeper gradient.
147 © 2011 by Taylor & Francis Group, LLC
3.6.2 Stiffness degradation for reducing loads As discussed in Section 3.4.3, for N > 2 two criteria are used to determine the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . The second of these criteria first determines the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , and then imposes a scale back factor to the predicted degree of softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure, pi-1 , that is less than pmax . The approach that we have used to assess the magnitude of the scale back factor is to consider the relationship between normalised stress (or strength) and the corresponding equivalent amount of damage (i.e. incremental cyclic shear strain) at varying levels of cyclic strain. The initial step in this assessment was to examine the available data from 2-way strain controlled cyclic simple shear tests performed on uncemented carbonate soil samples. The results of these tests were assessed in terms of the normalised stress ratio τ/τref (where τref is the peak shear stress attained during the first cycle) versus the stiffness degradation ratio Kdeg at selected cycle numbers during each test. The value of Kdeg is in turn defined as γ/γref (where γref is the change in γ between the first and second cycles, while γ is change in γ between any other two cycles). For displacement controlled tests the strain does not change between one cycle and the next, but the stiffness does and hence it is still possible to ascertain γ and γref in an indirect manner as shown on Figure 40. Curves were fitted to the τ/τref versus Kdeg response for each test according to the following relationship:
where the exponent n was found (empirically) to vary as a function of the cyclic shear strain. The experimental results have demonstrated an appropriate form for Kdeg in terms of stresses and strains. However, for application to the lateral pile analysis, we need to transform these results into an equivalent format but with stress replaced by the lateral pressure acting on the pile, and with strain replaced by pile lateral movement. For this exercise we have assumed that pressure may be directly substituted for stress; i.e. P/Ppeak acting on the pile is assumed to be directly proportional to the values of τ/τref acting throughout the soil. We have then interrogated the various FE/FD analyses in order to determine a value for Kdeg-ave , which is the average stiffness degradation factor value applicable to a pile rather than a specific value that applies for a single soil element. For specified values of P/Ppeak , the appropriate values of γcyc and τ/τref are determined on an element by element basis from the FE/ FD analysis results, with γcyc first being used to determine the appropriate value for the exponent n,
Figure 40. Calculation of stiffness degradation from strain controlled tests.
which is then inserted along with τ/τref into Equation (9) in order to determine element specific values of Kdeg . The weighted average value of Kdeg-ave is then obtained following a similar approach to that used for assessing y0.25 , as discussed in the previous section:
Using this approach it was found that the best-fit relationships obtained for any specific value of yapp /D varied by only a modest degree for different soil types and cycle numbers, but there was a systematic variation with varying yapp /D. This is demonstrated on Figure 41, which presents the best fit curves of the form described by Equation (9) for varying yapp /D but with τ/τref replaced by P/Ppeak . As might be expected it can be seen that the stiffness degradation factor is higher (ie. more degradation) when the cyclic displacement amplitude is large. With Equation (9) recast in terms of P/Ppeak , unique values for the exponent n were then determined as a function of yapp /D. This simple equation finalises the determination of the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . 3.7
A computer program named pCyCOS has been developed to incorporate the new p-y model that is presented here. This program is based on the platform that was developed for the CHIPPER program (Erbrich, 2004) since the new algorithm has many structural similarities to the CHIPPER algorithm. It was therefore
148 © 2011 by Taylor & Francis Group, LLC
Description of the pCyCOS program
– Pile geometry parameters (diameter, wall thickness, length, stickup, Young’s modulus). – Pile loads (various combinations of lateral load, head moment, lateral head displacement and pile head rotation are permitted). – Whether or not monotonic or cyclic loading is to be considered, and if the latter, how many cycles are to be applied. Post-cyclic monotonic loading can also be specified. Output obtained from pCyCOS includes full information along the pile length after the required number of cycles has been applied (displacement, bending moment, shear force and reaction pressure), and p-y data pairs along the pile. The latter data can be compiled from various analyses at different load levels (with a constant N) to give p-y curves that can be used as input into structural analysis packages such as SACS or SESAM. It should be appreciated that a single unique p-y curve does not exist for any particular depth in the soil, even for a constant N, since it is a function of the pile head fixity. Hence it is necessary to perform full pile analyses for each particular case, and an iterative approach will be required with the structural designer in order to ensure that the finally defined p-y curves are consistent with the pile head fixity arising out of the structural model.
Figure 41. Stiffness degradation for reducing cyclic loads.
relatively straightforward to incorporate the new p-y model into CHIPPER, while retaining much of the pre-existing code. The existing CHIPPER code has been extensively tested and verified and hence this approach also helped to minimise the risk of bugs being introduced. It should be understood that the new program does not perform a full time domain analysis, such as that used in the CYCLOPS program for axial pile design discussed in Part 1 of this paper. Cyclic loading in the pCyCOS program is assessed on a peak-to-peak basis, meaning that the lateral response is only calculated at the peak of each cycle. This approach is consistent with that generally used to determine the undrained cyclic bearing capacity of shallow foundations and suction caissons over the last 30 years (Hoeg, 1976). In principle, analyses of full storm histories could be performed but in practice there are several important reasons why this is not recommended. The most important of these is that the assumptions made about the p-y form are really only appropriate when considering relatively few cycles. The appropriate equivalent number of cycles should therefore be calculated from the full design storm history data using strain accumulation methods similar to those used for these other applications. The input parameters required for pCyCOS include: – Profile of monotonic undrained shear strength, su , as a function of depth and the effective unit weight of the soil, γ . – Envelope p-y curves for various values of su /σvo , sufficient to encompass the full range of monotonic strength ratios encountered in the design profile, and for various numbers of cycles between N = 1 to N = ne + 2 where ne is the number of equivalent cycles that need to be modelled. Linear interpolation is used to assess any intermediate values of su /σvo and N from the input data set. – Parameters to define the stiffness degradation function. – Parameters to define the p-y curve inflection point.
3.8
Two examples are presented here to illustrate the performance of the pCyCOS model. It is unfortunate that there are no suitable prototype scale lateral load tests in uncemented carbonate soil which can be used to validate the model presented in this paper. The limited available historical data are generally for fully drained conditions, which are quite different to the undrained conditions that are applicable for the typical large diameter piles used in offshore construction, even for quite sandy conditions (albeit the equivalent number of cycles for these more sandy soils may be only 1 to 3). The first example presented demonstrates the key characteristics of the cyclic degradation model, and are compared qualitatively with the results of some centrifuge model tests. The second example considers a ‘typical’ offshore pile, and the results from pCyCOS are compared to those predicted with other commonly used methods. 3.8.1 Example 1 The first example illustrates the global stiffness response of a single rigid pile subject to 2-way cyclic loading with fixed displacement amplitude. A rigid 2.83 m diameter by 20 m long pile subject to pure translation is considered for this case. A monotonic undrained strength profile that varies linearly with depth has been assumed, defined as su = 0.5 σvo . The response obtained for this pile in terms of pile head load versus number of applied load cycles is illustrated on Figure 42. Results are presented for a
149 © 2011 by Taylor & Francis Group, LLC
Example applications
Figure 42. Example 1 – Displacement controlled cycling.
range of pile head displacement amplitudes, varying between 50 mm and 3 m. At the smallest displacement amplitudes it can be seen that the pile head load in the first cycle is modest, but this load hardly degrades with continuing cycles. As the cyclic amplitude increases, the pile head load in the first cycle steadily increases but increasing levels of degradation occur during subsequent cycles. After 10 cycles, the highest head load that can still be sustained is found to be for a 200 mm head displacement, albeit the stiffness has degraded by around 40% from that obtained during the first cycle. As the pile head amplitude is increased even further, the 10 cycle response gradually degrades to a very low value; at 3 m head displacement over 5 MN can be applied during the first cycle but not much more than 100 kN during the 10th cycle. This sort of response is entirely consistent with expectation and implies that the soil is liquefied by the very large stresses imposed during the first cycle. The dramatic degradation of soil stiffness under cyclic loading, as simulated by the pCyCOS model, has also been demonstrated in the laboratory as shown on Figure 43. This figure presents p-y curves extracted from two tests (numbered 1111 and 1131) which are extracted from Dyson (1999). These were obtained from centrifuge model tests of laterally loaded piles in carbonate silt. The p-y curves are neither constant load nor constant displacement amplitude but still show the same mechanisms as revealed on Figure 42. For example, comparing the results at the shallowest depth of 0.5 pile diameters and the deepest depth of 8 diameters it may be noted that that a much higher p/su ratio was mobilised in the first cycle at the shallow depth than at the deep depth. However, after several load cycles, the resistance of the upper soil has degraded to almost nothing (disregarding a small negative offset, which is assumed to be an interpretation/ measurement error), whereas a reasonable resistance is still obtained (albeit reduced from the first cycle) at the deepest depth.
Figure 43. Typical p-y response in carbonate silts (centrifuge).
3.8.2 Example 2 The second example represents a typical large diameter offshore pile, 2.65 m diameter by 75 m long, with a wall thickness varying from 110 mm over the top 20 m, to 45 mm below a depth of 30 m, and 85 mm in between. We have performed analyses using pCyCOS, the API soft clay p-y criteria and the method proposed in Novello (1999). Lateral pile analyses were performed for the extreme cases of fully fixed or fully free pile heads and we have considered both monotonic and cyclic loading. The API soft clay procedure has often been considered for many of the silty and muddy carbonate soils found offshore Australia, particularly under monotonic loading where the complications of liquefaction under cyclic loading do not apply. The monotonic undrained shear strength profile used in both the pCyCOS andAPI analyses is presented on Figure 44, which is representative of a typical uncemented carbonate silt at most depths, interspersed with a few stronger and
150 © 2011 by Taylor & Francis Group, LLC
Figure 44. Undrained strength for example analysis.
more sandy layers. The API cyclic method is not tied to any specific number of cycles and simply represents an envelope of ‘residual’ cyclic response for clay. However, the pCyCOS response is a direct function of the number of applied load cycles. Hence for this example we have adopted 20 maximum load cycles, which is a typical value to represent a design storm in fine grained soil, where fully undrained conditions can be expected over the duration of a storm. The other parameters required for the API analysis are ε50 and J, for which values of 1% and 0.5 were adopted respectively. As discussed in Section 1.2 the Novello (1999) procedure is formulated in terms of cone resistance, qc , rather than undrained shear strength and hence it is necessary to derive an appropriate profile of qc for the analysis. For the current example we have adopted a cone factor, Nkt , of 12 which is a typical value for undrained carbonate silts. The required profile of qc is then simply determined as 12 times the undrained shear strength. The Novello procedure nominally incorporates cyclic degradation in accordance with Equation (3). It is suggested in Novello (1999) that by adopting a typical K0 of 0.4 for uncemented carbonate soil, Equation (3) can be simplified to pcyclic = (1-0.6U∗ )pstatic . A complication with this approach is that U∗ needs to be determined from a completely separate analysis, and hence retaining internal consistency is rather difficult. We have therefore not performed an explicit cyclic analysis with the Novello procedure. However, it is still important to note that, by definition, the Novello procedure implies a maximum softening of only 60% in a fully liquefied soil (ie. u/σvo = 1), which is not consistent with available cyclic laboratory test data on carbonate soils, nor with the centrifuge tests discussed earlier (Dyson, 1999), and nor with the predictions of pCyCOS, as demonstrated by Example 1. The pile head load displacement responses obtained from the various cases considered are presented on Figure 45. The maximum bending moments at the pile
Figure 45. Comparison of pile head lateral deflection.
head for the fixed head cases are summarised on Figure 46, while the maximum bending moments within the body of the soil are summarised for both fixed and free pile heads on Figure 47. It may be noted that the pCyCOS and API results are remarkably similar for monotonic loading, with the pCyCOS response being marginally stiffer at low load levels and marginally softer at higher load levels. The pile bending moments also exhibit the same trend. This is surprising and probably coincidental given that the API approach is essentially a purely empirical method derived from load tests on soft clays, whereas pCyCOS is a theoretically based model, with the p-y curves determined from FE/ FD analyses using laboratory measured stress-strain curves from carbonate soils. We do not anticipate that such good agreement would always be obtained; for example in denser mate rials with higher su /σvo ratios than considered in this example. The Novello (1999) procedure leads to a generally softer monotonic response than obtained from the other two methods. This is not entirely surprising since drained lateral pile load tests are the empirical basis for the method, and the high compressibility of the carbonate soils in those tests has led to a significantly softer
151 © 2011 by Taylor & Francis Group, LLC
Figure 46. Pile head bending moments.
response than would be expected under incompressible undrained conditions. Hence while the Novello procedure purports to be able to address undrained loading, with cyclic softening also included, we consider that the method, and all similar methods, are only applicable where fully drained loading conditions are expected to apply. Whereas close correlation was obtained between the pCyCOS and API methods for monotonic loading, a rather different result is obtained under cyclic loading. Except at the lowest load levels, a much softer response and significantly higher bending moments are obtained with pCyCOS compared to the API method. However, this is not unexpected since normal clays do not exhibit the significant cyclic degradation (‘liquefaction’) that occurs with carbonate soils. The softening behaviour can be clearly seen on Figure 48, which compares the pCyCOS reaction force distributions along the pile at 1 and 20 applied load cycles for a fixed head pile subject to a 12 MN head load. It can be seen that as the upper part of the soil softens, the load is shed down the pile. This inevitably increases the pile deflections and the pile bending moment as shown on Figure 49. The change in reaction force is also illustrated on a cycle by cycle basis on Figure 48, at three points on the pile.
Figure 47. Maximum bending moments below mudline.
Figure 48. Reaction forces.
The applied reaction pressure in each cycle is normalised by the monotonic ultimate resistance, pu , applicable at each specific depth on this figure. Once again it can be seen that shedding occurs over the
152 © 2011 by Taylor & Francis Group, LLC
Figure 49. Deflections and bending moments.
Figure 50. Monotonic and cyclic p-y curves.
upper section of pile, where high reaction pressures are mobilised in the first cycle. These are transferred down the pile into regions where lower, and hence nondegrading reaction pressures are mobilised in the first cycle, which allows the applied reaction pressures to increase with advancing cycles. Example p-y curves extracted from pCyCOS at selected depths for a fixed head pile are presented on Figure 50 for monotonic loading and for N = 20. These demonstrate nicely the substantial difference between monotonic and cyclic loading conditions. The resulting p-y curves can be input directly into structural analysis packages, although some manipulation of the cyclic p-y curves may be required to avoid numerical difficulties associated with their strain softening nature. 4
CONCLUSIONS
This paper has presented and demonstrated application of two new tools for the analysis and design of piles in carbonate soil: CYCLOPS for axial loading and pCyCOS for lateral loading. The procedures used in both
153 © 2011 by Taylor & Francis Group, LLC
are underpinned by a strong theoretical basis, but also include a number of factors determined from empirical calibrations in order to properly address observed soil behaviour. A key objective in the development of these methodologies was to simulate real behaviour as accurately as possible, but where the available data led to any doubt, conservative assumptions have been adopted. The procedures were also developed to be applicable to a wide variety of general situations, although as with all geotechnical design in carbonate soils, great care and good engineering judgement remain vital ingredients. With respect to axial pile design, we consider that the ‘CNS calibration approach’ embodied in CYCLOPS is an essential process for developing safe designs of cyclically loaded drilled and grouted piles in cemented carbonate soils.Actual sites exhibit highly variable properties; these may differ significantly from the specific characteristics encountered at Overland Corner, which remains the only site where a comprehensive program of cyclic pile tests has been undertaken in cemented carbonate soil. Direct scaling of the OC test results is therefore not recommended. The CNS device is the main tool available for differentiating the behavioural characteristics between different soil types and hence underpins the CYCLOPS method. With respect to lateral pile analysis, the pCyCOS procedure is also based on a rigorous link to laboratory monotonic and cyclic element testing, from which the basic p-y curves are developed. It is unfortunate that at this stage we do not have any equivalent of the OC tests for cyclic undrained lateral loading in uncemented carbonate soils. However, the centrifuge tests reported in Dyson (1999) show the same qualitative response as obtained with pCyCOS. It should also be appreciated that this method only represents a modest philosophical advance on the industry standard procedures that would be adopted for design of a shallow foundation or suction pile, albeit the practical implementation for lateral pile analysis proves to be necessarily complex. The reality of the destructive effects of cyclic loading on shallow foundations in carbonate soils has been amply demonstrated in full scale field situations (Erbrich, 2005). While the approach in pCyCOS may yield significantly more onerous results for cyclic loading than implied by current procedures (and this may have implications for existing structures), the paramount concern has been to ensure safe design, taking due account of the observed behaviour in laboratory testing. To date, the authors have utilised CYCLOPS and pCyCOS for the design of several major new facilities and all these designs have been accepted by third party certification bodies. Finally, it should be appreciated that while the discussions in this paper have centred on carbonate soils, the methodologies developed are of more general applicability. For example, pCyCOS could be used for any kind of uncemented soil, provided suitable laboratory test data is available to enable development of
the required p-y curves. In practice, this would probably only be a worthwhile activity where the soil in question might be subject to significant cyclic degradation under the design loading. Similarly, CYCLOPS has been developed specifically for weakly cemented carbonate rocks, but could also be applied to axial pile design in other kinds of soft rock, provided suitable CNS test data is available to allow determination of the necessary calibrated parameters. ACKNOWLEDGMENTS The authors would like to acknowledge their colleagues who made significant contributions to this work. In particular we would like to thank Professor John Carter who performed detailed independent analyses to verify the pCyCOS program, Edgard Barbosa Cruz, who performed the ABAQUS analyses used in the development of pCyCOS and to Mohamed Khorshid who provided support and encouragement throughout this work. We would also like to thank the many others in both AG and throughout our industry who provided constructive and valuable contributions to the development of these new design tools. REFERENCES ABAQUS Inc. (2006), ABAQUS Version 6.6, User Manual, Dsimuli Abbs, A.F. (1983). Lateral pile analysis in weak carbonate rocks. Proc. Conf. on Geotechnical Practice in Offshore Engineering, ASCE, Austin, 546–556. API (2000). API RP2A – Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms—Working Stress Design, 21st Edition, American Petroleum Institute, Washington. Coyle, H.M. and Reese, L.C. (1996). Load transfer for axially loaded piles in clay. J. Soil Mechanics and Foundations Division, ASCE, 92(SM2), 1–26. Dyson G.J. (1999), Laterally Loaded Piles in Calcareous Sediments, PhD Thesis, UWA. Dyson, G.J. and Randolph, M.F. (2001). Monotonic lateral loading of piles in calcareous sediments. J. Geotech Eng. Div, ASCE, 127(4), 346–352. Erbrich, C.T. (2004). A new method for the design of laterally loaded anchor piles in soft rock. Proc. Offshore Tech. Conf., Houston, Paper OTC 16441. Erbrich, C.T. (2005). Australian Frontiers – Spudcans on the Edge. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth.
Haberfield C.M. Paul D.R. Ervin M.C. and Chapman G.A. (2010) Cyclic Loading of Barrettes in Soft Calcareous Rock using Osterberg Cells. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth. Hoeg K. (1976), State of the Art – Foundation Engineering for Fixed Offshore Structures, Proc. BOSS76, Trondheim. ISO (2007). ISO 19902 – Petroleum and Natural Gas Industries — Fixed Steel Offshore Structures. International Standards Organisation, Geneva. Itasca (2005), FLAC Fast Lagrangian Analysis of Continua, User Manual, Itasca Consulting Group. Jeanjean, P., Watson, P.G., Kolk, H.J. and Lacasse, S. (2010). RP 2GEO: The new API recommended practice for geotechnical engineering. Proc. Offshore Technology Conf., Houston, Paper OTC20631. Kraft, L.M., Focht, J.A. and Amarasinghe, S.F. (1981). Friction capacity of piles driven into clay. J. Geotechnical Engineering Division, ASCE, 107(GT11), 1521–1541. Murff, J.D. & Hamilton, J.M. (1993), P-Ultimate for Undrained Analysis of Laterally Loaded Piles, Journal of Geotechnical Engineering Div. ASCE, 119(1), pp 91–107. Novello, E. (1999). From static to cyclic p-y data in calcareous sediments. Proc. 2nd Int. Conf. on Engineering for Calcareous Sediments, Bahrain. 1, 17–27. Randolph, M.F. (1986). RATZ: Load transfer analysis of axially loaded piles. Report No. Geo:86033, Department of Civil Engineering, The University of Western Australia. Randolph, M.F. (1988). Evaluation of grouted insert pile performance. Engineering for Calcareous Sediments, Perth, 2, 617–626. Randolph, M.F. (2003), RATZ Version 4.2 – Load Transfer Analysis of Axially Loaded Piles, User Manual. Randolph M.F., Jewell R.J. & Poulos H.G. (1988), Evaluation of Pile Lateral Load Performance, Proc. Eng. for Calcareous Sediments, ed. Jewell R.J. & Khorshid M.S. Randolph, M.F. and Jewell, R.J. (1989). Axial load transfer models for piles in calcareous soil. Proc. 12th Int. Conf. on Soil Mech. and Found. Eng., Rio de Janeiro, 1, 479–484. Randolph M.F., Joer H.A., Khorshid M.S. and Hyden A.M. (1996), Field and laboratory data from pile load tests in calcareous soil’, Proc. 28th Annual Offshore Tech. Conf., Houston, Paper OTC7992. Randolph, M.F. and Wroth, C.P. (1978). Analysis of deformation of vertically loaded piles. J. of Geotechnical Engineering Division, ASCE, 104(GT12), 1465–1488. Wesselink, B.D., Murff, J.D., Randolph, M.F., Nunez, I.L. and Hyden, A.M. (1988). Analysis of centrifuge model test data from laterally loaded piles in calcareous sand. Engineering for Calcareous Sediments, Perth, 1, 261–270. Wiltsie, E.A., Hulett, J.M., Murff, J.D., Brown, J.E., Hyden, A.M. and Abbs, A.F. (1988). Foundation design for external strut strengthening system for Bass Strait first generation platforms. Engineering for Calcareous Sediments, Perth, 1, 321–330.
154 © 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin Centre for Offshore Foundations Systems, UWA, Perth, Australia
E.C. Clukey BP America Inc., Houston, USA
J. Garnier Laboratoire Central des Ponts et Chaussées, Nantes, France
R. Phillips C-CORE, Newfoundland, Canada
ABSTRACT: Today, centrifuge modelling is a recognised tool, both within the academic and the industry community, to help understanding the behaviour of offshore structures and assist in their design. However, although expanding, its real impact on design practices is limited. The paper proposes, therefore, to revisit the perception of the advantages and disadvantages of centrifuge modelling and the technological developments achieved over the last few years, in the light of recent offshore projects undertaken. It aims to provide both academic and industry perspectives. Examples are presented that illustrate the difference between centrifuge testing and centrifuge modelling and which highlight the contribution that centrifuge methods can make in designing offshore structures and in providing new insights into soil-structure interaction. 1
INTRODUCTION
The initial use of centrifuge modelling in offshore geotechnics took place in Manchester University in 1973 where the behaviour and performance of gravity platforms to be used in the Gulf of Mexico was investigated (Rowe & Craig, 1981). The work encompassed a wide range of soil and loading conditions and provided pivotal insights into the failure mechanism taking place (Craig & Al-Saoudi, 1981). Even in the very early days of centrifuge modelling (when the technological capabilities were limited compared to present day), it was yet understood and acknowledged that centrifuge modelling could contribute significantly to design when novel or incompletely understood conditions (both in terms of soil and loading) prevailed (Craig, 1984). Since this pioneering work in 1973, a significant number of projects have been undertaken worldwide, on a wide variety of offshore geotechnical problems. A few centrifuge centres have emerged, in the wake of Manchester University, developing expertise on offshore geotechnics through collaboration with industry, such as Cambridge University in the UK (CUED), C-CORE in Canada, LCPC in France, Deltares (former Geodelft) in Netherlands, University of Colorado, Boulder in the USA and COFS at the University of Western Australia.
