LANDMARKS IN EARTH REINFORCEMENT VOLUME 1
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PROCEEDINGS OF THE INTERNATIONAL, SYMPOSIUM ON EARTH REINFORCEMENT FUKUOKA / KYUSHU / JAPAN / 14-16 NOVEMBER 2001
Edited by:
HIDETOSHI OCHIAI Kyushu University, Fukuoka, Japan
JUN OTANI Kumamoto University, Kumamoto, Japan
NORIYUKI YASUFUKU Kyushu University, Fukuoka, Japan
KIYOSHI OMINE Kyushu University, Fukuoka, Japan
VOLUME 1
/ EXTON(PA) / TOKYO A.A. BALKEMA PUBLISHERS LISSE/ ABINGDON
IS
us
1
Under the auspices of the Japanese Geotechnical Society and the Technical Committee for Geosynthetics and Earth Reinforcement of ISSMGE (TC-9)
Copyright 0 2001 Swets & Zeitlinger B.V., Lisse, The Netherlands All rights reserved. No part of this publication may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, photocopying, recording or otherwise, without the prior written permission of the publisher. Published by: A.A. Balkema, a member of Swets & Zeitlinger Publishers www.balkema.nl and www.szp.swets.nl For the complete set of two volumes, ISBN 90 265 1863 3 For Volume 1, ISBN 90 265 1864 1 For Volume 2, ISBN 90 265 1865 X Printed in the Netherlands
Table of contents
Preface Organization 1 Testing and materials Recommendations for the application of geosynthetics in a railway network modernisation project in Slovakia R. Baslik, M. Matys & L. Turinic
3
Clay-cement mix with reinforcing fibres for diaphragm walls A. Brinkmann, M. Benz, F. Bucher & P. Amann
7
Design and development of inclined plane test on geosynthetics A. Cancelli, P. Rimoldi, A. Moroni &A. Poltronieri
13
Grout injection in the laboratory C. Dano & N. Derache
21
A research on the tensile resistance of wired rope anchor H. Fujimura & Y. Taniguchi
27
Experimental evaluation of the factors affecting pull-out test results on geogrids V.N. Ghionna, N. Moraci & P. Rimoldi
31
Residual strength and its application to design of reinforced soil in seismic areas J.H. Greenwood, C.J.F. P. Jones & F. Tatsuoka
37
Safe and economical soil reinforcement using a new style geogrid G. Heerten, R. Floss & G. Brau
43
Installation survivability of flexible geogrids as earth reinforcement materials C. Hsieh & C.K. Lin
49
Effects of geogrid properties on pullout resistance J. Izawa, Y. Ishihama, J. Kuwano, A. Takahashi & H. Kimura
55
Geotechnical behavior of fiber reinforced fly ash S.R. Kaniraj & V. Gayathri
61
Partial factors for geosynthetics specific to limit state approach A.J. Khan
67
V
Mechanical properties in short-fiber mixture stabilized volcanic cohesive soil M. Kudo, H. Ochiai & K. Omine
73
Progressive pullout failure of geosynthetic reinforcement J. Mak & S-C.R. LO
77
Strain-induced toughness and shearing characteristics of short-fiber reinforced soils K. Makiuchi & K. Minegishi
83
Modelling the behaviour of geosynthetic reinforcements used to resist combined sustained and shock loading J. Kupec & A . McGown
89
Testing related to the introduction of a new geogrid with welded flat bars for use as a soil reinforcement G. Heerten, E. Reuter, A. McGown & J. Kupec
95
Prediction of failure stress of reinforced residual soil by simplified approach S.A. Mofiz & M.R. Taha
101
Effect of connection bar stiffness on failure strength of connected steel grid reinforcements Y. Nabeshima, S.G. Zhou, T. Matsui & N. Sakata
107
Improvement effect of composite geomaterial by utilization of plastic wastes K. Omine, H. Ochiai, N. Yasufuku, M. Yamamoto & Y. Inoue
111
Visualization of interaction behavior between soil and reinforcement using X-ray CT J. Otani, T. Hirai, K. Miyanzoto & T. Mukunoki
117
Laboratory testing of long-term performance of clay-geogrid interaction A. Pamuk, D. Leshchinsky, V.N. Kaliakin & H.I. Ling
121
Strength behaviour of lateritic soils randomly reinforced with jute fibre B.R. Phani Kumar & M.V.B. Ramana Sastry
127
Studies on geotextile/ soil interface shear behavior M. Salehi
131
Discussion of safety from a study of the creep rupture of polyester P. Segrestin & P. Orsat
135
Modelling of reinforced unpaved composite material using HISS model K.G. Sharma, K.K. Gupta & P. Agganval
141
Construction and monitoring of geotubes E.C. Shin, Y.I. Oh, B.M. Das & E.S. Lee
147
Modelling and instrumentation considerations of a geogrid B. V.S. Viswanadhain
153
Variation in creep rate at constant loading of PET geogrid strapping W. Voskamp & F. van Vliet
159
Residual strength of PET after more than 12 years creep loading W. Voskamp, F. van Vliet & J. RetzlafS
165
Static and dynamic strength of cement mixed soil reinforced by fibers S. Yasuda, H. Ishinabe & Y. Murasawa
171
Practical experience in small scale pullout tests H. Zanzinger, E. Gartung & G.L. Sivakumar Babu
177
2 Embankments Railroads on piled embankments in Germany: Milestone projects D. Alexiew & W. Vogel VI
185
Final design of an overbridging for railways endangered by cavities at Groebers W. Ast, J. Sobolewski & J. Haberland
191
A 12-m high geosynthetic-reinforced residual soil slope S.H. Chew, G.P. Karunaratne, S.A. Tan, Y.T. Seah, C.T. Ho, S.K. Lim, W.H. Ho, K.Q. Ho & J. Wei
197
Use of inclined test to assess stress mobilization of liner on slope J.P. Gourc, P. Villard, R.R. Ramirez, N. Feki, L. Briaqon & H. Girard
20 1
Performance of geotextile reinforced slopes subjected to seepage flow A. Hiro-oka, M. Kobayashi, H. Nagase, K. Shimizu & H. Fujiwara
207
Drainage effect of geosynthetics in high and very wet embankment S. Ito, Y. Yokota, M. Inagaki, A. Morikage, Y. Kumagai & K. Kawamura
213
A full-scale field trial of electrokinetically enhanced cohesive reinforced soil using electrically conductive geosynthetics C.J.F.P. Jones & R.C. Pugh
219
High embankment of clay reinforced by GHD and its utilities M. Kamon, T. Akai, A. Matsumoto, S. Suwa, M. Fukuda, T. Simonodan, S. Yanagihara, Y. Nambu, K. Iwata & M. Matsushita
225
Tension in geosynthetic liner based on hyperbolic interface response K.V.S. Krishna Prasad, M.R. Madhav, J. Kodikara & M. Bouazza
23 1
Trial construction of arching structure by using large-sized soilbags T. Kubo, Y. Yokota, S. Itou, L. Sihong & H. Matsuoka
235
Study on the critical height of fiber-reinforced slope by centrifuge test G.X. Li, Y.X. Jie & G.Z. Jie
239
Response of geosynthetic reinforcement to transverse force M.R. Madhav, H.B. Poorooshasb & N. Miura
243
Model experiment and analysis of sandwich earth fill reinforced with geosynthetics H . Nagashimn, Y. Tanabashi, H. I, N. Fujise & H. Nakahara
247
Reinforcing function of a liner system by geotextile and geogrid S. Nakamura, S. Inzaizumi & K. Kuzumaki
253
The design of steep slopes constructed from cohesive fills and a geogrid P.J. Naughton, R.A. Jewel1 & G.T. Kempton
259
Numerical analysis of reinforced embankments on soft soils C.T. Sa, E.M. Palmeira, L.M.A. Dellabianca & A.R.S. Fahel
265
Trial construction of the reinforced river dike and its performance K. Sawada, A. Yashima, Y. Sato, Y. Fujita, H. Maeda, N. Matsumoto & A. Hazama
27 1
Analysis of reinforced slopes and walls using horizontal slice method M. Shahgholi, A. Faklzer & C.J.F.P. Jones
277
Geogrid-reinforced road embankment over an old dump G. Stolarski & E. Gartung
28 1
Loading test of earth flow prevention embankment reinforced with geosynthetics N. Tatta, Y. Yokota, S. Ito, T. Kubo & K. Arai
287
Deformation analysis of PLPS GRS bridge pier during construction and in service T. Uchimura, M. Shinoda, M.S.A. Siddiquee & F. Tatsuoka
293
A diagrammatic evaluation of geo-composites for reinforcing cohesive soils K. Yasuhara, S. Murakami, C. Ghosh & J.R. Molina
299
Design of geosynthetic-reinforced veneer slopes J. G. Zornberg, S. Somasundaram & L. LaFountain
305
VII
3 Wall structures The performance of buried galvanized steel earth reinforcements after 20 years in service P. L. Aizderson & J , Saizkey
313
Performance of Mechanically Stabilized Earth walls over compressible soils R.A. Blooinfield, A.F. Soliman & A. Abraham
3 17
Flexible facing systems for reinforced soil wall structures - characteristics and performance M. Boyd & P. Segrestin
323
Shaking table and numerical modelling of reinforced soil walls M.M. El-Emam, R.J. Bathurst, K. Hatami & M.M. Mashhour
329
Reinforced earth ramps over flexible inclusions in Beirut J. B. Esta
335
Construction of a geogrid -reinforced earth-wall inside a ware-house F. Vie1
34 1
Bench-type wall with flat slabs and steel bars M. Fukuoka, K. Kondo, H. Kawahara, R. It0 & K. Misawa
345
Dynamic behavior of multi-anchored reinforced soil wall in large-scale shear box M. Futaki, N. Aoyama, K. Misawa, T. Konami, M. Sato, T. Tatsui & K. Mikanzi
35 1
Stability test of the multi-anchored reinforced soil wall constructed on soft ground H. Hashimoto, N. Aoyama, H. Miyatake & M. Hirasawa
359
Dimension analysis on reinforced soil walls by finite element method G. Huang, X. Yu & C. Yaizg
363
Analyses of a near-fault geosynthetic-reinforced modular block wall damaged during the 1999 Chi-Chi earthquake C.C. Huang, L.H. Chou & Y.H. Chen
369
A contribution to the design of flexible wire mesh facing J.-M. Jailloux, N. Freitag & P. Segrestin
375
A field instrumentation and FEM analysis for an isolated-reinforced earth wall Y. Y. Kim, K.J. Han & K.M. Kim
38 1
Limit analysis of soil structures subjected to constraints by reinforcement S. Kobayaslzi, A. Tanaka & T. Tamura
387
Seismic earth pressures acting on reinforced-soil and conventional type retaining walls J. Koseki, K. Wataizabe, M.Tateyama & K. Kojinza
393
Influence of reinforcement's inclination on bearing capacity of RS wall M. Kulczykowski
399
Effect of facing and construction sequence on the stability of reinforced soil wall Y. Lim, J. Jung, Y. Park & Y. Suh
405
Interface friction coefficient of extensible reinforcement and its influence on designing of retaining structures P. Michalski & K.M. Skarzynska
41 1
Reliability analysis of geosynthetics reinforced soil wall Y. Miyata, S. Shigehisa & K. Kogure
417
Experimental research of reinforced soil wall for rock-fall protection T. Nomura, S. Inoue, M. Fuchigami, Y. Obata, Y. Yokota, T. Kubo & K. Arai
42 1
Numerical analysis of a sheet pile mooring wharf having several tie rods S. Ohmaki, K. Saeki & M. Kiyozumi
425
VIII
Relation between wall displacement and reinforcement for reinforced retaining wall K. Okabayashi, K. Tagaya & M. Kawamura
429
A fully synthetic connection of polyester based strip reinforcement to concrete panels. Development, tests and first application P. Orsat & N. Freitag
433
Mechanical behaviour of soil reinforced by geocells N. Racana, R. Gouwks & M. Grkdiac
437
An analytical approach to compute design loads of reinforced earth embankments M. V. Ratnam & M.B. Rao
443
Evaluation of seismic performance in Mechanically Stabilized Earth structures J.E. Sankey & P. Segrestin
449
Use of a l g geogrid reinforced-wall model to evaluate the effectiveness of a FE numerical code M. Schiavo, P. Simonini, G. Gottardi & L. Tonni
453
Seismic stability of preloaded and prestressed reinforced soil structure against strong shaking M. Shinoda, T. Uchimura, F. Tatsuoka, M. Tateyama & T. Natsuki
459
Seismic behaviour of earth reinforcement walls R.A. Sofronie, C.A. Taylor & F. Iosif
465
Numerical analysis of soil nailed retaining wall B.R. Srinivasa Murthy, G.L. Sivakumar Babu & A. Srinivas
473
Ultra-high hybrid wire and concrete-faced Mechanically Stabilized Earth bridge abutments K.M. Truong, M.J. Berkebile & R.A. Gladstone
477
Combined reinforcement by means of EPS blocks and geogrids for retaining wall structures Y. Tsukamoto, K. Ishihara, H. Nakazawa, H. Kon, T. Masuo & K. Hara
483
Irregular shaking table tests on seismic stability of reinforced-soil retaining walls K. Watanabe, M. Tateyama, K. Kojima & J. Koseki
489
Issues in the use of clay in reinforced earth construction L.D. Wesley
495
Evaluation of confining effect in geogrid-reinforced retaining wall based on two dimensional model test N. Yasufuku, H. Ochiai, Y. Ninomiya, K. Omine, M. Nakashima & T. Kawamura
50 1
Applicability of an elasto-plastic model for reinforced soil structures T. Yoshida, K. Mori, A. Iizuka, F. Maegawa & S. Amano
507
4 Foundations The settlements of a continuous foundation footing resting on the geogrid-reinforced sand layer J. Adamczyk & T. Adamczyk
513
Field investigations on a soft ground of Bangladesh reinforced by granular piles M. Alaingir & S.M. Zaher
517
Bearing capacity considering stiffness of reinforcement material K. Arai, M. Kamon, S. Nomura & Y. Yokota
523
Reinforcement of soft sub-grade for high-speed railroads using geocell S.D. Cho, J.M. Kim, M. Chung, S.H. Yoon & Y. Y. Kim
529
Behaviour of footings on reinforced sloped fills A. K. Choudhary & B. P. Verma
535
The capacity of reinforced subsoil loaded by uplifted foundation E. Dembicki & R. Duszynski
54 1
IX
Practical aspects of the design of deep geotextile coated sand columns for the foundation of a dike on very soft soils M. Geduhn, M. Ruithel & H.-G. Kempfert
545
Bearing characteristics of clay reinforced by a sandwiched geogrid-sand system H. Ghiussiun & M. Juhunniu
549
Multi-layered reinforced granular soil resting on soft soil - tension membrane effect C. Ghosh, K. Yusuhuru & M.R. Mudhuv
553
Prediction of the behaviour of a geogrid reinforced sloped fill under footing load C.T. Gnunendrun
559
Performance of geotextile-reinforced shallow foundations used in Bangladesh M.A. Huque, M. Alunzgir, M. Sulim & M.H. Kubir
565
Effects of tensile and bending rigidities of reinforcement in reinforcing soil structures and ground N. Kotuke, F. Tutsuoku, T. Tunuku, M.S.A. Siddiquee & C.C. Huurzg
57 1
Collapse loads on reinforced foundation soils R.L. Michulowski & X. Xin
577
Model loading tests on the footing reinforced with prestressed micropiles K. Miuru, Y. Tsukudu, Y. Otuni, M. Ishito & G.-L. You
58 1
Mechanical properties of soilbags and their applications to earth reinforcement H. Mutsuoku, S. Liu & K. Yumuguchi
587
Behavior of reinforced foundation under uplift and push-in loadings - model tests and analyses T. Nukui, T. Terunishi, M. Hiizokio & K. Aduchi
593
Characteristics of geogrid reinforced cohesive soil and its analytical method E. Ogisuko & K. Ryokui
599
Field loading test on the footing reinforced with prestressed micropiles Y. Otuni, K. Miuru, M. Ishito, G.-L. You & Y. Tsukudu
605
Jet grouting application for quay restoration in Tunisia A. Sfur & M. Bouussidu
61 1
Use of technologies based on soil reinforcement when trunk pipeline repair Y d . Spector, N.M. Rukhmutullirz, V.V. Novoselov & N. F. Shchepin
617
Analysis of two layer soil system beneath rigid footings - a global approach A. Sridhurun, B.R. Srinivusu Murthy & P. Vinod
619
Design on limit equilibrium of foundation reinforced with geosynthetics Z. Wung & X.Q. Wung
625
The application of ground reinforcing materials for the pile foundation M. Yumudu, N. Iwugumi, H. Ochiui, Y. Muedu & Y. Iguse
629
Analysis of improved ground with geonet reinforced stone columns Z. Zhou, Q. Zhung & J. Zheng
635
5 Soil nailings Simulation of soil nailing facing walls in finite element analysis S. Bung & W. Nyuz
64 1
Reinforcing mechanism of anchors in slopes and numerical verification by FEM F. Cui & K. Ugui
647
Nailing a deep excavation in soft soil with jacked in pipe inclusions W.L. Cheung, S.A. Tun, K. Y. Yong & G.R. Dusuri
653
X
Study on reinforcement method for seismic slope stability T. Fukumasa, H. Murakami, R. Nishihara, H. Kimura, M. Yamaura & S. Razavi Darbar
659
Stabilization and surface protection of steep slopes using soil nails and prefabricated concrete part linings A.C. Lottmann, L.E. Wichter & W. Meiniger
663
An example of a high soil nailed wall in plastic clayey soil B. MariC, P. Kvasnitka, D. Radaljac & R. Mavar
669
Stabilization of historical retaining walls using soil-nailing methods W. Meiniger, L.E. Wichter, E. Joppa & R. Loer
675
An analytical study to the vertically reinforcing soil slopes M. Moradi & M. Davari
679
Countermeasure against the slope using difference of shearing mechanism on main slip layer M. Mukaitani, M. Hori & N. Kobayashi
685
Optimum design of nailed soil slopes C.R. Patra & P.K. Basudhar
69 1
Innovative solution for a Kei Cutting problem G. V. Price, A. C.S. Smith & A. Muhajer
697
Numerical analysis of the mechanical behaviour of rigid shallow foundations on geo-reinforced soil strata C. di Prisco, S. Imposimato, P. Rimoldi & M. Vecchiotti
703
Slope stabilization with high-performance steel wire meshes in combination with nails and anchors R. Riiegger, D. Flum & B. Haller
707
Behavior of angle cut cylinder excavation by cut reinforced earth work method A. Sato, S. Tayama, K. Ogata, M. Takemoto & U. Tanaka
713
Seismic ductility of cut slope reinforced by soil nail A. Takahashi, J. Izawa, 0. Kusakabe, S. Tayama & M. Takernoto
719
The resistance of jacked-in pipe inclusions in soft soil S.A. Tan, W.L. Cheang, K. Y. Yong & G.R. Dasari
725
Case study on engineering behaviors of the Simajiri mudstone for dam construction - slope reinforcement in dam reservoir Y. Uchimura, Y. Otsuka, F. Motida, T. Nakamura, Y. Tamaki & H. Uehara
73 1
Kinematics and failure of soil-nailed excavation models in dynamic centrifuge tests M. Vucetic, M. Doroudian & J. Kocijan
737
Stabilization of slopes and landslides using soil nailing methods L.E. Wichter & W. Meiniger
743
Stability analysis of reinforced slopes considering progressive failure T. Yamagami, J.-C. Jiang, S. Yamabe & M. Taki
749
Field measurement and numerical analysis of soil nailing on volcanic cohesive soil K. Yamamoto, K. Tabata & R. Kitamura
755
Soil nail design with respect to Hong Kong conditions K.C. Ye0 & S.K. Leung
759 765
Author index
XI
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Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 02001 Swets & Zeitlinger, ISBN 90 265 1852 8
Preface
Earth reinforcing techniques are increasingly becoming a useful, powerful and economical solution to various problems encountered in geotechnical engineering practice. Expansion of the experiences and knowledge in this area has succeeded in developing new techniques and their applications to geotechnical engineering problems. In order to discuss the latest experiences and knowledge, and with the purpose of spreading them all over the world for further development, the IS Kyushu conference series on the subject of earth reinforcement have been held in Fukuoka, Japan, every four years since 1988. The symposia were titled "Theory and Practice of Earth Reinforcement" for the first one in 1988, "Earth Reinforcement Practice" for the second one in 1992 and "Earth Reinforcement" for the third one in 1996. They have provided successful contributions towards the development of the earth reinforcement practice. This fourth symposium, entitled "Landmarks in Earth Reinforcement", is a continuation of the series of IS Kyushu conferences, and also aims at being one of the landmarks in the progress of the modern earth reinforcement practice. With this objective, one special lecture and five keynote lectures are arranged to deliver by internationally distinguished scholars, in addition to the presentations of the papers accepted for the symposium. A total of 212 abstracts from 32 countries were submitted for this symposium. The Scientific Committee carefully reviewed both the abstracts and full-length papers, and finally selected 137 papers for presentation during the symposium, some in technical sessions for oral presentations, and the rests in poster sessions. They were included in Volume 1 of the proceedings. The special session on "Soil Wailing" and the summary discussion session on "Design Procedure" have also been organized to summarize the results of fundamental and practical aspects on these areas developed in the recent years. These valuable reports will be included in Volume 2 of the proceedings to be published after the symposium. In the activities of the technical committees (TC) of the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE), the TC-9 covers the professional area on the subjects of geosynthetics and its applications to geotechnical engineering, and it had been called "Geotextiles and Geosynthetics" until 1997. But it was renamed as "Geosynthetics and Earth Reinforcement" in order to match the activities that are expected to develop in the future. This was due to the great successes of the previous IS Kyushu conferences. The report of TC-9 activities during the period of 1997-2001 will be included in Volume 2 of the proceedings. This symposium is being held under the joint-auspices of the Japanese Geotechnical Society (JGS) and the TC-9 of ISSMGE, and is being supported by the International Geosynthetics Society (IGS) and the Japan Society of Civil Engineers (JSCE). We would like to express our sincere gratitude to the members of the International Advisory Group for their encouragement and support to the symposium. It is our hope that this symposium will stand out as one of the notable landmarks in the progress of the modern earth reinforcement practice. Hidetoshi Ochiai Chairman of IS Kyushu 2001 XI11
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Landmarks in Earth Reinforcement, Ochiai et al (eds), 02001 Swets & Zeitlinger, ISBN 90 265 1852 8
ORGANIZATION
SYMPOSIUM COMMITTEE OF IS KYUSHU 2001 Prof. H.Ochiai (Chairman) Prof. S.Hayashi (Co-chairman) Prof. J .Otani*(Secretary General) Prof. G.Chen Prof. T.Esaki Mr. T.Fujii Dr. N.Fukuda Dr. H.Hazarika Prof. A.Hiro-oka Dr. H.Imanishi Prof. Y.Jiang Mr. K.Kasama Prof. S.Kato Mr. S.Kidera Prof. R.Kitamura Dr. N.Kotake Prof. T.Koumoto Prof. J.Kuwano Prof. Y.Maeda
Dr. K.Matsui Prof. Y.Mitani Dr. Y.Miyata Prof. N.Nagase Prof. M.Ohtsubo Dr. K.Omine Mr. Y.Saito Prof. K.Sato Mr. T.Sekiguchi Prof. Y.Tanabashi Mr. T.Tashima* Dr. H.Yakabe Prof. A.Yashima Prof. N.Yasufuku'K Prof. H.Yokota Prof. K.Zen
*Division Head
xv
SCIENTIFIC COMMITTEE OF IS KYUSHU 2001 Prof. H.Ochiai (Chairman) Prof. S.Hayashi (Co-chairman) Prof. J.Otani * (Secretary General) Prof. G.Chen Prof. T.Esaki Mr. T.Fujii Dr. N.Fukuda Dr. H.Hazarika Prof. A.Hiro-oka Dr. H.Imanishi Prof. Y.Jiang Mr. K.Kasama Prof. S.Kato Mr. S.Kidera Prof. R.Kitamura Dr. N.Kotake Prof. T.Koumoto Prof. JKuwano Prof. Y.Maeda
Dr. K.Matsui Prof. Y.Mitani Dr. Y.Miyata Prof. N .Nagase Prof. M.Ohtsubo Dr. K.Omine Mr. Y.Saito Prof. K.Sato Mr. T.Sekiguchi Prof. Y.Tanabashi Mr. T.Tashima'% Dr. H.Yakabe Prof. A.Yashima Prof. N.Yasufuku* Prof. H.Yokota Prof. KZen
*Division Head
INTERNATIONAL ADVISORY GROUP OF IS KYUSHU 2001 Prof. T.Akagi (Japan) Prof. A.S.Ba1asubramaniam (Thailand) Prof. R.J.Bathurst (Canada) Prof. M.Bouassida (Tunisia) Prof. F.Bucher (Switzerland) Prof. A.Cancelli (Italy) Ing. D.Cazzu€fi (Italy) Mr. R.K.S.Chan (Hong Kong, China) Prof. H.S.Chung (South Korea) Prof. E.Dembicki (Poland) Prof. R.Floss (Germany) Dr. A.Fourie (South Africa) Dr. E.Gartung (Germany) Dr. R.Gnanendran (Australia) Prof. J.P.Gourc (France) Prof. E.F.Guler (Turkey) Prof. C.J.F.P.Jones (U.K.) Prof. M.H.Kabir (Bangladesh)
Prof. G.P.Karunaratne (Singapore) Prof. S.K.Kim (South Korea) Prof. R.M.Koerner (USA) Dr. C.R.Lawson (Malaysia) Ir. C.Legrand (Belgium) Prof. D.Leshchinsky (USA) Prof. G.X.Li (China) Prof. M.R.Madhav (India) Prof. A.McGown (U.K.) Dr. Z.C.Moh (Taiwan R.O.C.) Prof. H.Ohta (Japan) Prof. E.M.Palmeira (Brazil) Prof. R.K.Rowe (Canada) Prof. J.Sun (China) Dr. F.Schlosser (France) Prof. ETatsuoka (Japan) Ir. W.Voskamp (The Netherlands)
XVI
ISSMGE/TC-9 (GEOSYNTHETICS AND EARTH REINFORCEMENT) CHAIRMAN Prof. H.Ochiai (Japan) SECRETARY Prof. J.Otani (Japan) CORE MEMBERS: Prof. J.P.Gourc (France) Prof. J.Kuwano (Japan) Mr. C.R.Lawson (Malaysia) Prof. D.Leshchinsky (USA) Prof. R.K.Rowe (Canada) MEMBERS: Dr. M.Abramento (Brazil) Dr. D.K.Atmatzidis (Greece) Dr. M.Bouassida (Tunisia) Prof. A.Cancelli (Italy) Ing. D.A.Cazzuffi (Italy) Dr. S.D.Cho (Korea) Dr. P.Delmas (France) Prof. E.Dembicki (Poland) Mr. G.Didier (France) KProf. R.Floss (Germany)
Dr. C .T.Gnanendran (Australia) Prof. E.F.Guler (Turkey) Prof. M.R.Hausmann (Australia) Dr. G.Heerten (Germany) Dr. S.Hermann (Norway) Prof. C.J.F.P.Jones (UK) Mr.C. Legrand (Belgium) Dr. M.D.G.A.Lopes (Portugal) Dr. M.Matys (Slovak Republic) Dr. 0.Murata (Japan) Prof. E.M .Palmeira (Brazil) Prof. L.Paredes (Chile) Dr. S.Paskauskas (Lithuania) Prof. G.V.Rao (India) Ms.Y.Rogbeck (Sweden) Prof. L.M.Timofeeva (Russia) Dr. S.Urie1 (Spain) Mr. W.Voskamp (Netherlands) Dr. A.Watn (Norway) Dr. K.C.Yeo (Hong Kong) Prof. J.G.Zornberg (U.S .A.)