The work performed, focusing first on phenomenological and site-specific studies, developed progressively towards more general studies, including the observation of failure mechanisms and the understanding of soil-structure interaction, eventually aiding in the development of predictive design methods in some cases. This transition from what was initially centrifuge modelling to centrifuge testing is discussed further in the paper. As centrifuge modelling was technically and scientifically developing, along with an increasing need for performance data and understanding of offshore soil structure interaction, the acceptance and awareness by the offshore community of the benefits of the use of centrifuge grew significantly. A key milestone in that process was certainly the keynote address given by Don Murff at OTC (Offshore Technology Conference) in 1996 to the wider offshore community, advocating the benefits of centrifuge modelling (Murff, 1996). Since then, the number of offshore projects benefiting from centrifuge modelling input has increased significantly and industry users have developed a strong expertise in analysing centrifuge modelling outcomes and incorporating them into their design approach. Nevertheless, as new needs arise from the increasing complexity of offshore projects, along with new possibilities from the technological development of new modelling techniques, it appears necessary to
155 © 2011 by Taylor & Francis Group, LLC
revisit the benefit and the contribution centrifuge testing and modelling may provide to the offshore industry in designing offshore structures subsea facilities and pipelines. This paper aims then to answer the fundamental questions any potential user has once facing a challenging offshore project: Why use a geotechnical centrifuge? Is it cheaper or more expensive than alternative investigation technique?; is it more or less reliable?; is it quicker to obtain relevant data?; does it account for behaviour aspects other techniques do not?; should it be replace or complement other techniques such as 1g reduced scale modelling or field testing? How to use a geotechnical centrifuge? Is it better to use natural or artificial soils?; should it investigate a wide range of options or focus on a particular solution?; should it aim at validating a design or assisting in the design; should it be modelling or testing? When to use a geotechnical centrifuge? Should it be considered at the early stages of the project or latter once some key parameters are already established?; at what stage within the design process is centrifuge modelling the most valuable? To answer these three questions, this paper proposes: 1. to revisit the advantages (and disadvantages) of the use of centrifuge in light of the scientific and technological developments of the past decade; 2. to detail the contribution of the use of centrifuge on industry practices, highlighting the difference between centrifuge testing and centrifuge modelling, an; 3. to evaluate the impact of centrifuge modelling/testing in industry practice and eventually to present some perspective for the future.
to obtain qualitative or quantitative information about the problem. The difference between centrifuge testing and modelling follows similar concepts. By performing centrifuge testing, one aims at validating or confirming a general set of assumptions about a geotechnical problem, while by performing centrifuge modelling, one aims at predicting or anticipating the behaviour or response of an actual geotechnical structure. The difference is significant as the requirements, the modelling techniques and the objectives for the modeller are different, as will be the outcomes for the user. 2.1
Focusing on soil-interaction structure, examples are presented from the authors’ experience, providing the perspective of both academic modellers and an industry user. Indeed, this presents a small part of the entire centrifuge work performed world wide. However, the authors believe that the variety of examples is representative of the contribution and possibilities that centrifuge testing and modelling can offer for the offshore industry.
2
ROLES OF CENTRIFUGE TESTING AND CENTRIFUGE MODELLING
The process of designing, performing, analysing centrifuge experiments and eventually integrating the outcomes into the design of offshore structures cannot be undertaken correctly without a clear understanding of the difference between centrifuge testing and centrifuge modelling. In engineering, it is common to refer to in-situ testing and numerical modelling. The former is usually used to validate an assumption (soil properties, design loads, soil response, etc) while the latter resorts to an idealisation of the problem investigated
156 © 2011 by Taylor & Francis Group, LLC
Centrifuge testing
The aim of centrifuge testing is to create an idealised representation of a problem in order to obtain quantitative and/or qualitative prediction about the mode of behaviour of the structure investigated, (Lee, 2001). It is therefore necessary to interpret the results in light of a model for soil-structure behaviour, and then use that model to simulate the design situation. Being an idealisation, the model does not necessarily replicate all the features of a real structure (commonly called prototype) and consequently, some flexibility is permitted regarding the representativeness of the model as long as the key features of the behaviour which are investigated are modelled correctly. The difficulty is to know which features are simulated exactly, approximately or not at all and to account for these differences in the interpretation of the results. The complementary use of other modelling and testing methods (such as finite element analysis) may help in this interpretative process. The idealisation relates mostly to the soil and the structure as follows: – The soil used in the model may be different from the prototype as long as some pivotal features of the behaviour are reproduced (such as the dilatancy of dense sand or the shear strength of cohesive soil, for instance). Local heterogeneity and non-critical stratigraphy or topography are often not simulated, but may still be accounted for to interpret dissimilar behaviour. – Some particular features of the prototype (geometric, mechanics, etc), which are believed to not have a significant effect on the phenomenon investigated, are not simulated. – Phenomenon aside of the ones investigated, and for which the influence is not believed to be dominant, are not simulated (a specific method of installation, a particular loading sequence, etc). In designing the model and choosing which features to replicate or not, the engineer needs to make assumptions about the behaviour to be modelled, to understand the consequences of the simplifications made and to define clearly the purpose of the model. Centrifuge testing may be conducted to: – Develop an initial understanding of the engineering concern.
– Identify a particular failure mechanism on which an analytical solution can be developed. – Observe a particular mode of soil behaviour (is the response drained or undrained, does the soil flow or collapse? etc). – Gather performance data which can be used to calibrate a numerical model. – Perform a sensitivity study to understand the relative importance of various parameters in the behaviour of the structure. – Perform parametric studies to generate design charts. – Determine soil properties by performing reduced scale testing of in-situ soil properties.
2. The requirements to represent the prototype are much stricter, particularly with regards to the soil characteristics and the loading conditions. Hence, as opposed to centrifuge testing, one, when designing centrifuge modelling should ensure that: – The same soil, featuring the same properties as the prototype, is used in the model. – The model considers all the details of the prototype (e.g. particular attention must be paid to the fidelity of the model). – All key phenomena involved in the response of the structure must respect the similitude principles. – Loading conditions must be the same between the model and the prototype (rate of loading, amplitude and frequency of loadings, etc).
Indeed, the objectives sought may affect the idealisation made as much as the assumptions and some judgement will be required to decide whether or not the outcomes may be directly extrapolated to prototype conditions and whether or not they need to be checked or validated by other means. The objectives presented could sometimes be achieved by other means, such as by numerical modelling, but the centrifuge offers the advantage to replicate specific features of the soil behaviour or the structure loadings. These include soil softening and reconsolidation, large deformation, cyclic loadings, collapse and the installation process. All of these features are particularly relevant for offshore geotechnics. Often, the complementary use of both numerical modelling and centrifuge testing accelerates the understanding of the phenomena, as the two methods are based on different sets of assumptions.
In other words, in centrifuge modelling, all aspects of the model must be properly scaled to that of the prototype. The objectives of centrifuge modelling may be various, although they are limited to the case studied. Extrapolation to other conditions should be undertaken with caution. Centrifuge modelling should help to: – Validate a technical solution (type and geometry of structure). – Validate or evaluate the performance of a final design (design loads, displacements within the limits established, etc). – Validate a mode of behaviour upon which the design was based (i.e. mode of collapse). – Observe the response of a given structure under a particular type of loadings, possibly different from the one used for design (e.g. check the performance of the structure under cyclic loadings).
2.2 Centrifuge modelling The aim of centrifuge modelling is to establish the validity/relevance of certain assumptions, designs or models. It aims to directly replicate the field conditions. The results provide an immediate representation of the design situation. What characterises centrifuge modelling is the uniqueness of the model investigated. It must refer to a specific prototype (as opposed to centrifuge testing) and focus on a particular aspect of the validation. This allows definitive conclusions about the prototype behaviour to be made. Note that the assumptions, design or model to be validated might have arisen from centrifuge testing (Lee, 2001). The reference to a particular prototype results in two fundamental differences with centrifuge testing: 1. The use of similitude principles and associated scaling laws (Garnier et al., 2007) is required to extrapolate the information gathered on the model (typically the measurements of dimensions such as loads, displacements, etc) to the prototype. The use of similitude principles in centrifuge testing is indeed required (this is the justification for the use of the centrifuge), but the scaling laws (e.g. the use of prototypes dimensions) are not necessarily required.
The objective presented here could be also achieved by field testing. However, centrifuge modelling offers the advantage (while ensuring the correct stresses are applied within the soil), of being less costly, provides a better control of the modelling conditions (e.g. known soil characteristics, better quality of data), and allows more data to be collected in a shorter period. It also provides the opportunity to repeat the experiments if required.
2.3 The use of centrifuge outcomes into design The design of an offshore foundation structure follows a series of steps, which are simplified in Figure 1. The soil and loading conditions will dictate a particular type of foundation whose geometry and characteristics will be determined using design methods, either empirical or based on constitutive soil models linking strains and displacements to the stresses and loads. Along with these different steps, the geotechnical engineer will use various tools to help him assess the soil conditions, choose the relevant technical solution, develop or use an existing design method and eventually validate the design.
157 © 2011 by Taylor & Francis Group, LLC
Figure 1. Steps involved in designing geotechnical structures and potential input of centrifuge modelling and testing (white shapes symbolises design steps, shaded shape symbolises centrifuge modelling or testing).
These tools can be divided in two categories; field testing and numerical/analytical modelling. Field testing is used to characterise the soil and help validate the design, while the modelling methods would be used to develop the design. The objective of this paper is to describe the potential centrifuge methods may have to assist in geotechnical design. As illustrated in Figure 1, by performing both testing and modelling, centrifuge methods may be a valuable tool for geotechnical engineers in every phase of the engineering process. Centrifuge testing may be used to develop a new foundation concept, a new technical solution, or to characterise soil properties. It may also be used to investigate soil behaviour and failure mechanism to establish constitutive models. In contrast, centrifuge modelling may be used to validate a new foundation concept, to validate a design and to calibrate and validate a specific design method. In offshore geotechnical engineering, where guidelines and design rules play a lesser role than onshore, engineers’ judgement is paramount in establishing the most technically sound and economically efficient design. Basing judgment on reliable and quality data from a range of modelling and testing methods may make a significant difference between a satisfactory and an optimal design.
The following sections will present various examples of centrifuge testing and modelling, highlighting the contributions to design. It is not implied here that centrifuge methods should necessarily replace more classical methods. Indeed, for routine design, classical methods are often preferred as they provide a safe and efficient path towards satisfactory solutions. However, for more challenging geotechnical design issues, which frequently occur in offshore engineering, geotechnical engineers should seriously consider the potential centrifuge methods may offer. 2.4
The offshore industry has come to accept the advantages of centrifuge methods as more and more results from various studies became available. ExxonMobil (former Exxon) were initial leaders in this area with a number of programs performed in the early 1980s (Murff, 1996). With the publication of the ExxonMobil results and, as the industry encountered more complex problems where 1-g testing was either difficult and/or expensive to perform, centrifuge testing or modelling became a more viable option for other operators. Joint Industry Programs involving a significant centrifuge testing component have become increasingly more frequent in the past 2 decades, such as
158 © 2011 by Taylor & Francis Group, LLC
Industry perspective
the SEPLA: suction emplaced plate loading anchor JIP involving 5 operators and 2 consulting groups in early 2000 at C-CORE. The Pressure Ridge Ice Scour Experiment (PRISE) performed at C-CORE attracted over 7 operators and 2 regulatory agencies and developed guidelines for design of buried offshore pipelines subject to ice scouring (Phillips et al., 2005). The INSAFE JIP, gathering 19 offshore operators, jack-up manufacturers, safety regulators and consulting companies, has been using centrifuge testing and modelling outcomes initially from Cambridge University, and more recently from COFS and NUS (National University of Singapore) to provide guidelines for design and best practice of jack-up installation and performance. Other examples of offshore JIPs using centrifuge techniques, in complement of more traditional techniques, includes the SAFEBUCK JIP for the design of on—bottom pipeline subjected to buckling, the MERIWA JIP, involving six major offshore companies, to model submarine landslide and their impact on pipelines. In France also, several centrifuge projects were carried out at LCPC, founded by CLAROM (www.clarom.com), an institution which gathers marine and offshore companies to carry out R&D programs aiming at better understanding offshore structures behaviour and improving design methods. The success of centrifuge techniques outside the offshore community has also helped to convince management to support various centrifuge efforts. The ability, for instance, of centrifuge tests to accurately define the failure mechanism for the levy failures in New Orleans during Hurricane Katrina is a recent example of the usefulness of the technique for defining failure modes. (Sasanakul et al., 2008 and Ubilla et al., 2008). The use of centrifuge testing by numerous earthquake researchers involved in a very large United States program (Wilson et al., 2010) also helped improve the credibility of the methodology which led to testing of large gravity based structures under ice and seismic loads (DeWoolkar et al., 2006 and DeWoolker et al., 2008). More recently the industry has used centrifuge techniques for a large number of deepwater foundation as well as subsea and pipeline issues, based on an operator-testing contractor partnership, rather than based on a JIP (Gaudin et al., 2010a). The timeliness in which results could be obtained was often a major reason why centrifuge methods were selected vs. alternative techniques such as 1g testing. This will be developed further in section 5 and 6.
3
PERCEPTION OF ADVANTAGES AND DISADVANTAGES OF CENTRIFUGE MODELLING AND TESTING
The advantages and disadvantages of centrifuge modelling are well known, and have been listed and commented in many publications, including Murff (1996) for offshore related problems.
However, centrifuge testing and modelling is an advancing science and the advantages and disadvantages are continuously changing with the problems they are investigating. Hence, some previously perceived disadvantages are disappearing due to progress made in centrifuge modelling theory and techniques, while others are appearing as centrifuge methods are applied to new geotechnical problems. This section presents advantages and disadvantages of centrifuge methods, as they relate to offshore geotechnics today. 3.1 Advantages The main advantages are as follows: 1. Centrifuge modelling properly simulates the prototype body forces and, thereby, maintains proper relationships with stresses and strains. 2. Because actual model tests are small, tests can be performed in a relatively short amount of time and at reduced cost, allowing the investigation of a wide range of parameters. 3. The field in-situ conditions are more easily replicated than is usually the case for 1-g model tests. Advancements in in-situ devices used to measure soil properties in the centrifuge, such as the piezocone and T-bar (Stewart and Randolph, 1994) have significantly enhanced the knowledge of the material being tested. 4. Pore pressures dissipate much more rapidly than in the prototype condition. This is clearly an advantage when preparing a model for testing, since it reduces consolidation time, and for testing, as cyclic sequences over several years can be modelled within hours in the centrifuge. For monotonic loading in clays, undrained conditions in clays can usually be maintained by loading to failure at a higher rate and correcting for strain rate effects. 5. Centrifuge test data can be used to validate numerical models. This method is very efficient since the soil characteristics, the boundary conditions and the applied loads are known. This is not always the case in full scale field testing. 3.2
Disadvantages are more difficult to list, as it is necessary to separate the perceived disadvantages resulting from a somewhat limited knowledge of the physical modelling principles and the real disadvantages, which the centrifuge modelling community clearly identifies (Garnier et al., 2007), and for which mitigation or correction measures may be applied. The following presents a list of the disadvantages and the measures to undertake to avoid or mitigate them: 1. Scale effects (different from size effects) were for a long time listed as the main disadvantage of centrifuge methods, supposedly preventing extrapolation of centrifuge modelling and testing outcomes to field problems. As the size of the particles may
159 © 2011 by Taylor & Francis Group, LLC
Disadvantages
not be negligible when compared to the dimensions of the model, some particular soil behaviour may be generated, that is different to that expected in-situ. This is a scale effect, first identified by Ovesen (1975) who observed different responses for footing of various widths loaded on sand. Ovesen concluded that, for flat footing resting on sand, the width of the footing should be higher than 30 times the mean diameter of the particles. This prevents any scale effects to occur (in this case, an increase of the prototype bearing capacity with a reduction of the footing width). Indeed, the identification of a scale effect as the ratio of a relevant model dimensions to the particle size indicate that this type of problem is more likely to be critical for sand than for clay. Over the years, various research has been undertaken to quantify the minimum model dimension to particle size ratio for a variety of problems. This includes a significant number of bearing related geotechnical problems and, maybe more importantly, interface shearing problems where the scaling of the interface roughness may lead to significant difference in behaviour between the model and the prototype (Garnier & Köenig, 1998). The outcomes of all this work and some significant progress made in establishing similitude principles (notably through dimensional analysis) have been gathered in a scaling law catalogue published by Garnier et al. (2007). This catalogue provides modellers with a set of rules to design valid modelling and testing, free of potential scale effects, or to account for them when extrapolating the results to prototype conditions if they cannot be avoided (e.g. increased shearing forces due to scale effect can be calculated following procedure developed by Lehane et al. (2005). In practice, as already pointed out by Murff (1996), scale effects are rarely an issue in offshore geotechnics, especially as marine soils are mostly constituted of small particles (clay and silt in majority). If scale effects are really a major concern (for instance, if prefailure displacements in sand need to be accurately scaled, Palmer et al., 2003), modelling of models (testing different size models at different accelerations to model the same prototype) is a relevant technique to validate the modelling outcomes. 2. Creep is a complex phenomenon which may not be modelled correctly in the centrifuge. If creep relates to volumetric compaction (e.g., the displacement of a flat footing subjected to sustained vertical loading), the rate of creep is governed by the rate water escapes or moves into a volume of soil. It is therefore scaled by n2 , similarly to consolidation. Consequently, any creep phenomenon would occur n2 time faster in the centrifuge and would be captured by the model within the time frame of the experiment. If creep relates to shear deformations (e.g. the displacement of a pile subjected to sustained tensile loading), the rate of creep is governed by the viscous resistance of the soil structure and displacements in the centrifuge would occur at
the same rate as that in the field (as the strain rate would be equivalent). This result has, for example, been observed in centrifuge loading tests of shallow footing resting on dry sand (Canepa et al., 1988). Consequently, creep phenomena related to shear strains are difficult to capture in a centrifuge model within the usual time frame of an experiment. 3. Inertia effects scale differently than body forces. Therefore, for earthquake problems the loading frequencies need to be significantly increased. Even with these enhanced frequencies the pore pressure response is much faster than the prototype case in cohesionless sandy soils and a more viscous pore fluid is required to properly model this response. This is now a commonly used technique as demonstrated in section 4.4. 4. For problems where shear strain localization issues are important, such as ice gouging of the seafloor, centrifuge test interpretation of discontinuities versus shear zones is important. The use of physical modelling of models and complementary numerical modelling has been found to be beneficial in correctly interpreting the observed behaviour (Phillips et al., 2010). 5. The acceleration field across the models are variable and dependent on the radius from the centre of rotation to any point on the model. Corrections are available to account for this effect. Although some of the potential problems can be overcome or mitigated, as described above, the different scaling relationships for body forces, pore pressures viscosity and inertia can make testing more complex and difficult to perform. It relies then on the experience of the modeller to clearly establish the limits of the validity of the model. 3.3
An implicit advantage of centrifuge testing is that similitude relationships are almost always directly addressed. Conditions where similitude is or is not achieved are more carefully considered. Therefore, the advantages and limitations of the technique are better defined. In contrast, 1-g model tests have significant similitude issues, requiring the modification of the analytical models used, to match the experimental model test data. However, if the experimental model tests do not properly account for all the key factors that influence the prototype behaviour, then the potential for errors when the analytical models are extended to the prototype condition is significant. A potential useful non-geotechnical analogy to illustrate the importance of considering similitude is the model testing performed for deepwater production facilities such as TLPs, SPARs and FPSOs. For almost any large scale offshore project that utilizes one of these facilities, extensive tests are performed in model test basins (Newman, 1977). The modellers and engineers who perform these tests are always aware of the key model parameters, such as the Froude number and
160 © 2011 by Taylor & Francis Group, LLC
Consideration of similitude
Reynolds number when interpreting results. Because all the key similitude parameters for their problem are usually not satisfied (Lyle Finn, personal communication), the model tests are augmented with extensive field measurements to further enhance measurements used to calibrate analytical and design models. Perhaps regrettably, comparable field measurements are not often performed on offshore foundations. Therefore, there is a greater reliance on the model test data for the foundation design procedures and a greater demand on centrifuge modellers to provide modelling validation. The laws of similitude need to be carefully considered when performing model tests. To assist modellers in validating their model tests, the centrifuge community has been working extensively to produce a scaling law catalogue, listing scale factors and dimensional analysis procedures for a wide range of geotechnical problems (Garnier et al., 2007). A particularly relevant example is the dimensional analysis carried out by Boylan et al. (2010a) to validate the centrifuge modelling of the behaviour of the runout of submarine slides. While 1g laboratory are commonly performed to investigate runout behaviour, the modelling of a phenomena which relies on both soil mechanics and fluid mechanics necessitated detailed similitude analyses. 4
CENTRIFUGE TECHNOLOGY DEVELOPMENT
Over the last 30 years, the progress made in miniaturised electronics, micro-computing, software engineering and digital imagery have been incorporated almost immediately in centrifuge technology. It results in more realistic simulations, more accurate measurements and more detailed observations, subsequently improving the benefit centrifuge modelling can yield to the geotechnical community. This is particularly relevant for offshore geotechnics where complex environmental conditions, emphasis in capacity and saturated soil conditions prevail. Hence, sophisticated motion control to apply environmental cyclic loadings, image acquisition techniques to observe failure mechanisms and enhanced data acquisition systems to measure accurately pore pressure developments and fast responses have been key developments in centrifuge modelling techniques. Some of the most noticeable developments recently undertaken in centrifuge technology are presented below, with examples highlighting the benefits obtained from these improvements presented in sections 5 and 6. 4.1
Data acquisition system
In the early days of centrifuge testing data acquisition was limited to post-testing measurements and observations. With the development of computing and electronic, on-board data acquisition systems appeared. A typical data acquisition system would
Figure 2. New high-speed data logging unit developed at COFS. The unit replaces standard PC-data acquisition card combination (after Gaudin et al., 2009a).
include transducers, a signal amplifier and an analog/digital converter system in a slave computer onboard of the centrifuge (Liu et al., 1988; Garnier & Cottineau, 1988). The signal would be transmitted through electrical slip rings to a master computer inside the control command room. In such a system, electrical slip rings constitute the weak point (Taylor, 1995). They are prone to failure and generate significant electrical and mechanical noise. Over the years, improvements in sampling, signal resolutions and signal conditioning were undertaken, with 128 channels sampled at 100 Hz with a resolution of 16-bits being the norm. The use of either optical slip rings or wireless transmission also contributed to improve the quality to the data acquired. Nevertheless, the overall architecture remained unchanged as one can see in the development of the newest centrifuge facilities (Ng et al., 2006). Recently, some significant breakthroughs in data acquisition technology have been made at COFS. An independent self-powered unit 150x60x40 mm in size, embarking processing and storage units and capable to monitor up to 8 transducers, has been developed for centrifuge application (Gaudin et al., 2009a). Each unit (see Figure 2) is capable of powering and monitoring eight instrument channels at a sampling rate of up to 1 MHz at 16-bit resolution. The data are stored within the logging unit in solid-state memory, but may also be streamed in real-time at low frequency (up to 10 Hz) to the centrifuge control room, via wireless transmission. Unlike PC-based data acquisition solutions, this system performs the full sequence of amplification, conditioning, digitization and storage on a single circuit board via an independent micro-controller allocated to each pair of instrumented channels. It results in a much cleaner signal and the faster sampling rate, coupled with automatic triggering features, proves to be extremely useful in acquiring data during cyclic events or fast events such as submarine landslide (Boylan et al., 2009a). 4.2
Actuation devices have followed an improving trend similar to data acquisition systems. Early devices
161 © 2011 by Taylor & Francis Group, LLC
Robotic control
such as hydraulic mono-axis actuators developed into bi-axis electric or hydraulic actuators, to eventually multi-axis robots, first in LCPC (Derkx et al., 1998) and then in HKUST (Ng et al., 2002). In parallel, the development of sophisticated motion control systems occurred (De Catania et al., 2010). Complex loading sequences, including varying control modes on various axes are now possible, mimicking more realistically typical offshore loadings and permitting the study of particular aspects of soil structure interaction. One example is the investigation of pipeline embedment during laying process, when the pipe is subjected to a combination of horizontal and vertical cyclic motion with varying amplitude (Gaudin & White, 2009). This example is developed in more detail in section 6. The last ten years have also witnessed the development of actuation devices to generate wave loadings and wave-induced phenomenon. The device commonly includes a wave paddle (Sassa & Sekigushi, 1999) or a plunger moved by a horizontal (Baba et al., 2002). In both cases, frequency and amplitude of the motion of the wave generator device may be controlled in order to generate the appropriate wave spectrum. Particular attention is paid to limit or control the wave refraction against the edges of the centrifuge container and the use of fluids with viscosity higher than water to satisfy the similarity rule with regard to laminar flow (e.g. conservation of the Reynolds number). The drum centrifuge, which features a high developed length, was initially used for such tests (Phillips & Sekiguchi, 1992) and is particularly appropriate to model waves (Gao & Randolph, 2005). The steel catenary riser tests at C-CORE described in section 5.7 utilise 2 synchronised servo-hydraulic actuators to provide the necessary broad band frequency surge and heave motions (Figure 3). The actuators are counter balanced to minimise driving forces to maximise the system frequency response. 4.3
Digital imagery
In the early ages of centrifuge modelling, computer image processing techniques were used in combination with digital cameras (Garnier et al., 1991). The subsequent development of the particle image velocimetry (PIV) and photogrammetry techniques, for soil mechanics, by White et al. (2003) have constituted a major breakthrough for centrifuge technology, expanding considerably the range of investigations. More detailed observations can be gathered and insight into failure mechanisms taking place around foundations can be obtained. This is particularly relevant for offshore foundations where failure often governs the design, as opposed to onshore foundations where designs are governed more by displacements or serviceability. The technique consists in processing digital images of the soil and structure investigated, placed against a transparent observation window (see Figure 4), in order to obtain information about the displacements of the soil and the structure (White et al., 2005).