ISSMGE/TC-9 SUPPORTING COMMITTEE OF THE JAPANESE GEOTECHNICAL SOCIETY CHAIRMAN Prof. H.Ochiai TC9 MEMBERS: Prof. J.Otani Prof. J.Kuwano Dr. 0.Murata MEMBERS: Dr. H.Abe Dr. N.Fukuda Prof. S.Hayashi Dr. T.Hirai Dr. M.Hirata Prof. S.Imaizumi Mr. K.Kojima Prof. K.Kumagai
Prof. K.Makiuchi Dr. H.Miki Dr. Y.Miyata Mr. H.Miyatake Mr. K.Nakamura Dr. J.Nishimura Mr. K.Ogata Dr. E.Ogisako Mr. Y.Otani Prof. T.Shimaoka Dr. S.Tayama Dr. T.Uchimura Mr. A.Yamamoto Prof. A.Yashima Mr. Y.Yokota
XVII
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1 Testing and materials
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Landmarks in Earth Reinforcement, Ochiai et al (eds), 02001 Swets & Zeitlinger, lSBN 90 265 1852 8
Recommendations for the application of geosynthetics in a railway network modernisation project in Slovakia R. Baslik Tectum Geosynthetic, Ltd., Bratislava, Slovak Republic
M. Matys Department of Engineering Geology, Faculty of Natural Sciences, Comenius University, Bratislava, Slovak Republic
L. Turinic Prodex, Ltd., Bratislava, Slovak Republic ABSTRACT: The paper deals with the use of geosynthetics in the railway network modernisation project. A new track subbase arrangement with stiff integral biaxial geogrids on a low bearing capacity soft subgrade is presented. The experimental tests have been carried out with the aim of investigating the effects of geogrid within reinforced subbase. In situ plate load bearing capacity control tests confirm the track substructure improvement using geosynthetic reinforcement.
1 INTRODUCTION
subbase layers in difficult foundation conditions. According to Table 1, in the design of reinforced subbase layers, it is recommended to take into consideration the maximum expected elongation of the geosynthetic reinforcement in the site. In dealing with the design according to STN 73 3041, the type of load, design life of the structure, the type of soil used in the subbase layer, the maximum elongation of geosynthetic reinforcement and its type, the position, number and the arrangement of the reinforcement should be assessed and reviewed. The bearing capacity and settlement (if required) of the reinforced subbase layer is reviewed or is determined experimentally. Finally, the overall (external) and internal stability of the designed reinforced sub-base layer is assessed.
For the application of geotextiles and geotextilerelated products in Slovakia, two basic standards (Slovak Technical Standard - STN) are used. STN 73 3040 introduces the selected technical parameters and to them appropriate required values for geotextiles and geotextile-related products. Geosynthetic products used in constructions of the Slovak Railway Authority must have permission from them. STN 73 3041 refers to the technical requirements of soil structures reinforced by geosynthetics. The standard introduces the requirement for the back fill soil, geosynthetic reinforcement and other materials, as well as the requirement for the facing of the soil structure reinforced by geosynthetics. The principle for the design according to the significance of the structure, and the adoption of different reduction coefficient values for the design of reinforced soil structures according to the I. and 11. limit states are presented. The designs are analogical with those of the European Standard drafts for geotechnics - ENV 1997 - 1 and others. STN 73 3041 presents the guide for the design of reinforced soil retaining walls and bridge abutments, reinforced steep slopes, embankments on soft subsoil and reinforced subbases. The paper will deal only with reinforced subbase.
3 GEOSYNTHETICS IN THE RAILWAY NETWORK MODERNISATION PROJECT 3.1 General The Central and Eastern European Countries joining the EU have made strong efforts to modernise their rail network. Impacted rail lines include two of the north-south and west-east TransEuropean corridors between Poland, Hungary and Ukraine, a corridor that runs across the west and north region of Slovak Republic. The new railway track will fulfil requirements for /1/ higher speed - 160 km/hr., /2/ higher load - 22.5 tons for axle, and /3/ higher traffic comfort. To fulfil these requirements, it requires not only new construction elements to the track superstructure, but also reconstruction of the track substructure. It was necessary to improve the bearing capacity of the railway subgrade. The current Slovak Railways Regulation for Railway Subgrade includes
2 REINFORCED SUBBASE ACCORDING TO STN 73 3041 The purpose of laying a reinforcement into the subbase is either to improve the strength and deformation characteristics of the subbase layer, to decrease its thickness, to bear the stress developed and its spreading and reducing differential settlement, or to create an alternative solution for the construction of
3
Table 1. Recommended maximum elongation of the geosynthetic reinforcement in the subbase layers. ,,,*,~, , , , , , , , , , , , ,, , , , , --,*
, , , -,* , , , , , ,, , , , , # -,--, , , , , , , ,
A,A*
Geotechnical category
, , , , *-, , , , , , ,-*, , , , ,en,w*,d*,,
, ,
Structure significance
Recommended maximum elongation of the geosynthetic reinforcement,
1.
Restricted
From 8 to 10
1. 2. 3.
Medium From 5 to 8 High From 3 to 5 Extraordinary From 1 to 3
-
E,,,,,
^
, , , , ,** , , , , , , , , , --*#,a,-*-,#,,*,I
~ # ,# -,~ * -,n ,# ,,~ * ,A ,,
, , , , ,, , , , A M , "" * * #
, , , , , #*, , , , * , I , n , * , # , * , n
, , , , ,
.,,, a,",
Examples of soil structures, where by, subbase layers are reinforced
(%I Simple, temporary structures with design life of 5 years (casual communications-roads, stock areas, working areas) Structures with design life between 5 to 25 years Structures with design life from 25 to 50 years Important structures with design life from 50 to 120 years (motorways, principal roads, railways, airports, bridge abutments, retaining walls, living houses, hydro-technical structures, industnal structures)
various types of soft subgrade improvementstabilisation such as lime, cement, chemical, and asphalt stabilisation only. The use of geosynthetic reinforcement in subbase is a new improvement method of track substructure on soft subsoil. The new recommendations of the Slovak Railway Authorities for the application of reinforcement techniques to the railway track foundation are based on the standards stated above, as well as the foreign experience analysis, laboratory and field tests and local experiences in the field of thin reinforced soil layers. According to the schedule of network realization the first section length of 50 km between Bratislava and Trnava began in 2000.
Table 2. Requirements for reinforcement application in truck substructure. ...,Proyerty _.._,.x.I...., ..x,.,...,.,,..I.....^.. ~
~
~
.... "I.,^,..,,... ..._.,....,... ~
Tensile strength (wide strip) Elongation (strain at failure) Junction strength (GRI:GG2 test method) Geogrid matrix stiffness (inplane torsion rigidity according to Kinney & Xiaolin test method) Agreement with application
Requirement > 30.0 kN/m < 15 % > 85 %
> 0.9 N.m/degree
Slovak Railways Authorities review and agreement prior to ap-
~plication#,,,~~_~_~_ , , , , ,
3.2 Subbase reinforcement
Reinforcing of the subbase provided to decrease an excavation soil volume. A small vertical distance between the rail level of existing line No.2 and a subgrade surface level of the reconstructed line No. 1 made possible to construct the new line No.1 without special support measures between lines. It was very important economical and time benefit. The only necessary there, was to reduce a speed to 30 km/hr. Geomat was installed on the slope as a erosion control system.
The subsoil investigation was performed to access strength and compressibility properties of subgrade. Extensive plate load bearing capacity and dynamic penetration tests were carried out. A cross section of the ground with bearing capacity values along the track axis was determined. The method based on bringing track grating down was used. The old track structure was excavated. After detailed geotechnical analysis of soft subsoil, a plate load bearing capacity test defines the modulus of deformation at the subgrade surface. A modulus of deformation approach based on in-situ tests in Melcice-Zlatovce track section (year 199697) and other empirical experiences were applied. Geosynthetic reinforcement used in the project shall conform to the properties of Table 2. These requirements fulfil the stiff integral biaxial geogrid (SIBG). One of the types of SIBGs was used to improve the performance of weak soils to support railway track foundation. The designer prepared a set of typical structure arrangements. Single or multi-layer geogrid reinforcement is used to increase the bearing capacity (synergistic effect) of the layer on the very soft subsoil. Onto the surface of soft subgrade the SIBG 30 x 30 kN/m was placed, as shown on Figure 1.
3.3 Nonwoven and woven geotextiles in track substructure The installation damage of nonwoven geotextile placed within the backfll material was found out.
Figure 1. Typical arrangement of the track superstructure, if E,, = I 1 to 21 MPa.
4
structure. Three kinds of soil were used in a series of tests. These were: gravel(G): particle size 2132 9mm mean particle size, D50 grading curve in Figure 3 classified as gravel poorly graded (GP) 2132 crushed stone(CS): particle size 10 mm mean particle size, D50 grading curve in Figure 3 classified as gravel poorly graded (GP) clay: moisture content: 37 % density: 15 84 kg.m-3 plastic limit: 27 % classified as high plasticity clay (CH) As reinforcing material a polypropylene SIBG was used. It has aperture size of 39 mm x 39 mm. With the purpose of checking the effectiveness of the geogrids, the cone was pushed through thin layer of soil, which was lying on geogrid. The test started with cone position of 50 mm above the geogrid. The test ended, when the cone intersected the starting geogrid plane and penetrated approx. 10 mm into the subgrade.
Properties of used geotextile were: mass per unit area 300 g/m2, CBR = 1620 N, tensile strength 8 kN/m, elongation at failure 150 %. Based on that situation and similar other experiences, Slovak Railway Authorities accepted new criteria defined by designer. The defined characteristics specify Slovak Railway requirements concerning the primary and secondary properties stipulated by the national standard STN 73 3040. Those stringent requirements on properties reflect in particular the method of application, use of crushed stone and cyclic load of railway traffic: Only nonwoven geotextile in the track substructure : Mass per unit area: min. 350 g/m2 CBR: min. 3.5 kN Short-term tensile strength: min. 15 kN/m Elongation at failure: max. 90 % Nonwoven geotextile under reinforcing geogrid: Mass per unit area: min. 250 g/m2 CBR: min. 2.5 kN Short-term tensile strength: min. 10 kN/m Elongation at failure: max. 115 % (recommended up to 90 %) Width of geotextile sheet is no less than 4.5m.
3.4 EJjcectiveizess of geogrid In order to verify the effectiveness of geogrid within subbase, one set of experimental tests was recently carried out at Comenius University in Bratislava. Figure 2 shows a sketch of model box developed to investigate the effect of the cone pushed into geogrid
Figure 2. Push test
Figure 4. Results of push tests.
5
Figure 5. Results of a plate load bearing capacity control tests.
moduli of deformation Esg were from 6.2 to 24.9 MPa. The regulation for new corridors requires a minimum modulus of deformation of 50 MPa on the subbase surface (Esb). The data presented in Figure 5 confirm that the designer succeeded in reaching the requested subbase bearing capacity with application of SIBGs. This type of geosynthetic, used in this manner in the new track substructure, can give significant technical, technological and economic benefits.
Figure 4 presents tests results. Figure provides the typical results of the relationship between static pushing force and the cone positions from h50 to hcg (see Figure 2). Symbols are as follows: CS/GG/C
The cone pushing can be interpreted as a large soil particle locally pushing into geogrid. From the data is evident that the increasing of the force is mainly due to geogrid. It is also interesting to consider the influence of the subgrade stiffness on the response of the cone. Despite the limited number of tests, the strong effect of the subgrade type on the pushing force can be noted.
REFERENCES Baslik,R. 1998. STN 73 3040 Geotextiles and geotextile-related products for construction purposes. Basic regulations. Bratislava: UNMS SR. /in Slovakl. Baslik,R., Matys,M., Turinic,L. 1999. STN 73 3041 Soil structures reirzforced by geosynthetics. Technical requirements. Bratislava: UNMS SR. /in SlovaW. ZSR. 1990. S4 -Railway substructure. /in SlovaM. Ex-Zeling s.r.0. 2000. Modernisation of railway track, CiferTrnava section. Control tests. Final Repor, 168p. /in Slovak/.
3.5 Control tests Quality assurance plan considers plate load bearing capacity tests as a Quality Control tests. In Figure 5 are presented the results of these tests. The subgrade
6
Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 0200 1 Swets & Zeitlinger, lS5N 90 265 1852 8
Clay-cement mix with reinforcing fibres for diaphragm walls A. Brinkmann, M. Benz, F. Bucher & P. Amann Swiss Federal Institute of Technology ETH Zurich, Switzerland
ABSTRACT: Diaphragm walls constructed by the slurry trench method are one of the most frequently used and proven possibilities to encapsulate landfills or old industrial sites. Thus, contaminant propagation into ground water may be hindered in cases where insufficient or non-existent bottom sealing capacity exists. In the case of external loading after the execution and hardening of the wall, the diaphragm wall's capacity to withstand these mechanical solicitations without a decrease in performance must be verified. As the most recent research results have shown, in the case of very small deformations, clay-cement mixes for diaphragm walls are only able to withstand low tensile stress levels. In the following article, research results obtained for fibre reinforced samples will be presented. Samples reinforced with polyvinyl alcohol fibres showed increased tensile strength, higher unconfined compressive strength and higher triaxial shear strength. The principal advantage of fibre reinforcing is most probably due to the fact that if tension resistance is exceeded due to bending, the formation of gaping cracks over the entire wall section is hindered.
1 INTRODUCTION
the tested mixture. Tests to investigate the general effect of soil reinforcement have previously been carried out by the Institute (Bucher 1984). The test results presented in this publication were obtained as part of a project for the diploma (Benz 2000).
At the Institute for Geotechnical Engineering (IGT) of the Swiss Federal Institute of Technology in Zurich, a clay-cement mixture for two phase diaphragm walls was developed (Hermanns 1993). During the two phase procedure, the trench is supported during excavation by a bentonite suspension (Phase 1). The second phase, in which the diaphragm wall is formed, is carried out immediately after the completion of the trench during the first phase. The abovementioned mixture is made up of 491 kg/m3 clay, 368 kglm' Portland cement, 123 kg/m' bottom filter ash and 654 kg/m' water. This results in a density of p = 1.63 t/m3. In order to be able to pump the mixture, the addition of a cement workability agent of approximately 1.5% by weight of the cement mixture was necessary. The feasibility of diaphragm walls using this mixture has been successfully tested on several occasions (Gunther et al. 1995, Brinkmann 2000). During the course of research on the mechanical behaviour of this mixture in a hardened state (Brinkmann & Amann 1999, Brinkmann 2001), it was able to be shown that only small tension stresses could be taken up. Large-scale tests at a 1:1 scale showed that the maximum strain under bending solicitations which may be withstood without damage is only about E = 0.01 %. Based on these results, the decision was made to undertake research to determine if fibre reinforcing is capable of improving the mechanical properties of
2 CHOICE OF FIBRES In the case of fibre reinforcing, attention must be paid so that the barrier effect of the mixture is not compromised. In addition, the workability of the mixture must be guaranteed. In this case, this means that the mixture must be able to be pumped. In addition, the fibres must not lose their mechanical properties in the alkaline medium of the hardened mixture or through contact with the contaminated leachate. For the tests carried out, polyvinyl alcohol fibres were chosen. These are used in place of asbestos in the production of fibre cement (Akers et al. 1989). Fibres 4 mm, 6 mm and 12 mm in length were available for the tests. The fibres were obtained from two producers: the 4 mm long fibres were provided by Eternit@ AG, Niederurnen, Switzerland, and were manufactured in Japan by the Kuraray Co. Ltd. The longer fibres came from the Schwarzwaelder TextilWerke, Schenkenzell, Germany. The material properties are presented in Table 1. The upper value in each row corresponds to the fibres from the Kuraray Company and the lower value to the fibres from the 7
Table I . Material property values for polyvinyl alcohol fibres (PVA), manufacturers’ values, upper value Kuraray Company and lower value Schwarzwaelder Textil-Werke. Property
Value
Property
Length
4 mm 6, 12 mm
Elastic 36’000 MN/m’ modulus 40’000 MN/m’ ~_-____ __ Limit strain 7.4% (tension) 6 -7%
___
~____
Diameter
14 pm 13 pm
Density
no data 1.3 g/cm’
Tensile strength
Value
~
cost
Through the addition of approximately 3-4% cement workability agent, the same consistency as that of the non-reinforced mixture may be obtained visually. Tests using the mixing and pumping equipment used on site should be carried out to test the pumpability under site conditions. The density of the fibre reinforced mixture remained virtually unchanged with respect to that of the non-reinforced mixture.
-
3.2 Unconfined compression tests
about 5 CHF/kg no data
Unconfined compression tests were carried out on 14 and 28 day old samples. Non-reinforced samples of the same age were tested to obtain reference values. It was seen that the unconfined compressive strength remained unchanged with the addition of 4 mm long fibres. The test results on samples containing 6 mm and 12 mm long fibres are presented in Figure 1. Each value shown represents the average of three individual tests. The values of the reinforced samples are presented with respect to those of non-reinforced samples, 100% being the strength of the nonreinforced samples. It may be observed that the unconfined compressive strength qu of the reinforced samples was higher than that of the non-reinforced samples. For identical fibre content, the samples prepared with 12 mm long fibres showed the greatest increase. The results showed that fibre content greater than 2.6% of cement weight would probably not result in further strength increases. The largest increase occurred for fibre contents between 1.3% and 2.6% of cement weight. The increase in strength seemed to decrease with increasing sample age. The 14 day samples reached an increase in strength between 10% and 23%, while for the 28 day samples an increase of only 7% to 17% was observed. The coefficients of variation v (v =
1’530 MN/m’ 1’830 MN/m’
Schwarzwaelder Textil-Werke. Longer fibres than those described here were not available from the above companies. An OH group is bound in the molecule of the polyvinyl alcohol fibres which synthesises with the available calcium bond in the cement. This results in a good bond between the matrix and the fibres. 3 TEST PROGRAM AND RESULTS 3.1 Mixing tests, workability, sample preparation Before carrying out tests on the hardened samples, it was necessary to investigate at which point in time during the mixing procedure the fibres should be introduced in order to guarantee the most even distribution. In addition, it remained to be seen how the consistency of the mixture was modified through the addition of the fibres. Mixing tests were carried out in accordance with the fibre quantities presented in Table 2. After the mixing tests, the samples for the laboratory tests were prepared with the given fibre contents. Sample heights were 100 or 120 mm, diameters 56 or 100 mm. Table 2. Fibre quantities used. Fibre content (% cement weight) Fibre length
0.3%
0.7%
1.3%
4mm
x
X
X
6 mm 12 mm
X
2.6%
3.9%
X
X
X
For the non-reinforced mixture, the components were introduced in the following order: water, clay, bottom filter ash and cement. A blade mixer with a speed of 1200 rpm was used for the mixing. The mixing tests showed that the distribution of the fibres was most favourable when they were introduced at the end of the mixing process with a reduced mixer speed of 300 rpm. The suspension became stiffer as the fibres were introduced.
Figure 1. Normalized unconfined compressive strength qu for various fibre contents.
8
average valuehtandard deviation) lay between v = 0.8% and v = 4.6% for the test series. The increased unconfined strength can be attributed to the reinforcing effect of the fibres. In compression testing, traction stresses occur above all perpendicular to the loading direction, which could then be taken up by the fibres. It was clear that the anchoring length began to be sufficient at 6 mm; samples with 4 m long fibres and fibre content of 1.3% cement weight showed no increase in strength. The absolute highest strength of all of the samples with 12 mm long fibres confirms this consideration. For a fibre content of 3.9% cement weight, the number of fibres was apparently so high that the matrix strength could no longer be developed in the same amount, as in the non-reinforced samples, and the positive effect of the reinforcing was no longer felt.
higher strength increases. For a fibre content of 2.6% cement weight, the samples with 6 mm long fibres and an age of 28 days reached an increase in tensile strength of 7%. The corresponding samples with 12 mm long fibres showed an increase of 14%. The coefficients of variation lay between v = 0.3% and v = 6.6% for this test series. In Figure 3, the stress development for individual samples as a function of piston displacement is shown. This figure clearly shows the sample behaviour after the exceeding of the tensile strength. The curve for a 28 day non-reinforced sample is included for comparison. This sample broke after failure in two halves, such that the sample offered no more resistance under the press. This was not the case for the reinforced samples. Even after exceeding the failure stress, the reinforcing held the samples together. No cracks were able to be observed with the naked eye after exceeding the tensile strength. The higher the fibre content and the longer the fibres, the less the stress decrease after reaching the maximum stress value. In addition, this decrease was less for 28 day samples than for 14 day samples. Therefore, if may be concluded that an improved anchoring of the fibres in the matrix occurred with increasing fibre length and sample age.
3.3 Indirect tension tests (Brazilian tests) It was also observed during the indirect tension tests that the addition of 4 mm long fibres did not result in any increase in tensile strength. In Figure 2, results from indirect tension tests on samples with 6 mm and 12 mm long fibres are presented. The behaviour was similar to that observed during unconfined compression testing. The tensile strength of the reinforced samples was higher than that of the non-reinforced samples for all fibre contents and lengths. It may be seen in these tests that for fibre contents greater than 2.6% cement weight no further improvement in strength properties occurred. 14 day old samples with a fibre content of 3.9% cement weight are an exception. A further trend may be observed where the strength increase for 28 day samples was lower than the increase for 14 day samples. Thus, 14 day sam ples with a fibre content of 2.6% cement weight (6 rnm long) showed a 15% strength increase while similar 28 day samples showed only 7%. The test series on 14 day samples with a fibre content of 3.9% cement weight also proved to be an exception here. Samples with 12 mm long fibres showed generally
3.4 Triaxial shear tests Based on the test results up to this point, it was decided to carry out drained triaxial shear tests on samples with a fibre content of 2.6% cement weight and 12 mm long fibres. The greatest improvement in behaviour appeared to occur for this fibre length and content. The test program for the triaxial shear tests is presented in Table 3. The triaxial shear tests were carried out under consolidated drained conditions. Isotropic consolidation was carried out over a time period of two days.
Figure 2. Normalized tensile strength o1for various fibre contents.
Figure 3. Results from indirect tension tests for various fibre lengths and contents.
9
Table 3. Test program: triaxial shear tests. Lateral Pressure
50
100 200
400
200 (without fibres)
400 (without fibres)
36
38
43
30
45
0 3
cally identical for both reinforced and nonreinforced samples up to a deformation of E = 0.7%. This means that the elastic modulus is identical in this section. In Figure 5 the results of the triaxial tests in a p'-q diagram are presented and interpreted. The interpretation is based on the hypothesis that during the shear process no pore water pressure developed. This hypothesis remains to be verified by further tests. The points lie on a line verified by the correlation coefficient R. This correlation yields the following shear parameters:
[kN/m'] Sample age [dl
41
The volume change of the sample during this time was less than A E ~=~ 1%. , The shearing process was deformation controlled. The stress ol was applied with a constant deformation speed of v = 0.002 mm/min. Drainage was permitted on both the upper and lower faces of the sample. The test results of the two reinforced samples with lateral pressures of o3= 200 kN/m2 and o3= 400 kN/m2 and the non-reinforced samples with identical lateral pressures are presented in Figure 4. The deviator ci, - o3 is presented as a function of sample deformation E. It may be stated as a result of this comparison that the results for reinforced samples reached higher deviators than those of the non-reinforced samples. However, it should be noted that the reinforced sample with a lateral pressure of q = 200 kN/m2was 11 days older than the non-reinforced sample and thus part of the increased strength was attained throughfurther hydration of the cement. For the samples with a lateral pressure of o3= 400 kN/m2, the reinforced sample reached a 14% higher deviator than the non-reinforced sample. It may also be observed that for the reinforced samples the maximum deviator was reached for a larger deformation than for the non-reinforced samples. For the samples with a lateral pressure of o3= 200 kN/m2, an increase from E = 1.4% to E = 1.9% occurred, while for o3= 400 kN/m2 the increase is between E = 2.7% and E = 4.2%. The stress-deformation curves were practi-
cp' = 39" and c' = 610 kN/m2 3.5 Permeability tests Figure 4. Results of triaxial shear tests under consolidated drained conditions for reinforced and non-reinforced samples.
The permeability behaviour of the fibre reinforced mixture was investigated in the triaxial cell at room temperature (20" C). The test equipment described by Hermanns (1993) was used. The gradient i with which the samples were percolated was 30 and was increased to 50 for 28 day samples. The samples were percolated from the bottom to the top. The tests were carried out with samples with fibres 6 mm and 12 mm long, for fibre contents between 1.3% and 3.9% cement weight (see Table 1). One sample was mounted for each fibre content. A non-reinforced sample served as a reference. In Figure 6, the results of permeability tests are presented. The tests were begun at a sample age of three weeks and ran approximately seven weeks. In four out of the five tests, clogging of the filter stone occurred, thus impeding further flow through it. Thus, the samples were remounted with new, smoother filter stones on the surface and the tests were restarted. The results of these samples are presented from the day of the new mounting. The reason for the clogging was attributed to the incomplete hydration of the cement, which in conjunction with the fine fibres led to the described situation.
Figure 5. p'-q diagram with test results for reinforced samples.
Figure 4. Results of triaxial ,.hear under consolidated drained conditions for reinforced and non-reinforced samples.
10
Concerning the unconfined compression and tensile strength values, for 28 day samples, an increase of approximately 15% was reached, It is important that the samples did not break after reaching failure but held together with a high resistance. For unconfined compression tests, this also occurred for deformations greater than 5%. Higher shear strengths and higher failure deformations were seen in drained traixial shear tests. The stiffness of the reinforced samples remained the same compared to that of non-reinforced samples. In order to confirm the results obtained, investigations on a larger number of samples are necessary. Tests on older samples are especially necessary in order to see if for aged material, improved mechanical properties can also be counted onpurther variant possibilities, in addition to fibre amount, are fibre length and diameter. During tests on longer fibres, attention should be paid that the samples are large enough in order to exclude any scale effects. In addition, it remains to be tested if longer fibres can lead to problems with the mixing machinery. The bond with the cement matrix should be checked for fibres with larger diameters in comparison to that with the finer fibres used here. It is possible that the use of thicker fibres could enable an even better distribution of the fibres in the mixture, and less fibres would be needed for identical fibre content. If fibre reinforcement is used, the principal advantage seems to be that after failure no breaking of the material occurs. This means for bending solicitations greater than the tensile strength of the material, no continuous, gaping cracks occur. Part of the wall cross section would be under pressure. It can be assumed that in this pressure zone the barrier effect of the material remains the same. For further investigations, small and large-scale tension bending tests are being planned.
Figure 6. Results of the permeability tests for various fibre lengths and contents.
For all of the samples, a decrease in the permeability coefficient k was observed for an increase in test duration, as was already observed by Hermanns ( 1993). The decrease in permeability over time was similar for all of the samples. All of the samples reached the usually required permeability value of k = 1 ~ 1 0m/s. - ~ For the sample with a fibre content of 3.9% cement weight (12 mm), this was first reached at a sample age between 35 and 55 days. Higher permeability values were observed for samples with higher fibre content for identical fibre lengths. For samples that were prepared with 12 mm long fibres, the difference was greater than an order of magnitude. It was clearly seen that, on the average, the addition of fibres in comparison with the non-reinforced mixture led to an increase in permeability.
4 CONCLUSIONS AND NEED FOR FURTHER RESEARCH
REFERENCES
It was seen that with the addition of polyvinyl alcohol fibres, quite homogeneous mixtures were able to be produced. An increased amount of workability agent was however necessary in order to render the mixture texture optically identical to that of the nonreinforced mixture. The pumpability of the reinforced mixture remains to be tested under site conditions. The permeability values of the reinforced mixture were on the average less than an order of magnitude higher than those of the non-reinforced mixture. It is to be expected that the required limit values be reached for the tested mixture. Higher fibre contents than those tested probably would lead to an distinct worsening of the permeability behaviour.
Akers, S.; Studinka, J.; Meier, P.; Dobb, M.; Johnson, D.; Hikasa, J. (1989). Long term durability of PVA reinforcing fibres in a cement matrix. Interizat~onalJournal of Cement Co~~posj~e.s and L ~ ~ h ~ eConcrete, i ~ h t 1 I(2): 73-78 Longman Scientific & Technical, Harlow. Benz, M. (2000). Verbesserung der mechanischen Eigenschaften einer tonzementgebundenen Dichtwandmischung durch eine Faserbewehrung. Diploma project at the Institute for Geotechnical Engineering, ETH Zurich ~unpublished). Brinkmann, A. (2000). Ton- zementgebundene Dichtwandmasse fir das Zweiphasenverfahren, Ergebnisse aus Forschung und Praxis. Mitteilung des Institutsjiir Grurzdbau und Bodenmechmik, Technische Un~versitatBraunschweig. No. 63. Brinkmann, A. (2001). Untersuchungen zum mechanischen Verhalten von ton-zement-gebundenem Dichtwandmaterial f i r das Zweiphasen-Verfahren. Publication of the Institute
11
for Geotechnical Engineering, ETH Zurich Verlag der Fachvereine Zurich. Brinkmann, A.; Amann, P. (1999). KIein- und grossmassstabliche Versuche zur Ermittlung des mechanischen Verhaltens einer ton- zementgebundenen Dichtwand. ~ ~ u j n g e r z i 74: eu~ 390-396. Springer VDI Verlag. Bucher, F. (1984). Verfestigung von Boden durch Fasem. Publication of the Swiss Society for Soil and Rock Mechanics. NO. 108,65-68
12
Gunther, K.; Kucera, P.; Kudla, W.; Lachler, W.; Mergelsberg, W. (1995). Die Sanierung der Sonderabfalldeponie Malsch. Buutechnik 72. 622-630, Emst & Sohn. Hermanns, R. (1993). Sicherung von Altlasten mit vertikalen mineralischen B a ~ e r e s y s t e m ~im n Zweiphasen- Schlitzwandverfahren. Publication of the Institute for Geotechnical Engineering, ETH Zurich, No. 204, Verlag der Fachvereine Zurich.
Landmarks in Earth Reinforcement, Ochiai et al (eds), 02001 Swets & Zeitlinger, ISBN 90 265 1852 8
Design and development of inclined plane test on geosynthetics A. Cancelli Professor. Director of Department of Geological Science and Geotecnologies, University of Milano-Bicocca; Milano, Italy
P. ~ i ~ o l d i Technical Director, Tenax SPA - Geosynthetics Technical OfJice, Milano, Italy
A. Moroni Graduate student, U n i v ~ rof~ ~i i~l u n o~, i l a n oItaly ,
A. Poltronieri Graduate student, University of Milano, Milano, Italy ABSTRACT: This paper presents experimental results from tests on geosynthetic/geosynthet~cand sandlgeosynthetic interfaces performed with an inclined plane apparatus designed in c o n f o r ~ t ywith the European Standard prEN I S 0 12957-2. More than 240 tests have been carried out, investigating almost all the interfaces that may occur in a landfill. The test results have been elaborated in order to make them comparable with direct shear tests results in a cs - z plane. It is shown that the friction angles obtained by inclined plane tests are always lower than the ones obtained by direct shear tests for the same applied normal stress.
1 INTRODUCTION
2 EXPERIMENTAL PROGRAM
In the design of geotechnical works there is always the need to analyze the stability conditions. This paper is focused on applications for geosynthetics where the potential sliding surfaces occur along interfaces between different materials placed on an inclined substratum. In particular, problems connected with landfill final cover systems stability have been considered: these structures present, from the upper surface layer to the lower gas collection layer, many possible surfaces along which sliding can occur. Therefore a fundamental point for the designers is the knowledge of frictional parameters between different types of geosynthetics and soil, derived from tests representing in situ conditions and variables. Many authors maintain that conditions characterised by low normal stresses cannot be represented by the direct shear test since it yields too high friction angle values, because of its inability to work well in this low range of forces. The need for a more suitable test method has caused the development of a new test called inclined plane. Therefore a new inclined plane apparatus has been designed and developed in the Tenax Laboratory in Viganb (Italy), with the aim of: 1) verifying the adequacy of the apparatus itself and in general of the test method; 2) comparing the results of the two tests (inclined plane and direct shear), and find their limits of applicability. The design of the inclined plane apparatus and the definition of test procedures have been carried out following CEN/TC 189 indications included in the European Standard prEN IS0 12957-2. Lacking any experience about this standard, during the development of the research a critical view has been maintained with the purpose to find any possible irnprovement to the inclined plane apparatus.
Products and interfaces to be tested have been chosen with reference to landfill final cover systems. For soil-geosynthetic interfaces the soil used is a standard sand in accordance with EN 196-1, dried to a moisture content of less than 2 % and compacted to a density of 1.7 Mglm'. The tested geosynthetics and their main characteristics are summarized in Table 1. 3 INCLINED PLANE TEST APPARATUS The apparatus (Figg. 1 to 5) consists of a 680 mm x 450 m x 15 mrn tilting table provided with two long screws which allow the plane to be set perfectly horizontal before the beginning of every test. Above it the lower box is placed: its internal dimensions are 400 mm x 325 mm x 100 mm; on its back side (see Fig. 3) there is the clamping system that allows to fix the geosynthetics: it consists of two bars with parallel series of holes where the rear side of the geosynthetic specimen has to be screwed down. The upper box has internal dimensions of 300 mm x 300 mm x 130 mm: it is fitted with rollers which bear on runners fixed on the lower box outer walls. The normal force application system consists of a rigid plate and a frame placed on it through a circular section guide (see Fig. 1 and 5) which carries the weights clasped to the supports at the bottom. This system is able to be always vertical and passing through the center of gravity of the upper box. The normal forces applied produce the required normal stresses of 5 , 10 and 25 kPa, The tilting table lifts up at an angular speed of 3"/min. The rising device consists of an hydraulic 33
Table I . Tested products Products Tessilbrenta Geotess TC/PP
Polymer PP
Tessilbrenta Geotess TC/PP
PP
Bidim S 61
PP
GTX 3
Geofabrics MP 200
PP
GTX 4
Du Pont Typar SF 40
PP
GTX 5
Amoco Propex 606 1
PP
GTX 6
Agru HDPE Smooth
HDPE
GMB I
Agru HDPE Rough
HDPE
GMB2
Agru HDPE Micro spikes
HDPE
GMB3
Agru HDPE Medium spikes
HDPE
GMB4
Agru HDPE Big spikes
HDPE
GMB5
Flag Flagon C/SL
PVC
GMB6
Cover Top 32
LDPE
GMB7 GCD 1
Tenax TNT 450
Symbol GTX 1 GTX 2
Tenax Multimat 100
HDPE(GNT) PP (GTX) HDPE (GNT) PP (GTXj HDPE (GN'r) PP (GTX) HDPE G N T ~ PP (GTX) HDPE (GNT) PP (GTX j PPCGEC) PET (GGR) PP
Tenax IT SAMP 045
HDPE
GGRl
Tenax TT SAMP 160
HDPE
GGR:!
Tenax R SAMP 20 1
HDPE
GGR3
Tenax TT SAMP 40 I
HDPE
GGR4
Tenax TNT 600 Tenax TNT 900 Tenax TNT 1200 Tenax Tendrain I300/2 Tenax Multimat R 110
Tenax LBO SAMP 220 Tenax CE 450 Tenax CE 600 Laviosa Geobent STD 50 Laviosa Geobent HT 2450
Figure 1. Inclined plane testing apparatus.
GCD 2 GCD 3 GCD 4 GCD 5 GEC 1 GEC 2
PP
GGR 5
HDPE
GNTl
HDPE
GNT2
PET(GTxw) PP (GTX NW) PP (GTX W j PP (GTX NW)
GCL 1
Figure 2. Start of an inclined plane test.
GCL 2
ram connected to the table by a shafting chain whose vertical movement is made horizontal by a cogwheel (Figure 2). The displacement measuring device is a transducer (LVDT) connected to the rear wall of the upper box (Figure 3) through an inextensible wire. The ram and the transducer are connected to a multi-axes servo-hydraulic digital actuator, able to handle both data processing and control, up to 4 axes, either independently or combined between themselves, by controlling them in a feedback closed loop. For tests involving Geosynthetics with full surface (without apertures), like Geotextiles and Geomembranes, the lower box has been filled not with sand, but with a 400 mrn x 325 m x 97 m block of wood, on
Figure 3. Detail of the clamping system and the movement transducer.
which some plates of different thickness (1-3 mm) are laid to maintain a 0.5 rnm gap between the fixed geosynthetic and the base of the upper box. The testing of specimens with open structures requires the filling of the lower box with the soil. Two values of internal soil depth H, in the upper box have been assumed: 100 mm and 50 mm, with the aim of observing its possible influence on the test results. 14
Figure 5. Plan view of the inclined plane test apparatus.
4 DIRECT SHEAR TEST APPARATUS locity of I mdmin. The movement device consists of an hydraulic ram connected to the front wall of the lower box by a threaded pin whose vertical movement is made horizontal by a cog-wheel (this hydraulic ram is the same which produces the tilting of the inclined plane apparatus). The dimensions of the lower box have been established considering the possibility to carry out also performance tests using site specific soils. The vertical force is applied by means of an hydraulic piston which provides constant loading. The normal forces applied produce the normal stresses of 25, 50 and 100 Wa, suitable for a
This apparatus consists of two rigid metal boxes, each one containing one of the two materials to be tested (Figure 6). The upper box remains steady during the tests, standing on four supports fixed to the plane of the hydraulic testing machine: its dimensions are 3 16 mm x 3 16 mm x 100 mm. The elevation of this box is adjustable by means of calibrated spacers which are inserted between the supports and the box itself. During the test, the 670 mm x 470 mm x 225 mm lower box moves on steel rollers standing on the plane of the testing machine at a constant ve-
15
Figure 6. Scheme of the direct shear test apparatus.
comparison with inclined plane test. Both the pistons are connected to the multi-axes servo-hydraulic digital actuator. Through transducers connected to the pistons, the system is able to control simultaneously their positions, loads and velocity, keeping them within very narrow variation intervals. To fix geosynthetics specimen to the lower box, a plywood board has been used. Geocomposites have been attached to the plywood board by means of selftapping screws distributed along three edges of the sample; the screws are placed outside of the contact area between the two materials and this assures that they cannot influence the results. Screws are not placed on the last side of the specimen, in order to prevent, in the case of lengthening of the geosynthetic, that it could bulge or wave and cause an increase of friction.
5 TEST RESULTS Figure 7. Typical results of inclined plane tests: top) “sudden death”; middle) “step curve”; bottom) “slow motion”.
5.1 Inclined plane Results are presented in As-P graphs (Figure 7), where As is the movement of the upper box during the test and p is the inclination of the tilting table which increases at a constant rate. From the analysis of these diagrams three kind of behavior can be distinguished, in accordance with results recorded in the preliminary research made for the development of the European standard (AA, VV, 1995). The graphs may be categorized in: a) “sudden death”: a long period of no displacement followed, after a movement at least of 1-2 mm, by a sudden and steep increase of the differential ratio ds/dp; b) “steps curve” where the angle of first movement and the angle of failure do not coincide and the graph is characterized by steps of variable dimension; c) “slow motion before sudden death” in which there is a larger displacement of the upper box before sudden failure. Graphs As-p allow to determine the angle of slippage (p), which is defined as the angle at which the displacement of the upper box is equal to 50 mm. From the value of the angle of slippage, the angle of
friction @ needs to be calculated. For this purpose normal and shear stresses at the moment of the sliding have to be determined by means of the following equations: c & , ~= (9.8 1 W CO$)/( 1000 A)
(1)
a = 9.8 1 (W senp + fr(B))/(1000 A)
(2)
tan@= donp (3) where: W = mass of soil, surcharge weights and any part of the upper box not supported on rollers, in kg; fr(p)= force required to restrain the empty upper box when the tilt table is inclined at an angle p; A = conAfter the calculation of tact area (300 x 300 m2). are the single values of @, the results (a and onP) plotted in the 0-a plane in order to compare in a better way inclined plane and direct shear tests. In a different way from others Authors (Wasti & Ozduzgun, 2001) the normal stress used for the 0-z
16
Table 3. Test conditions
Table 2. Test results
I
s/s s/s
I
@
a
(")
(Wa)
E E E E E E E G E G E G A E E A A (wet) E E D D
5-10-15-20-25 5-10-15-20-25 5-10 5-10-25 5-10 5-10-25 5-10 5-10 5-10-25 5-10 5-10-25 5-10 5-10 5-10 5-10 5-10 5-10 5-10 5-10-25 5-10 5-10-25 5-10-25 5-10-25 5-10-25 5-10-25 5-10-25 5-10-25
24.69 29.39 20.57 25.22 20.26 24.47 21.73 12.34 12.98 13.34 21.62 15.90 21.79 9.97 21.84 16.98 18.91 10.44 20.92 16.37 22.55 26.55 27.97 24.85 21.54 28.36 28.31
0.76 0.61 1.30 0.33 1.17 1.2 0.37 1.60 0.25 0.73 1.10 2.17 1.10 1.13 0.63 1.36 1.05 0.99 0.02 0.29 0.53 1.06 0.88 0.07 0.77 0.89 1.05
23.39 24.06 28.20 29.20 11.09 21.78 23.99 22.02 24.24 27.40 15.96 20.77
0.22
I L L A A I I
5-10-25 5-10-25 5-10-25 5-10-25 5-10 5-10 5-10-25 5-10-25 5-10-25 5-10-25 5-10 5-10
Test Interface (S = Sand)
S/GTXI S/GTX2 S/GTX3 S/GTX4 S/GTX5 S/GTX6 S/GMB 1 S/GMB 1, SIGMB2 S/GMB2, S/GMB3 S/GMB3, S/GMB4 S/GMB4, S/GMB5 S/GMB5, S/GMB6 S/GMB7 S/GCDl S/GCDl S/GCD 1 S/GCD3 S/GCD4 S/GECl S/GEC2 S/GGRl S/GGR2 S/GGR3 S/GGR4 S/GGR5 S/GNTl GCD 1 /GMB 1 GCD 1/GMB2 GCD2/GCLI GCD2/GCL2 GEC 1/GCD 1 GEC 1/GCD5 GNTl/GTX2 GNT2/GTX2
Applied
Conditions pressures (Wa)
H D E A E A
Test Soil Depth Hs conditions (mm) A 50 50 B 50 C D 50 E 100 F 100 G 100 H 100 I L
Lower box content Block of wood Block of wood Soil Soil Block of wood Block of wood Soil Soil Block of wood Soil
Fixing specimen Yes No Yes No Yes No Yes No Yes No
Tab. 2, where the different test conditions can be found in Tab. 3. 5.2 Direct shear tests The program of direct shear tests has included tests on three different interfaces, each representing one of the fundamental types of possible contacts which can be present in a real application, and in order to compare the results with the ones of the plane. The values of shear strength obtained by direct shear tests and the relative normal stresses have been represented in the B-T plane, where the friction angle and the adhesion or cohesion have been evaluated from the failure envelope line.
6 COMPARISON Previously published works offer some examples of the comparison between these two test methods. Matichard et al. (1991) present values of friction angles obtained from direct shear tests that are larger than the ones from inclined plane tests. Wasti & Ozduzgun (2001), in particular for tests on rough HDPE geomembrane-geotextile interfaces, find both friction angle and adhesion values to be almost always higher by varying degrees for the direct shear test. Therefore, direct shear test envelopes lie above those of the inclined plane. Also in the present work the values from direct shear tests are always larger than the ones from inclined plane tests: the results obtained from comparison of tests on soilgeosynthetic and geosynthetic-geosynthetic interfaces yield important differences of 10-12 Wa for shear strength values and of 11"-15" for friction angle values. Lower differences have been registered for the tests on soil-soil interfaces (respectively 2 kPa and 4").
0.74 0.69 0.05 1.52 1.19 0.54 0.33 -0.18 1.15 -0.02
plot is not the initial normal stress (for example 5 or 10 kPa), but the normal stress calculated with Eq. (1) at the instant of failure slippage, characterized by a lower value. In fact, to the authors'opinion, even though the initial CT may be considered as a "test condition", it is more significant to use the normal stress on the interface at failure for determining the Cp value. By getting three points for the three normal stresses, it has been possible to obtain the failure envelope line on the B-z graph and to evaluate the friction angle and the adhesion for soil-geosynthetic and geosynthetic-geosynthetic interfaces or cohesion for soil-soil interfaces. The results are summarized in
7 CONCLUSIONS Measurements of shear resistance along interfaces between different materials, significant from the ap17
plicative point of view, have been made using a new inclined plane apparatus and a more experienced direct shear device. By means of the inclined plane method more than 240 test and 38 interfaces have been evaluated at the three initial normal stresses of 5, 10 and 25 kPa; the direct shear method was used to test 3 interfaces at the normal stresses of 25, 50 and 100 kPa, in order to afford a comparison with the same ones investigated by the previous apparatus. From inclined plane tests the following conclusions can be drawn: the friction angle tends to decrease with the increase of the initial normal stress; in soil-geosynthetic tests the position of the failure surface varies depending on the type of geosynthetic surface: for smooth surfaces it develops just at the interface, while for rough specimens it develops in the soil above the interface; the geosynthetic surface influences also the behavior of the displacement of the upper box; all the interfaces tested always present the final failure within 50 mm of relative movement of the upper box; tests with soil depth in the upper box H, = 50 mm provide friction angles always higher than the tests with H, = 100 mm, probably because the procedure with H, = 100 mm creates an unbalanced distribution of the forces working on the interface, thus causing an advanced failure along the shear surface; taking into consideration the sensitivity of this apparatus and of this test method with regards to the boundary conditions, it seems fundamental to set all the test variables in order to assure repeatability and reproducibility of results. From the comparison of the results of the inclined plane and the direct shear tests, the following conclusions can be drawn: friction angle values obtained by inclined plane tests are always lower than the ones obtained by direct shear tests; inclined plane test results shall be considered in modeling load conditions on slope, while values from direct shear tests cannot be considered correct for this situation; direct shear test has to be used for conditions characterized by loads on horizontal interfaces, for which inclined plane test produce too much conservative results; from the previous point it follows that the stability analysis of particular projects, like landfill cover systems, shall be performed considering friction angle and adhesion values determined by inclined plane, since this apparatus can model the actual conditions in a correct way, yielding more realistic results, which put the designer in favor of safety.
Figure 8. Comparison of results in the 0-2 plane
8 ACKNOWLEDGEMENT This research program was carried out with the financial support of MURST (Italian Ministry for University and Scientific and Technical Research).