Figure 3. Servo hydraulic actuators developed at C-CORE.
The use of the PIV technique has been pivotal in a significant number of breakthroughs on the behaviour and performance of offshore structures. One may highlight: 1. The identification of the back flow mechanism, as opposed to the wall collapse mechanism previously assumed, governing the limiting cavity depth during spudcan penetration (Hossain et al., 2005). 2. The identification of the bearing capacity mechanism for spudcans penetrating sand overlaying clay, which showed the changes in overall failure mechanism due to varying geometric and strength conditions of the layered soil (Teh et al., 2008). 3. The identification of the spudcan extraction mechanism and the resulting insight provided about the development of suction at the spudcan invert (Purwana et al., 2006) or the identification of the ground movement generated by spudcan penetration leading to indirect loading of nearby piles (Leung et al., 2008). 4. The identification of the berm creation mechanism and its influence on the soil resistance during lateral pipeline motion (Dingle et al., 2008). 5. The identification of a Hill-type reverse end bearing mechanism during uplift of skirted foundation, due to the development of suction at the foundation invert (Mana et al., 2010).
162 © 2011 by Taylor & Francis Group, LLC
Other pivotal examples include the identification of the strain path followed by a soil particles nearby the tip of a pile which contributed to the development of theoretical solutions for the assessment of pile
Figure 4. Typical PIV setup in the COFS drum centrifuge (after White et al., 2005).
Figure 5. Soil displacement vectors for skirted foundation with embedment ratio = 0.2 subjected to uplift (Mana et al., 2010).
base resistance (White & Bolton, 2004) and the understanding of the phenomenon of friction fatigue which governs the shaft resistance along piles driven in sand (White & Bolton, 2001).
The PIV technique may provide insights, not only on the soil displacement pattern and the failure mechanism taking place, but may also reveal the particular behaviour of the structure investigated. A
163 © 2011 by Taylor & Francis Group, LLC
typical example of such insight of offshore structures behaviour is the identification and the understanding of the mechanism governing the behaviour of keying flap of suction embedded plate anchors. Recent experiments clearly demonstrated the inefficiency of the keying flap in reducing the loss of embedment during keying (Gaudin et al., 2010b). All these examples highlight the dramatic improvements in the utility of physical modelling in providing important insights in the mechanisms and behaviours which can subsequently be used to develop, improve or validate design and calculation procedures. 4.4
Soil sample reconstitution and characterisation
Improvements in soils sample reconstitution results more from the experience gathered by modellers and from the better understanding of soil behaviour, than from technological developments. Automatic air pluviation techniques to reconstitute sand samples and in-flight consolidation of normally consolidated clay samples or in-press consolidation of over consolidated clay samples are now well established techniques in every centrifuge centre, resulting in homogeneous and well controlled soil samples (Garnier, 2002). Nevertheless, recent projects have seen a significant increase in sophistication of the soil samples reconstituted, with natural soils sourced in-situ (as opposed to commercially available clay and sand) becoming more common, both in full-size samples, or in small boxes for PIV analysis. This has become common practice at, for example, LCPC (natural clay from North Sea and from West Africa) and at COFS (carbonate sediments from North West Shelf, soft clay from West Africa). The reconstitution of multi layered soils (typically a sand layer within a clay seabed) and the use of alternative pore fluid in granular material (such as silicon oil or aqueous methyl cellulose) to increase the pore fluid viscosity in order to generate and observe liquefaction phenomenon is also becoming common practice. A typical example is the reconstitution of a sample featuring a kaolin clay layer saturated with water overlaying a silica sand layer saturated with silicon oil. This technique has been used to replicate in-situ conditions offshore Hong Kong, where suction caissons to found wind turbine tripod will be installed (Gaudin & Randolph, 2008). 100 cst silicon oil (e.g. 100 times more viscous than water) was used as the pore fluid to model correctly the pore pressure regime resulting from environmental cyclic behaviour whose frequency could not be scaled correctly by a factor of 100. The silicon oil, therefore allowed proper scaling of the pore fluid. Unsaturated conditions may also be simulated in centrifuge models as demonstrated during the European project NECER (Rezzoug et al., 2000). In parallel to the increasing sophistication of reconstitution techniques using natural soils, alternative materials have been developed, replicating the geotechnical properties of natural soil, with the aim to either simplify and increase the reliability of the
reconstitution techniques or provide particular features to maximise testing outcomes. Artificial soil made from mineral ingredients, high boiling liquid and solvent and exhibiting shearing behaviour identical to soil, has been developed in order to overcome the existing limitations of reconstitution techniques (Sarma, 2006). However, the very good control of the reconstitution process of natural and kaolin clays and the improvement in soil characterisation techniques have dramatically limited the need of such artificial soil. Another artificial material which is seeing much use is the “transparent soil” (Iskander et al., 2001). By mixing flumed silica with paraffin oil and white spirit, one can obtain a transparent slurry, which upon consolidation, exhibits a strength profile similar to clay materials. The process permits the visual observation, through a window, of the offshore structure investigated. This is similar to the PIV technique, but it also permits accurate measurements of the loads applied, which is not possible with the PIV technique due to the contact of the structure with the observation window. An example offshore application using observation, through transparent soil, has been the investigation of the behaviour and trajectory of an embedded plate anchor (Figure 6) during keying and pullout (Song et al., 2009). Improved in-flight soil characterisation techniques may also be partly responsible for the improvement in sample reconstitution techniques. Reduced scale versions of cone penetrometer, piezocone and vane apparatus, identical to the in-situ tools, were used successfully since the early day of centrifuge testing to characterise soils. The last two decades have, however, seen the development of new characterisation tools in the centrifuge. This has proved to provide more accurate and more comprehensive information about the soil properties. This is notably the case of the T-bar penetrometer (Stewart & Randolph, 1994), initially developed for the centrifuge, which is now widely used offshore. The T-bar provides an accurate measurement of the soil undrained shear strength and offers a valuable alternative to cone penetrometer devices. The latter requires various corrections and have a disadvantage of a low ratio of the change in resistance to the ambient hydrostatic pressure acting on the cone tip (Randolph et al., 2005). In addition, more information can be extracted from T-bar results as illustrated by recent research. By performing successive cyclic sequences, one may access the strength degradation rate, the soil sensitivity and the post reconsolidation strength (Hodder, et al., 2010, Zhou & Randolph, 2009). By performing tests at various penetration rates, one may access information about the consolidation characteristics of the soil and the strength enhancement due to viscous effects (Chung et al., 2006, Lehane et al., 2009). More recently, a new analysis technique, based on the T-bar, has been developed which allows the accurate determination of the shear strength in the first meter of the
164 © 2011 by Taylor & Francis Group, LLC
Figure 7. O-bar developed at C-CORE.
Figure 6. Anchor model in transparent soil (after Song et al., 2009).
seabed, accounting for buoyancy of the apparatus and surface effects (White et al., 2010). This analysis technique is particularly relevant to assess the soils shear strength for soil-pipeline interaction. This is developed further in section 6.3. C-CORE has recently developed an O-bar for the assessment of near surface strength by reducing the bar diameter to 2.4 mm while increasing its length to 110 mm using the form of an annular ring (Figure 7). Other full flow penetrometer such as the ball penetrometer (Kelleher & Randolph, 2005) and more recently the piezoball penetrometer (Boylan et al., 2010b) have also been developed and have started to be introduced offshore (Figure 8). The centrifuge environment, with perfect control of the testing conditions, is the appropriate environment to develop and validate new soil characterisation tools. Near future will surely see new devices arising, such as the “doughnut” shearing device developed to measure peak and post peak pipe-soil friction factors and the associated pore pressures (Yan et al., 2010). Soil containers have also benefited from some technological developments. Thermally controlled strongboxes (Phillips et al., 2002) and ice generators (Lau et al., 2002) have been introduced to investigate the behaviour of offshore structures in artic regions. These
Figure 8. Piezoball penetrometer developed at COFS to investigate strength and consolidation properties of soft soils (after Boylan et al., 2010b).
are based on the similitude principles for soil freezing established by Yang & Goodings (1998). 4.5
165 © 2011 by Taylor & Francis Group, LLC
Models and instrumentation
The general improvement in instrumentation and sensing technology has indeed benefited centrifuge technology. Laser sensors offering high resolution are progressively replacing linear displacement transducers. The miniaturisation of commercially available load cells, pore pressure and total pressure transducers results in more accurate and more numerous measurements. This is giving access to information pivotal to understand detailed phenomena. One relevant example is the use of contact stress transducers to measure the radial stress changes on the outside of the 0.4 mm thick skirt of a suction caisson during installation, consolidation and subsequent extraction (Chen & Randolph, 2004). Measurements provided direct insight onto the skin friction of the caisson skirt, which was previously back calculated from penetration and extraction resistance. Jeanjean et al. (2006) used a double walled suction caisson to separate internal and external skin friction by nesting two stainless steel tubes and measuring the
Figure 10. Example of model suction embedded plate anchor (30 mm high, 53 mm wide) replicating all key features of prototype anchors, including keying flap and flap hinge eccentricity (from COFS internal report).
Figure 9. “Stroud” type load cell with protective sleeve removed (after Daiyan et al., 2010).
loads applied both to the inner wall and on the total system. There was a 0.4 mm gap between the outside wall and the inside wall of the anchor. Specific instrumentation for centrifuge applications has also been developed. For instance, Take & Bolton (2002) and Oung & Bezuijen (2002) have developed pore pressure tensiometers which overcome some of the limitations of the commercially available pore pressure transducer Druck PDCR-81. Daiyan et al., (2010) have used a miniature sealed ‘Stroud’ type load cell to measure the axial, transverse and moment loads transmitted through a pipeline subject to oblique loading in sands and clays (Figure 9). The use of bender element to measure in-flight shear wave velocity of a soil sample is also becoming regular practice (Brandenberg et al., 2006). The manufacturing of models has also improved significantly, with more precise foundation and anchor models now commonly available. An example of this improvement relates to drag and plate anchor models which feature particular details such as shank and fluke geometry and keying flap. The use of devices to replicate suction installation of caissons (Clukey & Phillips, 2002, Chen & Randolph, 2007b, Raines et al., 2005, Jeanjean et al., 2006, Colliat et al., 2010) or suction embedded plate anchors (Gaudin et al., 2006c), has also become common practice (Figure 10).
5 THE CONTRIBUTION OF CENTRIFUGE TESTING/MODELLING The contribution of centrifuge testing and modelling to the design of offshore structures and the understanding of offshore structure soil interaction has been addressed successively by Murff (1996), Garnier (2004) and Gaudin et al. (2006a). With the continuous development of new centrifuge technology and the great state of flux exhibited by
the offshore industry, continuously moving towards deeper waters and unusual soil conditions, it is anticipated that new areas will emerged in the near future. While the offshore industry will most likely use more and more centrifuge modelling techniques and benefits to its outcomes, it is as important to appreciate that the offshore industry will contribute significantly to the development of new modelling and testing techniques. The following provides a review and some additional examples of where centrifuge testing has enhanced offshore foundation engineering. 5.1
As opposed to field tests, centrifuge modelling provides homogeneous and well characterised soil conditions, known boundary conditions, accurate measurements of parameters and repeatable testing conditions. Therefore centrifuge testing offers reliable performance data for a given idealised problem which can be used to calibrate analytical and numerical models. These can subsequently be applied to specific field problems, which by nature are more complex. Some relevant examples, outside those already presented by Murff (1996), Martin (2001) and Gaudin et al. (2006a), include: 1. The calibration of numerical model for the design of the foundations of the Rion-Antirion bridge in Greece (Garnier & Pecker, 1999), a novel type of foundation including a sand raft resting on pilereinforced marine clay and subjected to seismic loading in addition to the more common environmental loadings. 2. The calibration of the geotechnical model used to design the pier foundation of the Confederation bridge in Canada (Phillips et al., 1998). The Confederation Bridge main pier foundation design
166 © 2011 by Taylor & Francis Group, LLC
Provide performance data to calibrate analytical or numerical models
was carried out using the LRFD approach and a probabilistic assessment (Becker et al., 1998). A reliable assessment of bearing resistance requires that the failure mechanism that is likely to develop in a foundation material be known and understood. There is limited field data regarding the performance of structures founded on ring footings, particularly in an offshore environment where high horizontal loads and moments are generated. Therefore, centrifuge model tests of the main pier foundations were carried out to investigate potential mechanisms of failure, to examine how these mechanisms vary with changes in foundation and loading conditions, and to evaluate the foundation design analysis carried out as described above. This testing also helped to reduce some of the inherent uncertainties associated with the analyses, such as limitations in constitutive models, and to improve the fundamental understanding of the geotechnical aspects of foundation design under high horizontal loads and associated eccentricities. 3. The calibration of the numerical model developed to design the sloping sea wall of the pier foundations of the Gwang Yang bridge in Korea against the ship impacts (Gaudin & Colwill, 2007). The Gwang Yang bridge is a 1.28 km long suspended bridge in South Korea. Its piers are funded on a 60.5 m long by 36 m wide raft resting on rubble, soft marine clay and sand compaction piles. In turn, the raft is supported by a network of 38 piles, 32.6 m long, embedded in the rock bed. The piers are protected against potential ship impact by a sloping sea wall, whose geometry was initially designed from empirical analytical models. In order to calibrate the numerical model used to finalise and validate the design, centrifuge tests were performed on an idealised model, where the rubble, soft marine clay and sand compaction piles were replaced by loose sand. Other features of the prototype such as the embedment of the piles into the rock bed through a layer of soft marine clay and the geometry and velocity of the bow and energy realised by the ship during impact were accurately replicated. The model was heavily instrumented and provided performance data such as the total pressure and the pore pressure in the front and at the invert of the raft, displacements and acceleration of the raft, and 3D profiling of the geometry of the sloping sea wall after impact. These data were used to calibrate a finite element model, which was subsequently improved to account for the properties of the rubble and sand compaction piles. The FE model was then used for design (Figure 11). 5.2 Providing qualitative insights into soil-structure interaction and mechanisms This aspect is particularly important when novel concepts or unusual conditions are encountered. Understanding the structure behaviour, observing the failure mechanism taking place, is a pivotal step into the
Figure 11. Model of the GwangYang bridge pier foundation and protecting seawall. (a) Footprint left by the impact of a ship at 30 degrees, and (b) comparison with the numerical model (b)(after Gaudin & Colwill, 2007).
development of sound design methodology. Some of the most noticeable breakthrough includes: 1. The identification of the shallow failure mechanism for gravity platform under horizontal and vertical cycling loading in comparison to the much deeper mechanism resulting from monotonic loading (Craig & Al-Sauodi, 1981). 2. The identification of the flow failure (Figure 12) taking place during spudcan penetration, as opposed to a cavity wall failure (Hossain et al., 2005). This has resulted in the amendment of the SNAME (2008) guidelines. 3. The observation during pipeline lateral breakout of the growth of a soil berm in front of the pipe and of the suction developed at the rear of the pipe, governing both the trajectory of the pipe and the peak breakout resistance (Dingle et al., 2008). This aspect is developed in more details in section 6.5. 4. The observation of the behaviour of the keying flap equipping suction embedded plate anchors. The flap was designed to rotate during keying and hence limit the loss of embedment. Centrifuge results demonstrated that the flap does not rotate (Figure 13) due to the rotational mechanism of the soil, resulting in the soil applying a bearing pressure at the back of the flap (Gaudin et al., 2010b).
167 © 2011 by Taylor & Francis Group, LLC
Figure 13. Plate anchor pulled out vertically. Demonstration of the non-activation of the keying flap during anchor keying (after Gaudin et al., 2010b).
with full-scale measurements to provide confidence to the end-users that this is a viable approach.
5.3 Validate the structure design for a particular site or a specific design approach By allowing the use of natural soils, sampled in-situ, by applying complex loading sequences directly relevant to design and by using a models replicating precisely prototypes, centrifuge modelling offers a valuable tool to validate or justify specific offshore structures. Among the relevant examples are:
Figure 12. Observation of the back flow mechanism during spudcan penetration (after Hossain et al., 2005).
Centrifuge modelling has also provided an efficient and cost-effective complement to conventional ice tank modelling to simulate the deformation of level ice and rubble. Lau et al. (2002) presented an overview of the ice-structure interaction research conducted at C-CORE. They discuss the advantages of centrifuge modelling of ice problems in terms of the scalability of ice properties, the controllability of test environments, and the quality of test data. There was an excellent agreement between data obtained from the centrifuge, an analytical algorithm and other conventional refrigerated model basins for a range of ice conditions. These included level ice, ridges and rubbles interacting with structures including conical piers with level ice sheets; unconsolidated ice rubble and ridges with cylindrical structures; and consolidated rubble ice with conical structures. Due to the small size of the model required, centrifuge modelling also has the advantage of simulating and maintaining a controlled test environment at a fraction of the cost of conventional tanks. Centrifuge modelling is relatively new to the ice engineering community and additional comparisons are still needed
1. The investigation of the response under cyclic horizontal loading of the Maari platform in New Zealand. Operated by OMV, the Maari platform is founded on a rectangular skirted mat (Figure 14a), installed in sand overlaid by impermeable silt (Gaudin et al., 2006b). Tests were used to investigate both the feasibility of suction installation through layered soil, and the performance of the platform under cyclic horizontal loading. Results demonstrated that suction installation was achievable, with limited silt plug uplift, provided that sufficient direct load is applied and the pumping rate is high enough to result in a fast installation, limiting the amount of plug uplift. Similar observations were made on another offshore project in Gulf of Mexico where suction installation on sand overlaid by clay was required (Watson et al., 2006). Results for the Maari platform, under horizontal cyclic loading highlighted the very stiff response of the top silt layer and the very limited vertical displacements experienced by the platform, even after several packages of cyclic loading (Figure 14b). 2. The investigation of the performance of a dolphin made of sheet pile walls for the protection of the Korena Incheon bridge foundation against ships
168 © 2011 by Taylor & Francis Group, LLC
Figure 14. (a) The model Maari platform and (b) Accumulation of vertical displacements under cyclic horizontal loading (after Gaudin et al., 2006b).
impact (Bezuijen et al., 2007). A total of 31 tests were performed at 200g investigating the response of the dolphin under static and dynamic tests loadings. Other parameters investigated included the density of the soil, the diameter of the dolphin and the location of impact. The results show that for dynamic tests (as opposed to static tests) suction was generated underneath the dolphin increasing its stability. 3. The validation of design of the Genesis Spar platform installed in the Gulf of Mexico, and operated by Chevron. The 70 m long piles anchoring the platform were modelled in a series of centrifuge tests performed at LCPC to investigate notably the performance of the piles under combined vertical and horizontal loading, and cyclic storm loading. The bending moment profiles (see Figure 15a) along the piles were monitored to optimise the pile thickness and to develop p-y reaction curves (see Figure 15b) these were subsequently used to validate the pile design.
Figure 15. (a) Bending moment (MN.m) vs. depth (m) recorded during the Genesis pile loading tests ranging from 10 to 23 MN, and (b) experimental P-y reaction curves derived from the bending moment profiles (depth ranging from 21 m to 58.8 m). (From internal LCPC report).
5.4
The very good control of the testing conditions, the possibility of measuring a significant number of parameters, such as displacement, loads, pressures and even strains from digital imagery and the relatively low cost of centrifuge testing compared to field testing (even in reduced scale) make the centrifuge an ideal tool to develop new concepts and investigate the feasibility of particular foundations, which have not been tested in-situ yet. The modelling does not intend to replicate precisely a prototype structure, but rather puts emphasis on a particular aspect which requires validation or deeper understanding. Clukey & Morrison (1993) investigated the efficiency of multi-cell suction caissons as anchoring system for tension leg platforms (Figure
169 © 2011 by Taylor & Francis Group, LLC
Investigating the feasibility or developing new foundation concept
Figure 16. Model multi-cell suction caisson (source LCPC).