18
REFERENCES
Matichard, Y., Delmas, PH., Soyez, B., Girard, H., Mathieu, M. (1991). Stability of lining systems on slopes. Proceedings Third lnternarional Landfill Symposium AA, VV. (1 995). Research and intercomparison tests necessary ‘91”. Cagliari, Italy. for the harmonisation of standards in geotextiles. EC MeasWasti, Y. & Ozduzgun Z.B. (2001). Geomembrane-geotextiles urement and Testing Prograinme Project 0169. interfaces shear~properties ~as determined board l CEN/TC I 89 (1999) p r ISO ~ 12957-2: ~ ~ ~and gee- ~ ~ ~ by inclined i and direct shear box tests. Geotextiles and Ceoinembranes textile-related products - Determination of friction charac19 (2001)45-57. teristics - Part 2: Inclined plane test. Bruxelles, Belgium. ‘lSardinia
19
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Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 02001 Swets & Zeitlinger, ISBN 90 265 1852 8
Grout injection in the laboratory C . Dano Rkgie Autoizome des Transports Parisiens, RATP, Paris, France
N. Derache Irztrafor Company, Research Laboratory, France
ABSTRACT: Many grout injection devices have been described in literature. The design of our own injection procedure was strongly inspired by these previous works. However, we made it original by the tests systematically added to check its performances concerning the homogeneity and the reproducibility of the grouted samples. After it had been validated, three parameters were examined: the cement to water ratio, the rate of discharge of the grout and the nature of the soil. Mechanical tests were also carried out to control the quality of the procedure and to assess the effect of the parameters on the properties of the grouted soils, in particular the elastic properties in the small strain range. We proved that the cement to water ratio and the nature of the soil were the main influential parameters. American Standard D 4320-93). All of them are mainly composed of a pump that allows to inject the grout and of a transparent cylindrical tube inside which the soil is deposited. Those devices can be classified in three groups according to the height to diameter ratio (WD) of the tube and the objectives of the grout injection (Table 1). A wide range of diameter, from 22 to 100 mm, was used by the practitioners. Nevertheless, it has been recognised that the smooth face of the tube and the wall effect could influence the filtration process by creating preferential paths for the grout into the soil (Azzar 1997). Consequently, the use of the greatest possible diameter D was advised. The main part of our experimental set-up, as previous ones, is therefore a cylindrical and transparent tube made of a rigid Plexiglas with an internal diameter of 100 mm and an height of 900 mm (Fig. 1). We noticed that the grouted sand was not homogeneous over the complete volume of the tube when the soil was in direct contact with the rubber stoppers. We have assumed that the sudden reduction of the outlet diameter acted as a funnel where the finest particles concentrated, causing the soil in the upper section to be more reinforced. Another evidence was that relatively high injection pressures were recorded at the end of the injection step. Consequently two 50 mm thick filters composed of gravel with a diameter between 4 to 8 mm are hence interposed between the soil and the stoppers. Sieve nets are also set in place to separate the soil and the gravel in order to create a laminar flow of grout through the tested soil.
1 GROUTING AND INJECTABILITY The injection, under pressure, of a microfine cement grout is a suitable technique in order to improve the geotechnical characteristics of a soil with an initial permeability between 10-5and 10'3 m/s. The injection and the setting of the cement involve both the reduction of the porosity and the increase of the stiffness and the strength of the initial soil. Unfortunately, in situ conditions do not usually lend to an extensive study of the behaviour of the grouted soils. So the technique has been reproduced in the laboratory. In many cases, the main purpose of an injection test in the laboratory is to assess the injectability of a grout in a given soil. Some useful injectability rules have therefore been proposed. A relatively complete review was already done by Benhamou (1994). Rather than injectability, injection tests in the laboratory now aim at the understanding of the physlcal or chemical mechanisms that occur whereas the grout permeates the soil. Another objective is to prepare homogeneous samples of grouted soils.
2 FORMER INJECTION EQUIPMENTS Many injection devices have been described in literature (Zebovitz et al. 1989, Di Prisco et al. 1992, Benhamou 1994, Schwarz et al. 1994, Bennabi et al. 1995, Azzar 1997, Tailliez 1998, Ismail et al. 2000) or in Standards (French Standard NF P 19-891,
21
Height to diameter ratio WD WD=2
6<WD<10 H/D > 10
Objective of the in.jection References Preparation of one homogeneous grouted sample Standards, Bennabi et a1 1995 Ismail et a1 2000 Benhamou 1994 Study of the filtration process and of the meSchwarz et a1 1994, Tailliez 1998 chanical behaviour of several grouted samples Filtration understanding Bouchelaghem 2000
Figure 1. Injection device.
When no hammer strokes are applied, the relative density index is close to 40 % but a density gradient then appears inside the tube. When 30 strong hammer strokes are applied, the relative density index is close to 95 %. In this case, the reproducibility is good. We also established a procedure in order to obtain intermediate values of the relative density index but it has not been validated yet. Once completely filled, the column is closed and tightened by means of compressed rubber gaskets. The sand is then saturated with water by an upward flow of 6 cm3/s (or 21.6 l/h). This step can be removed if necessary.
3 EXPERIMENTAL PROCEDURE The injection procedure consists of six successive steps: the setting of the equipment, the filling of the tube with the granular material, the preparation of the microfine cement grout, the grout injection itself, the storage of the tubes and finally the mechanical tests. 3.1 Filling ofthe tube Two methods could be used to fill the tube with sand: by sand raining or by successive layers. The pluviation method (Levacher et al. 1994, Ismail et al. 2000) did not seem to be suitable for columns with high height to diameter ratios. So we have adopted the second method that consists in dropping the sand through a 1 m long hose surmounted by a funnel, placed inside the Plexiglas tube and slowly pulled up. Each 10 cm thick sand layer is then compacted by hammer strokes on the perimeter of the tube. The number and the intensity of these hammer strokes influence the homogeneity of the sand and the relative density index Id, such as: Id - Ymax
Y
3.2 Preparation of the microfine cement grout The patented microfine cement grout used in this research program is composed of water, cement with particles size less than 12 pm, an inert additive and a superplasticizer which role is to disperse the cement particles. Precise contents of each component are prepared and mixed together according to a prescribed order. The suspension is simultaneously energetically stirred in an agitator supply tank in which the grout batch is 4000 cm3. The Rayneri high-speed paddle mixer has not been changed for the duration of the study since Schwarz et al. (1994) have enhanced the influence of the mixer on the rheological behaviour of the grout.
Y-Ymin Ymax -Ymin
where y, Ymin and ymax are respectively the unit weight, the minimal unit weight and the maximal unit weight.
22
Table 3. Properties of soils tested.
Since the grout is ready, we determine its density
pgroutand its rheological properties. The density is
Soil
measured with a Baroid scale. As a microfine cement grout has often been considered as a binghamian fluid, we also characterise it by its plastic vis. viscosity cosity qp and its yield shear stress T ~ The is measured by means of a Fann viscometer or evaluated from a Marsh flow cone in which a predetermined volume of grout is permitted to escape through a precisely sized orifice, the time of efflux being used as the indication of the grout consistency (ASCE 1980). These properties mainly depend on the cement to water ratio C/W. Table 2 shows the properties of the grout for 3 mix designs corresponding to cement to water ratios of 0.172, 0.299 and 0.437. Finally we check that these properties are constant over the working time, namely about 20 minutes, provided that the stirring action with a special propeller is maintained to avoid flocculation and segregation of the cement particles.
Unit SF AAM AAG
Ds0 pm 220 410 1300
D6O/
1.4 2.1 5.9
DlO
ymin
kN/m-i 14.0 14.9 16.4
y,ll;,x kN/d 16.5 17.9 19.4
Figure 2. Grading curves of the sands and the cement grout. Table 2. Rheological properties of the grouts.
4.2 Follow-up of the injection parameters CIW
Pgrout
TP
,CP
TMmh
g/cnz3 1.OO 1.17 1.23 1.27
CPO
Pa
S
Pure water 0.172 0.299 0.437
1 1 1
27 29 29.5 29.8
2.5 3 3.5
The monitoring of the injection parameters, namely the pumping rate q, the injection pressure P, the mass M and the position of the grout h in the tube, as a function of the time t, allowed to qualitatively appraise the quality of the grout permeation. As the pumping rate q was set to a constant value, the mass of the grout injected into the soil had to linearly evolve with time, which was confirmed for all the tests. More, the velocity of the injection front, easily obtained by noting the height of the grout in the tube and the corresponding time, was a good indicator of the homogeneity of the soil density. Irregular velocities revealed an heterogeneity in the sand due to segregation of the soil particles or a larger size of the grains for instance. In the same way, the grout permeation had to occur with a regular pressure increase until the grout reached the top of the column. A pressure fall means either a leak or a breakdown in the soil. We observed such an horizontal breakdown in a soil with a low relative density index. It very likely corresponded to an auto-compaction phenomenon. In general, due to the low pumping rate and the rheological properties of the grout, injection occurred without any problem. Figure 3, with the corresponding characteristics of the tests in Table 4, shows the effect of the cement to water ratio and the nature of the soil in the development of the injection pressure. The relative density index was 95 % for the three soils. We noted that the higher the cement to water ratio was, the greater the injection pressures were. Smaller mean diameters of soil particles also seemed to induce stronger injection pressures.
3.3 Grout injection A fixed volume of grout is injected from the base to the top of the column with a constant rate of discharge. We record the evolution of the mass, of the position of the injection front and of the injection pressure. We also control the rheological properties of the grout when it left the column. After injection, the tubes are stored in water, 28 days at least, until the cement setting provides a sufficient strength to the grouted sands. Three samples are finally cut out in each column. Their faces are perfectly aligned. Their height to diameter ratio is close to 2. 4 INJECTION ANALYSIS 4.1 Granular soils tested Following the procedure described above, three different granular soils were tested: a standard siliceous fine Fontainebleau Sand (noted SF) and two silicocalcareous alluvial deposits (noted AAM and AAG). Their characteristics are reported in Table 3 (D, is the grain size corresponding to a passing of x %). The shape of the particles was quite similar for the three soils. Their grading curve (Fig. 2) was cut to a particle size of 1 cm.
23
Figure 4. Result of a gammadensimeter test. Figure 3. Follow-up of the grout pressure.
5.1 Experimental equipmen2
Table 4. Characteristics of injection tests. Test
Soil
c/w
1 2 3 4 5 6
SF AAM AAG SF SF SF
Pure water 0.172 0.172 0.172 0.235 0.172
Two laboratory testing devices were used to determine the dynamic elastic properties of both the grouted sand and the pure grout in the small strain range: the Grindosonic apparatus (Allison 1987, Allison 1988, French Standard NF P 18-414) and an Ultrasonic Concrete Tester (French Standard NF P 18-4 18). Both are non-destructive methods. The Grindosonic apparatus consists in exciting the sample by a slight mechanical impulse. The fundamental vibration frequency is deduced from the analysis of the resulting transient vibration pattern and the elastic parameters are then calculated. Accurate performances are obtained for Young’s moduli varying from 100 MPa to 840 GPa. They mainly depend on the shape of the samples and of the physical characteristics of the materials. Two pulse mode among the three following mode (bending, torsion, compression) are required to determine the Young’s modulus E,, the Poisson’s ratio v, and the shear modulus G,. Allison (1988) also indicated that the dynamic elastic parameters depend on the water content of the tested material. For this reason, our samples were stored in the same conditions of relative humidity and temperature. They were dried in the open air before any measurement. The Ultrasonic Concrete Tester E46 consists in measuring the time T corresponding to the propagation of an ultrasonic longitudinal wave through the tested material. Two transducers are placed on each end of the samples. A good contact is assured by means of a scan gel. An accurate determination of the length L of the sample is also required to calculate the velocity V,, of the ultrasonic wave. Strains are small enough to assume a linear elastic behaviour of the material. From this, it can be shown that:
(3
cm-%F 6 6 6 6 6 8.77
4.3 Homogeneity of the grouted sands Schwarz et al. (1994) have noted a decrease of the unconfined compressive strength with the growing distance from the point of injection. This was all the more obvious as the cement to water ratio was high. Consequently, we have assessed the degree of heterogeneity of the grouted sand density in a tube by means of a gammadensimeter. Figure 4 shows such a measure on a grouted Fontainebleau sand sample (Id=95 %, C/W=0.172). The mean unit weight of the grouted sand was 20.4 kN/m3. The succession of humps and hollows was clearly due to the method for filling the tube with sand. However, the variation around the mean value was relatively low, about f.2 %.
5 MECHANICAL TESTS The elastic properties of grouted soils in the small strain domain (E < 10-5)are of great importance for structural design. Their determination requires the use of special testing methods among which wave propagation testing devices. Unconfined compressive tests are also of practical interest for a first estimate of the mechanical improvement of the grouted sands. In our case, these tests also allowed to check, after injection, the homogeneity of the grouted samples.
2 (l+v)(1-2v) E=PVus
24
L
(I+v)(l-2v) l-v
(2)
where p is the density of the grouted sand. As the Poisson’s ratio can not be determined, we only compare the longitudinal wave velocity.
5.2 Experimental results We show in Table 5 the experimental results as a function of the cement to water ratio C/W, the rate of discharge of the grout q and the nature of the soil. G, is the shear modulus determined by the GrindoSonic apparatus, E, the Young’s modulus, V,, the velocity of the ultrasonic wave measured with the Ultrasonic Tester, y the dry unit weight. The subscript m stands for the mean value and A for the difference between the minimal and the maximal values of a given property.
5.3 Comments Grouted sands prepared in the same conditions led to very similar elastic properties as shown in Table 5. In the same way, the mechanical tests confirmed the previous observations about the homogeneity of the grouted sand in a tube. Indeed, the dynamic elastic properties of three samples stemmed from the same column seemed to be quite similar, except for the test No. 5 for which the preparation probably caused damage in the samples. The small variations can be attributed to a small variation in the local density or in a more important content of coarser grains that favour the propagation of mechanical waves. If we consider that the value of the Poisson’s ratio determined by the GrindoSonic apparatus is acceptable and if we report this value in Eq. 3, the dynamic elastic moduli measured by the GrindoSonic apparatus and by the Ultrasonic Tester present differences between 12 to 52 % in favour of the latter. This can be attributed to the stratification observed in Figure 4. Indeed, the GrindoSonic apparatus provides a fundamental frequency typical of the whole sample. On the contrary, the U1trasonic Tester provides a propagation time dependent of the internal structure of the samples, in particular of the anisotropy induced by the preparation method. That is the reason why we hence recommend to take precautions as for the interpretation of the ultrasonic unidirectional test results. Some experimental results are shown in Figure 5. W e briefly indicate that the rate of discharge of the grout has little influence on the shear modulus. The nature of the soil also appears to have an effect on the shear modulus, but inversely of the unconfined compressive strength, through the stiffness of the grains themselves, their size, their roughness. Finally, for a given soil, the shear modulus and the unconfined compressive strength linearly depend on the cement to water ratio.
Figure 5 . Evolution of the elastic parameters
6 CONCLUSIONS The experimental results described in this paper represent a first step in a more ambitious research program related to the behaviour of grouted sands. Our first preoccupation was therefore to establish a reliable method to reproduce improved soils. The injection procedure was consistent with our expectations in terms of homogeneity and reproducibility that have been controlled by different ways, during the injection as well as after the setting of the cement. We performed some mechanical tests that proved these statements. Further step will be to compare elastic properties of grouted sands prepared in the laboratory and the ones of grouted sands injected in situ by means of sleeved grout pipes.
25
Table 5. Experimental elastic parameters.
2 3 4 5 6 7 8 9 10 11
SF SF SF SF SF SF SF AAM AAG Pure Grout
6 6 6 2.84 8.77 8.77 8.77 6 6
0.172 0.235 0.299 0.172 0.172 0.172 0.172 0.172 0.172 0.172
3.9 5.2 6.1 3.7 3.6 3.6 3.7 4.4 5.2 0.47
2.6 1.6 5.4 23.4 3.6 4.4 2.4 9.8 2.7
9.1 12.4 15.7 8.4 8.6 8.4 8.5 11.1 11.9 1.35
0.19 0.18 0.22 0.15 0.19 0.15 0.16 0.20 0.14 0.44
2603 2895 3087 2397 2503 2545 2568 2778 2790 1870
4.8 2.9 2.1 20.1 2.6 3.2 2.0 6.0 10.9
20.54 20.55 20.68 20.4 1 20.5 1 20.37 20.46 20.62 21.60 11.48
0.8 1.o 1.2 0.2 0.8 0.7 0.8 0.5 0.8
D. Gouvenot 1990. Les nouvelles techniques de reconnaissance et de traitement des sols. Cornptes-rendus des Journkes d ’Etudes Internationales, A R E S , Lille, pp 253-261 M.A. Ismail, H.A. Joer & M.F. Randolph 2000. Sample preparation technique for artificially cemented soils. Geotechnical Testing Journal, Vol. 23, No. 2, pp 171-177 D. Levacher, J. Garnier & P. Chambon 1994. Reconstitution d’kprouvettes de sable - Appareils de pluviation. Revue FranGuise de Gkotechnique, No. 68, pp 49-56 A.E. Miltiadou 1991. Etude des coulis hydrauliques pour la rCparation et le renforcement des structures et des monuments historiques en mqonnerie. Etudes et Recherches des L. P. C., SCrie Ouvrages d’art, OA8 L.G. Schwarz & R.J. Krizek 1994. Properties of cementgrouted sand with distance from injection. Proceedings of the 13th International Corlference on Soil Mechanics and Foundation Engineering, Vol. 1, pp 287-290 L.G. Schwarz & R.J. Krizek 1994. Effect of preparation technique on permeability and strength of cement-grouted sand. Geoteclzniral Testing Journal, Vol. 17, No. 4, pp 434-443 S. Tailliez 1998. Etude expkrimentale du comportement mkcanique des sols granulaires inject&. P1z.D. Thesis, Ecole Centrale de Paris (France) S. Zebovitz, R.J. Krizek & D.K. Atmatzidis 1989. Injection of fine sands with very fine cement grout. Journal of Geotechtzical Engineering, Vol. 115, No. 12, pp 1717-1733
7 ACKNOKEDGEMENTS The authors have appreciated the technical and financial support of the Rkgie Autonome des Transports Parisiens (RATP, France), the Intrafor Company and the Civil Engineering Laboratory of Nantes Saint-Nazaire within the framework of a common research program on the mechanical behaviour of grouted sands. REFERENCES R.J. Allison 1987. Non-destructive determination of Young’s modulus and its relationship with compressive strength, POrosity and density. From M.E. Joizes & R.M.F. Preston (eds), Deformation of Sediments and Sedimentary Rocks, Geological Society Special Publication, No. 29, pp 63-69 R.J. Allison 1988. A non-destructive method of determining rock strength. Earth Suiface Processes and Lnndfonns, Vol. 13, John Wiley & Sons (eds), pp 1-8 L. Arenzana, R.J. Krizek & S.F. Pepper 1989. Injection of dilute microfine cement suspensions-into fine sands. Proceedings of the 12th International Conference on Soil Mechanics arid Foundation Engineering, Vol. 2, pp 1331-1334 G. Azzar 1997. Modtlisation des injections de coulis de bentonite-ciment dans les sols. Ph.D. Thesis, Ecole Nationale SupCrieure d’Arts et MCtiers, Bordeaux (France) 0. Benhamou 1994. Comportement rheologique des coulis de liants hydrauliques ultrafins destinCs B l’injection. Ph.D. Thesis, Ecole Nationale SupCrieure des Mines de Paris (France) A. Bennabi & D. Levacher 1995. Application d’un produit POlymkre B la consolidation des sables carbonatks. Revue Frarzcaise de Gkotechnique, No. 72, pp 55-66 F. Bouchelaghem, L. Laloui, L. Vulliet & F. Descoeudres 1999. Numerical model of miscible grout propagation in deformable saturated porous media. 7th International Syniposium on Numerical Models in Geomechnaics, NUMOG 99, pp 243-248, Grasz (Austria) C. Dano, P.Y. Hicher & S. Tailliez 2000. Laboratory and field study of the behaviour of grouted sands. Submitted to the Journal of Geotechnical and Geoenvironmental Engineering C. Di Prisco, R. Matiotti & R. Nova 1992. A mathematical model of grouted sand allowing for strength degradation. Numerical Models in Geomechanics, Pande & Pietruszczak (eds), pp 25-35
STANDARDS AFNOR NF P 18-414, 1993. Essais non destructifs - Mesure de la frkquence de rCsonance fondamentale. AFNOR NF P 18-418, 1989. Auscultation sonique - Mesure du temps de propagation d’ondes soniques dans le bCton. AFNOR NF P 18-891, 1992. Produits B base de risines synthCtiques ou de liants hydrauliques pour injections dans des structures en bCton - Essai d’injectabilitk B la colonne de sable en milieux sec et/ou humide. ASCE Committee on grouting of the Geotechnical Engineering DiTrision 1980. Preliminary glossary of terms relating to grouting. Journal of the Geotechrzical Engineering Division, Vol. 106, No. GT7, pp 804-815 ASTM D 4320-93, 1993. Standard test method for laboratory preparation of chemically grouted soil specimens for obtaining design.
26
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
A research on the tensile resistance of wired rope anchor H. Fujimura Department of Civil Engineering, Tottori University, Japan
Y. Taniguchi Eito Consultants Company Limited, Tottori, Japan
ABSTRACT: This research is an attempt to propose a new method of stabilizing soil using wired rope anchor as tension reinforcement embedded in soil mass. The proposal is supported by experiments involving the pulling of wired ropes placed in sand. Resistance against pull was measured experimentally using ropes and rods separately. Currently steel rods are usually being used for this purpose. However, during this research, comparative experiments using the conventional steel rods and wired rope were conducted. It was found that wired ropes were more effective than their conventional counterparts.
1 INTRODUCTION
ties; and the parameter varied for sand was its density. In one case, experiments were conducted by using wire ropes or rods, and in another case, ropes or rods were tied together with a wire-netted frame (Figure 1). This wire frame was 38 mm by 60 mm in dimensions, and two types of frame, one with 7mm by 7mm mesh and the other with 9mm by 9mm mesh, were used. During the experiment, a constant surcharge load was applied at top of reinforced sand layer, and tensile force applied at the end of reinforcement bars. Displacement, 6, and pull out force needed to pull out the reinforcement, Pt were measured.
I n civil construction, anchors of various forms and material are in use for various purposes such as strengthening of retaining walls, supporting cut slopes, stabilizing weak soils etc. It is important to consider the effectiveness of the anchor type and materials selected for a particular application. Studies on the use of wired rope anchor as reinforcement in embankments are almost non-existent. This study tries to fill this gap. Laboratory experiments were conducted to examine the effectiveness and safety of using wired rope anchors as reinforcement in soil embankment. The experiment was conducted in sand. Separate sets of tests were performed with wired ropes and steel rods as soil reinforcement. Parameters describing the wired ropes or rods were their number, arrangement, size and shape of cross-section, and material proper-
2 METHODOLOGY 2. I Experiment set up Sand was placed in a container, made of acrylic acid resin, with inside dimension 570 mm long, 200 mm high and 50 mm wide. After placing sand to about half of container depth in case of using ropeshods (Figure 3(a), 3(b)), or one-third of container depth in case of using wire frame, the ropehod or frame was placed over sand and more sand poured to fill up the container. Sand was poured into the container in three layers of equal depth, each layer being compacted by 20 blows of 660 g hammer before another layer was poured. Wire ropes were placed in sand mass as shown in Figure 2. Figure 3 shows the schematic layout of the experiment set up. Initialy,the embedded length of rope or rod was 50cm. In case (a), a single rope; in case (b), three ropes; and in case (c), four ropes in two layers were placed.
Figurel. Anchor structure.
27
varying diameters of 1.0, 1.5, and 2.0 mm of ropes and rods. 2.3 Physical properties of material Sand used in the experiment was sampled at Tottori sand dune. Its physical properties are shown in table 1 below. Table 1. Properties of sand. Sample Tottori sand dune
I Max.grain.size I 0.85mm
D50 0.32mm
I U,
I Unit weight
I 23.2kN/m3
I 2.00
Grain size distribution is shown in Figure 4. The tensile load vs elongation relations for wire rope of different diameter used in this experiment are shown in Figure 5. The relationship is linear up to a load of around 23.8N.
Figure 3. Rope layout.
2.2 Loading and measurement
A 5 kg constant surcharge load was applied to sand mass through a plate placed on top. Under this loading, tensile force was applied to ropes or rods through rotary handle, resulting in the gradual pulling out of the rope at a rate of 2.0 &min. Displacement (6) and the corresponding tensile force (P,) were measured during the interval of pulling out. The measurement were taken at a maximum pull-out length of 20 mm. These measurements were auto recorded at data logger. measurements were taken with varying sand density of 1.60g/cm3 and 1.71g/cm3, and with
Figure 5. Elongation property of wire undertension.
28
Figure 6. Relation between displacement of rope and tensile load.
Figure 7. Relationship between displacement of wirekteel rods and tensile load, for varying number of ropes/rods.
3 RESULTS AND DISCUSSION
3.2 Comparison with respect to number of ropes/rods
Experiment results obtained with different cases were compared as follows.
In this case, for a sand density of 15.7 kN/m3 and for all diameters, wire rope exhibited an increasing resis-tance with increasing number of ropes. In the case of steel rod, the resistance was smaller when four rods were used than with only three rods.This pattern in the case of steel rods is thought to have been due to the difference in rod spacing (see Figure 3) in the two cases. Figure 7.
3.1 Comparison with respect to reinforcement type Case a) Single rope/rod: In this case, pull out resistance was higher with wire rope than that with steel rod, and this difference was higher for larger diameter ropedrods and for denser sand. This is shown in Figure 6. Case b): Three ropes/rods: As in case (a), resistance was higher for rope. However, this pattern was found only for a rope/rod diameter of lmm. For 1.5 and 2.0 mm diameter wire ropes, there was no difference in resistance with variation in sand density. Case c): Four ropes/rods (in two 1ayers):The result in this case was almost similar to case (b) mentioned above. The reason for the difference in resistance behavior between wire rope and steel rod is the difference in surface friction. Compared to steel rod, surface roughness is high in wire rope due to the twisting of rope strands. Coefficient of form unevenness (FU) was calculated for rope and rod, using the relation:
3.3 Comparison with respect to diameter Case a): Single ropehod: In case of wire rope, resistance was higher for larger diameter. On the other hand, no difference in resistance was noted with varying diameter. Case b): Three ropes/rod: In this case, resistance pattern for wire rope was same as in case (a). However, for steel rod, resistance was higher with larger diameter. Case c): Four ropes/rod: In this case, resistance was higher with larger diameter for wire rope as well as steel rod. This pattern can be explained by the fact that larger diameter ropes/rods will result in bigger surface area in contact with sand. This would mean stronger friction. Figure 8.