16). Tests were performed on kaolin clay and particular attention was paid on the potential for suction to be developed at the caisson lid invert to resist monotonic and cyclic tensile loads. Gaudin & Cassidy (2007) investigated the feasibility of suction-induced preloading of a skirted mat as an alternative to spudcans as a foundation for jack-up platform in intermediate water depth, where ballast tank may not provide enough preloading. Tests did not model any particular type of foundations, but focused on the preload level which could be achieved by suction and the drainage conditions within the foundation. Results demonstrated the feasibility of the concept but also the mechanism governing suction-induced preloading. It was notably observed, the absence of drained plug failure during the application of suction-induced preload, and the dominant effect of the consolidation time versus the preload level on the post preloading bearing capacity of the foundation (Figure 17). Another typical example of a novel design that has greatly benefited from centrifuge methods is the development of dynamic anchors, commonly called torpedo anchors. Torpedo anchors are usually 1 to 1.2 m in diameter and 10 to 15 m high. They are released from a height of 50 to 100 m above the seabed, achieving a free fall a velocity of 10 to 30 m/s, before impacting soft seabed sediments and embedding by 2 to 3 times their length. Key uncertainties relate to (i) the embedment depth resulting from the dynamic penetration and (ii) the soil setup after installation and the resulting holding capacity. Centrifuge techniques were particularly appropriate to conduct model tests, using the high g environment to achieve a fast free fall velocity from limited drop height. They played a significant role in both providing key insight about the soil-anchor interaction and establishing techniques to predict anchor embedment and anchor capacity. Early centrifuge work, reported notably by O’Loughlin et al. (2004) and Richardson et al. (2006) focused on anchor installation and highlighted the effect of the very high strain rate generated by the
Figure 17. (a) Hybrid foundation model with the suction caisson in the centre and the skirted mat around the caisson (b) Increase of bearing capacity due to suction induced preloading (after Gaudin & Cassidy, 2007).
penetration high velocity. This results in an increase in shear strength due to viscous effect, potential entrainment of water at the soil anchor interface and a decrease of skin friction. The necessity to account for a drag factor, in addition to the soil bearing factor, in calculating the anchor embedment depth was also demonstrated. More recent work, reported by Richardson et al. (2009) focused on the soil strength recovery after installation and the subsequent holding capacity (Figure 18). Results demonstrated the significant increase of capacity with time (by a factor of 5 from immediate extraction to full consolidation) and the significant time required to achieve full consolidation, compare to piles or suction caisson (up to 7 years in the case presented here). 5.5
As highlighted by previous examples, the range of benefits from centrifuge application to offshore
170 © 2011 by Taylor & Francis Group, LLC
Developing new design methods
the pile head lateral displacement may be estimated by:
Figure 18. (a) Typical model torpedo anchors, without fluke and with four flukes (b) Illustration of the gain in capacity with consolidation time after penetration. QS refers to static installation, DY to dynamic installation (after Richardson et al., 2009).
geotechnics offers the opportunity of providing valuable information at all stages of foundation design. Centrifuge modelling can also be an appropriate means to develop empirical design methods for some particular cases, where the phenomenon taking place are too complex to be captured by numerical or analytical models. One example is the development of a new design method for piles under lateral cyclic loading, which was poorly accounted for in common design guidelines. Centrifuge tests have been carried out by Rosquöet et al. (2007) and Rakotonindriana (2009), where up to 75000 lateral cycles were applied to model piles installed in dry sand. By normalising these cyclic loads by the maximum applied load Fmax and by introducing the amplitude of the load variation DF, Rosquoet et al., 2007 demonstrated that, for service conditions, the effect of the number of cycles n on
where y1 and yn are the pile head displacement under the maximum load Fmax , at respectively the first loading and at the nth cycle. Cyclic p-y curves were determined from bending moment measurements during loading (Figure 19). The degradation of the soil lateral resistance at different depths was then evaluated and P-multiplier coefficients r were established. These permits the determination of the cyclic p-y curves from the static ones (Table 1). Another typical example is the empirical method developed by Bienen et al. (2009) to determine the flow rate required to successfully extract spudcans in soft clay using on-bottom jetting (Figure 20a). The method was based on insights provided by a series of centrifuge tests, investigating various parameters such as the extraction rate, the jetting flow rate and the jetting pressure (Bienen et al., 2009; Gaudin et al., 2010c) and the demonstration that the efficiency of the jetting related to its capability to fill the gap at the spudcan invert created by its uplift. Hence, jetting flow was shown to be a dominant factor over jetting pressure, in contrast to generally accepted belief in the jack-up industry. A jetting methodology was then established based on (i) the pullout force resulting from the buoyancy of the jack-up hull, and a filling ratio f, defined as the ratio of the jetting flow rate to the product of the extraction rate to the spudcan area (i.e., the volume at the spudcan invert created by the extraction) (Figure 20b). The Pressure Ridge Ice Scour Experiment (PRISE) developed the capability to design pipelines and other seabed installations in regions gouged by ice, taking into account the soil deformations and stress changes, which may be caused during a gouge event (Phillips et al., 2005). This capability has been used to design offshore pipelines in regions such as the US Beaufort Sea and off Sakhalin Island, Russia. One of the main activities for the PRISE program was to assess the significance of subgouge deformations imposed on buried pipelines during ice gouge events. The PRISE joint industry research program conducted analytical, numerical, experimental and field studies. Comparison of field investigations on relict gouge events with centrifuge experiments demonstrated that subgouge deformations should be considered. Analytical and numerical models were developed to assess the gouge forces in sands and clays, based on the experimental failure mechanism observations. The implications of the models on ice keel ablation, spoil heap development and the development of subgouge deformations are discussed by Phillips et al. (2005). Eulerian based finite element models for clay have also been validated against the centrifuge model test results (Phillips et al., 2010).
171 © 2011 by Taylor & Francis Group, LLC
Figure 19. Cyclic p-y curves at different depths obtained from one-way lateral loading tests (after Rosquoet et al., 2007).
Table 1. P-multipliers r at different depths for determining the cyclic p-y reaction curves (after Rosquoet et al., 2007). Depth z
P-multiplier r
0 < z < 1.5B 1.5B < z < 3B 3B < z < 5B
r = 0.7 − 0.12 DF/Fmax r = 0.94 − 0.058 DF/Fmax r = 0.97 − 0.029 DF/Fmax
5.6
5.7
Characterising in-situ soils
The improvement of soils reconstitution methods allows the replication of complex soil stratigraphy using natural soils exhibiting the same properties (such as permeability, compressibility, void ratio, water content, unit weight and shear strength) as found in the field (see development in section 1.3). The centrifuge offers the possibility of performing a large number of tests at a reduced cost, using characterisation tools identical to the ones used in-situ. Also, in addition, the techniques provides extra information by using newly developed characterisation tools not yet commonly used in-situ. One such example is the recently developed piezo-ball penetrometer presented in section 4.4, for characterisation of soft soils. The test, in addition to the measurement of the penetration resistance (and hence undrained shear strength), provides information on the drainage characteristics of the soil and potentially the consolidation behaviour (Boylan et al., 2010b). With the acquisition of large soil samples taken with box core becoming more standard practice, it
172 © 2011 by Taylor & Francis Group, LLC
is also possible to perform typical soil characterisation tests in a relative undisturbed state (White & Gaudin, 2007). Centrifuge testing helps augment these tests. Providing data for assessment of fatigue for conductors and SCRs
Designing for fatigue of conductors (uppermost outer pipe in a well) and Steel Catenary Risers (SCRs) is an important design challenge in deepwater. The interaction of the conductor and riser with the soil is an important part of both these fatigue assessments. Jeanjean (2009) describes a case where the API p-y curves were modified based on finite element analyses and centrifuge testing to satisfy fatigue requirements for a number of offshore conductors. For this particular case the small displacement lateral soil response was critical for predicting the expected fatigue life. Figure 21 shows a comparison between theAPI recommended lateral soil response and the response derived from FE analyses and centrifuge testing. In this problem a stiffer soil response reduced fatigue life. Cyclic centrifuge tests did show some reduction in the soil stiffness. However, based on the design methods used to assess fatigue, during the unloadreload soil response appropriate to cyclic loading, the soil stiffness was never less than the tangent stiffness from the monotonic test results used in the analyses (Jeanjean, 2009). The fatigue of an SCR in the soil touchdown point region (Figure 22) depends on both small and large amplitude displacements. Hodder et al. (2008) describes a series of centrifuge tests to investigating
Figure 20. (a) Model spudcan with jetting in action and (b) conceptual chart of jetting extraction efficiency. The chart indicates, for a given extraction load, the required flow rate for a successful jetted extraction (after Gaudin et al., 2010c).
the response of a small portion of pipe in the soil touchdown point region. The tests show that the soil stiffness can be dramatically reduced by the large displacements incurred from the riser motions, especially when
173 © 2011 by Taylor & Francis Group, LLC
the riser pipe separates from the seafloor. However, a significant portion of the stiffness loss is regained when the motions stop and the soil has an opportunity to recover from thixotropic and consolidation effects
Figure 21. Comparison of API p-y curves versus p-y curves obtained with centrifuge testing and FE analyses (after Jeanjean, 2009).
Figure 22. Schematic of offshore facility and SCR interacting with soil in the touch down area.
(Hodder et al., 2009). The large motions of the SCR will also enhance trench formation which will also affect fatigue life. Centrifuge tests to investigate the response of a complete riser through the touchdown point region are now being performed at C-CORE. A special actuator has been developed for these tests to simultaneously provide both heave and surge motions to the riser. The motions are applied to both these motions at two frequencies to capture the overall riser motions expected in the field. Comparisons of trenches observed in the field and in the centrifuge are shown on Figure 23 and Figure 24. 5.8
Figure 23. SCR trench developed in centrifuge test.
distribution of loads. One such example is the work performed by Gaudin & Landon (2008) and Gaudin et al., 2009b who investigated the efficiency of a rock cover to protect a buried offshore pipeline against the dragging of anchors. The assessment of the clearance between the anchor and the pipeline over a series of centrifuge tests resulted in the determination of frequency of failure of the pipeline which was subsequently introduced into a probabilistic analysis on the frequency of hazard affecting the pipeline. 6
Providing data for probabilistic analyses
One recognised advantage of centrifuge modelling is the ability to conduct parametric studies in a well controlled environment. It is therefore possible to focus the study on a particular parameter, such as the natural variation of soil characteristics or the statistical
6.1
Suction caisson technology development
The history of the development of suction caisson as anchoring system for deep water solutions illustrates particularly well the contribution of centrifuge
174 © 2011 by Taylor & Francis Group, LLC
EXAMPLE OF CONTRIBUTIONS
Figure 26. Evolution of penetration resistance and external radial total stress of a suction caisson in centrifuge during jacking and suction installation (After Chen & Randolph, 2007a).
Figure 24. SCR trench in field.
Figure 25. Results of static pullout tests. Comparisons between in-situ 1g tests and centrifuge tests performed on natural (Lysaker) and artificial (kaolin) clays. (after Morrison et al., 1994).
testing and modelling in investigating behaviour and validating design procedures throughout their development. Centrifuge results have impacted installation procedure and assessment of the capacity, first under monotonic vertical loading and then under combined cyclic loading. In contrast to the development of driven piles technology for offshore structures, which was primarily based on a series of 1-g model tests, suction caisson technology has achieved significant advancements through centrifuge testing. The initial tests, however, for suction caissons used in deepwater were in-situ 1-g tests for the Snorre TLP in the North Sea. (Dyvik et al., 1993). These tests were subsequently replicated with centrifuge tests at LCPC (Morrison et al., 1994). When the undrained shear strength profile is well simulated, the load-displacement behaviour between the in-situ and centrifuge tests are very close, both in peak resistance and stiffness (Figure 25), and provided added confidence in the use of centrifuge testing for suction caisson applications.
Additional testing for TLP applications (Clukey & Morrison, 1993, Clukey et al., 1995) provided additional information on suction caisson behaviour under both monotonic and cyclic loads for TLPs, notably about the cyclic ratios resulting in failure and about the increase of post-cyclic static capacity. With ever increasing water depths, the design trends moved from multi-cell configurations systems, to single cell arrangements with the mooring line attached about 2/3 down the side of the caisson to enhance the holding capacity. The utilization of these single cell suction caissons expanded considerably as the industry moved into water depths too deep for conventional offshore pile driving. Centrifuge testing at a number of facilities then began to provide important information on the behaviour of these foundations both during installation and under a variety of loading conditions. Andersen et al. (2003) showed that suction caissons could potentially be installed to over ten times their diameter without failing the internal soil plug. Clukey & Phillips (2002) described monotonic tests which showed a potential for the reduction in external skin friction for suction installed versus jacked in caissons. Subsequent tests by Raines et al. (2005), however, showed no difference in external skin friction for suction versus jacked installed suction caissons. Similar results were obtained by Chen & Randolph (2007a) (see Figure 26) and Jeanjean et al. (2006). The latter used a double walled suction caisson to separate internal and external skin friction. Their results did show, however, a reduction in external skin friction versus API recommended values for driven piles. The reverse end bearing observed by Jeanjean et al. (2006) had a peak bearing capacity factor (Nc ) of 12 at large displacements and 9 when the external skin friction and total holding capacity reached peak values. Similar magnitudes of design parameters were obtained from research on suction caisson response carried out by Chen & Randolph (2007a). More recently, the effect of stiffeners in both the installation resistance and the pullout capacity has been investigated by Westgate et al. (2009) and Colliat
175 © 2011 by Taylor & Francis Group, LLC
verified the effectiveness of design tools in predicting suction caisson holding capacity in stiff overconsolidated soils in the Gulf of Mexico. It is most noteworthy that the results were obtained in a time frame, at a cost and with a level of details that could not have been matched by field testing alone. 6.2
Figure 27. Effects of stiffeners and consolidation on normalized pull-out capacity (after Colliat et al., 2010).
et al. (2010) at LCPC. The former investigated various stiffener configurations in kaolin clay, while the latter, using in-situ deepwater Nigeria clay, focused on the effect of consolidation time following installation, for caissons with and without stiffeners. Results, from both studies, demonstrated the significant contribution of the stiffeners to the pullout capacity. For undrained pullout, the increase ranges from 20% shortly after installation, up to 33% when full consolidation is permitted (Figure 27). Combined loading was investigated by Randolph et al. (1998), who described tests that demonstrated the potential for gapping behind the caisson, in soft carbonate sediments, due to interactions between horizontal and vertical loadings. Clukey et al. (2003), further investigated combined loading and demonstrated that limit analysis techniques could be used to predict these interactions. Clukey et al. (2004) also described a combination of independent centrifuge test programs at UWA and C-CORE that showed that reverse end bearing could be maintained for several months without reduction in capacity from pore water pressure dissipation. These results were also further corroborated with finite element analyses (Clukey et al., 2004). These results were important in assessing the performance of suction caisson for long term loop current load. As demonstrated above, the use of centrifuge techniques from various centres in the world, in collaboration with offshore operators have resulted in significant breakthrough in understanding the behaviour of suction caissons, both during installation and under various conditions of loadings. These various research programmes had significant impact in operators’ practice, both validating and optimising innovative designs. Hence, centrifuge model tests formed part of the design validation process for applications in Australia, off the west coast of Africa (Randolph et al., 1998), as well as in the Gulf of Mexico. Jeanjean et al. (2006) also described results obtained at the University of Colorado, Boulder, which
1. Accelerated time frame. Centrifuge modelling and testing require a limited volume of soil, accelerating sample preparation. For soft soil, the process may be further accelerated by in-flight self-weight consolidation. Similarly, testing sequences are considerably shortened while ensuring the correct drainage conditions. This allows collection of a significant number of data in a short time frame and at a reasonable cost. 2. Accurate loading sequences. New motion control techniques permit accurate replication of complex loading sequences, including cyclic vertical and/or horizontal motion under either load or displacement control. This is particularly relevant when mimicking the complex motion of the pipe at the touchdown zone during laying, or the large deformations resulting from lateral buckling.
176 © 2011 by Taylor & Francis Group, LLC
Pipeline-soil interaction
Offshore pipelines are laid on the seabed and used as flowlines or trunklines tied back to shore. They are becoming more and more common as oil and gas extraction is taking place in deeper and more remote areas. Geotechnical design procedures for offshore pipeline and risers have not yet reached the maturity exhibited by design procedures for piles, shallow foundations and anchoring systems. Reasons are multiple and relate mainly to the unknown geometry of the problem (the final embedment of the pipe is uncertain), the uncertainties of the soil conditions (both difficult to assess at very shallow depth and significantly affected by the installation process) and the very large deformations (and associated post-failure behaviour) the pipeline may experience (notably during controlled or unexpected lateral buckling). Whilst the on-bottom stability of pipelines (i.e. its stability against hydrodynamic forces) has been extensively investigated since the early 60s, notably in the USA, lateral buckling and axial walking (due to changes in internal pressure and temperature in the pipeline) have only emerged recently as a major research topic. Significant advances have been made over the last 5 years, using data from 1g model tests (performed notably in Cambridge, Cheuk et al., 2007, and NGI, Dendani & Jaeck, 2007), in-situ tests (notably using the SMARTPIPE device developed by Fugro, Jacob & Looijen, 2008) and also centrifuge tests. While each method has its own advantages and disadvantages (see Hill & Jacob, 2008 for a comparison of each method as applied to pipelines), centrifuge tests have certainly boosted knowledge of pipeline-soil interaction and generated significant breakthroughs. The main reasons are:
3. Use of natural soils. By requiring a limited volume of soil, it is possible to use in-situ soil with reasonable supply costs. The use of natural soils for model tests is maybe more important for pipelines than for other geotechnical structures, because of the heavy remoulding and the large deformations experienced by the soil. This may trigger specific behaviour that would not be captured by common artificial laboratory soils. 4. Accurate seabed characterisation. The good control of the soil reconstitution process results in a homogeneous sample. By using standard soil characterisation tools (such as the T-bar) or dedicated ones (such as the O-bar) in controlled conditions, it is possible to determine the soil characteristics accurately. 5. Enhanced instrumentation. By using image acquisition systems and pore pressure measurements, in addition to the measurements of load and displacements, one gains access to particular features of the pipe-soil interaction, such as the development of lateral berms during buckling, the creation of a trench during dynamic laying and the drainage conditions around the pipe during the pipe motion. All this information is pivotal in understanding and describing pipe-soil interaction. 6. Case-specific study. Centrifuge testing can be used to provide insight in particular issues related to pipeline design. This includes site-specific storm loading conditions or geohazard conditions, such as scarp crossings. The rest of this section aims to illustrate by a few relevant examples, some of the insights and breakthroughs, obtained from both centrifuge testing performed to gain insight into specific aspects of pipeline interaction, and from centrifuge modelling performed to assist in the design of specific pipeline projects. It is not a state-of-the-art review of pipeline behaviour and design. Such reviews have been presented by Cathie et al. (2005) and White & Cathie (2010).
6.3 Seabed characterisation Improving the reliability of design predictions of pipeline embedment during laying and lateral buckling relies on accurate measurements of the seabed characteristics in the upper 0.5 m of the seabed, which is the zone relevant to pipeline-soil interaction. The T-bar penetrometer, originally developed for the geotechnical centrifuge (Stewart & Randolph, 1994) is increasingly favoured in the field for characterising soft soils. The interpretation of T-bar penetration tests to assess the strength of soft seabed soils in the upper 0.5 m can be refined beyond the use of a constant NT-bar = 10.5 factor (Stewart & Randolph, 1994). These refinements have been described by White et al. (2010) and account for the soil buoyancy and the reduced bearing factor arising from the shallow failure mechanism mobilised prior to full flow of soil
© 2011 by Taylor & Francis Group, LLC
Figure 28. Example the influence of near surface and buoyancy correction on the determination of the strength at shallow embedment of lightly overconsolidated clay. (after White et al., 2010).
around the bar. These corrections, which are marginal for assessing the strength of clay at moderate depths (>1 m) can be significant at shallow depth in soft materials. White et al. (2010) demonstrated that the omission of these effects may result in an underestimation of the shear strength in the depth range relevant to pipeline analysis (Figure 28). More recently, the use of T-bar penetrometers has been extended to investigate in the centrifuge phenomena relevant to pipeline behaviour, including the successive cycles of consolidation and remoulding experienced by the soil (Hodder et al., 2010) and the potential entrainment of water (and consequent swelling) during cyclic events. Figure 29 illustrates the loss of strength (expressed as the ratio of the remoulded shear strength su,r to the intact shear strength su,int ) experienced by different soils during cycles of remoulding for artificial clay and naturals clayey silts typical of the North West Shelf region, off the coast of Australia. The most striking feature of this cyclic test is the ten-fold reduction in strength during cycling for the Western Australian carbonate silt A in contrast to the two-fold reduction in strength for normally consolidated Kaolin. Further strength loss may be experienced by the soil as the pipeline remoulds and softens the soil, and water is entrained, increasing the moisture content of the soil. This is illustrated in Figure 30, which compares the strength loss for normally consolidated Kaolin clay resulting from deep cycles, from cycles 22 mm deep, breaking at the surface and from cycles 50 mm deep, breaking at the surface (using model scale units). The deep cycles do not break the surface, so water entrainment does not occur. For the kaolin test to 50 mm depth, the strength loss has been quantified at four different depths, from 10 to 40 mm, while for the kaolin
177
test to 22 mm, the loss of strength has been quantified at a depth of 20 mm. This test clearly indicates that the entrainment of water contributes to a continuous strength loss down to a value 5 times lower than the fully remoulded shear strength determined from deep cycles (without water entrainment). For typical normally consolidated kaolin, this strength reduction corresponds to an increase of water content by about 30%. The cycles to 50 mm depth indicate a similar trend with the magnitude of strength loss decreasing with depth. At 10 mm depth, the strength loss reaches a value similar to the one exhibited by the 22 mm cycles, but at a much faster rate. At deeper depths the degree of strength degradation reduces to eventually reach a value similar to the one obtained from deep cycles. As the T-bar penetrates deeper, the backflow of soil gradually limits the water entrainment. Even at very shallow depth, the effect of the water entrainment seems to reach a maximum, resulting in the soil still exhibiting some shear strength, even after a significant number of cycles. The simple examples presented above illustrate (i) the wide range of sensitivity exhibited by different types of offshore soils and (ii) the significant loss of strength resulting from water entrainment – a phenomenon likely to occur during pipe-soil interaction but which is not accounted in current design
approaches. In both cases, simple characterisation tests performed on in-situ reconstituted samples in the centrifuge may provide valuable and reliable information for design. 6.4
The knowledge of the pipe embedment is pivotal for subsequent on-bottom stability and lateral buckling design. The effect of dynamic laying on pipe embedment has been first observed by Lund (2000) who concluded on the necessity to account for pipeline laying history in subsequent on-bottom stability design, as it results in excessive embedment compared to a static laying process. During the lay process, an element of pipe moves through the touchdown zone, from an initial contact with the seabed to a stationary position, where the pipe weight is supported by an equal upwards seabed reaction force. The dynamic behaviour of the pipe through this process is complex and difficult to replicate. It is a function of the following parameters:
Figure 29. Degradation of strength during cyclic T-bar penetrometer tests (after Gaudin & White, 2009).
Figure 30. Strength loss at different depths in normally consolidated clay during deep and surface-breaking cyclic T-bar penetrometer tests (after Gaudin & White, 2009).
178 © 2011 by Taylor & Francis Group, LLC
Dynamic lay embedment
– The profile of stress concentration in the touchdown zone due to the catenary shape, resulting in a vertical load on the soil varying along the pipe and higher, at some locations, than the as-laid pipeline weight. These pipe stresses can be estimated, based on a structural analysis of the hanging pipeline using the planned lay tension and hang-off angle. – The horizontal and vertical oscillation due to the motion of the laying vessel, resulting in a sweeping and damping of the soil at the touchdown zone. It can be estimated from the analysis of the motion of the vessel under for the wave motion considered. – The number of oscillations during the entire lay process. It can be assessed, based on the estimated pipeline laying rate, the touchdown zone length and the wave-induced oscillation frequency. Figure 31 presents a typical pipeline setup used to model dynamic laying and lateral buckling. The pipe is modelled as a short section (with an aspect ratio of at least 6 to prevent any end effect) and may be sandblasted to achieve a particular roughness. It is connected to a VHM loading arm, used to measure the horizontal loads, via a shear load cell, insensitive to bending moment. The purpose of this load cell is to provide a very accurate reading of the vertical load applied (permitting a resolution of 1 N at model scale) that is essential for pipelines, which impose stresses on the seabed that are orders of magnitude lower than typical geotechnical structures. Additional instrumentation may include pore pressure transducers at the pipe invert, as shown in Figure 31. A purposely developed motion control system (De Catania et al., 2010) is used at COFS to (i) apply a targeted vertical load depending on the stress concentration profile in the touchdown zone, via a feedback loop on the axial load cell and (ii) apply lateral oscillation of varying amplitude, linked via a second feedback loop to the pipe embedment.
The following example presents centrifuge modelling performed recently to determine the final embedment of a flowline to be installed offshore Australia (Gaudin & White, 2009). In-situ soil was used and the dynamic motion applied accurately replicated the motions determined from a numerical analysis of the wave, vessel and pipeline motion. The pipeline was divided into six sections, from the initial contact point to a far away position when the pipe was considered to be unaffected by the laying motion. The six sections featured six different loading sequences. Each sequence included a cyclic vertical loading of specified amplitude (from 0 to 1.55 times
Figure 31. Typical pipeline model setup.
the as-laid pipeline weight Vlay ) concurrent with horizontal cyclic motion, also of specified amplitude, as presented in Figure 32. The first sequence, modelling the first segment of the pipe at the touchdown zone, features a specific cyclic motion in which the pipe was pushed into the soil to a specified load of 0.7Vlay before being lifted up until separation between the soil and the pipe occurred. The resulting accumulation of pipe embedment is presented in Figure 33 for each sequence as a function of the cumulative number of imposed cycles. The key observation from Figure 32 and Figure 33 is the much larger pipe embedment (about 0.24 m) compared to a static embedment (about 44 mm using the formulation presented by Randolph & White, 2008a), even accounting for the stress concentration. Other important observations are the suction developed at the pipe invert during the first sequence of loading, where the pipe was lifted up from the soil (which potentially affects the stresses in the pipeline), and the dominant effect of the lateral motion amplitude compared to the cyclic vertical load. The first sequence, which featured the largest amplitude of lateral motion, resulted in a deeper embedment that the subsequent sequences, which featured a higher maximum vertical load but smaller lateral motion amplitudes. In other words, for these lay conditions and this soil type, the action of pushing the soil to either side of the pipeline during lateral motion (and concurrently remoulding it) has a dominant effect on
Figure 32. Dynamic lay simulation. Cyclic vertical loading and the associated pipe invert trajectory (after Gaudin & White, 2009).
179 © 2011 by Taylor & Francis Group, LLC
Figure 33. Accumulation of pipe embedment resulting from the dynamic lay simulation of Figure 32 (after Gaudin & White, 2009).
the pipe embedment in comparison to any remoulding or penetration of the soil resulting from the vertical cyclic loading. These results stress the necessity to model accurately dynamic laying motion, in order to determine the pipe embedment, but also the remoulding of the soil resulting from the laying process, which are going to govern both on-bottom stability and lateral buckling. Further, the results demonstrate the capability of centrifuge modelling for simulating accurate laying motions, capturing particular soil behaviour features and delivering useful performance results. 6.5
Lateral buckling
To reduce the structural load resulting from thermal expansion, pipelines are permitted to buckle at targeted location, resulting in lateral displacements of the order of 10–20 times the pipeline diameter (Bruton et al., 2006). Whilst lateral bucking reduces the axial load on the pipeline, it also generates bending moment in the buckling zone, which can ultimately lead to local bending failure. The structural analyses used to assess lateral buckling in design require as input the soil resistance when the pipeline moves laterally. Unlike on-bottom stability design, overestimating the lateral resistance is not necessarily conservative as a high ‘friction’ factor can result in a more onerous structural load. The determination of an accurate friction factor is therefore pivotal for a safe and sound design. This requires the understanding of the soil behaviour at large displacements and through many cycles of loading, well beyond the point of failure, all features that can be well investigated by centrifuge methods. Two different examples are presented of centrifuge studies of lateral buckling. One of centrifuge
testing performed to understand mechanisms taking place during lateral breakout, and one of centrifuge modelling to provide friction factors for a particular soil and loading conditions, to assist in the design of a pipeline offshore Australia. In the first example a specific device was developed to observe and analyse, using PIV techniques, the deformation of the soil around the pipe by placing it behind a Perspex window (Dingle et al., 2008). In order to accommodate the large displacements occurring during lateral breakout, the camera was placed on a second actuator, which followed the displacement of the pipe, so the pipe remained constantly at the centre of the image. The corresponding setup is presented in Figure 34. Lightly over consolidated kaolin was used, and the model pipe was lowered onto the seabed to a targeted depth, hence ignoring any dynamic process and consequent soil remoulding. The test setup aimed at replicating the lateral breakout from known initial conditions. It deliberately ignored side aspects, which although important, are not believed to affect the qualitative response of the soil. Hence it qualified as testing rather than modelling. Dingle et al. (2008) presented a thorough analysis of the testing results, covering notably, the soil heave resulting from penetration and the assessment of the mobilised soil shear strength during lateral breakout. Two pivotal observations for heavy pipes (i.e. penetrating significantly into the seabed) were made, unknown at the time of project, regarding lateral breakout. First, the soil behaviour exhibited a brittle response, characterised by a sudden drop in resistance once suction at the rear of the pipe was lost. (Figure 35a) and resulting in the pipe rising up to shallower embedment. Hence, the peak lateral resistance is governed by the available tensile resistance at the rear of the pipe. Note that theoretical solutions for the two-sided mechanism
180 © 2011 by Taylor & Francis Group, LLC
Figure 34. Centrifuge testing of pipe lateral breakout (after Dingle et al., 2008).
and for the one-sided mechanism were derived from upper bound plasticity limit analysis by Cheuk et al. (2008) and Randolph & White (2008b), respectively, following the observations from the centrifuge testing. Second, the post peak lateral response is governed by the growth of a soil berm at the front of the pipe (see Figure 35b; this can also be seen in Figure 31 and Figure 34). Further analysis presented by White & Cheuk (2008) indicated that the lateral resistance could be expressed as a conventional frictional term plus a berm resistance term proportional to the volume of the berm. This finding has critical consequences for design as it can lead to an overall friction factor, combining both phenomena, higher than unity while conventional friction factors used for design lied in the range 0.2–0.8. The second example presents centrifuge modelling performed to provide design data (dynamic embedment and friction factors) to assist in the design of a pipeline offshore Australia (White & Gaudin, 2008). Natural soils sampled from the site were used and reconstituted so they exhibited the same strength characteristics and index parameters as the in-situ soil. The dynamic laying process was accounted for, replicating the change of vertical contact force along the touchdown zone during laying. It was followed by lateral breakout and a series of lateral sweeps. By replicating soil conditions and key phenomena of the soil-pipe interaction, and by aiming to provide data directly applicable for design, the project qualifies as modelling rather than testing. Detailed results are presented by White & Gaudin (2008). Some key observations, related to initial breakout in fine grained soils are presented in Figure 36. Load developments, pore pressure measurements at the pipe invert and vertical displacement with pipe lateral displacements from Figure 36, and the knowledge obtained from centrifuge testing from Dingle et al.
181 © 2011 by Taylor & Francis Group, LLC
Figure 35. Instantaneous velocity fields during pipe lateral breakout. Illustration of the tensile crack at the rear of the pipe (left) and formation of the soil berm at the front of the pipe (right) (after Dingle et al., 2008).
(2008), indicate the mechanism operating during initial breakout. Prior to separation the failure mechanism is two-sided, with soil being pulled behind the pipe in addition to being pushed ahead. The opening of a crack
Figure 36. Centrifuge modelling of lateral breakout (replotted from White & Gaudin, 2008).
at the rear of the pipe was demonstrated by the pore pressure measurements. After separation, a one-sided mechanism (e.g. with a berm forming at the front of the pipe) is operative. After breakout, the pipe rises to a shallow embedment. During this phase the lateral resistance also drops, reflecting the shallower failure mechanism that is mobilised. Once the pipe reaches a steady elevation the lateral resistance remains approximately constant and is created by shear at the base of the pipe and passive resistance against the berm of soil being pushed ahead. It is noteworthy that the pipe-soil interaction features observed during centrifuge testing, from which a design method has been refined (Bruton et al., 2008) were also observed during centrifuge modelling. Hence, the friction factor extracted from modelling and applicable to the specific site conditions could be confidently used for design using enhanced design methods developed from centrifuge testing. The two examples illustrate particularly well the combined contribution of centrifuge modelling and testing. The knowledge about a particular feature of soil-pipe interaction obtained from well controlled centrifuge testing helped interpret the data obtained from centrifuge modelling on specific soil and loading conditions. 6.6
Integration into design
The integration of physical model methods within the design process for geotechnical aspects of pipeline design such as lateral buckling has been discussed by White & Gaudin (2008) and is summarised in Figure 37. Several recent projects worldwide have included
182 © 2011 by Taylor & Francis Group, LLC
centrifuge testing of pipe-soil interaction as part of the design process, making pipeline design the offshore area where centrifuge methods are the most commonly used. As, for any other offshore site specific project, centrifuge modelling needs to be planned in the early stages of the project so sufficient volume of soil can be recovered from the seabed during the geotechnical survey. The modelling programme may include specific centrifuge soil characterisation test to complement results from the geotechnical survey and/or validate the soil sample reconstitution process. A preliminary design of the pipelines will have been undertaken, from which the likely pipe weights, laying motion and in-service behaviour can be extracted in order to provide the required parameters to develop the model and instrumentation and to specify the number of tests needed to provide the required information for design. The centrifuge modelling programme should then be tailored to replicate the anticipated lay process and the in-service pipe movements (with an appropriate range to capture uncertainties) aiming for the results to provide a suitable representation of the design situation. Best practice is for the reporting to follow the same pattern as the conventional geotechnical investigation, with a sequence of factual and interpretive reports. The factual report provides a full record of the modelling activity, typically presenting pipe load-displacement responses during laying, lateral breakout and lateral sweeping, as well as the development of excess pore pressure if measured. The interpretive report analyses the results in light of the calculation model that the design team is anticipating to use for the pipe-soil behaviour. Key values from the testing, such as the as-laid embedment and
Figure 37. Integration of pipe-soil centrifuge modelling within the pipeline design process (after White & Gaudin, 2008).
the friction factors at key stages, should be extracted from the load displacement responses, allowing the uncertain aspects of the calculation model to be calibrated for the particular site. The calibrated calculation model, which may have been originally developed from centrifuge testing, is then used to simulate the particular design situation. This generalisation is usually required because the scope of the centrifuge modelling cannot encompass all patterns of pipe movement during operation. These vary longitudinally along the pipeline. It is therefore necessary to interpolate between the patterns of pipe movement simulated in the model tests. Similarly, the project may involve various weights and diameter of pipeline, which cannot all be simulated in the centrifuge study. Depending on the complexity of the project, outcomes from centrifuge modelling may be completed using finite element analysis calibrated from the centrifuge results.
7
DIRECTIONS FOR THE FUTURE
7.1 Hybrid modelling Real time substructure testing is an emerging technology within civil engineering modelling (Blakeborough et al., 2001). In this approach, the physical model represents only part of a larger dynamic system, with the remainder being simulated numerically in real time, during the experiment.
This permits the modelling of geotechnical structures under realistic and complex loading sequences which will be updated in real time from numerical analysis looped to the motion control system. This is particularly relevant for offshore structures, such as pipelines or jack-up, where the structural response and the geotechnical response are often dependant. This will surely constitute a major breakthrough in physical modelling techniques in the near future. One such example is the hybrid actuator, currently in development at COFS for centrifuge application, to investigate the behaviour of jack-up legs and spudcan when penetrated close to an existing footprint. The actuator models a single leg of a jack-up unit and can apply or measure independent or combined vertical V, horizontal H and moment M loading. In order to include the structural response of the jack-up unit, a stiffness matrix is incorporated at the top of the actuator leg, governing the leg displacement response as a function of the loads developed within the leg. The stiffness matrix is updated in real time, depending on the structural response of the jack-up unit, using a finite element programme. This provides a more realistic modelling of the footprint interaction problem. An even more advanced development is the UKNEES project (Madabhushi et al., 2010). The UKNEES project is a distributed testing network, where testing facilities in one site can be controlled by numerical or physical models located in a different site. In addition, the network features tele-participation with video feed, allowing participants to communicate
183 © 2011 by Taylor & Francis Group, LLC
Figure 38. A pile featuring a wireless transmission module at its head: Mounted on the centrifuge actuator (left), details of the wireless transmission module (right) (after Deeks et al., 2010).
and share information simultaneously. In the example provide by Madabhushi et al. (2010), different parts of a bridge structure with shallow foundations were tested at different locations. The shallow foundation was tested in the Cambridge centrifuge facility. Loading on the foundation was achieved by an actuator, whose input came from physical and analytical models for the pier and sub-structure of the bridge deck tested at Bristol and Oxford. This type of network opens new possibilities for soil-structure interaction, with potential applications in the field of offshore geotechnics. A distributed testing network could for instance integrate the real-time numerical calculation of environmental loadings in one facility, which would feed the demand of a motion control system of a physical model in another facility. With the rapid development of networking facilities and the need for more realistic model testing, such networks will likely become more common in the future. 7.2
Enhanced instrumentation
The continuous development in computing and sensing technology will likely result in improved modelling techniques and new types of investigations. Two trends are identified, which will augment the benefit provided by centrifuge methods, namely image acquisition and intelligent sensors. As described in section 4.3, the development of PIV techniques has already constituted a major
breakthrough in physical modelling. Further improvements are expected with the increasing resolution and miniaturisation of cameras, leading to enhanced soilstructure investigations. High definition high speed cameras have started to be used more commonly for model testing at 1-g (Chow et al., 2010) and in the centrifuge (Boylan et al., 2009b), providing valuable information on high speed events. A good example is provided by Boylan et al. (2010a), where the usual geotechnical analysis is augmented by a geomorphologic analysis of submarine landslide, providing key information about the runout behaviour, and notably about the fluidisation of material and the transition from debris flow to turbidity current. This information could not be captured by usual data monitoring devices. The second trend relates to the development of intelligent sensors. One such example is the system presented by Deeks et al. (2010) to model pile installation by rotary jacking. The challenge associated with the continuous rotation of the pile during installation required the development of a wireless miniaturised data acquisition system to be located within the pile (Figure 38). The system is approximately 90 mm × 40 mm × 40 mm in size, and provides logging, conditioning and wireless transmission of up to 8 channels of data, eliminating the need of a data acquisition card on-board a computer. Such development opens new possibilities in term of data monitoring opening the field for new type of investigations.
184 © 2011 by Taylor & Francis Group, LLC
8
CONCLUSIONS
The paper presented a review of typical offshore centrifuge applications, with a particular emphasis on the contribution of centrifuge methods to offshore design and the benefit of collaboration with industry. The paper lists the advantages and disadvantages of the use of centrifuge methods. It details the difference between centrifuge testing and modelling, with an aim to assist potential centrifuge users and existing modellers in optimising the outcomes of their model tests by defining why, how and when centrifuge methods should be used. The authors advocate the use of centrifuge methods as a valuable tool to assist the offshore industry in developing and designing solutions for a wide range of geotechnical problems. Indeed, centrifuge modelling and testing should not been considered as the unique tool to be used, but as a particular one whose outcomes may be maximised if integrated in a global approach. It is believed that while technological and scientific developments will, without doubt, increase the utility of centrifuge methods, the most spectacular improvement will be in understanding centrifuge methods contribution and their integration at key stages into a global design procedure, incoporating in-situ, numerical and analytical methods. ACKNOWLEDGEMENTS Centrifuge modelling and testing are experimental techniques which rely upon the expertise and dedication of technical staff. The authors would like to gratefully acknowledge the contribution of the technical teams at COFS, C-CORE and LCPC which were invaluable in carrying out successfully the projects presented in this paper. The authors would also like to thank BP for allowing the 2nd author to participate on this paper and funding several studies discussed herein. Progress of centrifuge techniques have greatly benefited from industry incentive and support. The contribution and financial support of the various industry partners involved in the projects presented in the paper are gratefully acknowledged. Fruitful and thought provoking discussions with various colleagues have helped shape this paper. The authors would like to thank particularly Prof. Mark Cassidy, Prof. Dave White and Dr Philippe Jeanjean for their contribution. REFERENCES Andersen, K., Jeanjean, P., Luger, D., Jostad, H.P. 2003. Centrifuge tests on installation of suction anchors in soft clays”. Proc., Intern. Symposium, Deepwater Mooring Systems, Houston, TX. Baba, S., Miyake, M., Tsurugasaki, K., Kim, H. 2002. Development of wave generation system in a drum centrifuge. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 265–270.
Becker, D.E., Burwash, W.J, Montgomery, R.A., Liu,Y. 1998. Foundation design aspects of the Confederation Bridge. Canadian Geotechnical Journal, 35(5), 750–768. Bezuijen, A., Kim, J.H., Schaminée, P.E.L., Lottum H. Van 2007. The protection efficiency of dolphins tested in a geotechnical centrifuge. Proc. 17th Intern. Conf. Offshore and Polar Eng., Lisbon, Portugal, 2, 1609–1625. Bienen, B., Gaudin, C., Cassidy, M.J. 2009.The influence of pull-out load on the efficiency of jetting during spudcan extraction. Applied Ocean Research, 31, 201–211. Blakeborough, A., Williams, BM.S., Darby A.P., Williams, D.M. 2001. The development of real-time structure testing. Earthquake Eng. Struct. Dyn. 30, 431–448. Boylan, N., Gaudin, C., White, D.J., Randolph, M.F. 2010a. Centrifuge modelling of submarine slides. Ocean Engineering. (submitted in May 2010) Boylan, N., Randolph, M.F., Low, H.E. 2010b. Enhancement of the ball penetrometer test with pore pressure measurements. Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. Boylan, N., Gaudin, C., White, D.J., Randolph, M.F., Schneider, J.A. 2009a. Characterisation of submarine slide runout in geotechnical centrifuge. Proc. of the 28th Intern. Conf. Ocean, Offshore and Artic Eng, Honolulu, Hawaii, USA, OMAE2009-79059. Brandenberg, S.J., Choi, S., Kutter, B.L., Wilson, D.W., Santamarina, J.C. 2006. A bender element system for measuring shear wave velocities in centrifuge models. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 165–171. Bruton, D.A.S., White, D.J., Carr, M.C., Cheuk C.Y. 2008. Pipe-soil interaction during lateral buckling and pipeline walking: the SAFEBUCK JIP. Offshore Technology Conf., OTC19589. Bruton, D.A.S., White, D.J., Cheuk, C.Y., Bolton, M.D., Carr, M.C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Offshore Technology Conf., OTC17944. Canepa,Y., Garnier, J., Amar, S. & Corté, J.F., 1988. Comparison between field and centrifugal model tests on shallow foundations, Centrifuge 88, Paris, France, 313–323. Cathie, D.N., Jaeck, C., Ballard, J.-C., Wintgens, J.-F. 2005. Pipeline geotechnics – State-of-the-art. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 95–114. Chen, W. & Randolph, M.F. 2004. Radial stress changes around caissons installed in clay by jacking and by suction. Proc. 14th Int. Offshore and Polar Eng., 2, 493–499. Chen, W. & Randolph, M.F. 2007a. Radial stress changes and axial capacity for suction caissons in soft clay. Géotechnique, 57(6), 499–511. Chen, W. & Randolph, M.F. 2007b. Uplift capacity of suction caissons under sustained and cyclic loading in soft clay. ASCE J. of Geotech. and Geoenv. Eng., 133(11), 1352– 1363. Cheuk, C.Y., White, D.J. Dingle, H.R.C. 2008. Upper bound plasticity analysis of a partially-embedded pipe under combined vertical and horizontal loading. Soils & Foundations. 48(1), 137–148. Cheuk, C.Y., White, D.J. and Bolton, M.D., 2007. Large-scale modelling of soil-pipe interaction during large amplitude cyclic movements of partially embedded pipelines. Canadian Geotechnical Journal, 44(8), 977–996. Chung, S.F., Randolph, M.F., Schneider, J.A. 2006. Effect of penetration rate on penetrometer resistance in clay. ASCE J. of Geo-tech. and Geoen. Eng, 132(9), 1188–1196.
185 © 2011 by Taylor & Francis Group, LLC
Chow, S.H., Nazhat, Y., Airey, D.W. 2010. Applications of high speed photography in dynamic tests. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Clukey, E.C. & Morrison, M. J. 1993.A centrifuge and analytical study to evaluate suction caissons forTLP applications in the Gulf of Mexico. ASCE Proc. Conf. on Foundations, Geotechnical Special Publications, 38, 141–156. Clukey, E.C. & Phillips, R. 2002. Centrifuge model tests to verify suction caisson capacities for taut and semi-taut legged mooring systems. Proc. Deep Offshore Technology Conf., New Orleans. Clukey, E.C., Aubeny, C.P., Murff, J.D. 2003. Comparison of analytical and centrifuge model tests for suction caissons subjected to combined loads. Proc. 22nd Int. Conf. on Offshore Mechanics and Artic Engineering. OMAE-37503. Clukey, E.C., Morrison, M.J., Garnier, J., Corté, J.F. 1995. The response of suction caissons in normally consolidated clays to cyclic TLP loading conditions. Proc. Offshore Technology Conf., OTC 7796. Clukey, E.C., Templeton, J.S., Randolph, M.F., Phillips, R.A. 2004. Suction caisson response under sustained loopcurrent loads. Proc. Offshore Technology Conf. OTC 16843. Colliat, J.L., Dendani, H., Jostad, H.P., Andersen, K.H., Thorel, L., Garnier, J., Rault, G. 2010. Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners end consolidation time, Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. Craig, W.H. 1984. Preface. Proc. Intern. Symp. Application of Centrifuge Modellingto geotechnical Design, Manchester, UK, 1. Craig, W.H., & Al-Sauodi, N.K.S. 1981. The behaviour of some model offshore structures. Proc. 10th Intern. Conf. on Soil Mech. And Found. Eng., Stockholm, Sweden, (2), 541–556. Daiyan, N., Kenny, S., Phillips, R. and Popescu, R. 2010. Investigation of axial/lateral interaction of pipes in dense sand. Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. De Catania, S., Breen, J., Gaudin, C., White D.J. 2010. Development of a multiple axis actuator control system. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Deeks, A.D., White, D.J., Breen, J., Hortin, P., De Catania, S., Brown, T.A. 2010. The use of a wireless data acquisition system for monitoring rotary pile jacking (‘Gyropiling’) in the centrifuge. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Dendani, H., & Jaeck, C. 2007. Pipe-soil interaction in highly plastic clays. Proc. 6th Intern. Conf. Offshore Site Investigation and Geotech. pp. 115–124. Derkx, F., Merliot, E., Garnier, J., Cottineau, L.M. 1998. On-board remote-controlled centrifuge robot. Proc. Centrifuge 98, Tokyo, Japan, 97–104. DeWoolkar, M.M. , Hwang, J. and Ko, H. Y. 2006. Lateral stability of offshore skirted gravity structures. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong Kong, (1), 631–637. DeWoolkar, M.M. , Hwang, J. and Ko, H. Y. 2008. Physical and finite element modelling of lateral stability of offshore skirted gravity structures subjected to iceberg impact load. Int. Ocean Engineering, 35(16), 1615–1626. Dingle, H.R.C, White, D.J., Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout in soft clay. Canadian Geotechnical Journal, 45, 636–652. Dyvik, R., Andersen, K.H., Hansen, S.B., Christophersen, H.P. 1993. Field tests of anchors in clay.