FU= 4 ~ ~ ~ 1 1 ~ Where, a: cross-section area I: cross-section perimeter.
3.4 Comparison with respect to use of anchor net
This coefficient, FU, will have a value of 1.O for a perfectly even one, and less than 1.0 for uneven cross-section. A smaller value would mean higher degree of unevenness FU values of wire rope and steel rod were obtained as OS6 and 1.0 respectively. This difference in FU values has a direct influence upon the difference in pull out resistance mentioned above.
When anchor net was used, the resistance was higher which is obvious i.e. ropedrods would get anchored at the net making them difficult to pull out. However, the effect of grid size of the anchor net was observed. Resistance was higher when mesh size was smaller. As shown in Figure 9 when anchor net was used, resistance increased continuously with is placement. When it was not used, there was a peak after which resistance stayed almost uniform. 29
Figure 9. Relation between displacement and tensile load usinghot using anchor net
Figure 8. Relation between displacement of steel rod and tensile load.
3) In general, resistance was higher for larger number of ropes or rods. However, in case of steel rods, resistance was higher with three rods than with four rods. 4) Resistance was higher when an anchor net was attached to the ropes or rods.
3.5 Comparison with respect to sand density In all cases, stronger resistance was observed when a denser sand was used in the experiment.
4 CONCLUSION 1) Compared to steel rod, resistance against pulling was higher in case of rope. 2) Resistance was higher for larger diameter of rope as well as rod.
REFEWNCES Japanese Geotechnical Society, 1980. Basic course instructure und fouizdation:286-291. Japanese Geotechnical Society, 1982. Design nzethodof retairzing structures:31 1-3 18.
30
Landmarks in Earth Reinforcement, Ochiai et al (eds), 0 2001 Swets & Zeitlinger, ISBN 90 265 1852 8
Experimental evaluation of the factors affecting pull-out test results on geogrids V.N. Ghionna “Mediterranea ” University of Reggio Calabria, Italy
N. Moraci “Mediterranea University of Reggio Calabria, Italy
P. Rimoldi Tenax S.p.A. - Milano, Italy ABSTRACT: A new pull-out test apparatus has been designed, which is similar in dimensions to the ones already under discussion (that is 1.5 m x 0.6 m x 0.7 m) in different Standardization Bodies, but with peculiar features in terms of measurement of the internal displacement along the reinforcement and for a special clamp which is placed inside the soil in order to investigate the confined failure in pull-out condition. Tests have been performed on a granular soil and an HDPE extruded geogrids. The discussion of the results of this vast research program allows to evaluate the influence of test parameters on pull-out test results.
1 INTRODUCTION
nal development of an international test standard, which could provide performance values for the design of geogrid reinforced soil structures.
The apparatus and procedure for pull-out tests on Geosynthetics are currently under discussion in several Standardization Bodies all over the world. Pull-out tests are necessary in order to study the interaction behaviour between soil and Geosynthetics in the anchorage zone, hence these properties have direct implications in the design of reinforced soil structures. To be really usable for design, pull-out tests shall be performed in such a way as to reproduce as close as possible the actual conditions that a geosynthetic undergoes when embedded in soil in a reinforced soil structure. In order to provide a positive contribution to the development of a proper standard, which could be used for design purposes, a research program has been carried out for investigating the influence of the main parameters on pull-out test results on geogrids embedded in granular soils. Therefore a new test apparatus has been designed, which is similar in dimensions to the ones already under discussion (1.5 m x 0.6 m x 0.7 m), but with peculiar features in terms of measurement of the internal displacement along the reinforcement and for a special clamp which is placed inside the soil in order to investigate the confined failure in pull-out loading condition. More than 25 tests have been performed, varying both the width of the specimens and their length, at vertical confining pressures equal to 10, 25, 50 and 100 kPa. The discussion of the results of this research program allows to evaluate the influence of different test parameters on pull-out test results. Therefore the results of the research can be very useful for the fi-
2 TEST EQUIPMENT AND PROCEDURE The test apparatus, shown in Figure 1, is composed by a pull-out box, a vertical load application system, a horizontal force application device, a special clamp, and all the required instrumentation. The box dimensions and the clamping device were designed on the base of the interaction mechanisms between soil and geogrid in pull-out loading conditions (Palmeira and Milligan, 1989; Farrag et al., 1993). The height of the box (640 mm) comes from the shape of the passive failure surface, which is supposed to involved a soil thickness equal to 40 times the thickness of the geogrid bars. The need to avoid wall effects at the front of the box suggested to use smooth steel profiles (sleeves), 200 mm long (Fig. l), connected to the front wall.
Figure I . Scheme of the test apparatus.
31
Figure 2. Schematic plan view of the test apparatus.
The width of the box (600 mm) comes from the need to include an adequate number of geogrid ribs in order to reproduce properly the behaviour of the reinforcement when embedded in the soil. Finally the box length (1500 mm) was selected with the aim of investigating the soil - geogrid interaction for different reinforcement length, particularly at low confining pressure, when the potential of geogrid pull-out is more evident. For high confining pressures, instead, the failure mechanism becomes the tensile rupture of the geogrid. The confined failure has been studied by using a clamp placed inside the soil, beyond the sleeve, in order to keep the geogrid specimen always confined in the soil during the whole test duration (Fig. 2). The vertical load has been applied through a hydraulic jack, controlled by a servo-hydraulic multi axial closed loop digital controller type Instron 8580. The hydraulic jack push against a rigid steel plate, which lays on water filled cushion, used to distribute the load evenly on the soil surface. The horizontal force has been applied through another hydraulic jack, controlled by the same Instron 8580, connected to the clamp through a chain. During the test the following parameters have been measured and recorded (by the Instron digital controller): vertical load, vertical displacement, pullout displacement rate, pull-out force, displacement of the transversal bars of the geogrids in at least six different points. This last measurement has been made through inextensible steel wires, cased into rigid PVC micro tubes, connected to the geogrid bars at an end and to the electrical displacement transducers at the other end, outside the pull-out box. The transducers are connected to the Instron controller, which digitally records all the data at defined time intervals. This system allows to follow the pull-out dynamics along the whole specimen length (Fig. 3). All tests have been performed on one type of HDPE extruded mono-oriented geogrid (Tenax TT 090 SAMP). Wide width tensile tests (EN IS0 10319) on this geogrid have been carried out at different displacement rate (1, 10, 100 mdmin); the test results are reported in Figure 4 and Table 1.
Figure 3. The electrical displacement transducers.
Figure 4. Tensile tests at different displacement rate. Table 1 . Results of tensile tests at different displacement rate.
V (mm/min) I00 10 1
T (&=2%) (kN/m) 29.47 23.33 18.93
T (&=5%) (kN/m) 53.85 43.27 35.97
Tmax (kN/m) 95.19 79.96 66.73
The soil was tested for the main geotechnical parameters: results of classification tests indicate that the soil is a medium sand (SP according to USCS system), with uniformity coefficient Cu=D6o/D 10 =I .5 and average grain size Dso=0.22 mm. The Standard Proctor compaction test indicates a maximum dry unit weight ?/dmax= 16.24 IdV/m3 at a water content wept= 13.5%. Direct shear tests, performed at an initial unit weight equal to 95% of Ydmax (obtained at a water content of 9.3%), yield very high single values of the peak shear strength angle QtP, in the range between 48" and 42", where the higher values correspond to the lower confining pressures, while the shear strength angle at constant volume $'cv results equal to 34". The pull-out test procedure was the following: 1) preparation of the surfaces of the pull-out box: in order to minimize the friction between the soil 32
Table 2. Pullout test results.
and the box, all the box walls have been covered with adhesive Teflon film; filling and compaction of the soil in the lower half of the box: the soil, previously dried in oven at 105 "C for 24h, has been prepared at a water content of 9.3%; after it was laid in the box in 100 mm thick layers and manually tamped till a final thickness of 0.265 m; positioning of the clamp and connection to the geogrid specimen; the parallelism of the specimen with the box length and the perfect horizontality have been carefully checked; a small preload has been applied to the geogrid in order to avoid any waving; insertion of the inextensible wires into the PVC tubes and connection to the geogrid bars and to the electrical transducers; filling and compaction of the soil in the upper half of the box, for a final thickness of 0.265 m; placing of the water filled cushion and the steel plate on top of soil; connection of the jacks and the instrument to the Instron controller; setting of the horizontal testing speed and the vertical applied force, and starting of the test. More than 25 tests have been performed, varying both the width and length of the specimens. The a p plied vertical pressures were equal to 10, 25, 50 and 100 Wa. The horizontal displacement rate has been equal to 1.0 mm/min for all tests. All tests have been performed, until geogrid rupture or till a total horizontal displacement of 100 mm i n this way the geogrid specimen remains always confined in the soil for its whole length. The friction between the clamp and the test soil has been evaluated, for each confining stress, by performing pull-out tests on the clamp embedded in soil without the geogrid specimen. The pull-out force values obtained have been subtracted, at each displacement level, from the pull-out forces measured i n the tests with the geogrids at the same displacement and at the same confining stress.
Confining stress 0,' (Wa) 10 25 50 100 10 25 50 100 10 25 50 100 10 25 50 100 10
Specimen lenght LR(m) 1.15 1.15 1.15 1.15 1.15 1.15 1.15 1.15 0.90 0.90 0.90 0.90 0.90 0.90 0.90 0.90
Specimen width W (m) 0.40 0.40 0.40 0.40 0.58 0.58 0.58 0.58 0.40 0.40 0.40 0.40 0.58 0.58 0.58 0.58
Maximum Pullout Force Tmax,, (kN/m) 15.80 32.18 55.08 Geocrid ruDture 14.43 25.70 50.58 Geogrid rupture 12.84 29.06 43.79 65.60 9.77 24.83 37.64 57.22 "
0.40
0.40
4.87
0.40 0.40 0.40 0.40 0.40
0.40 0.40 0.40 0.58 0.58
13.59 22.99 29.52 4.68 10.93
50
0.40
0 58
15 94
100
0.40
0.58
27.69
25 50 100 10 25
I
These values are very close to the tensile strength (66.73 kN/m) obtained by the wide width tensile tests performed at the same rate (v = 1 m d m i n ) of the pull-out tests (Tab. 1). Being the short term tensile properties of geogrids practically unaltered by the soil confinement, at least for the type of soil and reinforcement used in this research, there is now the need to evaluate whether the long term strength, commonly used for the design of reinforced soil structure, remains unaltered as well when the geogrids are embedded in a compacted soil. Since the confined condition produces a variable tensile stress level along the reinforcement due to the reinforcement extensibility and to the different interaction mechanisms (Moraci and Montanelli, ZOOO), in-soil creep tests may produce substantially different results than in-air creep test, where the ten-
3 EXPERIMENTAL RESULTS The results obtained in the first phase of this research are synthetically reported in Table 2. Thanks to the clamp inserted into the soil the mechanism of pull-out failure, with the specimen constantly and wholly confined by the soil, has been properly studied. From Figure 5, representing the pull-out force versus the first bar displacement for the two tests carried out with the same specimen length, LR= 1.15 m, but different specimen width (i.e. W = 0.40 m and 0.58 m) at a vertical pressure of 100 kPa, it can be noted that tensile rupture of the geogrid occurred at 66.55 kN/m tensile force, for W = 0.40 m, and at 62.27 kN/m for W = 0.58 m.
Figure 5. Tensile rupture in pull-out loading conditions.
33
sile stress is constant along the whole specimen. Recent studies on instrumented reinforced walls seems to confirm this fact (Carrubba et al. 2000). The apparatus here presented seems to be suitable to run such long term in-soil pull-out tests in laboratory controlled conditions. Analysing the pattern of the pull-out resistance versus the displacement of the first embedded geogrid bar (Fig. 6), it is evident that the interface behaviour is strongly influenced by the embedded geogrid length. In fact, tests performed on “long” specimens (LR = 0.90 m and LR= 1.15 m) show a strain hardening behaviour, with a progressive increase of the pullout resistance with the increase of the displacement of the first embedded geogrid bar; while tests on “short” specimens (LR=0.40 m) show a strain softening behaviour, with a progressive decrease of pull-out resistance after the peak. It is therefore possible to say that the pull-out interaction mechanism is progressively developed along the long reinforcement layer, while it is developed almost at the same time along the whole length of short layers. In this latter case the pull-out resistance shows a pattern similar to stress-strain curves of compacted soils. Such result is confirmed by the analysis of the bar displacements along the specimen for different ap plied tensile forces. Here it is possible to observethe two different phases that characterize the pull-out of
the reinforcement from the soil: first the tensile force is progressively transferred to the geogrid (Fig. 7), until the whole embedded length become under tension and also the last point of the geogrid start to be displaced; in the second phase (Fig. 8) the pull-out resistance increases until a peak (for short reinforcements) or up to maximum pull-out force or up to geogrid tensile rupture (for long specimens). In details, there is a non-linear distribution of the displacements for long geogrid layers (LR= 0.90 m e LR= 1.15 m), hence a markedly non linear interface behaviour, due to the deformability of the reinforcement and to the various other factors affecting the complex interaction between soil and open grid structures (Moraci and Montanelli, 2000). With short layers (LK=0.40 m), after the “loading” phase (Fig. 9), displacements become practically constant along the geogrid (Fig. lO), therefore showing a similar behaviour to very stiff reinforcing elements like steel bars. Comparing the tests with different specimen width (Fig. l l ) , we can note that the pull-out resistance is higher for narrow specimens, at equal vertical pressure and specimen length: the difference of pull-out resistance between the 0.40 m wide and the 0.58 m wide specimens vary from 4 % (for short length and high pressure) to 43 % (for long length and low pressure). This differences can be explained by the effects of constrained dilatancy at the geogrid
Figure 7. First phase: tensile force transfer.
Figure 6. Pull-out curves: a) long specimen; b) short specimen.
Figure 8. Second phase: pull-out.
34
soil interface along the geogrid edges: such effects have large influence on specimen narrower than the box width. For 0.40 m wide specimens the tendency of the soil to dilatancy develops a three-dimensional effect (Hayashi et al.1996). In fact, the non dilating zone in the soil surrounding narrower geogrid specimens (zone a in Fig.12) behave as a restrain against soil dilatancy in the dilating zone (zone b in Fig.12). This in turn generates shear stresses at the border between the two zones and produces an increase of the effective normal stress on the soil geogrid interface and, consequently, an increase of pull-out resistance. By increasing the specimen width such effect is reduced because the soil area that block the dilatancy decreases, and the shear stresses cannot be generated anymore, thanks to the smoothness of the box walls lined with Teflon film (Fig. 12). The pull-out factor fpo can be calculated directly from the tests results: it depends clearly on anchorage length and width, on the vertical pressure, and on the value assumed for the shear strength angle of the soil. In Figure 13 the experimental results are compared with values obtained by Jewell (1984 ,1985) and Matsui (1996), through theoretical approaches: it can be noted that the theoretical approaches tend to underestimate fpo. In particular, with the use of the constant volume shear strength angle, suggested by Jewell, (1984, 1985), the theoretical expressions yield constant values of fpo, independent of the geogrid length and the vertical pressure, which is in contrast with experimental evidence. This is probably due to the Fact that the theoretical expressions proposed so far don’t take into considerations all the factors influencing the pull-out phenomenon. A correct way of evaluating fpo would be to use values of the soil shear strength angle related to the reinforcement behaviour and to the applied vertical effective stress. Considering the two different modes of pull-out resistance mobilization, it seems more appropriate to evaluate the pull-out factor fpousing values of the shear strength angle close to the peak one (evaluated for the applied vertical effective stress), for short reinforcement layers, and intermediate values between the constant volume and the peak shear strength angle for long geogrid layers.
Figure 9. First phase: tensile force transfer.
Figure 10. Second phase: pull-out.
Figure 12. Scheme of the interaction for narrow and wide specimens.
Figure 11. Pull-out curves for specimens of different width.
35
5 ACKNOWLEDGEMENTS The authors are grateful to eng. D. Gioffrk and eng. G. Romano for their cooperation to the experimental work reported in the paper and for the cooperation in the preparation of this paper. This research program was carried out with the financial support of MURST (Italian Ministry for University and Scientific and Technical Research). REFERENCES Figure 13. Pull-out factor v s . ~ ’ ,at different specimen lenghts.
Carrubba P., N. Moraci, F. Montanelli 2000. Long-term behaviour of an instrumented wall reinforced with geogrids. Second European Geosynthetics Conference: Eurogeo 2000: 125-129. Bologna: Patron. EN-IS0 10319 1992. Geotextile Wide-Width Tensile Test. International Organization for Standardization, ISO, Ginevra. Farrag K., Y.B. Acar, I. Juran 1993. Pull-out Resistence of Geogrid Reinforcement. Geotextile and Geomembranes, 12: 133-159. England: Elsevier. Gioffri: D. 200 1. Studio sperimentale dell’interazione terrenorinforzo in condizioni di sfilamento. Degree Thesis in Civil Engineering. “Mediterranea Univ. of Reggio Calabria, Italy. Hayashi S., M.C. Alfaro, K. Watanabe 1996. Dilatancy effects of granular soil on pullout resistance of strip. Proc. Earth Reinforcement, Osaka, Japan: 39-44. Rotterdam: Balkema. Jewell R.A., G.W.E. Milligan, R.W Sarsby, D.D. Dubois 1985. Interactions Between Soil and Geogrids. Proc. from the Symposium on Polynier Grid Reinforcement in Civil Engineering : 18-30. London: The Institution of Civil Engineers. Matsui T., K.C. San, Y. Nabeshirna, U.N. Amin 1996. Bearing mechanism of steel reinforcement in pullout test. Proc. Earth Reitgorcement, Osaka, Japan: 10 1-105. Rotterdam: Balkema. Moraci N., F. Montanelli 2000. Analisi di prove di sfilamento di geogriglie estruse installate in terreno granulare compattato. Rivista Italianu di Geotecnica 4: 5-21. Bologna: P d tron. Palmeira E.M., G.W.E. Milligan 1989, Scale and Other Factors Affecting the Results of Pull-out Tests of Grid Buried in Sand. Geotecliinique 39, 3: 5 1 1-524. London: Thomas Telford. Romano G. 2000. Prove di sfilamento su geogriglie di rinforzo installate in terreno granulare compattato. Degree Thesis in Civil Engineering. Uiiiv. c,fReggio Calabria, Italy.
Finally it must be noted that tests performed with the specimen width narrower than the box width yields values of fpothat are higher (hence less conservative) than the ones obtained with specimens covering the full box width. 4 CONCLUSIONS The main conclusions of the present work are the following: In the experimental conditions the influence of soil confinement on the short term tensile strength of geogrids is negligible; The interface behaviour is strongly influenced by the embedded geogrid length: tests performed on “long” specimens show a strain hardening behaviour, while tests on “short” specimens show a strain softening behaviour; For the design of reinforced soil structures, where generally the geogrids cover the whole horizontal area, pull-out tests shall be performed with specimen width equal to the box width, in order to avoid three-dimensional effects that cannot occur in reality; Theoretical expressions (Jewell, 1984, 1985; Matsui et al. 1996) tends to underestimate the pullout factor fpo;a proper evaluation of fpo,shall be made on the base of pull-out test results using the appropriate value of the soil shear strength angle.
”
36
Landmarks in Earth Reinforcement, Ochiai et a/ (eds), 0 2001 Swets & Zeitlinger, ISBN 90 265 1852 8
Residual strength and its application to design of reinforced soil in seismic areas John H. Greenwood ERA Technology Ltd, Leatherhead, United KiJi gdoJn
Colin J.F.P. Jones University of Newcastle, Newcastle upon Tyne, United Kingdom
Fumio Tatsuoka University of Tokyo, Japan ABSTRACT: In designing for seismic applications it is critical to know the strength of a reinforcing geosynthetic at all stages of its service life. Residual strength tests performed on two polyester geosynthetics using the stepped isothermal method and conventional creep-rupture testing demonstrated that the strength of the geosynthetic is retained as far as the creep-rupture region. The modulus appears to increase. It is argued that, in contrast to current design methodologies, design should be based upon factored lifetime rather than on factored load. life. The design load is equal to this unfactored strength divided by a safety factor to allow for the variability in material properties. Under this lesser load the tensile strength of the geosynthetic remains at a higher level up to and beyond the end of the design life, Figure l . The ratio of the strength of the geosynthetic to the design load is thus higher than the intended safety factor. The structure is overdesigned. This material behaviour is not recognised in many design codes. Most geosynthetic reinforced soil structures are designed using stress-rupture curves, Figure 2. These do not recognise the existence of residual strength. The difference between Figures 1 and 2 in respect of design philosophy and the assumed and actual design strength of the structure is profound.
1 INTRODUCTION At the 1996 Kyushu meeting one of us presented a discussion contribution concerning residual strength (Greenwood 1997), which was developed into a further publication (Greenwood 1998). The point made was that the creep-rupture diagram depicts sustained load against the lifetime under that load. It is not a diagram of reduction in strength against time, even though this may appear to be so. The strength of a geosynthetic is in fact maintained until late in its service life. This was demonstrated by Orsat et al (1998). The unfactored strength derived from the stressrupture diagram is the sustained load which is predicted to lead to failure on the last day of the design
Figure 1. Schematic diagram showing the reduction in strength with time of a geosynthetic under a sustained design load. The unfactored strength is reduced by a safety factor to give the design load. The residual strength at the design life is now much greater than anticipated.
37
Figure 2. Current design assumptions relating to geosynthetic reinforcement under a sustained design load.
Many soil structures are designed not just to retain a margin of safety under a sustained load due to soil loading, but also to withstand occasional higher loads. Nowhere is this more important than in seismic loading. For effective seismic design, it is essential to know how the soil reinforcement will react to the additional seismic load. In recent earthquakes reinforced soil structures have proved highly stable, Tateyama et aZ(l995). During seismic conditions a short term increase in the design strength of the reinforcement is accepted, Jones (1996). The increase in required strength to counteract seismic forces can be in the order of 50100 percent of the design strength. Inspection of Figure 2 suggests that an increase in the design strength of polymeric reinforcement from, say, 40 percent of the characteristic strength (the manufacturer's guaranteed tensile strength) to 60/80 percent could be accommodated early in the design life, but could not be sustained late in the life of the structure. This raises concern over the long term viability of geosynthetic reinforced soil structures. This concern is resolved if the residual strength of the reinforcement is considered, Figure 1. The purpose of this work was to determine the residual or reserve strength of a number of commercially available soil reinforcements. Since tests at room temperature must either be performed at very high loads or for very long times if they are to produce useful data, time-temperature acceleration was used in addition to conventional room temperature
testing. For this purpose the stepped isothermal method (SIM) proved ideal. This paper describes the materials used, the methods applied, and the results of the tests. It also makes some suggestions for the manner in which safety factors should be applied in future.
2 MATERIALS Several different soil reinforcements were used in this study, of which the three listed in Table 1 have been selected for the purposes of illustration in this paper. 3 METHODS 3.1 Grips and extensometry All tests on geogrid R1 were performed on single ribs held in 50 mm diameter roller grips. If the rib was wound round the roller by more than one turn, a strip of nonwoven material was inserted to prevent the rib catching on itself. The tests were performed using SIM which is described in more detail in the next section. Lengths of strip R2 were held in large roller grips with knurled surfaces to increase the friction between the grip and the polyethylene sheath, and tested in a room controlled to 20 rt 2"C, 65 f 5% relative humidity.
Table 1. Materials selected Tensile strength 58.4 ? 1.7 kN/m
Elongation at break (%) 1 1.9 + 0.6
strip consisting of polyester yam bundles sheathed in low density polyethylene
40.9 2 0.4 kN
12.6 2 0.3
polyester yam bundles stitched to a nonwoven polypropylene backing
45.0 2 0.6 kN/m
11.1 2 0 . 3
RI
coated polyester geogrid
R2 R3
38
The method has been validated against ERA’s long-term tests for polyester reinforcements (Thornton et al, 1998b). An example of ERA’s measurements together with comments on the method were presented by Greenwood and Voskamp (2000). In these tests the creep was accelerated using SIM but the temperature of the geosynthetic was reduced to the starting temperature of 20°C for the measurement of residual strength. Load was applied to the specimen without interruption.
Extension was measured by a pair of linear variable differential transformers (LVDTs) parallel to the loading axis but placed at opposite corners to compensate for any rotation of the extensometer mounting. The extensometry was calibrated at the relevant test temperatures. 3.2 The stepped isothermal method The stepped isothermal method (SIM) was developed by Thornton and co-workers (1998a). The temperature of a conventional creep test is increased in steps, using a programmable oven whose temperature control is such that the change occurs within minutes. The sections of creep curve are plotted as creep modulus (loadstrain) against the logarithm of the time after the temperature change and are then shifted along the log (time) axis. Corrections, which allow for shrinkage of the fibre on heating and for the thermal history of the sample, enable the sections of curve to be aligned to form a smooth continuous master curve. Thanks to the high level of time-temperature acceleration for polyester fibres - increasing the temperature by 10°C speeds up the rate of creep by a factor of about 8 - and the fact that even as high as 90°C the basic mechanism of creep is unchanged, durations as long as the service life of a reinforced soil structure, typically 75-120 years, can be simulated in less than a day’s testing.