1: Description. ASCE J. Geot. Eng. Div., 119(10), 1515–1531. Finn, Lyle, 2010. Personal communication to Ed Clukey. Gao, F.P. & Randolph, M.F. 2005. Progressive ocean wave modelling in drum centrifuge. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 583–588. Garnier, J. & Cottineau, L.M. 1988. The LCPC centrifuge : means of model preparation and instrumentation, Centrifuge 88, Paris, France, 83–92. Garnier, J., Chambon, P., Ranaivoson, D., Charrier, J. & Mathurin, R. 1991, Computer image processing for displacement measurement, Centrifuge 91, Boulder, USA, 543–550. Garnier, J., König, D. 1998. Scale effects in piles and nails loading tests in sand. Proc. Centrifuge 98, Tokyo, Japan, 205–210. Garnier, J. 2002. Properties of soil samples used in centrifuge models, Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada (1), 5–19. Garnier, J. 2004. Centrifuge modelling in offshore engineering. Proc. 14th Int. Offshore and Polar Eng, Toulon, France. Garnier, J., Gaudin, C., Springman, S.M., Culligan, P.J., Goodings, D., Konig, D., Kutter, B., Phillips, R., Randolph, M.F., Thorel, L. 2007. Catalogue of scaling laws and similitude questions in centrifuge modelling. Intern. J. of Phy. Mod. in Geotech., 7(3), 1–24. Garnier J. & Pecker A. 1999. Use of centrifuge tests for the validation of innovative concepts in foundation engineering. Proc. 2nd Intern. Conf. on Earthquake Geotech. Eng. Lisbonne. Gaudin, C., White, D.J. Bezuijen, A., Schaminée, P.E.L., Garnier J. 2010a. Physical modelling with industry – Overview of practices and benefits. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Gaudin, C., Simkin, M., White, D.J., O’Loughlin, C.D. 2010b. Experimental investigation into the influence of a keying flap on the keying behaviour of plate anchors. Proc. 20th Intern. Conf. Offshore and Polar Eng., Beijing, China. Gaudin, C., Bienen, B., Cassidy, M.J. 2010c. Investigation of the potential of bottom water jetting to ease spudcan extraction in soft clay. Géotechnique. (accepted). Gaudin, C., White, D.J., Boylan, N. Breen, J., Brown, T.A., De Catania, S., Hortin, P. 2009a. A wireless data acquisition system for centrifuge model testing. Measurement Science and Technology, 20, No 095709. Gaudin, C., Landon, M.M., Pederson, V., Kiat, C.H., Lequin, W. 2009b. Design of rock berm cover for offshore pipeline LNG project in Hong Kong. Proc. of the 19th Intern. Conf. Offshore and Polar Eng., Osaka, Japan, 546–553. Gaudin, C. & White, D.J. 2009. New centrifuge modelling techniques for investigating seabed pipeline behaviour. Proc. of 17th Intern. Conf. of Soil Mech. and Geotech. Eng., Alexandria, Egypt, 1, 448–451. Gaudin, C. & Landon, M.M. 2008. CLP Power Ltd Hong Kong LNG project – Rock berm centrifuge testing. The University of Western Australia, GEO 08437, 67 p. Gaudin, C. & Randolph, M.F. 2008. CLP Power Hong Kong Ltd metrological mast. Laboratory centrifuge compression loading tests for Buildings Department in Hong Kong. The University of Western Australia, GEO 08460, 85 p. Gaudin, C. & Cassidy, M.J. 2007. Feasibility and performance of a hybrid foundation system. Report to Keppel Offshore and Marine Technology Centre PTE LTD. The University of Western Australia, GEO 07401, 85 p (confidential). Gaudin, C. & Colwill, R. 2007. Centrifuge modelling of a ship impact against a sloping seawall. Proc. 17th Intern.
186 © 2011 by Taylor & Francis Group, LLC
Conf. Offshore and Polar Eng., Lisbon, Portugal, 4, 3616–3622. Gaudin, C., Randolph, M.F., O’Loughlin, C.D. 2006a. New insights from model tests of foundation and anchoring systems in offshore geomechanics. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 47–62. Gaudin, C., Randolph, M.F., Acosta, H.E., Tran, M. 2006b. Maari field development – Centrifuge modelling of suction foundations. The University of Western Australia, Centre for Offshore Foundation Systems, GEO 06374, 135 p. Gaudin, C., O’Loughlin, C.D., Randolph, M.F., Lowmass, A.C. 2006c. Influence of the installation process on the behaviour of suction embedded plate anchors. Géotechnique, 56(6), 381–391. Hill, A.J. & Jacob, H. 2008. In-situ measurements of pipe-soil interaction in deep water. Offshore Technology Conference, OTC 19528. Hodder, M.S., White, D.J., Cassidy, M.J. 2008. Centrifuge modelling of riser-soil stiffness degradation in the touchdown zone of a steel catenary riser. Proc. Intern. Conf. on Offshore Mechanics and Arctic Engineering, Estoril, Portugal. Hodder, M.S., White, D.J., Cassidy, M.J. 2009. Effect of remolding and reconsolidation on the touchdown stiffness of a steel catenary riser: Observations from centrifuge modeling. Offshore Technology Conference, OTC-19871. Hodder, M.S., White, D.J., Cassidy, M.J. 2010. Analysis of soil strength degradation during episodes of cyclic loading, Illustrated by the T-Bar penetration test. ASCE Inter. J. Geomechanics, (10)3, 117–123. Hossain, M.S., Hu, Y., Randolph, M.F., White, D.J. 2005. Limiting cavity depth for spudcan foundations penetrating clay. Géotechnique, 55(9), 679–690. Iskander, M., Liu, J., Sadek, S. 2001. Transparent amorphous silica to model clay. ASCE J. Geotech. And Geoenv. Eng., 128(3), 262–273. Jacob, H. & Looijen, O. 2008. Development of a deep water tool for in-situ pipe-soil interaction measurements and its benefits in pipeline analysis. Offshore Pipeline Technology Conf., Amsterdam, Netherlands. Jeanjean, P.J., 2009. Development of p-y springs for fatigue analyses. Proc. Offshore Technology Conf., OTC 20158. Jeanjean, P., Znidarcic, D., Phillips, R., Ko, H.Y., Pfister, S., Schroeder, K. 2006. Centrifuge testing on suction anchors: Doublewall, stiff clays, and layered soil profile. Offshore Technology Conf., OTC 18007. Kelleher, P.J. & Randolph, M.F. 2005. Seabed geotechnical characterisation with a ball penetrometer deployed from the Portable Remotely Operated Drill. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 365–371. Lau, M., Phillips, R., McKenna, R.F., Jones, S.J. 2002. Modelling ice-structure interactions in a geotechnical centrifuge. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 325–330. Lee, F.H. 2001. The philosophy of modelling versus testing. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, MonteVerita, Switzerland, 113–131. Lehane, B.M., Gaudin, C., Schneider, J. 2005. Scale effects on tension capacity for rough piles buried in dense sand. Géotechnique. 55(10), 709–719. Lehane, B.M., O’Loughlin, C.D., Gaudin, C., Randolph, M.F. 2009. Rate effects on penetrometer resistance in kaolin. Géotechnique, 59(1), 41–52. Leung, C.F., Xie, Y., Chow, Y.K. 2008. Use of PIV to Investigate Spudcan-Pile Interaction. Proc. 18th Intern.
Conf. Offshore and Polar Eng., Vancouver, Canada, 721–728. Liu, L., Jiang, W., Ma, J. 1988. Automatic data acquisition and processing of centrifugal model tests. Proc. Centrifuge 88, Paris, France, 93–96. Lund, K.M. 2000. Effect of increase in pipeline soil penetration from installation. Proc. Intern. Conf. Ocean Offshore Artic Eng. New Orleans, USA, OMAE2000/PIPE-5047. Madabhushi, S.P.G., Haigh, S.K., Ali, A., Williams, M., Ojaghi, M., Lamata, I., Blakeborough,T.,Taylor, C., Dietz, M. 2010. Distributed testing of soil-structure systems using web-based applications. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Mana, D., Gourvenec, S.M., Hossain, M.S., Randolph, M.F. 2010. Kinematic failure mechanisms of skirted circular foundations under tension and compression. Submitted to journal. Martin, C.M. 2001. Impact of centrifuge modelling on offshore foundation design. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, Monte Verita, Switzerland, 135–154. Morrison, M.J., Clukey, E.C., Garnier, J. 1994. Behaviour of suction caissons under static uplift loading. Proc. Centrifuge94, Singapore, 823–828. Murff, J.D. 1996. The geotechnical centrifuge in offshore engineering. Proc. Offshore Technology Conf., OTC 8265. Newman, J.N. 1977. Marine Hydrodynamics, MIT Press, 354 p. Ng, C.W.W., Zhang, L.M., Wang, Y.H. 2006. Physical Modelling in Geotechnics – 6th ICMPG06, Taylor&Francis. Ng, C.W.W., Van Laak, P.A., Zhang, L.M., Tang, W.H., Zong, G.H., Wang, Z.L., Xu, G.M., Liu, S.H. 2002. Development of a four-axis robotic manipulator for centrifuge modelling at KHUST. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 71–76. O’Loughlin, C.D., Randolph, M.F., Richardson, M.D. 2004. Experimental and theoretical studies of deep penetrating anchors. Offshore Technology Conference, OTC 16841. Oung, O. & Bezuijen, A. 2002. Development of selective pore pressure transducers and their use to check the validity of classical capillary pressure-saturation curves. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 95–100. Ovesen, N.K. 1975. Centrifuge testing applied to bearing capacity problems of footings on sand. Géotechnique, 25(2), 394–401. Palmer, A.C., White, D.J., Baumgard, A.J., Bolton, M.D., Barefoot, A.J., Finch, M., Powell, T.A., Faranski, A.S., Baldry, J.A.S. 2003. Uplift resistance of buried submarine pipelines: comparison between centrifuge modelling and full-scale tests. Géotechnique, 53(10), 877–883. Phillips, R., Barrett, J.A., Al-Showaiter, A.S. 2010. Ice keel-seabed interaction: numerical modelling validation. Offshore Technology Conference, OTC 20696. Phillips, R., Clark, J.I. and Kenny, S. 2005 PRISE studies on gouge forces and subgouge deformations. 18th Int Conf on Port & Ocean Engineering under Arctic Conditions, 1, p75–84. Phillips, R., Clark, J.I., Hanke, R. 2002. Pipeline frost heave modelling. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 313–318. Phillips, R. & Sekiguchi, H. 1992. Generation of water wave trains in drum centrifuge. Proceedings of Techno-Ocean ’92, Yokohama, Japan, volume 1, pp 29–34. Phillips, R., Walter, D.J., Kosar, K.M. 1998. Physical modelling of foundations for Confederation Bridge, Canada. Proc. Centrifuge 98, Tokyo, 447–452. Purwana, O.A., Leung, C.F., Chow, Y.K. 2006. Breakout failure mechanism of jackup spudcan extraction. Proc.
187 © 2011 by Taylor & Francis Group, LLC
6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 667–672. Raines, R.D., Ugaz, O., Garnier, J. 2005. Centrifuge modelling of suction piles in clay. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 303–308. Rakotonindriana, J. 2009. Comportement des pieux et groupes de pieux sous chargement latéral cyclique, PhD Thesis, Ecole Nationale des Ponts et Chaussées, Paris, 381 p. Randolph, M.F. & White, M.J. 2008a. Pipeline embedment in deep water: processes and quantitative assessment. Offshore Technology Conference, OTC 19128. Randolph, M.F. & White, D.J. 2008b. Upper bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4), 297–301. Randolph, M.F., Cassidy, M.J., Gourvenec, S.M., Erbrich, C. 2005. Challenges of offshore geotechnical engineering. Proc. of 16th Intern. Conf. of Soil Mech. and Geotech. Eng., Osaka, Japan, 1, 123–176. Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. 1998. Performance of suction anchors in fine-grained calcareous soils. Proc. Offshore Technology Conference, OTC 8831. Rezzoug, A., König, D., Triantafyllidis, T.. 2000. Scaling laws in centrifuge modelling for capillary rise in soils, Proc. Int. Symp. On Physical Modelling in Environmental Geotechnics, La Baule, France, 217–224. Richardson, M.D., O’Loughlin, C.D., Randolph, M.F. and Gaudin, C. 2009. Setup following installation of dynamic anchors in normally consolidated clay. ASCE J. of Geotech. and Geoenvironmental Eng., 135(4), 487–496. Richardson, M.D., O’Loughlin, C.D., Randolph, M.F., Cunnigham, T.J. 2006. Drum centrifuge modelling of dynamically penetrating anchor. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, 673–678. Rosquöet, F., Thorel, L., Garnier, J., Canepa, Y. 2007. Lateral cyclic loading of sand-installed piles. Soils and Foundations, (47)5, 821–832. Rowe, P.W. & Craig, W.H. 1981.Applications of models to the prediction of offshore gravity platform foundation performance. Proc. Intern. Conf. on Offshore Site Investigation, London, 269–281. Sasanakul, I., Vanadit-Ellis, W., Sharp, M., Abdoun, T., Ubilla, J., Steedman, S., and Stone, K. 2008. New Orleans levee system-performance during Hurricane Katrina. Proc. 17th Street Canal and Orleans Canal North. ASCE J. Geotech. Eng. Div., 134(5), 657–667. Sassa, S. & Sekiguchi, H. 1999. Wave-induced liquefaction of beds of sand in centrifuge. Géotechnique, 49(5), 621–638. Sarma, D. 2006. Geotechnical modelling by artificial soil. A new approach in model study. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 235–240. SNAME. 1997. Guidelines for site specific assessment of mobile jack-up units. Society of Naval Architects and Marine Engineers, Technical and Research Bulletin 5-5A Rev 1, New Jersey. Song, Z., Hu, Y., O’Loughlin, C.D., Randolph, M.F. 2009. Loss in anchor embedment during plate anchor keying in clay. ASCE J. of Geot. And Geoenv. Eng., 135(10), 1475–1485. Stewart, D.P. & Randolph, M.F. 1994. T-bar penetration testing in soft clay. ASCE J. Geotech. Eng. Div., 120(12), 2230–2235. Take, AW.A. & Bolton, M.D. 2002. A new device for the measurements of negative pore water pressure in centrifuge models. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 89–94. Taylor, R.N. 1995. Geotechnical centrifuge technology. Blackie Academic and Professional.
Teh, K.L., Cassidy, M.J., Leung, C.F., Chow, Y.K., Randolph, M.F. and Quah, C.K. 2008. Revealing the bearing capacity mechanisms of a penetrating spudcan through sand overlying clay. Géotechnique, 58(10), 793–804. Ubilla, J., Abdoun, T., Sasanakul, I., Sharp, M., Steedman, S., Vanadit-Ellis, W., and Zimmie, T. 2008. New Orleans levee system performance during Hurricane Katrina: London Avenue and Or-leans Canal South Canal and Orleans Canal North. ASCE J. Geotech. Eng. Div., 134(5), 668–680. Watson, P.G., Gaudin, C., Senders, M., Randolph, M.F. 2006. Installation of suction caisson in layered soil. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 685–692. Westgate, Z., Tapper, L., Lehane, B.M., Gaudin, C. 2009. Modelling the installation of stiffened caissons into overconsolidated clay. Proc. 28th Intern. Conf. Ocean, Off. and Artic Engineering, Honolulu, Hawaii, USA. OMAE200979125. White, D.J. & Hodder, M. 2010. A simple model for the effect on soil strength of episodes of remoulding and reconsolidation. Canadian Geotechnical Journal, In Press. White, D.J. & Cathie, D.N. 2010. Geotechnics for subsea pipelines. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth, Australia. White, D.J., Gaudin, C., Boylan, N., Zhou, H. 2010. Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils. Canadian Geotechnical Journal, 47(2), 218–229. White, D.J. & Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement. Marine Structures, 21, 59–79. White, D.J. & Gaudin, C., 2008. Simulation of seabed pipeline behaviour using centrifuge modelling. Proc. Intern. Deep Offshore Tech. Conf., Perth, Australia. White, D.J. & Gaudin, C. 2007. Browse LNG Development Phase 2C: Centrifuge modelling of cone penetration tests. Report to WorleyParsons and Woodside Energy Ltd. The University Of Western Australia, GEO 07422, 21 p. (confidential). White, D.J., Take, W.A., Bolton, M.D. 2003. Soil deformation measurement using particle image velocimetry PIV and photo-grammetry. Géotechnique, 53(7), 619–631. White, D.J, Randolph, M.F., Thompson, B. 2005. An imagebased deformation measurement system for the geotechnical centrifuge. Int. J. of Physical Modelling in Geotechnics, 5(3), 1–12. White, D.J. & Bolton, M.D. 2004. Displacement and strain paths during plane-strain model pile installation in sand. Géotechnique. 54(6), 375–397. White, D.J. & Bolton, M. D. 2001. Observing friction fatigue on a jacked pile. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, Monte Verita, Switzerland, 347–355. Wilson, D.W., Kutter, B.L., Boulanger, R.W. 2010. A summary of the Davis centrifuge facility, recent history and current research, NEES@UC Davis. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Yan, Y., White, D.J., Randolph M.F. 2010. Investigations into novel shallow penetrometers for fine-grained soils. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth, Australia. Yang, D. & Goodings, D.J. 1998. Climatic Soil Freezing Modeled in Centrifuge. ASCE J. of Geot. And Geoenv. Eng., 124(12), 1186–1194. Zhou, H. & Randolph, M.F. 2009. Numerical investigations into cycling of full-flow penetrometers in soft clay. Géotechnique, 59(10), 801–912.
188 © 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert The University of Texas at Austin
J.D. Murff Consultant
E.C. Clukey BP America
ABSTRACT: Operating in the frontiers of offshore energy production poses significant risk. The objective of this paper is to present lessons learned in managing risk on the frontier of offshore geotechnics. Case histories are described to underscore the following themes: (1) Achieving an appropriate risk requires balancing risk and conservatism; (2) Managing risk requires understanding the loads on our designs as well as the capacities; (3) Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions; and (4) Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The paper concludes with guidance to improve the state of practice. Better communication between all parties involved in design, construction and operation and earlier application of risk and reliability principles in the life cycle of a project will enhance the practical value of these principles.
1
INTRODUCTION
Operating in the frontiers of offshore energy production poses significant risk. The consequences of a failure are severe and the costs to mitigate risks are enormous. The offshore industry has been a leader in explicitly considering risk in design and decision making. One of the first reliability-based design guidance documents was developed for offshore facilities. In addition, this industry has been at the forefront in articulating and managing risk levels. Most importantly, due to the extreme conditions under which we operate, we have had the opportunity to learn from experience to innovate and improve practice to better manage risk. The objective of this paper is to present lessons learned in managing risk. Case histories are described to underscore the following themes: 1. Achieving an appropriate risk requires balancing risk and conservatism. 2. Managing risk requires understanding the loads on our designs as well as the capacities. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The case histories presented represent a wide range of facilities, operators and locations. While what
is described is derived from real projects, it only represents a small and geotechnically-biased window into the numerous decisions and factors being considered in making those decisions. The paper concludes with guidance for improving the state of offshore geotechnical practice to better manage risk in the future. 2 ACHIEVING APPROPRIATE RISK Risk is the possibility of a loss. An appropriate risk is one that balances the cost of reducing the risk against that of accepting the risk.Appropriate risk is never zero risk, and excessive conservatism can be as troublesome as excessive risk. Risk is quantified as the expected consequence of loss. In the simplest case, if the possibility of loss is a binary event that either does or does not occur, then the risk is equal to the probability of the loss multiplied by its consequence. Therefore, risk is commonly represented by the probability and consequence of a loss. Consequences can include human safety, environmental and economic losses. General guidance for striking the balance between reducing versus accepting risk has been developed by different governments and industries. Examples include USNRC (1975) for nuclear power plants, AIChE (1989) for chemical process facilities, ANCOLD (1998) and USBR (2003) for dams, and Bea (1991), Stahl et al. (1998) Goodwin et al. (2002) and Gilbert et al. (2001) for offshore production facilities.
189 © 2011 by Taylor & Francis Group, LLC
ballast would provide an acceptable reliability, even if the riser was installed immediately after pile driving. 2.2
Figure 1. Required ballast versus set-up time to achieve target probabilities of overload during installation for a riser tower foundation.
or on the capacity side of the design check (increasing the capacity),
The following case histories illustrate how the cost of reducing risk against that of accepting risk can be balanced in practice. 2.1
where rn is the nominal design capacity of the suction caisson in uplift due to side shear and reverse end bearing, sn is the nominal design load in uplift, and ϕR and γS are resistance and load factors, respectively. The ratio γS /ϕR is greater than 1.0, meaning that a larger nominal design capacity would be required if the weight of the caisson were considered to be a component of the capacity (Equation 1b). This case underscores the arbitrariness implied typically in different design formulations. The design-build contractor proposed using Equation (1a) versus Equation (1b). The basis for their proposal was that there was relatively less uncertainty in the weight of the caisson than the components of side shear and reverse end bearing. In addition, this proposal provided for a less expensive foundation. In order to evaluate the appropriateness of this proposal, a reliability analysis was performed. This analysis has the advantage that the results do not depend on a particular formulation for the design check. The event of interest was failure of the suction caisson in uplift under the long-term sustained load from the riser tower (short-term dynamic loads were very small in comparison to the total uplift load). The calculated probability of uplift failure in the 20-year design life was between 0.01 and 0.1. This probability of failure was considered to be intolerably high, and the design standard was subsequently clarified and revised in order to provide for a higher level of reliability.
Driven pile foundations for riser towers
The riser towers for an oil production system in deep water were to be held in place with driven pile foundations. Since the preliminary design was based on suction caissons, the soil borings were relatively shallow. The driven pile lengths were subsequently constrained by the depth of the deepest soil boring due to concerns about driveability. In addition, the largest load would be applied during riser installation, which potentially could occur before significant set-up after pile driving. Based on a design check using nominal loads and capacities, the axial uplift capacity of the piles alone was not sufficient. In order to increase the capacity, a ballasting system was developed, with boxes at the head of the piles that could be filled with steel ballast to increase the uplift capacity. A reliability analysis was performed to assess the need for the ballast boxes and the optimal amount of ballast (Fig. 1). The event of concern was overload during riser installation when the uplift force would be greatest because the riser is empty and the pile capacity would not necessarily have reached its maximum value due to full set-up. Based on a consideration of the consequences of a failure (primarily economic), the target threshold for the probability of overload was set between 0.001 and 0.0001. The results in Figure 1 provide a range of possible means to achieve an appropriate level of risk, including using larger ballast with a shorter set-up time or smaller ballast with a longer set-up time. Based on the proposed check, in which nominal or conservative values were used both for the uplift load and pile capacity, a ballast weight that was nearly 70 percent of the design load was selected (Fig. 1). The reliability analysis indicated that less than half that amount of
3
UNDERSTANDING LOADS
The focus in offshore geotechnics is generally on the capacity of geotechnical systems. However, the reliability of these systems can be influenced as much by the loads they experience as by their capacity. The
190 © 2011 by Taylor & Francis Group, LLC
Suction caisson foundations for riser towers
A suction caisson was designed to anchor a bundle of risers for an oil production system in deep water. This design was unconventional in that the maximum load was a sustained tension load and not a storm load and the capacity was primarily from the weight of the foundation and not the shear strength of the soil. This particular situation meant that conventional factors of safety in the design standard did not necessarily apply. Specifically, the question was whether to include the weight of the foundation on the load side of the design check (reducing the total uplift load),
Figure 2. Schematic of wave-induced mudslides.
loading
mechanism
for
following case histories highlight the importance in understanding loads.
Figure 3. Results of slope stability analyses for offshore site in 130 m of water.
3.1 Wave-induced mudslides Wave-induced mudslides have caused considerable damage to offshore facilities in the Gulf of Mexico, particularly to pipelines that are an essential link in the supply of oil and gas. In Hurricanes Ivan (2004) and Katrina (2005), more than 50 pipelines and one platform were damaged or destroyed by mudslides in the Mississippi Delta (OTRC 2008). A schematic of the loading mechanism for a wave-induced mudslide is shown in Figure 2. The differential pressure on the sea floor underneath the waves (Fig. 3) imparts shear stresses in the soil. If these shear stresses exceed the shear strength of the soil, then a slope failure will occur. Results of stability analyses for a site subjected to large waves in Hurricanes Ivan and Katrina are shown in Figure 3 (OTRC 2008). The analysis assumes that the soft clays are sheared under undrained conditions by the waves and uses a profile of undrained shear strength versus depth from a nearby soil boring. Mudslides occurred here in both hurricanes, which is consistent with the stability analyses (Fig. 3). While the maximum wave heights in Hurricane Ivan at this location were nearly 20 percent smaller than those in Hurricane Katrina, the bottom pressure loading was actually higher in Hurricane Ivan and the factor of safety was smaller due to the longer wave period (Fig. 3). This result is significant because the eye for Hurricane Ivan was nearly 150 km to the east of the Delta (in contrast to Katrina’s eye, which passed over the Delta); the relatively small wave heights for Ivan in the Delta would not have been expected to induce mudslides, particularly in water depths of 130 m. However, the magnitude of bottom pressures, and therefore the risk of mudslides, depends both on the wave height and period. A comparison of wave heights and periods in the Delta for these two storms is shown in Figure 4. The conventional approach to characterize sea states in a hurricane is to relate wave period to wave height
Figure 4. Comparisons of wave periods and wave heights for waves in Mississippi Delta during recent hurricanes (from OTRC 2008).
based on empirical data collected from the largest waves in a storm (i.e., those waves near the eye). The curve labeled “Average” in Figure 4 is typically used to associate wave periods with wave heights for design purposes, such as in designing a production platform. However, the average curve under-represents the potential for long period waves away from the eye; the periods for the waves in the Delta during Ivan were similar to those near the eye, 150 km to the east, even though the wave heights were smaller (Fig. 4). The risk for mudslides in the Delta would be underestimated if this potential for longer period waves were not considered. Since platform loads are not very sensitive to the wave period, this issue was not recognized by and is not captured by the standard design guidance documents developed for platforms. However, it is an important consideration for mudslides. This case history underscores the significance of having a careful understanding of loads and not using information from design guidance documents without knowing its source and limitations.
191 © 2011 by Taylor & Francis Group, LLC
Figure 6. Conceptual relationship between design conservatism, information and acceptable reliability.
of designers to apply conservatism at every decision point, it can clearly result in costly designs.
Figure 5. Annual probability that the actual load on the foundation will exceed the design load.
4
3.2 Tendon loads There can be significant conservatism embedded in the design loads that geotechnical engineers use to size foundations. To illustrate this point, the following conservative assumptions were made in establishing the design tendon load for a driven pile foundation of a Tension Leg Platform. First, an unlikely combination of wave height and period was assumed to establish the maximum tendon load. The 100-year design load corresponded to a significant wave height having a 0.01 annual probability of exceedance (the “100-year” wave height) combined with an associated peak spectral period with only a 5-percent probability of exceedance. Since the maximum tendon load increases with decreasing wave period due to the natural period of the structure, the resulting design load can be significantly greater than the load that will be exceeded with an annual probability of 0.01 (the “100-year load”). In addition, a conservative estimate of the tide and the worst possible wave loading direction were used for design, giving the maximum possible tendon load for a sea state. However, this combination of tide and wave loading direction was relatively unlikely in the conditions expected to produce the largest waves. Second, conservative assumptions were made concerning pre-tension measurement error in the tendon, mis-positioning of the pile head, and subsidence of the sea floor. The result of these compounded conservatisms was a design load with an extremely small probability of being exceeded. Figure 5 shows a probability distribution for the maximum annual tendon load on the pile. The annual probability that it will be exceeded is less than 1 in 100,000 (Fig. 5). Therefore, the “100-year” design load was really a “100,000-year” design load, which was subsequently factored up to establish the required design capacity. Therefore, the probability of failure for this design was many orders of magnitude smaller than what would generally be considered tolerable. This approach, selecting a series of conservative parameters in a design, is often referred as “double dipping.” Although it may be the natural tendency
A common decision point in offshore design is whether or not additional data are needed or would be beneficial enough to justify the cost of acquiring the data. A conceptual schematic illustrating how reliabilitybased design can be used to guide this decision is shown in Figure 6. This figure shows the cost of a foundation versus the cost of the information that is used to design the foundation. An acceptable level of reliability can be achieved by a variety of combinations of design information to reduce uncertainty in the foundation performance and design conservatism to limit the effect of uncertainty on the performance of the design. Design conservatism is reflected in the load and resistance factors or factors of safety as well as the nominal values used for loads and capacities. Increasing the cost of design entails increasing the difference between the expected capacity and the expected load. The level of information used in developing a design is rarely expressed explicitly in a code; however, it is typically implied through conventional practice. For example, design codes for foundations typically assume that a site-specific geotechnical investigation has been conducted to develop the design. Once a reliability-based design code is calibrated, the combination of the design information and the design conservatism presumably gives foundations with an acceptable level of reliability. The value of additional information depends on how much that information is expected to reduce design conservatism. This potential cost savings can then be compared against the cost of obtaining additional information, which is often governed by the time it will take to acquire it and the subsequent delay in development. Maximizing the value of information entails finding the optimal combination of design information and conservatism, such as the combination that will minimize the total expected cost. The following case histories illustrate projects where risk and reliability principles were used to improve the value of information.
192 © 2011 by Taylor & Francis Group, LLC
MAXIMIZING THE VALUE OF INFORMATION
4.1 Design without site-specific soil boring A mature field had a rather substantial amount of geotechnical data, with more than 50 soil borings and geophysical survey lines. In addition, the cost and time required to obtain site-specific borings had been steadily increasing. Therefore, the question was whether or not a site-specific soil boring was worth its cost for every new facility in this field. The conceptual approach shown in Figure 6 can be captured in the design checking equation for a reliability-based code as follows:
where rn is the nominal design capacity and sn is the nominal design load; ϕR and γS are the conventional load and resistance factors, respectively, to account for the uncertainty in these quantities even if a new soil boring were available at the site location; and ϕspatial is a partial resistance factor to account for the added uncertainty if the amount of design information is less than the convention. As the magnitude of added uncertainty increases due to having less design data, the quantity (1/ϕspatial ) increases, meaning that a more conservative design is required to achieve the conventional reliability. In order to relate the value of (1/ϕspatial ) to the amount of design information, a geostatistical model was developed for this offshore field and calibrated with the available geotechnical data. The details of this model are described elsewhere (Gambino and Gilbert 1999 and Gilbert et al. 2008), and its important features are summarized here: • The model accounts for horizontal and vertical
correlation in the pile capacity. • The model incorporates data from modern borings (pushed sampling methods) as well as data from older borings where samples were obtained using wire-line percussion methods and were more disturbed. The effect of the method of sampling was included both in the expected pile capacity as well as the standard deviation in the capacity. • The model predictions reflect uncertainty both due to spatial variations as well as systematic uncertainty in the model due to the limited amount of data available to calibrate it. The output from this model is shown in Figure 7 for a small section in this field. The mean capacity at the location of an available, modern boring is equal to the design capacity obtained directly from that boring (Fig. 7a). The mean capacity at the location of an available, older boring is adjusted from the boring data to reflect what would be expected if a modern boring were drilled at that location. The unconditional mean or average for the field is 30 MN (Fig. 7a); the data at both the modern and the older boring suggest that the pile capacity within this section is higher than that for the field on average. As the location of the structure moves away from the location of the borings, the
Figure 7. (a) Expected value (b) and coefficient of variation (c.o.v.) for estimated pile capacity versus location for a 100-m long, 1-m diameter steel pipe pile (from Gilbert et al. 2008).
expected value for the design capacity approaches the unconditional mean (Fig. 7a). The added uncertainty in the capacity, expressed as the coefficient of variation or standard deviation divided by the mean, is zero at the location of the modern boring because this data point represents the conventional practice (Fig. 7b). The added uncertainty in the capacity is greater at the location of the older boring, reflecting the greater uncertainty associated with these data compared to modern practice. The ceiling of Figure 7b corresponds to the unconditional variance for the field. As the location of the structure moves away from the boring locations, the added variance approaches the unconditional variance (Fig. 7b). The required value for (1/ϕspatial ) is related to the added uncertainty due to not having a modern boring at the site location. Figure 8 shows this relationship for this field, where the added uncertainty is expressed as a coefficient of variation. In this application, the coefficient of variation for the added uncertainty was typically between 0.05 and 0.08 (Figure 7b); therefore, the required safety margin is about 10 percent greater than for the conventional case (i.e., 1/ϕspatial is about 1.1 in Fig. 8). These results provided the owner of the structure with guidance in making a decision between drilling a
193 © 2011 by Taylor & Francis Group, LLC
Figure 8. Partial resistance factor versus coefficient of variation for spatial variability.
Figure 9. Probability distributions for tendon load and tensile pile capacity.
new boring or relying on the existing data and increasing the safety margin by about 10 percent. If the cost of increasing the pile length was less than the cost of a site-specific boring, then the alternative of moving ahead without another soil boring (i.e., “Less Data” on Fig. 6) would be preferred. The value of information from a site-specific boring depends on the geology; the greater the spatial variability, the greater the penalty to be paid by not having a site-specific boring (Fig. 8). The value of information also depends on the use of the information. For example, decisions about pile driveability may be more or less sensitive than those about pile capacity to the information from a site-specific boring. 4.2
Updated design using pile driving data
The pile foundation for a tension leg platform in a frontier area was designed based on a preliminary analysis of the site investigation data. The geotechnical properties of the site were treated in design as if the soil conditions were similar to other offshore areas where the experience base was large. The steel was then ordered. Subsequently, a more detailed analysis of the geotechnical properties showed that the soil conditions were rather unusual, calling into question how the properties should be used in design and leading to relatively large uncertainty in the estimated pile capacity. With three different commonly-used methods for estimating the tensile capacity of the piles, the factor of safety ranged from less than 2.5 to more than 4 (Fig. 9). For reference, the target factor of safety was 3. The owner was faced with a series of decisions. First, should they stay with the original pile design or change it given the uncertainty in the pile capacity, considering that changing the pile design after the steel had been ordered would substantially impact the cost and schedule of the project? Second, if they decided to stay with the original pile design, should they monitor the installation to confirm the capacity was acceptable, considering that this approach required a flexible contract where the pile design may need to be updated after the pile is installed?
194 © 2011 by Taylor & Francis Group, LLC
Figure 10. Expanded view of probability distributions for tendon load and tensile pile capacity.
A reliability analysis was performed to provide guidance in answering these questions. One of the first findings from this analysis was that the uncertainty in the maximum tendon load was small compared to that in the pile capacity (Fig. 9). The second finding from this analysis was that the reliability of the pile foundation was governed by the possibility of the capacity being far below what was predicted by any of the design methods being considered. Figure 10 shows how the most probable combination of load and capacity leading to failure (the “Most Probable Failure Point”) is well below the estimated axial capacity assuming that the side shear is equal to the remolded undrained shear strength, which is a reasonable lower-bound on the available capacity. To support this concept of a lower-bound capacity, the pile load database that was used to develop and calibrate the API Design Method is shown in Figure 11 together with the calculated lower-bound for each data point. In all cases, the measured capacity exceeds the calculated lower-bound.
Figure 12. Probability of foundation failure in design life versus a lower-bound estimate of the pile capacity.
Figure 11. Comparison of measured capacity for driven piles in clay soils with estimated lower-bound capacity calculated assuming the side shear is equal to the remolded undrained shear strength of the clay; measurements are shown as a range to account for uncertainty in interpreting the load tests results (adapted from Najjar 2005).
A reliability analysis that explicitly accounted for the existence of a lower-bound or minimum value of the pile capacity was then performed. The lower bound was modeled as uncertain, with a normal distribution having a coefficient of variation equal to 0.2. The results of this analysis are shown in Figure 12, where the estimated value for the lower bound represents the most likely lower-bound value. If the estimated lower bound was greater than 0.25 times the estimated (or median) pile capacity after set-up, then the probability of foundation failure in the design life was smaller than the target of 0.001 (Fig. 12). The estimated value for the lower bound, obtained using the side shear equal to the remolded undrained shear strength, was about 0.4 times the estimated capacity after set-up, meaning that the existing design would have an acceptable reliability. Pile driving data during and after driving were used to update the reliability of this foundation. In general, pile monitoring information for piles driven into normally to slightly overconsolidated marine clays cannot easily be related to the ultimate pile capacity due to the effects of set-up following installation and uncertainties in the pile driving model. The capacity measured at the time of installation may only be 20 to 30 percent of the capacity after set-up. However, the pile capacity during installation does arguably provide a lower bound on the ultimate pile capacity, which governed the reliability of the foundation in this case. The estimated capacity based on the soil resistance to driving indicated that the reliability of the foundation was acceptable (Fig. 12). A re-tap after several days of set-up further supported this conclusion (Fig. 12). 4.3 Communication of uncertainty due to lack of information An important aspect in evaluating information is to clearly communicate uncertainty. As geotechnical
Figure 13. Stratigraphic cross-section with locations of soil borings and proposed platform (adapted from Gilbert et al. 2006).
engineers, we are rarely in the position of making decisions directly; rather we are conveying what we do and do not know as guidance to the decision makers. If a decision maker does not understand the magnitude of uncertainty, then they may not appreciate the potential value of additional information. A common offshore example is shown in Figure 13 where the final location of the platform does not correspond to the locations of the available soil borings. The geologic setting is a normally consolidated marine clay that is interbedded about every 30 m with overconsolidated clay crusts. In addition, there are buried alluvial channels that are filled with a variety of clay, silt and/or sand. A channel filled with dense sand was encountered at the location of Boring B. The risk is whether the driven piles at the platform location will encounter a thick enough layer of dense sand to cause driving problems. The conventional means to convey the available information is shown in Figure 13. A cross-section is drawn between the borings that includes geophysical data together with the data from the borings. The difficulty is that this cross-section conveys what is
195 © 2011 by Taylor & Francis Group, LLC
Figure 14. Uncertainty multiples showing uncertainty in straigraphy at platform location (adapted from Gilbert et al. 2006).
expected, not what might occur. Even though there are question marks on the channel location away from the boring, a decision maker will tend to assume that there is no possibility of sand at the platform location because it is not shown below the platform location in the cross-section (Fig. 13). However, there is considerable uncertainty in whether or not and how much dense sand lies below the platform location. The available geophysical survey lines do not match up directly with the location of the platform. In addition, there is considerable uncertainty in the interpretation of the geophysical data, time-slice (seiscrop) plots from high resolution, surface-towed boomers. Comparisons of geophysical interpretations with subsequent boring data in this field indicated that the geophysical interpretations were not very accurate at identifying and locating channels buried with sand. An alternative depiction of the uncertainty in this information is shown in Figure 14. This depiction makes use of multiples, which are small-scale images positioned within the eye span on a single page or screen (Tufte 1990). In this case, “uncertainty multiples” (Gilbert et al. 2006) are used to convey the range of possible interpretations of the information based on the collective judgment of the geotechnical engineers and geologists, with each image representing an equally probable possibility. Figure 14 shows that there is a reasonable possibility (4/9 probability) of encountering sand at the platform location. It shows that if sand is encountered, its thickness could range from several meters to tens of meters. It shows that if a channel does lie below the platform location, it may or may not be the same channel or filled with the same material as the channel encountered at Boring B. It even shows that the thickness of the sand layer at the location of Boring B is uncertain; there is a 1/9 probability that it is thinner due to sand running into the borehole when the layer was encountered. The intent of Figure 14 is that a decision maker, who is most likely not a geotechnical engineer or a geologist, will be readily able to understand the uncertainty in the subsurface conditions. Whether or not a new site-specific soil boring is warranted will depend
Figure 15. Schematic plan view of mooring system for study spar.
on the cost of the boring, the cost of being prepared to drive through a dense sand layer if it is encountered, and the expected cost (or risk) of pile driving difficulty. 5
Most design considerations for foundations are based on individual components. However, these components generally perform within a system of foundation and structural elements. Ultimately, it is the performance of the system that is of most interest and concern. The following case histories underscore the importance of considering the system as well as the components. 5.1
Mooring system
The mooring system for a spar is shown in Figure 15. This spar design was developed by industry to be representative of typical practice for the purposes of studying and assessing that practice (OTRC 2006). There were fourteen lines, each composed of a section of chain connecting the line to the hull, a section of rope from the upper chain to just above the mudline, a section of lower chain connecting the rope to the suction caisson anchor at a padeye below the mudline. Three different water depths were considered for this system: 1,000 m with a semi-taut mooring spread and wire rope lines; 2,000 m with a taut mooring spread and polyester lines; and 3,000 m with a taut mooring spread and polyester lines. The probabilities of failure for individual components in the most-heavily-loaded line are shown in Figure 16. For each design (water depth), the probability of failure for the anchor is more than two orders of magnitude smaller than that for the rope and chain segments of the line; hence, failure of a mooring line during a storm is expected to be a break in the line itself versus a pull-out of the anchor. There is a potential to make the anchor designs more efficient (e.g., using a lower factor of safety) without jeopardizing the performance of the mooring system. Second, the consequence of a failure may depend on how the
196 © 2011 by Taylor & Francis Group, LLC
CONSIDERING SYSTEMS AS WELL AS COMPONENTS
Figure 16. Probability of failure for components in most-heavily-loaded line.
failure occurs. Failure in the lines means that the hull could move off station by hundreds of kilometers during a storm and collide with other offshore facilities or coastal facilities. Failure by pull-out of the anchors means that the hull may not move off station as far due to the restoring force provided by the weight and dragging resistance of the anchors. However, the dragging anchors could damage seafloor facilities such as well heads and pipelines. The results on Figure 16 also show that the reliabilities of the components (chains, rope and anchor) and of the system of components that make up a single line (total) depend on the type of mooring system. The probabilities of failure are notably larger for the semi-taut system in 1,000 m of water compared to the taut systems in 2,000 and 3,000 m of water. These differences in reliability are due to using the same design recipe for all mooring systems, which is the standard of practice. Uncertainty in the maximum line load for a taut system is smaller than that for a semi-taut system because a greater proportion of the total load is pre-tension and controlled. Hence, the probability of failure for a taut system is smaller since it is less likely that the actual load will exceed the design capacity when the same factors of safety or load and resistance factors are used. Redundancy in the mooring system is shown in Figure 17 as the probability that the mooring system will fail (i.e., loss of station keeping for the spar) in the event that a single line has failed during a hurricane. A measure of redundancy is the inverse of the probability of system failure given component failure; for example, this redundancy factor for the semi-taut system is about 1.6 (or 1/0.6). Again, there is a significant difference between the different designs (water depths). For the semi-taut system, there is relatively little redundancy between failure of a single line and failure of the system. Conversely, the taut system distributes the environmental load more evenly between lines, and the corresponding redundancy factor is between 15 and 20. For comparison, the redundancy for an eight-leg fixed jacket in shallow water is between 30 and 125, depending on the loading direction (Tang and Gilbert 1993).
Figure 17. Probability of failure of the mooring system in the event that one line fails during a hurricane.
5.2
197 © 2011 by Taylor & Francis Group, LLC
Jacket pile system
System effects can be even more pronounced for fixed jackets compared to floating production systems. To illustrate interactions between the jacket and the foundation system, the capacity of the foundation system for an 8-pile jacket that experienced large loads in Hurricane Ike is shown in Figure 18. The curves in Figure 18 represent capacity envelopes for various combinations of base shear and overturning moment applied to the foundation system; a combination of shear and moment that is outside of the envelope is expected to cause system failure. These system capacity curves were obtained using an upperbound, plasticity model of the pile system (Murff and Wesselink 1986 and Gilbert et al. 2010). The case study platform in Figure 18 is of particular interest because while the maximum load in Hurricane Ike was greater than the estimated capacity for the foundation system, the foundation survived the hurricane. The structural engineers assessing this structure after the hurricane increased the undrained shear strength of clay layers along the pile sides by a factor of three times in order to “explain” its survival. This practice is common place, and has led to a perception that foundation designs are excessively conservative (e.g., Gilbert et al. 2010). This case also points out the fallacy of trying to adjust an overall result by dramatically varying a parameter that has a relatively weak effect. However, the survival of this foundation system can more plausibly be explained by considering the entire structural system. The ultimate capacity of the system will be reached at large lateral deformations of the individual piles; therefore, any degradation of lateral soil resistance due to cyclic loading at lower load levels are not expected to affect the ultimate lateral capacity. In addition, the steel members in the structure, including the piles and jacket legs, are expected to have a higher yield strength than the nominal value used in design. Using expected lateral soil resistance and steel yield strength alone are enough to explain the
Figure 18. Increase in foundation system capacity of jacket platform in the hurricane loading direction due to higher lateral soil resistance (static versus cyclic p-y curves), increased steel yield strength (fY ) and modeling jacket leg stubs (adapted from Gilbert et al. 2010).
survival of this foundation in Hurricane Ike (Fig. 18). In addition, the jacket legs actually extend 3 to 4 m below the mudline, which increases the shear capacity of the system by forcing the plastic hinges in the piles deeper below the mudline. Accounting for this system effect further increases the capacity of the foundation system beyond the loading experienced in Hurricane Ike. Therefore, the performance this platform does not necessarily suggest that foundation designs are excessively conservative when the entire structural system is considered. A comparison of system effects for two jacket pile systems is shown in Figure 19. These figures were developed by assessing the sensitivity of system capacity to variations in the capacity of individual piles. The axial and lateral capacities of each pile were varied by ±30 percent from the design capacities to reflect possible variations in soil properties, installation conditions and structural capacities between piles. The maximum and minimum capacities obtained from this variation in any single pile are shown in Figure 19 to assess robustness of the pile system (i.e. the ability of the system to accommodate variations in component strengths). Both jacket pile systems are robust to variations in individual pile capacities for a shear-dominated failure (i.e., smaller overturning moments). These results complement those on Figure 18, where the pile system is more sensitive to structural factors than geotechnical factors for a shear-dominated failure. Conversely, both jackets are less robust to variations in individual pile capacities for an overturning-dominated failure (Fig. 19).Therefore, these jacket pile systems are much more sensitive to uncertainties in axial pile capacity than to uncertainties in lateral pile capacity. The three-pile system (Fig. 19a) is less robust in overturning than the six-pile system (Fig. 19b). For the same ±30-percent variation in axial capacity for individual piles, the three-pile system exhibits
© 2011 by Taylor & Francis Group, LLC
Figure 19. Robustness in pile system capacity for (a) a three-pile jacket and (b) a 6-pile jacket (b) (adapted from Chen et al. 2010).
about a ±25-percent variation in system overturning capacity while the six-pile system exhibits less than ±10-percent variation in system overturning capacity. This difference in system behavior is real. Both pile systems experienced hurricane loads near their system capacities. The three-pile system failed in overturning while the six-pile system survived in shear without detectable damage. This difference in system behavior is not explicitly reflected in design practice. The same design recipe is used to calculate the loads and required design capacities for individual piles, regardless of whether the piles are in a single-pile system (like a caisson or a riser tower), a three-pile system, or a six, eight or more pile system. Therefore, the resulting reliability for these various systems will not necessarily be consistent. 6 THE FUTURE The keys to better managing risk in the future on the frontier of offshore geotechnics are embodied in the preceding lessons and case histories. Achieving an appropriate level of risk requires that we pay as much attention to the consequences of failures as to the probabilities of failures and costs associated with reducing the probabilities of failure.