4 RESULTS 4.1 Tensile strengths The tensile strengths and elongations at break were measured to I S 0 10319 but on the specimen widths described. The results, which are used as the basis for the creep and creep-rupture tests, are presented in Table 1. 4.2 Creep-rupture Figure 3 shows the creep-rupture curve for geogrid R1 using SIM plotted as percentage of tensile strength in Table 1 against the logarithm of time to failure in h. The load leading to failure after 106
Figure 3. Creep rupture and residual strength of polyester geogrid RI.
39
(114 years) is 69.2% of tensile strength. Figure 4 shows a creep-rupture curve derived from earlier measurements on an earlier sample of strip R2 (Greenwood, Kempton et al, 2000), superimposed by several measurements made using R2 itself. The load leading to failure after 106hours (1 14 years) is 68.6% of tensile strength. Both diagrams show the creep-rupture curve with its upper and lower (two-
sided) 90% confidence limits, together with the range of tensile strengths with the same confidence limits inserted at the left hand edge of the diagram. The measured ruptures for R2 agree with the creeprupture curve for the similar strip. Figure 5 shows the creep-rupture curve for geosynthetic R3 measured using SIM.
Figure 4. Creep rupture and residual strength of polyester strip R2.
Figure 5. Creep rupture and residual strength of geosynthetic R3.
40
Three measurements were made of the strains during measurement of residual strength at room temperature. One, performed on R1, showed that the additional strain during the residual strength measurement was 1.5%; two on R2 showed additional strains of 1.5% and 1.0%. Considerably larger strains would have been expected from the stress-strain diagram. These results indicate that during the period under sustained load the modulus of the polyester increases, leaving a lower strain margin available in response to any additional seismic load. This is shown schematically in Figure 6 for a high sustained load. The increase in modulus would not be expected to be as pronounced after exposure to a lower load.
4.3 Residual strength Residual strength measurements on geogrid R 1 were performed by loading a specimen at 69.2% of tensile strength, performing a SIM test to a particular simulated lifetime, stopping the test, cooling the sample under the same load to room temperature, and measuring the tensile strength. The results are shown in Figure 3. The sustained load is shown as a horizontal bar. When the durations of tests under sustained load extend into the creep-rupture scatter band, some tests fail before their residual strength can be measured. This was expected and additional tests were included in the test plan to allow for it. Residual strength tests were performed on strip R2 at a load of 76.6% of tensile strength, corresponding to a lifetime of 1500 h. The results are plotted in Figure 4 in the same way. Figure 5 shows the results for geosynthetic R3. The results show that the residual strength is retained over the lifetime of the material with little detectable reduction.
5 DISCUSSION The results show that in response to an additional load the full strength of a geosynthetic is available but that the strain response may be less than predicted from the original stress-strain curve. The residual strengths all lie above the lower confidence limit of tensile strength and/or to the right of the lower confidence limit for creep-rupture for a particular load, such that if these two limits are used in design no further determination of residual strength is necessary.
4.4 Strain at rupture Analysis of the strains at creep-rupture strains showed no clear dependence on applied load. The strains at rupture for R1 averaged 13.1 rt O S % , marginally higher than the 11.9 rt 0.6% elongation at break in the tensile tests. The strains at rupture for R2 averaged 11.2 rt 1.0%, rather lower than the 12.6 t 0.3% elongation at break in the tensile tests.
Figure 6. Schematic diagram showing the strain response of strip R2 after a period under high sustained load (creep test) followed by determination of residual strength. The strains in the creep test differ marginally from those in the tensile test because the loading i s slower. The durations illustrate that even at this high load (77% of tensile strength) the post-construction strain is small compared with the strains on loading.
41
The use of safety factors based on material strength is fundamental to limit state design. It has been shown here that the strength of the reinforcement is retained as far as the creep-rupture region. Since there is no gradual reduction in strength, the application of a reduction factor to design strength is not appropriate. It is more appropriate to predict the lifetime and apply a reduction factor to time (or the logarithm of time) based on the statistical likelihood of premature failure. If the life predicted from the creep-rupture diagram is 200 k 50 years and a 95% one-sided confidence limit is required, then the geosynthetic should not be relied on to last for more than 200 - (1.64 x 50) = 1 18 years, a factor of safety on lifetime of 1.7. If this approach is applied to design based on continuous sustained loading, i.e. on creep-rupture, it will lead to the same reduction in applied load regardless of whether the safety factor is applied to the load or time axis. If however the design is for residual or reserve strength, then the requirement is that under the applied load the full strength of the material should be retained over the design lifetime. The design life for that applied load, and its standard deviation, should be determined from creeprupture measurements as before. A service life should be calculated, an appropriate reduction factor applied as in the example above, and the design load calculated accordingly. As far as the geosynthetic is concerned no further reduction factor is then required. The same approach based on lifetime rather than reduction in strength applies to any form of degradation that occurs abruptly, for example oxidation when the antioxidant is exhausted.
It is proposed that the design load should be based on lifetime predictions alone and should not include any further factorisation.
7 ACKNOWLEDGEMENTS We thank Engtex NB, Linear Composites Ltd, Polyfelt Ges.m.b.H, Teletextiles N S , Tensar International, and Terram Ltd for supporting this work. J H Greenwood thanks J M Palmer for performing the measurements and the Directors of ERA Technology Ltd for permission to publish. C J F P Jones thanks Dr S Glendinning and L Moore. REFERENCES Greenwood J H. 1996. Residual strength: an alternative to stress-rupture for earth reinforcement. International Symposium on Earth Reinforcement, Fukuoka, Kyushu, Japan, published as Earth Reinforcement, eds Ochiai H, Yufuku N, Omine K. Balkema, Rotterdam: 1081-1083. Greenwood J H. 1998. Designing to residual strength of geosynthetics instead of stress-rupture. Geosynthetics International, 4(1): 1-10. Greenwood J H, Kempton G T, Watts G R A, Bush D I. 2000. Twelve year creep tests on geosynthetics reinforcements. Proceedings of the Second European Geosynthetics Con~ference,Bologna: 333-336. Greenwood J H, Voskamp W. 2000. Predicting the long-term strength of a geogrid using the stepped isothermal method. Proceedings of the Second European Geosynthetics Coizference, Bologna: pp 329-33 1. Jones, C J F P. 1996, Earth reinforcement and soil structures. Thomas Telford: 379. Orsat P, Khay M, McCreath M K . 1998. Study on creep-rupture of polyester tendons: full scale tests. 6"' International Con,ference on Geosynthetics, Atlanta, USA: 675-678. Tateyama, M, Tatsuoka, F, Koreki, J and Harii, K. 1995. Damage to soil retaining walls for railway embankments during the Great Hanskui-Aweji Earthquake, January 17 1995. 1st International Conference on Earthquake Geotechnical Engineering, IS-Tokyo '95, Tokyo. Thornton, J. S., Allen, S. R., Thomas, R. W., Sandri, D. 1998a. The stepped isothermal method for time-temperature superposition and its application to creep data on polyester yam. 6"' International Coizfereizce on Geosynthetics, Atlanta, USA: 699-706. Thornton, J. S., Paulson, J. N., Sandri, D. 1998b. Conventional and stepped isothermal methods for characterising long term creep strength of polyester geogrids. 6"' International Conference on Geosynthetics, Atlanta, USA: 69 1-698.
6 CONCLUSION The results show that in response to a seismic load the full characteristic strength of a geosynthetic is available but that the strain response may be less than predicted from the original stress-strain curve. The reduction in strength with time implied by Figure 2 is misleading, and geosynthetic reinforced soil structures are safe against seismic loads which occur late in their design life.
42
Landmarks in Earth Reinforcement, Ochiai et al (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 852 8
Safe and economical soil reinforcement using a new style geogrid G. Heerten Naue Fasertechnik GmbH & Co. KG, Liibbecke, Germany
R. Floss Lehrstuhl und Priifamtf i r Grundbau, Bodenmechanik und Felsmechanik, TU Miinchen, Germany
G. Brau Lehrstuhl und Priifamtfur Grundbau, Bodenmechanik und Felsmechanik, TU Miinchen, Germany
ABSTRACT: It's a vision of today that in future the soil reinforcement with synthetic geogrids will be as common as traditional measures. This trend is supported by economical and ecological advantages. By developing this technology also new products are entering the market. This paper will give the background of the development of a 'hew style geogrid" and is presenting test results from pull-out tests, miniature steep slope tests and first field application experience.
1 INTRODUCTION
3 DIFFERENT GEOGRIDS
The reinforcement of soils and the construction of soil reinforced structures is a big challenge for the future of earth construction activities. Different products with different properties are offered on the market. All these products have to reinforce soil structures and therefore to fulfil1 some basic requirements which are explained below.
Stretched, woven and laid geogrids can be distinguished (FGSV 1994). Up to now stretched geogrids are made of extruded synthetic sheets (polyolefine raw material). Holes are punched into these membrane sheets which are then stretched in machine direction or in machine direction and cross direction at the same time. Through this process the polymer molecules are orientated in direction of yield. This increases the strength of the material and reduces elongation. Stretched geogrids are made of polypropylene (PP) or high density polyethylene (HDPE) resulting in homogeneous integrated joints, which cannot be displaced, which are torsionally rigid in and vertical to the geogrid plain and which are normally 1.5 to 3 times as thick as the bars are. The manufacturer cannot directly influence the junction strength, it is always as solid as allowed by the production techniques. Disadvantages of stretched geogrids are the limited strength (subject to raw material), which means that the short term strength is limited right now to about 150 kN/m, and the different degree of stretching in the bar and junction area, that there is hardly no stretching at the joints (due to production). Due to the strongly developed creep characteristics of polyolefine raw materials the short-term strength can only be used to a low degree as long-term strength. Woven geogrids are wovens with openings exceeding 10 mm. They are flexible due to cross-laid warp and weft threads usually made of polyester filament yarns. The filaments are often less than 0.1
2 THE GEOGRID-SOIL-INTERACTION If geogrids with mesh sizes exceeding 10 mm are used as reinforcing elements, immediate interlocking interaction can take place between the reinforcement and the filling. The interaction can be intensified the higher the form stability and the modulus of the used geogrid and the better the geogrid matches the filling related to geometry and grain size. Depending on the selected geogrid type - woven, extruded, laid - considerable differences can occur regarding deformations of constructionsin use. For this reason, product properties have to be defined which help to distinguish products related to their reinforcing efficiency with deformations as low as possible. According to Lopes et al. 1999, optimum interlocking is given if at least 20 % of the filling's grain size ranges between the thickness of the geogrid bars and the geogrid mesh size. The interlocking results in transfering strengths from the geogrid's crossbars to the longitudinal bars and requires strong, rigid junctions.
43
mm thick and have a large specific surface. Thus they are possibly more susceptible to installation stresses and environmental influences than for example monolithic stretched and laid geogrids (BMBV 1990 and Brau 1999). As protective measure they are often equipped with a PVC or PVA coating. For all flexible and not prestressed reinforcing products, especially all woven geogrids, a so-called structional elongation occurs with the initial stress. The product elongates by the product specific structural elongation before carrying any reinforcing load. This leads to an inefficient utilisation of the strength of the reinforcing elements and results in unnecessary deformation at site within the geosynthetic soil composite. Laid geogrids are made of coated strips respectively strip-like elements (FGSV 1994). Products of different manufacturers are offered in this product group. There are flexible products, for example bundled polyester or aramide yams which are fixed at the junctions by polymer coating. These products are less significant on the market than woven or extruded/stretched geogrids. Looking at the given market situation, the main challenge for the development of a more effective soil reinforcing synthetic geogrid is the best combination of highest long-term design strength and highest initial modulus to achieve the highest reinforcing success. Parallelly, the geogrid structure has to be as rigid as possible to withstand installation stresses with limited and little damages and all environmental influences of the application.
Based on this approach, Naue Fasertechnik GmbH & Co. KG, a leading manufacturer of geosynthetics in Germany, has developed a new style geogrid. It is made of prestressed monolithic flat bars (raw material polyester / PET) with welded joints. It provides very low elongation at high strength and has very low creep characteristics. Figure 1 shows this new style geogrid which is available as Q type with the same strengthhtrain characteristics in machine and cross direction (biaxial grid) and as R type with higher strength in machine direction (uniaxial grid) with short-term strength up to 600 kN/m. As a result of the monolithic bar geometry (cross section some mm2) the geogrid has a high resistance to mechanical and chemical/biological damage. Figure 2 is showing the success of this development with the welded PET grid having the highest initial modulus compared to other products (woven grids from PET, PVA or even aramide yarns, extruded stretched grids, woven fabrics). 4 COMPARISON OF PULL-OUT BEHAVIOUR As pull-out tests and the interpretation of the results are topic of discussions for many years up to now, the following results are not intended to be used for calculations, but for the comparison of different products under the same conditions. A shear box with the dimensions l/b = 50 cm / 50 cm was used. The box consists of two soil containers with a height of h = 25 cm each. In the tests, always a rate of dispIacement of v = 10 mm/h was used. For the pull-out tests three different geogrid types (Table 1) and two different soil types (sand and crushed stones) were used. More details of the soil parameters are given by Floss et al. 2000.
Figure 2. Different initial modulus of various geosynthetic reinforcement products (measured with a pre-load of 1% of the max. tensile strength and shown with standardized max. tensile strength = 100%)(Heerten, 2000). Figure 1. The new style geogrid.
44
Table 1. Characteristic values of geogrids for pull-out test. product
raw material
extruded geogrid
PP
laid geogrid (LGl) woven geogrid (WG1)
PET
mesh size (mm)
39
tensile strength
(MD, W/m) 30
0 Geosynthetic reinforcement OEx%ensometer €3 Bracing @ Hydraulic piston 8Loadingarea
tensile strength (CD, kN/m) 30
40
40
40
PETyarns, 20 PVC coatine
35
30
[I
@
I
&
n
r
1
I
Figure 3. Relation of the tangent modulus at 5 mm pull-out length of three different geogrid types in two different soil environments.
All 3 products are biaxial geogrids. From the numerous tests the following results are obtained from product specimens which always have two longitudinal bars (1 = 70 cm) in the distance of the grid aperture and all transverse bars with a length up to the middle of the grid aperture following the longitudinal bars. The results show that under the same conditions the EGl reaches the highest values of the maximum pullout force. WG1 only reaches about 50 %, LG1 about 90 % of these values for sand and crushed stone. But LG1 has 90 % of the maximum pull-out force of EG1 at only half the pull-out length and therefore a much higher initial modulus as shown in Figure 3, where the relation of the initial modulus at 5 mm pull-out length is presented with the EG1 value being set to 100%.
Figure 4. Cross section and picture of MSS.
The construction consists of two soil layers (h = 40 cm each), where the geosynthetics were wrapped around at the front. Between the two soil layers the products are in direct contact without connection. The reinforcement length is 1 m. The loading area consists of a rigid steel construction with 1 m width (same as MSS) and 0.5 m in length direction. The loading is done by a hydraulic piston. The loading is done in several steps each documented with measuring results. When the deformations increased too much or when it was not possible to raise the load anymore, the test was stopped. It was difficult to fix a definite maximum bearing capacity, because usually there was no clear failure of the MSS. This was a problem for the test, but is of course an advantage of those constructions in practice. The series of tests again was carried out with two coarse-grained soil types, a poor-graded sand and a well-graded gravel.
5 FULL-SCALE MODEL TESTS For studying the geosynthetic/soil interaction in a full-scale model test, several two-layered "miniature steep slopes" (MSS) were built and loaded. Although it is no direct simulation of a real steep slope, using the same conditions for all the products gives clear advice on the different behaviours. Figure 4 shows a cross section of the MSS construction.
45
For the MSS tests three different geogrid types, one mechanically bonded nonwoven and one thermally bonded nonwoven were used. The relevant characteristic values are shown in Table 2. For more details of the test set-up and soils used in the test see Floss 2000. EG2 is an uniaxial geogrid. The LG2 is assembled from flat bars made of PET black with a cross section of 8 mm x 0,95 mm. The following Table 3 shows the maximum loads of the tests carried out. MNW and TNW have not been tested with sand because of the expected very high deformations and settlements. These results show that MSS with gravel have higher failure loads and less settlement than those with sand. Within the gravel tests the effect of reinforcement is quite obvious comparing No. 7 (unreinforced gravel) with the other tests. Slightly surprising are the failure loads for MNW and TNW in the same range as with geogrids, but with higher deformations, especially in horizontal direction.
The horizontal deformations are measured along a vertical line in the middle of the construction. Fol lowing the loading steps it was found in the gravel tests that WG2 had already great deformations at lower load levels, which occurred from one step to the other. The absolute deformations were within the range of the MNW. The deformations of EG2 and LG2 tests had a continuous and similar development at lower and medium load levels. At higher load levels the EG2 has higher values of deformation than LG2. For all products the deformations with sand are 5 to 6 times higher than those with gravel. There are results for all loading steps (Floss et al. ZOOO), but for comparison it seems best to look at the horizontal deformations at a certain high load for all tests. For loading (T = 850 kPa, Figure 5 shows the results for MSS tests with sand and Figure 6 those with gravel. For MSS tests with sand the horizontal deformations are similar for all 3 products with the smallest deformations at the LG2. For MSS with gravel great differences are between the LG2 and EG2 tests and the tests with the other products. While LG2 and EG2 tests show a parallel movement of the front, the WG2 and the nonwovens have great deformations at the lower layer. The range of deformations at WG2 is the same as with MNW. For sand and gravel tests the horizontal deformations are in following order: LG2 < EG2 < WG2. The tests with gravel show according to the very small deformations at the front also low values of strain. Only the TNW has significant values that increase to values of more than 30 %. The LG2 and EG2 have values of 2 to 3 % even at the highest load level, whereas WG2 has values similar to the MNW. At maximum loads for most products the strains reached failure states which led to ruptures that were found at excavation. Only with LG2 strain values of about 3 % at (T = 1600 kPa were measured and a visually undamaged product was found on excavation. All test results show clear differences depending on the soil type and the production method of the geosynthetics with best results for the LG type, the new style laid geogrid. Finally, a comparison test with a non-reinforced slope (gravel) is showing the high reserves of bearing capacity of geosynthetic reinforced systems with announcing failure mechanism instead of sudden crashes of systems with soil only. The very high levels of loadings (up to 1500 Wa) and bearing capacity in the MSS tests are in congruence with the experience of field loading tests (Brau et al. 2000).
Table 2. Characteristic values of geogrids for MSS test. product
raw material
geogrid $G2) laid geogrid (LG2) woven geogrid (WG2) mechan. bonded nonwoven
tensile
strain at
(MDICD,
(MD, %)
40140
60160
7
PETyarns, 23/23 PVC coating ___ PET
55/30
12.5
18
65
PET
mesh size
Table 3. Maximum loads and vertical deformations of MSS. No.
soil
geosyn- test final thetic state max. load(kPa)
1 2
sand sand sand gravel gravel gravel gravel gravel oravel
WG2 LG2 EG2 WG2 LG2 EG2
3 4 5 6 7 8 9
---
MNW TNW
830 960 1080 1330 1580 1600 330 1460 1710
max. settlement (cm) 28.3 26.3 25.9 10.35 8.65 11 1.2 11.5 24.6
geosynthetic was
ruptured ruptured ruptured rutxured
46
kN/m MD, 60 kN/m CD) for in-situ tests to determine the consolidation and deformation behaviour. In order to gain suitable soil stability, the highway route over the soft ground requires both compaction and consolidation treatment. This is to be done by placing an overburden soil layer of approx. 4 m in height on the geogrid until the required settlement is achieved (vertical drains are additionally installed to accelerate this process). The overburden soil will then be lowered to a height of approx. 2.5 m and the road will be constructed on top of this base. Since mid September 1999 the overburden soil layer has been placed to a height of 2.00 m. Final results of the elongation, consolidation and deformation behaviour will be available only after a longer period of loading. But the deformation measuring devices already show the expected activation of tensile forces in the geogrid.
Figure 5. Horizontal deformations of MSS tests (sand, CT = 850
6.2 Heightening of a railway dam First Certificates of Approval have been granted by the Eisenbahnbundesamt, Bonn, for the newly developed geogrids made of welded PET flat bars. Therefore, geogrids of this type can be generally used for reinforcing applications in railway construction. The first application of these new products in 1999 at the Deutsche Bahn AG is described in the following. In the city of Hanover, the raise of a railway dam was necessary due to the multitrack extension of the railway section Hanover-Berlin. Since an additional purchase of land for the broadening of the dam was not possible, the existing railway dam had to be raised. To stabilize the railway dam, a geogrid made of welded PET flat bars (400 kN/m MD, 60 kN/m CD) was installed over the whole crest width, thus ensuring safety against slope failure. The use of geosynthetics was very advantageous for the realization of the construction work. First, one line was removed and the existing layer of crushed stone excavated. The geogrid was laid on the formation level at right angles to the dam's centre line, rolled up and stored in the area between the two lines, and then the first line was set up again. After excavation and construction of the second line, the geogrid stored between the two lines could be carried on. Thus, a continuously laid geogrid reinforces the dam safely and provides tensile interlocking.
Figure 6. Horizontal deformations of MSS tests (gravel, 0
6 APPLICATION EXAMPLES OF THE "NEW STYLE GEOGRID" Since production started in summer 1999, already more than 1.5 million m2 of the laid PET geogrids have been delivered to construction sites and have been successfully used and installed. Among others, some projects should be mentioned to show the variety of already existing experience. 6.1 Federal motorway A31 (Germany) The Road Construction Agency Aurich in North West Germany extended the existing city highway by adding two lanes to the federal motorway A31. However, the route of this motorway section led over soft soil layers which have a low bearing capacity up to 7 m below the road. After the investigation of the Federal Agency for Road Engineering (BAST) which also included the assessment of the costs, the owner decided upon a reinforced road construction which showed cost savings of approx. 5 million DM compared to a bridge. Due to the stability calculations, a geosynthetic with a short-term tensile strength of 400 kN/m was required to prevent base failure. The Road Construction Agency Aurich and the BAST supported the use of a geogrid made of welded PET flat bars (400
6.3 Noise barrier application motorway A12, Netherlands To reduce traffic noise levels in a new housing development, the city of Vleuten/De Meern planned a 10 m high noise barrier with an additional, approx. 3 m high wall on the wall crest at the motorway A12. The noise barrier runs parallel to the motorway and is approx. 1500 m long. 47
Figure 7. Raised railway dam near Hanover with geogrid reinforcement.
Figure 8. Installation of the laid PET geogrid as toplayer of the capping system.
The barrier was built with approx. 650,000 tons of cinder from incineration plants. According to the Dutch Guideline for Construction Materials the cinders may be considered as a secondary construction material. When using it, various requirements for the protection of the environment must be fulfilled, e.g.:
(Fig. 8). For anchoring, the geogrid was embedded into the fill crest. Approx. 1000 m2/day of the capping system were installed. REFERENCES
'P The stability of the construction must be ensured in the long term. 'P After final consolidation, the secondary construction material must be at least 0.5 m above the average highest groundwater level. 0 No precipitation may get into the noise barrier. P An expert company certified according to IS0 9001 should carry out the sealing with approved materials and in accordance with the quality control measures.
BMBV 1990. Wirkungsweise von Geotextilien unter intensiver aus den1 dynamischer Beanspruchung. F~~rschuizgsberichte Forschutigsl~rogrammdes Bundesmirzistersfiir- Verkehr und der F~~rsclzurzgsgesellsclzaft ,fiir StrajTen- und Verkehrswesen e. V. (595). Eigenverlag. Brau, G. 1999. Einbaubeanspruchung von Geokunststoffen Stand der Erkenntnisse. Schrifterzreihe der Naue Fczsertechnck: Geokunststoffe in der Geotechnik (I), Gekunststoff-kolloquium am 22. und 23. Januar 1999. Eigenverlag. Brau, G. & Floss, R. 2000. Geotextile structures used for the reconstruction of the motorway Munich-Salzburg. Proceedings of the Second European Geosynthetics Conference in Bologna (1): 373-377. Bologna: Patron Editore. FGSV 1994: Merkblatt fur die Anwendung von Geotextilien und Geogittern in1 Erdbau des StraJenbaus. Aufgestellt vom Arbeitsausschuss "Anwendung von Geotextilien und Geokunststoffen iin StraBenbau" und dem Arbeitskreis "Anwendung von Geogittern im StraRenbau" in der Arbeitsgruppe Erd- und Grundbau der Forschungsgesellschaft fiir StraRen- und Verkehrswesen. Eigenverlag. Floss, R., Brau, G. & Bauer, A. 2000. In-soil testing of geogrids with low construction deformations. Proceedings of the Second European Geosynthetics CoiQerence in Bologna (2): 945-950. Bologna: Pgtron Editore. Heerten, G. 2000. Geokunststoffe bei Verkehrsprojekten im Raum Hannover-Berlin. Vortriige der Baugrundtclguizg 2000 in Hannover: 425-43 1. Deutsche Gesellschaft fur Geotechnik. Essen: Verlag Gluckauf GmbH. Lopes, M.J. & Lopes, M.L. 1999. Soil-Geosynthetic Interaction - Influence of Soil Particle Size and Geosynthetic Structure. Geosynthetics lnternational(6): 26 1-282.