198
A thoughtful examination of the consequences of failures can also highlight useful means to reduce risks. It may be more cost-effective to reduce risk by mitigating consequences compared to increasing reliability. A fundamental understanding of what affects the performance of a geotechnical system, including both the loads and the capacities, is far more effective in engineering appropriate solutions than simply being conservative to manage risk. Realistic estimates for loads and capacities are needed as well as the “nominal” values used in design. Realistic estimates include what is most likely, how much uncertainty or variation there is, what the sources of uncertainty are (e.g., temporal variations, spatial variations, inference errors, model uncertainties, installation tolerances, human errors, etc.), and any lower or upper bounds that physically limit what is possible. Furthermore, being “conservative” can have unintended consequences that result in taking on greater risk. Difficulty installing a “conservatively” designed foundation that is larger or longer than usual may leave a facility with less capacity in the foundation than if a “less conservative” design had been used. There is ample potential to improve the benefitto-cost ratio for the information used in offshore geotechnics. Realizing this potential will mean focusing on acquiring data that will be most informative or have the greatest impact on project-specific decisions. This approach is in contrast to always getting similar information for each project. It requires a frontend investment to develop preliminary decisions and designs based on available information so that the value of additional information can be assessed. It requires developing decisions and designs that are adaptable so that they can be updated throughout their lifetime based on new information. It requires considering how information from an existing project may provide value in future projects. Adopting a system-wide perspective can pay large dividends in better managing risks. Designs need to implicitly consider how the performance of geotechnical components will affect the overall performance of the system. These system considerations are not necessarily captured in design codes and guidelines, and they are project specific. These system considerations extend beyond the physical facilities to the factors driving decisions on the project. An elegant and economical foundation design using new technology may not be useful if the contractor selected to install it is unwilling to use different technology. These system considerations require that geotechnical engineers interact closely and communicate clearly with other disciplines, from conceptual design through construction and operation. Finally, humans are an integral part of managing risk. Humans collect and interpret the available information, make decisions and develop designs, construct and operate facilities, and are impacted directly and indirectly when failures occur. Since geotechnical engineers are not formally trained in psychology, sociology or communication, we need to reach out to
professionals in these non-technical areas in order to more effectively manage risk. 7
The intent of this paper was to provide case histories to guide practitioners in managing risk on the frontier of offshore geotechnics. The major themes are: 1. Achieving an appropriate risk requires balancing risk and conservatism. We need to become as involved in assessing and mitigating the consequences as we are assessing and minimizing the probabilities of failures, particularly in the context of the larger systems. 2. Managing risk requires understanding the loads on our designs as well as the capacities. We need to be careful to convey realistic ranges of possibilities and to avoid inadvertently compounding conservatisms. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. We need to identify where information will be and will not be most valuable on a project-specific basis, develop designs and decisions that can readily be adapted based on new information, and convey uncertainty as clearly as possible to the decision makers. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. We need to interact closely with other disciplines, from conceptual design through construction and operation. The use of risk and reliability principles has become an important part of offshore geotechnics, especially on the frontier. Better communication between all parties involved in design, construction and operation and earlier application of these principles in the life cycle of a project will enhance the practical value of the principles. ACKNOWLEDGMENTS We would like to acknowledge the following organizations that have directly or indirectly supported the material in this paper: US Minerals Management Service, American Petroleum Institute, National Science Foundation, Offshore Technology Research Center, BP, ExxonMobil, Sonaghal, Chevron, Shell, and numerous other operators and contractors that constitute the offshore oil industry.The views and opinions presented herein are ours alone and do not necessarily reflect those of any sponsor or employer. REFERENCES AIChE. 1989. Guidelines for Chemical Process Quantitative Risk Analysis. Center for Chemical Process Safety of the American Institute of Chemical Engineers. New York, New York.
199 © 2011 by Taylor & Francis Group, LLC
CONCLUSIONS
ANCOLD. 1998. Guidelines on Risk Assessment, Working Group on Risk Assessment. Australian National Committee on Large Dams. Sydney, New South Wales, Australia. Bea, R.G. 1991. Offshore platform reliability acceptance criteria. Drilling Engineering, Society of Petroleum Engineers. June Issue, 131–136. Chen, J-Y., Gilbert, R.B., Murff, J.D.,Young, A. and Puskar, F. 2010. Structural factors affecting the system capacity of jacket pile foundations. Proceedings of International Symposium on Frontiers in Offshore Geotechnics. in press. Gambino, S.J. and Gilbert, R.B. 1999. Modeling spatial variability in pile capacity for reliability-based design. Analysis, Design, Construction and Testing of Deep Foundations, Roesset Ed., ASCE Geotechnical Special Publication No. 88, 135–149. Gilbert, R.B., Najjar, S.S., Choi, Y.J. and Gambino, S.J. 2008. Practical application of reliability-based design in decision making, Book chapter in Reliability-Based Design in Geotechnical Engineering: Computations and Applications, Phoon Ed., Taylor & Francis Books Ltd., London. Gilbert, R.B., Chen, J.Y., Materek, B., Puskar, F., Verret, S., Carpenter, J., Young, A. and Murff, J.D. 2010. Comparison of observed and predicted performance for jacket pile foundations in hurricanes. Proceedings of Offshore Technology Conference. OTC 20861. Gilbert, R.B., Tonon, F., Freire, J., Silva, C.T. and Maidment, D.R., 2006. Visualizing uncertainty with uncertainty multiples. Proceedings, GeoCongress 2006, ASCE, Reston, Virginia. Gilbert, R.B., Ward, E.G. and Wolford, A.J. 2001. A comparative risk analysis of FPSO’s with other deepwater production systems in the Gulf of Mexico. Proceedings of Offshore Technology Conference. OTC 13173.
Goodwin, P., Ahilan, R.V., Kavanagh, K. and Connaire, A. 2000. Integrated mooring and riser design: target reliabilities and safety factors. Proceedings, Conference on Offshore Mechanics and Arctic Engineering. 185–792. Murff, J.D. & Wesselink, B.D. 1986. Collapse analysis of pile foundations. 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France: 445–459. Najjar, S.S. 2005. The Importance of Lower-Bound Capacities in Geotechnical Reliability Assessments. Ph.D. Dissertation, The University of Texas at Austin, 347 pp. OTRC. 2006. Reliability of Mooring Systems for Floating Production Systems. Final Report for Minerals Management Service, Offshore Technology Research Center, College Station, Texas, 90 pp. OTRC. 2008. Mudslides During Hurricane Ivan and an Assessment of the Potential for Future Mudslides in the Gulf of Mexico. Final Project Report, Offshore Technology Research Center. Prepared for US Minerals Management Service. 190 pp. Stahl, B., Aune, S., Gebara, J.M. and Cornell, C.A. 1998. Acceptance criteria for offshore platforms. Proceedings of Conference on Offshore Mechanics and Arctic Engineering. OMAE98-1463. Tang, W.H. and Gilbert, R.B. 1993. Case study of offshore pile system reliability. Proceedings of Offshore Technology Conference. 677-686. USBR. 2003. Guidelines for Achieving Public Protection in Dam Safety Decision Making. Dam Safety Office, United States Bureau of Reclamation. Denver, Colorado. USNRC. 1975. Reactor Safety Study: An Assessment of Accident Risks in US Commercial Nuclear Power Plants. United States Nuclear Regulatory Commission, NUREG75/014. Washington, D. C. Tufte, E.R. 1990. Envisioning Information, Graphics Press. Cheshire, Connecticut.
200 © 2011 by Taylor & Francis Group, LLC
2 Geohazards and gas hydrates
© 2011 by Taylor & Francis Group, LLC
Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7
Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney The University of Western Australia, Perth, Australia
ABSTRACT: Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and reverse movement on reactivated normal faults; and (3) the Murchison region has clear evidence of Quaternary tectonic deformation, repeated surface rupturing events, and a record of two large magnitude historical earthquakes. The evidence of Neotectonic deformation in northwestern Australia indicates that a number of seismic sources are present and these sources have the potential to produce moderate to large magnitude earthquakes such as the Mw 7.1 Meeberrie event. Future seismic hazard assessments should implement refined seismic source models that treat distinct seismic sources and the epistemic uncertainty associated with those sources. These seismic sources should be considered in the selection and engineering of seabed infrastructure such as manifolds, flowlines, export pipelines, and anchorage systems, as well as onshore LNG plants and port facilities. 1
INTRODUCTION
Western Australia is commonly viewed as a “Stable Continental Region” (SCR). It is largely composed of Archean age terranes, such as the Yilgarn and Pilbara cratons, Proterozoic and Phanerozoic age basins such as the Kimberly, Canning, Carnarvon, and Eucla Basins, and intervening deformed belts such as the Albany Frazier, Capricorn, King Leopold, and Halls Creek orogens. There is no orogenesis (active mountain building) occurring in WesternAustralia (WA) and the landscape is severely weathered with deep regolith attesting to long term stability on a regional scale (Anand and Paine, 2002). However, the occurrence of large magnitude historical earthquakes such as the 1885 ML 6.6 Mt. Narryer, 1941 ML 7.1 Meeberrie, and 1967 ML 6.7 Meckering events, and geomorphic evidence of crustal deformation and fault scarps from previous earthquakes indicate that parts of WA are being actively deformed. EPRI (1994) established criteria to define SCRs, which include: (1) evidence for no tectonic activity younger than early Cretaceous (∼100 million years before present {Ma}); (2) no deformed forelands or orogenic belts younger than Cretaceous (∼65 Ma); (3) no anorogenic intrusions younger than Cretaceous; and (4), no rifting or significant extension younger than Paleogene (∼35 Ma). SCRs are divided into domains composed of extended and non-extended crust, and domains underlain by extended crust are further divided into continental margins and failed rifts. According to EPRI (1994), SCRs that are underlain by extended crust have greater seismogenic potential than
those underlain by non-extended crust.Although Western Australia broadly meets these criteria, regional tectonic warping as evidenced by coastal submergence in the north, folding, the presence of numerous fault scarps, and the occurrence of several large magnitude earthquakes suggest that tectonic deformation is occurring across parts of WA. Active crustal dynamic processes occurring in SCR’s remain an enigma within the earth sciences. Understanding these processes is important for characterizing the location, severity, and frequency of earthquake occurrence, and assessing potential triggers for submarine landslides, liquefaction, and site response for structural design of both offshore and onshore facilities. 2
The northern margin of the Australian Plate is involved in a complex collision with the Sunda and Philippine Sea plates (Figure 1). Relative motion of the Australian, Sunda, and Philippine Sea plates is constrained from repeated geodetic surveys that use Global Positioning System (GPS) satellites to measure the precise positions of survey points located throughout a region. The repeated surveys provide direct measurements of the rates and directions of motion of points on different tectonic plates. Reoccupation of GPS sites in the Pacific, Australia, Indonesia, and Southeast Asia between 1991 and 2003 indicate that: (1) theAustralian Plate is moving along an azimuth of 015◦ and is converging with the Sunda Plate/Banda Arc at a rate of
203 © 2011 by Taylor & Francis Group, LLC
REGIONAL TECTONIC SETTING
Figure 1. Regional tectonic setting showing major tectonic structures and relative motion vectors of tectonic plates.
67 to 75 mm/yr relative to a fixed Eurasian reference frame (Figure 1) (Bock et al., 2003; Nugroho et al., 2009); (2) the Philippine Sea plate is moving northwestward at a rate of approximately 110 mm/yr relative to Eurasia (DeMets, et al., 1994); and (3) the Sunda Plate is moving east-northeastward at a rate of about 7 to 11 mm/yr relative to Eurasia (Simons et al., 2007). The northern boundary of the Australian Plate follows the Sunda Arc subduction zone and the Banda Arc collision zone (Figure 1). The Australian Plate consists of two main parts including Australian continental crust and oceanic crust of the Indian Ocean, and the differences in crustal type control the nature of processes occurring along the plate boundary. Subduction, in the past, has occurred along the entire Sunda and Banda plate boundary, but now is limited to that part of the plate boundary where oceanic crust of the Indian Ocean is colliding with continental crust of the Sunda Plate (McCaffrey and Nabelek, 1984). Here, the thinner, denser oceanic crust is subducted northward beneath the thicker, less dense continental crust. Subduction extends from the Andaman Islands in the northwest to approximately 120–121◦ east longitude near the island of Flores (Audrey-Charles, 1975; Karig et al., 1987). East of this location, the oceanic crust of the Indian Ocean already has been consumed and subduction of oceanic lithosphere has ceased (Silver et al., 1983; Genrich et al., 1996). The plate boundary from Flores to East Timor is now characterized by collision of the Australian continental crust with fragments of the former island arc (e.g. Sumba, Rote, and Timor islands) and accretion of those fragments to the Australian Plate. The main deformation front now occurs along the northern side of the island arc on a system of south-dipping north-verging reverse faults that are referred to as the Bali-Flores and Wetar thrusts (McCaffrey and Nabelek, 1984). The formation of a south dipping thrust system on the north side of the island arc and similarities in motion vectors for both Australia and Timor (Genrich et al., 1996), indicates that the former island arc is being accreted to the
Australian Plate and the subduction zone has reversed polarity (compared to the Sunda Arc to the west). The collision of the Australian continental crust along the southern Banda Arc has caused profound changes in the style of deformation including cessation of north-directed oceanic subduction, accretion of the former island arc, and reversal of subduction polarity along the Flores and Wetar thrusts (east of 120◦ east). The collision of the Australian continental crust with the South Banda Arc also is causing warping and deformation of northern and northwestern Australia. Therefore, although the main active plate boundary lies on the north side of the Banda Arc, northern Australia is responding to the effects of the collision. We speculate that the transition from oceanic subduction to continental collision at approximately 120◦ east longitude is generating stresses in the Australian crust that are a source of strain energy for earthquakes observed in northern WA. Furthermore, although the main deformation front now lies on the north side of Timor, structures along the Timor trough may still be active earthquake sources. 3 TECTONIC DEFORMATION IN A STABLE CONTINENTAL REGION 3.1 Tectonic flexure Northeast directed convergence of the Australian Plate with the Banda Arc is causing downward flexure of the continental lithosphere in response to the collision. The flexure is most evident along the Kimberly and has formed an anomalously wide continental shelf and sinuous shoreline morphology consistent with crustal subsidence and coastal submergence. The continental shelf, shown on Figure 2, is over 500-km wide (at the position of East Timor) and decreases to less than 100-km wide near Exmouth (Cape Range). From Exmouth southward along the West Australian coast the continental shelf is typically between 40- and 100-km wide.
204 © 2011 by Taylor & Francis Group, LLC
Figure 2. Digital elevation model of the northwest Australian continental shelf. Note wide shelf between Australia and Timor and narrowing of the shelf to the southwest. Elevation data from Geosceince Australia (2002).
elevations to sea-level high and low-stands. Figure 3 shows a strong correlation of surface elevations to the sea-level curve and using the maximum ages predicts minimum subsidence rates of 0.22 to 0.25 mm/yr for the shelf area south of the Timor trough (Table 1 and Figure 4). The hinge line of this flexure lies south of Broome, where coastal morphology changes from wave erosional (stable) to sinuous (drowning), and last interglacial (Marine Isotope Stage 5e) shoreline deposits (circa 120–130 ka) are present to the south and appear to be absent to the north. If the last interglacial deposits are now submerged the minimum coastal subsidence rate in the vicinity of Broome would be approximately 0.05 mm/yr. This indicates that the tectonic flexure associated with Australia’s collision with the Banda Arc diminishes from north to south and reaches zero north of the Cape Range where coastal uplift is observed and Last interglacial marine units are at their expected heights (Kendrick et al., 1991). The tectonic flexure documented by subsidence of geomorphic surfaces indicates that the continental lithosphere of northwestern Australia is being actively deformed due to the collision with the Banda Arc. This deformation will cause strain in the crust, and therefore the area of tectonic flexure may have greater seismogenic potential than areas not undergoing flexure.
Figure 3. Sea-level prediction diagram showing correlation between surface elevations and sea-level stages. From top to bottom, geomorphic surface elevations are −25 m, −40 m, −50 m, −70 m, −85 m, and −125 m. Elevation from Van Andel & Veevers, 1965. Sea-level curve from Lambeck and Chappell, 2001. Upper and lower curves indicate uncertainty.
The anomalously wide shelf in the north is inferred to be the result of marine erosion during progressive tectonic subsidence. Van Andel & Veevers (1965) recognized a series of submergent shoals, shelfs and terraces and postulated that these could have formed through a combination of tectonic subsidence and sealevel fluctuations. However, at the time, the theory of plate tectonics had not yet emerged, data on sea-level fluctuations were sparse, and thus it was not possible to document the nature and rate of deformation. Van Andel & Veevers (1965) recognized six main submergent surfaces at elevations of −120 to 140 m (at the shelf break), −85 m, −70 m, −50 m, and −40 m (along the continental shelf), and −25 m at a series of submergent atolls on the shelf break (Karmt Shoals) (Figure 2).The geomorphic surface elevations are plotted on a sea-level prediction plot (Figure 3), which is tentatively used to correlate the geomorphic surface
3.2 Folding and faulting in the Cape Range The 320-km long section of coast from the Cape Range to Cape Cuvier exhibits evidence of Neogene uplift and folding.The folding and uplift have strong regional expression, giving rise to the Cape Range, Cape Rough and Giralia Ranges, as well as the Lake MacCleod basin. Each of these ranges is mapped as an anticlinal
205 © 2011 by Taylor & Francis Group, LLC
Table 1.
Correlation of geomorphic surface elevations and ages.
OIS stage
Sea-level (m)
Surface elevation (m)
Subsidence (m)
Age min. (ybp)
Age max. (ybp)
5e 5d 5b 3 3 2
3 −16 −30 −57 −75 −118
−25 −40 −50 −70 −85 −125
28 24 20 13 10 7
118000 106000 82000 54000 31000 19000
123000 108000 92000 60000 43000 28000
OIS = Oxygen Isotope Stage; ybp = years before present; m = metre
Figure 4. Subsidence rate diagram showing net subsidence and age values used in the rate calculation, and linear regression.
fold with intervening west-dipping reverse reactivated normal faults (Myers and Hawking, 1998). Evidence of uplift is spatially associated with the anticlinal structures. A series of at least four marine terraces occur along the west coast of the 120-km long N-NE trending Cape Range anticline. The presence of these elevated shoreline deposits indicates long-term emergence of the coast and active fold growth. The growth of the Cape Range anticline implies crustal shortening and movement along associated thrust faults. Coastal uplift also is expressed in the anomalous height of Pleistocene marine deposits exposed in coastal sections west of Lake MacLeod. Uplift of this section of coast appears associated with anticlinal structures, which has prevented a number of rivers from reaching the coast, and resulted in formation of the Lake MacLeod evaporate basin. In the Lake McLeod area, folding of Miocene and younger sedimentary deposits has formed the Gnargoo Range and diverted the Lyndon and Minilya rivers around the resulting anticlines (Figure 5). Development of these supercedent streams may indicate Quaternary uplift rates in excess of incision rates and can provide constraints on rates of fold growth and associated fault slip rates.
Figure 5. Neogene folding at Lake MacCleod. Black lines are fold axes. Streams flow around the noses of folds (supercedent) indicating fold growth controlled stream position. Modified from GSWA, 1985.
The crustal shortening and fold growth is likely associated with reverse movement along former extensional structures and is more pronounced in this region than elsewhere in the Carnarvon Basin. The geomorphic evidence for reactivation of former extensional structures indicates that these are potential seismogenic sources that should be explicitly considered in seismic hazard assessments. These structures follow the continental shelf break and cross much of the Exmouth Plateau (Myers and Hawking, 1998) and therefore may be near-field sources of ground shaking for offshore facilities. 3.3
The Mt. Narryer fault zone in central WA is located near the northwest margin of the Yilgarn craton and the eastern margin of the Southern Carnarvon Basin
206 © 2011 by Taylor & Francis Group, LLC
Surface faulting and historical earthquakes
Figure 6. Digital terrain model (30 m digital elevation data reproduced by permission of the Western Australian Land Information Authority, 2010) showing oblique view to northeast along the Mt. Narryer fault scarp at the Roderick River. Arrows point to fault scarp and line shows position of topographic profile on Figure 7.
Figure 7. Topographic profile across the Mt. Narryer fault showing offset of alluvial valley surface (2009 data reproduced by permission of Western Australian Land Information Authority, 2010).
(Williams, 1979). The fault is approximately 120-km long and strikes in a northeast direction. Historical reports of earthquake strong ground shaking suggest that the fault zone may have produced the 1885 ML 6.6 Mt. Narryer earthquake (Clark, 2006). The epicenter for the 1941 ML 7.1 earthquakes also is located near the fault zone and so it is likely that the Mt. Narryer fault zone has produced two historical large magnitude earthquakes. The 1941 ML 7.1 Meeberrie earthquake is the largest earthquake to have been recorded in Australia. The Mt. Narryer fault zone includes at least four left-stepping en-echelon fault segments. From north to south the segment lengths are 11 km, 33 km, 40 km, and 35 km. The northern fault segments are expressed by strong vegetation alignments and fault scarps on the order of 1 to 1.5 m high (Clark, 2006).The linear nature and subvertical dip of the northern scarps suggests a significant strike slip component of motion. The two southern segments of the fault zone are expressed by a west-side up reverse sense of displacement and have formed east facing scarps across the Roderick and Sanford river alluvial valley deposits (Williams et al., 1983; Myers, 1997; Clark, 2006). Analysis of imagery and digital terrain models indicate that the scarps across the Sanford and Roderick rivers have
captured and diverted active stream flow, formed sag ponds, and impounded Lake Wooleen.The alluvial surfaces in both valleys are uplifted, warped and incised (Figures 6 and 7). Folding in the hanging wall of the fault zone has caused uplift and abandonment of the main river channel and incision of the river through the fold. Where the river cuts through the fold the channel pattern changes from braided to incised. Scarp heights of 3 to 8 m (Figure 7) suggest that the fault has experienced multiple surface rupturing events in Quaternary time. Preliminary analysis of drainage patterns and stream profiles west of the Mt. Narryer fault zone suggest that additional fault scarps may be present. 4
Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and movement on reactivated