Against precipitation, a composite seal, comprising a geosynthetic clay liner and a 2 mm thick structured geomembrane made of chemically resistant high-density polyethylene, was installed directly on the compacted cinders. The structure on both sides of the geomembrane ensures the necessary friction performance in the shear plain. The polypropylene nonwoven protection geotextile (installed on the sealing system) has two functions. Primarily, it serves to safely protect the underlying geomembrane against mechanical damage. Secondarily, it enables seepage water to discharge from the topsoil layer. To ensure the stability of the 1.5 m thick top-soil layer and to take up the forces directed down the slope, a polyester (PET) geogrid with welded flat bars (120 kN/m MD, 40 kN/m CD) was installed
48
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 2001 Swets & Zeitlinger, ISBN 90 2651 852 8
Installation survivability of flexible geogrids as earth reinforcement materials Chiwan Hsieh Associate Professor, Department of Civil Engineering, National Pingtung University of Science and Technology, Taiwan
Chien Kuei Lin Graduate Student, Department of Civil Engineering, National Pingtung University of Science and Technology, Taiwan
ABSTRACT: An in-situ installation survivability study was performed for the purpose of evaluating the amount and degree of installation damage of PVC-coated PET flexible geogrids placed in five types of commonly used backfill materials. The reduction factor for tensile strength due to installation damage is significantly related to soil type and varies from 1.01 to 1.74. The reduction factors for junction strength due to installation damage varies from 1.00 to 1.40 for different type of soils. Additionally, 2% of the maximum load is recommended as the preload for single rib tensile strength test (GRI-GGl). In comparison of the test results, wide width tensile strength (ASTM D4595) of the test geogrids is about 10% to 20% lower than those obtained from single rib tensile strength tests (GRI-GGl ). The difference of tensile strength obtained from these two test methods increases as the construction damage effect increases.
I INTRODUCTION
ity values is less than 1000g-cm measured using ASTM D1388 test method).
Geogrids are currently being used in a number of different soil reinforcement applications, such as retaining walls, steep soil slope stabilization, and improve bearing capacity of foundation soils. All of these applications require design procedures that are based on the tensile strength of geogrid. The allowable strength can be related to one another on a sitespecific basis as follows (Koerner 1998):
2 FIELD SURVIVABILITY TEST PROGRAM The field survivability study was performed during the development of an industry park near ShinChun, Taiwan in the middle of September 1999. Five different test pits were prepared for each test backfill materials. The test soils included a low plasticity silty clay (CL), a poor graded fine sand (SP), a clayey gravel (GC), a poor graded silty gravel (GPGM), and a well-graded gravel (GW). PVC-coated PET geogrids produced by a local manufacture are used in the study. Four types of uniaxial geogrids with different tensile strengths were used in the study. The tensile strengths of the test geogrids are 60 kN/m, 100 kN/m, 150 kN/m, and 200 kN/m. The manufacture roll width is 3.8 meters. A number of 124 warp ribs are counted for the raw samples (equivalent 32 warp ribs per meter). The rib opening is about 2 cm by 2 cm, and the opening area is about 47%. The dimension of test sample is 1.9 meters by 2.9 meters. In order to evaluate the effect of compaction roller traveling direction on the strength of geogrid, the warp ribs of test sample were placed parallel (MDC test) and perpendicular (XMC test) to the roller traveling direction for these four types of geogrids. Totally 8 pieces of test samples were placed within each test pit. Thus, the minimum dimension of test pit is 8 meters by 11 meters.
Where TClrrow= allowable wide width tensile strength for use in design; T,,lt = ultimate wide width tensile strength on the as-received material; RFID= reduction factor for installation damage; RFCR= reduction factor for creep deformation; RFCD= reduction factor against chemical degradation; RFBD= reduction factor against biological degradation. Up to now, a number of studies related to the survivability of geotextiles and geogrids had been performed (Koerner and Koerner 1990, Koerner et al. 1993, Rainey and Barksdale 1993, Richardson 1998, and Troost and Ploeg 1990). Recent work by FHWA and IFAI has led development of reduction factors for installation damage of Geosynthetic used in reinforcement applications (Suits, 1996). Thus, the objective of the study is to fill the gap of the current database and to provide the test data of the installation damage of flexible geogrids (The flexural rigid-
49
3 TEST DATA AND RESULTS Upon exhuming the installed geogrids, a visual damage survey was made. The number of ribs broken per square meter was recorded. The exhumed geogrid samples were then labeled and shipped back to laboratory for testing. The single rib tensile strength test, junction strength test, and wide width tensile test according to GRI-GGl, GRI-GG2, and ASTM D4595 test standards were performed. For each geogrid sample, 20 specimens were tested for GGl and GG2 test methods, and 5 specimens were used for wide width tensile tests. For simplicity, the standard test method GRI-GG1 was used as the primary test method to evaluate the installation survivability of the test samples. By comparing the test values obtained from pre-construction and postconstruction test samples, the retained percentage for each test conditions can be obtained. By inversing of this value, the reduction factor for installation damage will be obtained. In addition, the difference between the test results obtained from single rib tensile tests and wide width tensile tests is also discussed in the subsequent section of the paper.
Figure 1. Typical single rib tensile strength test results for 150kN/m geogrid under various preload conditions.
and compaction roller traveling direction are summarized in the Table 2. The average retained percentages for the ultimate and 5% strain tensile strength for XMC test condition with 30 cm cover thickness are 86.0% and 85.0%, respectively. The average of elongation at failure is about 90.0% of pre-construction sample. In addition, as shown on the table, the compaction direction has no significant influence on rib tensile strength. However, the percentages of retained strength associated with 15-cm lift thickness are slightly less than those associated with 30-cm lift thickness. The percentages of retained strength are ranging from 8 1.4% to 86.0% for various conditions.
3.1 Test data for pre-construction samples In order to understand the effect of preload on tensile strength of geogrid, a series of single rib tensile tests with various preloads was performed for these four types of geogrids. Typical test results for various preload conditions are shown in Figure 1. As shown on the figure, the tensile strength versus elongation curves for 1% and 2% preload conditions are quite similar. In addition, the results of statistical analysis for the single rib tensile tests of 150-kN/m geogrid under various preload conditions are shown in Table 1. It is very clear to us, the test results obtained from 2% preload condition consist the lowest measurement uncertainty and 95% confidence interval. Thus, 2% preload was used for the rest of the single rib tensile tests.
3.3 Comparison of tensile strength for five diferent backfill materials The average test values and percentages of strength retained for the 60 kN/rn woven geogrids placed within the test soils for XMC and MDC test conditions are summarized in the Table 3. Since the effects of soil cover thickness on rib tensile strength were not significant for the tested conditions, the data shown in the table are the average values for conditions associated with both 15-cm and 30-cm cover thickness. The data were obtained based on the results from single rib tensile tests and junction strength tests. The percentages of strength retained for the 60-kN/m geogrid are about 95% for the silty clay, 94% for the poor graded fine sand, 69% for the
3.2 Tensile strength of the geogrid placed in the poor graded silty gravel
As mentioned earlier, 20 test specimens were used for the rib tensile tests of each test condition. The results of 150 kN/m geogrid samples placed under the poor graded silty gravel for different cover thickness
Table 1. The results of statistical analysis of typical single rib tensile tests for 150-kN/m geogrid under various preload conditions. Preload Tensile Strength (kN) Standard Deviation (MV) Measurement Uncertainty (MV) Confidence Level (kN) Elongation (%) 5%Strain Strength (kN)
0% 5.937 0.120743 0.038 182 0.086374 11.628 2.034
1% 5.945 0.143159 0.045271 0.1024 10 11.237 2.3 14
50
2% 5.991 0.142474 0.045054 0.10 1920 11.361 2.305
5% 6.052 0.173 192 0.054768 0.123894 1 1.476 2.315
Table 2. Summary of the test results between the pre-construction and installed samples (150 kN/m) placed under the poor graded silty gravel for various test methods and test conditions.
type MDC XMC
Or
'Over
Thickness
Tensile Strength Retain Percentage
Elongation Retain Percentage
5% Strain Strength Retain Percentage
Junction Strength Retain Percentage
81.386 84.958 83.402 85.960
88.050 89.969 89.979 90.03 1
80.304 83.587 79.840 84.957
9 1.926 75.1 16 76.086 87.355
15cm 30cm 15cm 30cm
Table 3. The test result and retained percentages of the 60-kIWm geogrid under five different soils for XMC and MDC test conditions. Tensile Strength Retained Test Soil type condition Test Value Percentage (kN)
GW
GC GP-GM
1.060 1.037 1.330 1.376
GEE :At:
95.30 94.89 93.74 94.3 1 69.5 1 68.02 87.24 90.25 74.82 67.20
Elongation Retained Test Percentage (%)
11.479 10.950 11.900 11.992 9. I05 9.062 10.938 12.178 9.464 9.207
(%I 95.34 90.95 98.84 99.60 75.63 75.27 90.84 101.14 78.60 76.47
5% Strain Strength Test Value Retained Percentage
Junction Strength Retained Test Percentage
(W
(a)
(N)
0.483 0.515 0.444 0.43 1 0.472 0.461 0.459 0.426 0.482 0.460
96.60 103.01 88.85 86.15 94.35 92.10 91.78 85.25 96.34 91.95
139.48 132.65 139.70 138.18 114.19 117.18 125.61 125.63 116.47 110.02
(%)
95.81 91.12 95.96 94.92 78.44 80.49 86.28 86.29 80.00 73.70
terials are also shown in Table 3. Based upon the test results, the average elongation values at failure for the various type pre-construction samples range from about 11.4% to 14.5%. The elongations at failure for various types geogrids and different test conditions are also shown in tables 3 to 6. For the great majority conditions, the data shown in the tables indicated that the installation process would reduce the elongation of test sample at failure. The amount of reduction of strength could be related to the type of damage of geogrid rib due to installation damage. By further analyzed the data, it is found that the compaction roller traveling direction is also having no significant effect on the single rib elongation at failure.
well graded gravels, 87% to 90% for the clayey gravels, and 67% to 75% for the poor graded silty gravels. The test values and percentages of strength retained for 1OO-kN/m, 150-kN/m, and 200-kNlm geogrids installed within the test soils are summarized in Tables 4, 5 , and 6, respectively. Results showed that these geogrids had similar installation damage behavior as the 60-kN/m geogrid. 3.4 Tensile strength at 5% strain The variation of the tensile strength at 5% strain for the 60-kN/m geogrid installed in the test backfill materials for XMC and MDC test conditions are shown in Tables 3. As shown in the table, the effect of installation process on the tensile strength at 5% strain is relatively less than that on ultimate rib tensile strength. Typically, the percent of strength retained varies from 85% to 96% for the conditions tested. By evaluating the data shown in tables 4 to 6, the effect of installation process on tensile strength at 5% strain various for different conditions.
4 JUNCTION STRENGTH Commonly, the opening area of geogrid is also an important physical property that controls the interlocking behavior of the soil/geogrid system. In addition to provide surface friction, geogrid junction strength is another mechanism that will transfer the pullout resistance of geogrid from soil to geogrid. Therefore, junction strength is another important mechanical property for geogrid. Tables 3 to 6 also consisted the junction strength for the tested geogrid samples installed in the five different backfill materials for XMC and MDC test conditions. The average junction strength for the 150kN/m preconstruction geogrid samples is about
3.5 Elongation at failure Compatibility is an important principle in the geogrid reinforcement application. Therefore, the rib elongation and tensile strength at desire strain are the important mechanical properties of geogrid. The elongations at failure of the 60-kN/m geogrid samples installed in the five different backfill ma
51
Table 4. The test result and retained percentages of the 100-kN/m geogrid under five different soils for XMC and MDC test conditions
Table 5. The percentages and retained reduction factors of the 150-kNh geogrid under five different soils for XMC and MDC test conditions Tensile Strength
Elongation
5% Strain Strength
Junction Strength
Table 6. The percentages and retained reduction factors of the 200-kN/m geogrid under five different soils for XMC and MDC test conditions
Soil type
Tensile Strength Retained Percentage
T:iE- Test Value (N)
SP GW
;: ;:;;: z:;
GC
GP-GM
;FE
7.052 7.386
z:
5% Strain Strength Elongation Retained Retained Percentage Test Value (kN) Percentage
Test Value
(G7-1 \'"I
(%I
(qn)
90.75 92.63 89.38 88.83 81.66 80.23 86.10 90.19 83.83 79.46
12.625 12.634 12.844 13.358 11.622 10.254 11.824 13.046 45.486 44.172
87.07 87.13 88.58 92.12 80.15 70.71 81.54 89.97 90.26 81.20
Junction Strength Retained Percentage (%> \
(WO> \ '"I
\ '"I
2.099 2.206 1.844 1.855 1.977 2.387 2.092 1.896 1.735 1.916
105.99 111.41 93.12 93.66 99.84 120.53 105.67 95.73 87.61 96.78
4 - 1
669.19 743.46 707.01 725.08 669.44 688.14 733.50 695.89 663.75 741.07
89.25 99.15 94.29 96.70 89.28 91.78 97.83 92.81 88.52 98.84
5 WIDE WIDTH TENSILE TESTS
693.8 Newtons. As shown on Table 6, the average junction strength for the 150-kN/m geogrid samples was generally decreased from 693.8 Newtons to the range between 555 Newtons and 592 Newtons. In addition, the behavior of junction strength for XMC and MDC test conditions was found to be quite similar to each other. And the junction strength behavior for others type of geogrid samples is quite similar to that associated with the150 kN/m geogrid samples.
Since geogrids commonly consist of high tensile strength, a pair of roller grips is required to perform the wide width tensile test. Typically, a pair of roller grips is weighted over 100 kilograms. Therefore, a wide width tensile test is more complex and time consuming to perform in comparing with performing single rib tensile test. In order to evaluate the difference between these two test methods, a series of 52
addition, the wide width tensile test results of preconstruction and exhumed samples under wellgraded crushed stone gravel for different test conditions are shown on Figure 3. As shown on the figure, the tensile strength and elongation of the exhumed samples are significantly less than the test data for the pre-construction samples. The comparison of the retained percentages for the 150-kN/m test samples placed under well-graded crushed stone gravel for single rib and wide width ten-sile tests is shown on Table 7. As shown on the table, the retained percentages of wide width tensile tests are generally 2% to 22% less than those for single rib tensile tests. Since the retained percentages for different test conditions are quite similar to each other, the average retained percentages for the test backfill materials are summarized in Table 8. As shown on the table, the retained percentages of wide width tensile strength, elongation at failure, and tensile strain at 5% strain are about 3% to 10%, 2% to 14%, and 0% to 17% less than the values for single rib tests, respectively.
Figure 2. Comparison of single rib and wide width tensile test results for pre-construction geogrid samples (60 kN/m).
6 CONCLUSIONS AND SUGGESTIONS The results of the study have indicated that installation damage of a flexible geogrid is a function of grain size distribution and angularity of backfill materials, and lift thickness. It is also clear that geogrid placed within the angular crushed stone gravel shown greater damage than other backfill materials. Due to the time constrain, the degree of compaction, the type and weight of compaction equipments are not examined as the variables in the study. It is also believe that, the PVC coating thickness and method of coating has some effect on the ultimate tensile strength and the survivability of the geogrid. However, based upon the limited database, these effects are not analyzed in this study. The wide width tensile test (ASTM D4595) is the most representative of the loss of strength due to installation procedure. However, if wide width roller grip is not available, single rib tensile test with more than 10 specimens could be used to evaluate the tensile strength of geogrid also. A 2% of the maximum load is recommended as the preload for single rib tensile test.
Figure 3. Tensile strength versus elongation curves (60-kN/m) of the pre-construction and exhumed samples under wellgraded gravel for different test conditions.
wide width tensile tests of 60-kN/m, lOO-kN/m, and 150-kN/m pre-construction and exhumed geogrid samples is performed. Figure 2 shows the typical test results of single rib and wide width tensile tests for the pre-construction samples. For the test geogrid, the wide width tensile test specimen consists of 7 ribs, the GRI-GGldata shown on the figure equivalent the single rib tensile test data multiplied by 7. As shown on the figure, the tensile strength versus elongation curves is quite similar to each other. In
Table 7. Comparison of the retained percentages of single rib and wide width tensile tests for lSO-kN/m geogrid samples placed under well-graded gravel for different test conditions. Tensile Strength Retained Percent- Elongation Retained Percentage (%) 5% Strain Strength Retained PerTest Cover thickness condition age (%) centage (%) Wide Width Test Single Rib Test Wide Width Test Single Rib Test Wide Width Test Single Rib Test 93.675 89.1 18 80.674 87.905 76.186 78.465 MDC 15cm 82.205 76.482 86.522 XMC 76.035 80.668 69.02 1 8 1.342 61.185 83.022 MDC 52.563 73.940 75.659 30 cm XMC 74.246 81.886 78.822 88.438 73.761 82.065
53
Table 8. Comparison on the average retained percentages for single rib and wide width test tests placed under five different backfill materials for different test conditions. Tensile Strength Retained Percentage
Elongation Retained Percentage
Wide Width Test 92.1 1 88.55 68.57 80.22 78.13
Wide Width Test 97.58 9 1.34 77.83 80.7 1 80.62
Soil type CL SP GC GW GP-GM
Single Rib Test 95.39 91.221 79.94 87.63 83.93
Single Rib Test 99.60 95.398 86.41 94.01 89.51
5% Strain Strength Retained Percentage (%) Wide Width Test Single Rib Test 71.30 87.84 74.68 85.034 83.07 75.14 8 1.58 80.61 72.87 82.17
Table 9. Typical recommended reduction factor for installation damage of flexible geogrids ’Oil Type CL SP GW GC GP-GM
Tensile Stren th Junction Stren th Range of Reduction Factor AveFage Reduction Factor Range of Reduction Factor Aver&e Reduction Factor 1.0101.20 1.10 1.ooo 1.21 1.11 1.04ll1.18 1.11 1.0001.31 1.16 1.1401.74 1.44 1.0001.41 1.21 1.05ul.19 1.12 1.OOo 1.25 1.12 1.13ll1.52 1.33 1.OOo 1.40 1.20
arranged by the President Tang-Hui Chang of AllToffs Industrial Co., Ltd. The Graduate and Undergraduate students Jen-Han Wu, Ming-Wen Hsieh, Ming-Cher Wong, and Chain-Kwue Lin of the National Pingtung University of Science and Technology performed the field installation study and laboratory-testing program. The authors express sincere appreciation to all the members having great contribution to the study
In addition to provide the installation damage database for this type of flexible geogrids, the other goal of the study was to quantify a reduction factor for geogrid installation survivability. Based upon the results of this study, the recommended typical and average reduction factors for flexible geogrids placed within various types of soils are listed in Table 9. As shown in the table, geogrid placed in wellgraded crushed stone gravels showed more severe damage than gravel with some fine grain soils. The typical reduction factors of single rib tensile strength for clayey soil, sandy soil, clayey gravels, poorgraded silty gravels, and well-graded crushed stone gravels are about 1.01 to 1.20, 1.04 to 1.18, 1.05 to 1.19, 1.13 to 1.52, and 1.14 to 1.74, respectively. In addition, the average reduction factors of junction strength for clayey soil, sandy soil, clayey gravels, poor-graded silty gravels, and well-graded crushed stone gravels are 1.11, 1.16, 1.12, 1.20, and 1.21, respectively. The recommended values appear to agree with those values recommended by FHWA. In general, wide width tensile strength of flexible geogrid is about 10% to 20% lower than that obtained from single rib tensile test. The difference of tensile strength obtained from wide width tensile test and single rib tensile test increases as the construction damage of the geogrid samples increases.
REFERENCES Bonaparte, R., Ah-Line, C., Charron, R., & Tisinger, L. 1998. “Survivability and Durability of a Nonwoven Geotextile,” Proc. Geosynthetics for Soil Improvement, Geotech. Spec. P ~ b l 18, . ASCE, pp. 68-91. Koerner, G. R. & Koerner, R. M. 1990. The Installation Survivability of Geotextiles and Geogrids, Fourth Zizteriiational Confereizce on Geotextiles, Geomembranes, and Related Products. The Hague, Balkema, Rotterdam. pp. 597-602. Koerner, G. R., Koerner, R. M. & Elias, V. 1993. Geosynthetic Installation Damage Under Two Different Backfill Conditions, Geosynthetic soil Reiiforceinent Testing Procedures, ASTMSTP 1190, West Conshohocken, PA, pp. 163-183. Koerner, R. M. 1998. Designing with Geosynthetics, qLh.Edition, Prentice Hall. Rainey, T. & Barksdale, R. 1993. Construction Induced Reduction in Tensile Strength of Polymer Geogrids, Geosynthetics ’93 Conference, St. Paul, Minn. Richardson, G. N. 1998. Field Evaluation of Geosynthetic Survivability in Aggregate Road Base, Geotecknical Fabrics Report. Vol. 16, No. 7, Sep. pp. 34-38. Suits, L.D. and G.N. Richardson, 1996, “M288-96: The Updated AASHTO Geotextile Specifications”, Geotechnical Fabrics Report, Vol. 16, No.2, March pp.39-41. Troost, G. H. & Ploeg, N. A. 1990. Influence of Weaving Structure and Coating on the Degree of Mechanical Damage of Reinforcing Mats and Woven Geogrids, Caused by Different Fills, During Installation, Proc. 4‘” ZCS Conference, The Hague, The Netherlands, pp. 609-614. Troost, G. H. & Ploeg, N. A. 1990. Influence of Weaving Structure and Coating on the Degree of Mechanical Damage of Reinforcing Mats and Woven Geogrids, Caused by Different Fills, During Installation, Proc. 41h ZGS Conference, The Hague, The Netherlands, pp. 1 1 19-1 124.
7 ACKNOWLEDGEMENT The study is part of the generic research on geogrids by the Geosynthetic Laboratory of National Pingtung University of Science and Technology at Taiwan. This laboratory has been recognized by the Council of Chinese National Laboratory Accreditation as an accredited laboratory for the test methods of GRI-GG1 and GRI-GG2. The study was partially supported by Seven States Enterprise Co. Ltd. The field installation damage program was assisted and 54
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lS5N 90 2651 863 3
Effects of geogrid properties on pullout resistance J. Izawa, Y. Ishihama & J. Kuwano Department of Civil Engineering, Tokyo Institute of Technology, Japan
A. Takahashi Department of International Development Engineering, Tokyo Institute of Technology, Japan
H. Kimura Maeda Corporation, Japan Cformerly Tokyo Institute of Technology)
ABSTRACT: Study on the pullout resistance was made on the model geogrids. The first sense, special attention was directed to see the effects of the shape of geogrids. All the model geogrids had the same horizontal surface area, but the dimensions of longitudinal and transverse elements were changed to investigate the effects of two vertical areas, i.e. longitudinal and transverse vertical areas, on the pullout resistance were investigated. It turned out that the geogrids of the same horizontal area, stiffness and thickness had the same friction angle between soil and geogrid, but apparent cohesion, displacement and strain distribution were different. The effects of stiffness of the geogrid were investigated in the second sense. Three model geogrids had the same shape and surface friction, but the stiffness was different. The peak pullout resistance of the stiff geogrid was obtained at the small pullout displacement. Difference in stiffness especially affected at the beginning of the pullout test. 1 INTRODUCTION
2 OUTLINE OF PULLOUT TEST
Various geogrids are widely utilized and contribute the improvement of stability of embankments. It has been recognized that pullout resistance is important when the stability of reinforced embankment is examined. Although material properties as well as pullout resistance of respective geogrids have been investigated, the properties of those geogrids have not been rarely compared each other and little is known about the effects of different material properties on pullout resistance. The material properties and pullout resistances of eleven maj or geogrids used in Japan were summarized and discussed by Kuwano et al. (1999) based on the data presented in the reports published by Public Works Research Center. It was found from the summarized data of the major geogrids that pullout resistance seems to depend more on the strength of soils than the geogrid properties such as tensile strength, stiffness etc. Strains at pullout failure are smaller for the stiffer grids and for the lower vertical stress. However, as the above-mentioned geogrids were of various types in the materials, shapes and so on, the geogrid properties affecting the pullout resistance were not clearly identified. In this study, model geogrids made of polycarbonate plate were used for pullout tests. Special attention was directed to look at the effects of the shape and the stiffness of geogrids.
Figure 1 illustrates the pullout test apparatus used in this study. The geogrid was laid in the model ground of 300mm in length, 150mm in width and 178mm in height. It was holded by a clamp, which was connected to a jack through a load cell to measure the pullout force. The width of the geogrid was made to be lOmm less than that of the soil container to reduce the effects of the side walls. The vertical stress was applied uniformly by the air pressure in the rubber bag put on the soil. Four wires were attached to the geogrid. They were used to measure the displacements of the respective points of geogrid by LVDTs. Points of measurements in this study are shown in Figure 2. The model ground of Toyoura sand was made by air pluviation to achieve the relative density of 80%. The area of contact between the geogrid and the soil was kept constant during the pullout test by leaving the 50mm geogrid behind the soil as shown in Figure 4. Pullout tests were made at a rate of l d m i n under the vertical stress, (r ", of 15, 25 and 35 kPa. The tests were finished at 45mm because of the limitation in the stroke of the jack. 3 EFFECTS OF GEOGRID SHAPE 3.1 Model geogrid Model geogrids used for pullout tests in this study were made of polycarbonate plate. They were
55
Figure 2. Points of measurement.
Figure 1. The pullout apparatus. Figure 3 . Schematic view of model geogrid.
Figure 4. Configurations of model geogrid.
140mm in width and 420mm in length, i.e. as schematically shown in Figure 2, 20mm for clamping, 50mm for the front of a soil specimen, 300mm in the soil and 50mm behind the soil. Five different types of geogrids were prepared to have different opening, rib width and thickness. However, the total horizontal surface area was kept to be same, 39,000mm2,for all model geogrids. Shapes of model geogrids are summarized in Table 1. In the table, AT denotes the total vertical area of the openings in the transverse direction, and AL is that in the longitudinal direction as schematically illustrated in Figure 3. AT and AL of standard type (S-type) geogrid were 900mm2 and 2,600mm2 respectively. AT and AL of other type geogrids were changed from S-type by changing the number and the width of the ribs as seen in Table 1.
AT of T2-type geogrid is the double of that of S-type. AT of T0.6-type geogrid is 60% of that of S-type. AL of L2-type is about the double of S-type. In the case of T2L2-type, shape of the opening is the same as that of S-type, but the thickness is double. Configurations of model geogrids are shown in Figure 4. Tensile tests on respective geogrids were carried out and the results are summarized in Table 2. 3.2 Pullout characteristics All the geogrids were pulled out without rupture in this sense. Therefore, the pullout resistance was calculated from the total contact area between the 0th sides of geogrid and the soil as shown by the following equation. 56
Table 1. Properties of model geogrid.
Thickness Pitch between transverse ribs Pitch between longitudinal ribs Width of the longitudinal rib Width of the transverse rib AT
AL
S 1 60 32.5 10 10 900 2600
(mm) (mm) (mm) (mm) (mm) (mm!) (mm')
T2 1 30 32.5 10 5 1800 2600
T0.6 I 100 32.5 10
16.7 540 2600
L2 1 60 15 5 10 900 5100
T2L2 2 60 32.5 10 10 1800 5200
Table 2. Results of tensile tests on respective geogrids.
S T2 T0.6 L2 T2L2
z=-
Tensile strength (kN/m) 20.22 19.87 20.08 21.64 34.64
F
2BLG
Tensile strain at rapture (%)
Tensile stiffness (kN/m)
5.869 5.935 5.890 6.146 5.916
449 446 539 473 845
Table 3. 6 ,, and cp.
(1) 61,
where T is the pullout resistance, F is a pullout force, B is a geogrid width, and LG is a length of geogrid in the soil. Figure 5 is the relationship between front displacement and pullout resistance of the model geogrids under (r ,=35kPa. Although all the geogrids except T2L2 had almost the same total horizontal surface area, tensile strength and stiffness, the ultimate pullout resistance, called as pullout strength, was different. Pullout strength of T2, L2, S, and T0.6-type geogrid was large in that order. It seemed that the increase in the pullout strength was attributed to the increase in the vertical area of transverse element, AT, and that of longitudinal element, AL. Further discussion on this will be made later.TO.6type geogrid required larger pullout displacement to mobilize the certain pullout resistance than other geogrids, though its pullout strength was almost the same as that of S-type. It is concluded that the pullout characteristics depend on the shape of the geogrid even if the surface friction, the tensile strength and the stiffness are the same.
S T2 T0.6 L2 T2L2
3 I .2" 32.3" 33.3" 3 1.4" 37.3"
cp (kPa) 3.365 2.764 1.093 5.281 6.564
3.3 Friction angle between soil and geogrid Pullout strengths of the respective geogrids are plotted in Figure 6 against the vertical stress. Strength parameters, cp and &, in the following equation were determined from the approximate straight lines shown in Figure 6.
zf = c, + cs,, tan&,
(2)
where Tf is the pullout strength, cp and aP are the apparent cohesion and the friction angle between soil and reinforcement respectively.
Figure 6. Pullout strength vs vertical stress.
57
tan 6, /tan 0.673 0.703 0.73 1 0.679 0.849
4
Values of cp and ?jp together with the friction angle ratio, tan?jp/tan@,are summarized in Table 3, where @ is the internal friction angle of Toyoura sand and 42" for the model ground in this study. It is seen that there is not much difference in the friction angles of the respective geogrids, though the relationship between pullout displacement and pullout resistance heavily depends on the type of geogrid as seen in Figure 5. The difference of ?jP among S, T2, T0.6, and L2-type geogrids, is just one to two degrees. Those geogrids have the same total horizontal surface area, tensile strength and stiffness. Therefore, it is thought that the friction angle is not affcted much by the shape of geogrid, though the friction on the surface should be the major component of ?ip. T2L2-type geogrid, which has a double thickness and high stiffness, showed higher friction angle than other geogrids. Further study is needed on the effects of thickness and stiffness of a geogrid on the pullout resistance.
Figure 7. Displacement distribution.
3.4 Deformation of geogrid Displacements of the respective points of geogrid shown in Figure 2 were measured through the atached wires. Displacements of the respective tests under oV=35kPaare shown in Figures 7 at the pullout shear stress of 15kPa. Closer the front, larger the displacements as usually observed. The displacements of T0.6-type geogrid, whose pullout strength was the lowest, were much larger than other geogrids. The displacement was large not only at the front but also at the back. Displacement patterns of S, T2 and L2-type grids were very similar and large in this order. It was the order of small pullout strength. In the case of T2L2-type geogrid, the slope of the displacement distribution is gentler than the other geogrids. It implies that the T2L2 geogrid was pulled out with small tensile strain in it. Tensile strains were calculated from the displacements at the respective points of the geogrid and shown in Figure 8. They were obtained at a pullout shear stress of 15kPa under the vertical stress of 35kPa. Tensile strains in the geogrids are increasing from the back to the front, and therefore tensile stresses. The strains in S, T2 and L2 geogrids are
Figure 8 Tensile strain distribution.
almost he same and lower than that in T0.6 geogrid, though the stiffness is the same for these geogrids. Strain in the T2L2 geogrid, which has higher stiffness than the others, is smaller and rather uniform. 4 EFFECTS OF GEOGRID STIFFNESS 4.1 Model geogrid and outline of Pullout test Effects of stiffness on pullout resistance were then studied. Three types of model geogrids were prepared. They had same shape, i.e. S-type, and same surface friction, but different tensile stiffness. Poly-
Figure 9. Model geogrid of different stiffness.
58
carbonate sheets of 0.5mm in thickness were pasted on both sides of l.0mm thick three different types of sheets, i.e. vinyl chloride, polycarbonate and steel, as shown in Figure 9. The results of tensile test of these geogrids were summarized in Table 4. Pullout tests were conducted with the same procedure as the first test series under the vertical stress of 15, 25, 35 and 5OkPa.
PSP, PPP-Type, whose stiffness were (1). PSP, PPPType, whose stiffness were higher than PCP-Type, showed the clear peak pullout resistance, and the peak pullout resistances were higher than that of PCP-type. But pullout resistances of the respective geogrids reached almost the same value at the pullout displacement of 25 30mm. Pullout strengths of the respective geogrids are plotted in Figure 11 against the vertical stress. Values of 6p together with the friction angle ratio, tan6p/tan@,are summarized in Table 5. It is seen that there is not much difference in the friction angle of PSP and PCP-Type. And that of PCP-Type is smaller than the other types. Figure 12 shows pullout resistance under oV= 25kPa until the 3mm pullout displacement. In this test series, 50-80% of the peak pullout resistance were obtained at 3mm of pullout displacement. PSPType, which has highest stiffness, showed the large pullout resistance at the small pullout displacement. It seems that the effects of stiffness appear also at the small pullout resistance. Displacements in the geogrids under o,=25kPa are shown in Figure 13 at the pullout resistance of 15kPa. The displacements of PSP-Type geogrid of the highest stiffness were much smaller, and the slope of the displacement distribution is gentler than
-
4.2 Effects of stifjCness on pullout resistance Figure 10 is the relationship between pullout resistance and pullout displacement under oV=25kPa.The pullout resistance was calculated by equation (1). Table 4. Properties of model geogrid.
PCP PPP PSP
Intermediate material Vinyl Chloride Polycarbonate Steel
Stiffness (N/m) 5.14 10' 1.03 10' 7.50 104
Figure 10. Pullout resistance vs pullout displacement.
Figure 12. Pullout resistance vs pullout displacement.
Figure 11. Pullout strength vs vertical stress. Table 5.6, and cp.
~~
PCP PPP PSP
6p 31.5 " 38.4" 35.0 O
c,(kPa) 2.34 0.547 2.26
tan Gp/tan 4 0.682 0.88 1 0.778
Figure 13. Displacement distribution.
59
same shape, but materials of geogrid were changed. The followings were obtained in this study.
1. The geogrids of the same horizontal area, stiffness and thickness had the same friction angle between soil and geogrid, but apparent cohesion, displacements and strain distributions in the geogrid were different. 2. AT, the total vertical area in the transverse direction, was more effective to increase the pullout resistance than AL, the area in the longitudinal direction. 3. Displacements of S, T2 and L2-type geogrids were large in this order. It was the order of small pullout strength. However, the displacement patterns were very similar for the geogrids whose stiffness were the same. 4. The stiff geogrid showed the peak pullout resistance at the small pullout displacement. On the contrary, the soft geogrid needed large pullout displacement and deformation of geogrid to mobilize the peak pullout resistance. 5. Effects of stiffness affected the pullout characteristics especially at the beginning of pullout test.
Figure 14. Tensile strain distribution.
the other geogrids. In the case of PCP-Type geogrid, the displacements were larger and the slope of the displacement distribution is steeper than the other geogrids. Figure 14 show strain distributions at the pullout resistance of 5kPa and 15kPa under oV=25kPa.Tensile strains of PSP-Type were almost 0. In the cases of PCP and PPP, tensile strains increased with the increase in the pullout displacement.
REFERENCES 5 CONCLUSIONS
Kuwano, J., A. Takahashi & H. Kimura 1999. Material properties and pullout characteristics of geogrids used in Japan. Geosynthetics Engineering Journal, Vol. 14: 19.5-204. Izawa, J., H. Kimura, J. Kuwano, A. Takahashi, Y. Ishihama 2000. Effects of geogrid shape on pullout resistance. Geosynthetics Engineering Journal, Vol. 15: 28-37. Japanese Society of Soil Mechanics and Foundation Engineering 1994. Determination of soilgeotextile frictional behavior by direct shear test andlor pull-out test. Tsuclzi-to-Kiso JSSMFE, Vol. 42, NO. 1: 93-102.
Study on the pullout resistance was made on the model geogrids. Special attention was directed to see the effects of the shape and stiffness of geogrids. In the first test series, all the model geogrids had the same horizontal surface area, but width of longitudinal and transverse members was changed. In the second test series, all the model geogrid had the
60
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Ceotechnical behavior of fiber reinforced fly ash Shenbaga R. Kaniraj & V. Gayathri Department of Civil Engineering, Indian Institute of Technology, New Delhi, India
ABSTRACT: The influence of randomly oriented fiber inclusions on the geotechnical behavior of two Indian fly ashes was studied. Polyester fibers and a constant fiber content of 1 % (by dry weight) were used in the experiments. The paper presents the results of the compaction tests, triaxial shear tests, and other geotechnical characterization tests on the raw and fiber reinforced fly ashes. The fiber inclusions increased the strength of the fly ashes and changed their behavior from brittle to ductile. dian fly ashes reinforced with randomly oriented fibers. The type and composition of coal, the method of burning and other factors affect the characteristics of fly ashes. Therefore, fly ashes can vary from plant to plant or even within the same plant over time. Similarly, the characteristics of fibers such as their type, diameter, length, tensile strength, and tensile modulus also vary widely. Therefore, in order to comprehend the influence of fiber inclusions it is necessary to compare the behaviors of fly ash-fiber mixtures of different combinations. The paper presents the results of an experimental study carried out with this as the objective.
1 INTRODUCTION AND SCOPE Gray (1970), Brown & Sheu (1975), Wu & Erb (1988), Wu et a1 (1988) observed that the presence of plant roots improved the strength of the soils and the stability of natural slopes. This led to more detailed studies on soils reinforced with artificial and natural fibers. While Gray & Ohashi (1983), and Shewbridge & Sitar (1989), carried out laboratory tests on soils in which the fibers were oriented in particular directions, Hoare (1979), Hoover et a1 (1982). Maher (1988), Gray & Maher (1989), Maher & Gray (1990), Maher & Ho (1994), Nataraj & McManis (1996), Gopal Ranjan et a1 (1996), Michalowski & Zhao (1996), and Consoli et a1 (1998) carried out tests with the fibers oriented randomly in the soils. Fly ash is a silt-size residual waste produced from burning of coal in thermal power stations. The main geotechnical applications of fly ash are in building highway embankments, fills, landfill liners, and covers. The effect of fiber inclusions on the geotechnical characteristics of fly ashes have not been studied in as much detail as that of soils. Chakraborty & Dasgupta (1996), and Kaniraj & Havanagi (2001) have carried out experimental studies on some In-
2 MATERIALS USED Two Indian fly ashes were used in the present study. They were collected in dry state from the electrostatic precipitators of the Dadri and Rajghat thermal power stations. The Rajghat fly ash was the same as that used by Kaniraj & Havanagi (2001) in their study. Polyester fibers were mixed with the fly ashes. The properties of the polyester fibers used in the present study and in the experiments of Kaniraj & Havanagi (2001) are given in Table 1.
Table 1. Properties of fibers used in different studies."" Study
Type Color Present study Polyester Black Kaniraj & Polyester Havanagi (200 1 ) Grey
Diameter mm 0.0203"
Length
Specific gravity 1.38
Tensile strength MPa 510
Tensile modulus Mpa
6
Aspect ratio 313
0.075+
20
267
1.3
80- 170
1450-2500
miii
*Approximate values, as the fibers tend to split. 'K"Values as given by the manufacturers HI-TECH FIBERS, South Carolina, U.S.A.
61
3 TESTPROGRAM
Table 2. Physical properties of fly ashes.
3.1 Test program
Property
The test program consisted of experiments for the determination of a) the physical, chemical, and geotechnical characteristics of the Dadri fly ash, and b) the geotechnical characteristics of the Dadri and Rajghat fly ashes mixed with randomly oriented 6 mm long polyester fibers.
Specific gravity, G Loss on ignition, % Specific surface area, cm'/g
Dadri fly ash 2.2 0.4 3520
Rajghat fly ash:!: 2.19 1.4 4020
*All results in the paper for Rajghat fly ash raw and 20mm fly ash fiber mixtures are from Kaniraj & Havanagi (2001)
4.2 Chemical composition
3.2 Fly ash-fiber mixture preparation
The chemical compositions of the Dadri and Rajghat fly ashes are summarized in Table 3. Both the fly ashes are classified as class F and pozzolanic materials as per ASTM C 618 (1993) specifications.
The fly ash was first dried under heating lamps at approximately 40" C. The total dry weight of the mixture required to prepare a specimen, W, is known from the specimen's dimensions and dry unit weight, yd. w is also expressed as, w = w, + W', where W, and Wf are the weights of the dry fly ash and fibers, respectively. Wf is expressed as, Wf = fc Wp, where fc is the fiber content. Since a constant fiber content of 1 % was used the weight of dry fly ash is given by, W, = (W/.Ol) = 0.99 W. First, the required amounts of fly ash and fibers were measured and mixed together in the dry state. As the fibers tended to lump together, it required considerable care and time to separate them to get an even distribution of the fibers in the mixture. The dry fiy ash-fiber mixture was then mixed with the required amount of water. All mixing was done manually and proper care was taken to prepare homogeneous mixtures at each stage of mixing. Kaniraj & Havanagi (2001), however, found that the fibers could be mixed with the fly ash more efficiently in the moist state than in the dry state. Therefore, they first mixed the dry fly ash with water and then mixed the moist fly ash with the fibers. This method of mixing was not satisfactory with the fibers used in the present study.
4.3 Geotechnical classification The results of the experiments for grain size distribution and Atterberg limits are summarized in Table 4. Both the fly ashes can be classified as ML type belonging to the non-plastic silt category. 5 FIBER REINFORCED FLY ASH The effect of fibers on the geotechnical characteristics of fly ashes was investigated by conducting standard Proctor compaction tests, unconfined compression tests, unconsolidated undrained and drained triaxial shear tests on raw fly ashes and fly ashes blended with 1 % polyester fibers by weight. Table 3. Chemical compositions (%) of fly ashes. Composition Silica (SiO2) Alumina (A120.?) Iron oxide (Fe2O.j) Lime (CuO) Magnesia (MgO) Titania (Ti02) Soda ( N u 2 0 ) Potash (KzO) Sulphates (SO3)
3.3 Fly ash-fiber specimen preparation
For triaxial shear tests, cylindrical specimens were prepared by static compaction of the fly ash-fiber mixture at the standard Proctor MDD-OMC of the raw fly ash. A 37.7 mm inner diameter and 73.5 mm long mould with additional detachable collars at both ends was used. To ensure uniform compaction, the entire required quantity of material was placed inside the mould-collars assembly and compressed in three steps alternately from the two ends till the specimen reached the dimensions of the mould.
Dadri fly ash 60. I2 30.16 6.36 1.oo 0.53
Rajghat fly ash 61.21 30.07 4.17 0.10 0.40 2.60 < 0.01 0.02 < 0.01
-__
0.06 0.007 0.10
Table 4. Geotechnical classification test results. Properties Grain size distribution Fine sand, 0.475-0.075 mm, % Silt size, 0.075-0.002 mm, % Clay size. < 0.002 mm, % Uniformity coefficient, C,d Coefficient of curvature, C, Atterberg limits Liquid limit, w,,, % Plastic limit. WI, %
4 FLY ASH CHARACTERISTICS 4.1 Physical properties The physical properties of the Dadri and Rajghat fly ashes are summarized in Table 2.
62
Fly ash Dadri Rajghat
5 82 13 4.82 1.01
20 77 3 5.65 0.9
30.5 NP
48-50 NP
Table 5. MDD & OMC of fly ash-fiber mixtures.
5.1 Compaction tests Figure 1 shows the compaction curves of the fly ashes with and without fiber inclusions. The values of maximum dry unit weight (MDD) and optimum moisture content (OMC) are summarized in Table 5. The results show that the Dadri and Rajghat fly ashes are distinctly different in their compaction characteristics. The small fiber content has not affected the MDD and OMC of the Dadri fly ash appreciably. In Rajghat fly ash, however, the effect was a little more marked; the fiber inclusions increased the MDD and decreased the OMC. All the specimens for the shear tests were prepared at the MDD-OMC of the raw fly ashes. The effects due to differences in the unit weight and water content of the raw and reinforced fly ash specimens were thus avoided. The differences in the behavior will be only due to the fiber inclusions.
Fiber Dadri Rajghat Rajghat Rajghat
1 1 1 !h
6 6 20 20
MDD, kN/m' (OMC %) raw fly ash fiber reinforced 13.8 (21) 13.8 (22) 10.5 (37) 10.7 (34) 10.5 (37) 11.0 (33) 10.5 (37) 11.0 (32)
Table 6. UCS of fly ash-fiber specimens cfc= I %). Fly ash
Fiber
UCS, kPa (q,%) Fiber raw ash reinforced
even at 15% axial strain. The unconfined compressive strength (UCS) also increased. Table 6 summarizes the UCS and failure strain, ~ y ,of the unreinforced and fiber reinforced fly ash specimens. The 6 mm long fiber inclusions have increased the UCS of the Dadri and Rajghat fly ash specimens by 55% and 174%, respectively. The increase in UCS is, therefore, more significant in the Rajghat fly ash-fiber specimens than in the Dadri fly ash-fiber specimens. Kaniraj & Havanagi (2001) reported that the increase in UCS depends on the UCS of the unreinforced specimen. Lower the UCS of the unreinforced specimen higher is the increase in UCS due to the fiber inclusion. The results of the present study confirm to this observation. Further, the increase in the UCS of the Rajghat fly ash specimens is more due to the shorter (6 mm) fiber inclusions than due to the longer (20 mm) ones.
5.2 Unconfined compression tests A minimum of three specimens of the fly ash-fiber mixtures were prepared and tested at a deformation rate of 0.4064 m d m i n . Figure 2 shows the stressstrain curves of the raw and fly ash-fiber specimens. The fiber inclusions had a significant effect on the stress-strain behavior. In raw fly ash specimens, a distinct failure axial stress was reached at an axial strain of about 1.5%-2.5% following which the specimens collapsed. Whereas, the fiber reinforced specimens exhibited more ductile behavior. After reaching a peak axial stress at a small strain, the specimens continued to deform under declining axial stress. Kaniraj & Havanagi (2001) observed that with the 20 mm fiber inclusions in the Rajghat fly ash there was no distinct reduction in axial stress
Figure 1. Compaction curves for fly ashes and fly-ash fiber mixtures.
63
Figure 2. Stress-strain curves for specimens in unconfined compression tests.
5.3 Unconsolidated undrained (UU) tests Unconsolidated undrained triaxial shear tests were carried out at confining stresses in the range of 98.1 to 490.5 kPa. In raw fly ash specimens, the deviator stress attained a peak value and thereafter remained almost constant. In Dadri fly ash the strain to attain the peak deviator stress varied between 11-14%. In fiber-reinforced specimens, no peak deviator stress was reached even at 15% axial strain. This may be a manifestation of the ductile behavior induced by the fiber inclusions. For determination of total stress the failure deshear strength parameters cllI, and hU, ~ ) ~taken , as the peak deviator viator stress, ( c T I - cwas stress for raw fly ash specimens and as the deviator stress at 15% axial strain for fly ash-fiber specimens. Figure 3 shows the p-q [p = (01+43)f/2,q = (ol-g3)y/2] plots for the UU tests. The values of c~,~, and @,,,,are summarized in Table 7. There is an increase in both due to fiber inclusions. Both the short cNIC and and long fibers have caused an increase in the c,, and &,,,, of the Rajghat fly ash specimens. 5.4 Drained triaxial shear (CD) tests
E.=1%)
Fly ash
length mm
raw fly ash
Dadri Rajghat Rajghat
6 6 20
12.6 (31") 43.2 (30.1") 43.2 (30.1")
6 CONCLUSIONS Laboratory experiments were carried out to study the influence of randomly oriented fiber inclusions on the geotechnical characteristics of two Indian fly ashes. Polyester fibers and a constant fiber content of 1% (by dry weight) were used in the experiments. The following are the main conclusions from the present experimental study and the comparison of the results with the previous studies.
1. The small amount of 1% fiber content in the form of randomly oriented fiber inclusions improved the stress-strain response of the fly ash specimens in all triaxial shear tests and changed their behavior from brittle to ductile. 2. The 6 mm long fiber inclusions increased the unconfined compressive strength of the Rajghat fly ash-fiber specimens more significantly than the UCS of the Dadri fly ash-fiber specimens. The effect of longer (20mm) fibers on the UCS of Rajghat fly ash is less than that of the shorter fibers. 3. The fiber inclusions also increased the total stress in the UUtshear strength parameters cI,ll and ests and the effective stress parameters c'and 4' in the drained triaxial shear tests.
Drained triaxial shear tests were carried out only on the combinations of Dadri fly ash and 1 % fiber content of 6 mm long fibers. Prior to consolidation, the specimen was saturated in the following way. A small confining stress of about 20 kPa was applied and water was allowed to flow from a burette Table 7. UU test results
through the specimen, from bottom to top, for 24 hours. The CD tests were conducted at confining stresses in the range of 98.1 to 490.5kPa. The deviator stress-axial strain behavior in CD tests was also found to be the same as in the UU tests. Using the same procedure as for the UU tests, the values of c' and @' were determined. Figure 4 shows the p '-q'plots. The fiber inclusions increased the c'of the Dadri fly ash specimens from 0 to 55.9 kPa and the @'from 29.3" to 33.6".
fiber reinforced 93.7 (32.5") 102.8 (36") 128.6 (36")
64
o Dadri fly ash o Rajghat fly ash
Dadri fly ash (1% 6mm fibers) Rajghat fly ash (1% 6mm fibers) Rajghat fly ash (1% 20mm fibers)
Figure 3. p-q plots for unconsolidated undrained tests
Gray, D.H. (1 970). Role of woody vegetation in reinforcing soils and stabilizing slopes, Proc. Symp. on Soil Reirzfoi-ceineizt and Stabilizing Techniques, Sydney, Australia, 253306. Gray, D.H. & Maher, M.H. (1989). Admixture stabilization of sand with discrete randomly distributed fibers, Proc. XII Int. Conf on Soil Mech. Found. Eng., Rio de Janeiro, Brazil, 1363-1366. Gray, D.H. & Ohashi, H. (1983). Mechanics of fiber reinforcing in sand, J Geotech. Eng. Div., ASCE, 109(3), 335-353. Hoare, D.J. (1979). Laboratory study of granular soils reinforced with randomly oriented discrete fibers, Proc. Int. Cotif: on Soil Reiizforcetneizt, Paris, France, 1, 47-52. Hoover, J.M., Moeller, D.T., Pitt, J.M., Smith, S.G. & Wainaina, N.W. (1982). Performance of randomly oriented fiber reinforced roadway soils, Iowa DOT Pi-oject-HR-21I , Department of Transportation, Highway Division, Iowa State University, U.S.A. Kaniraj, S.R. & Havanagi, V.G. (2001). Behavior of cement stabilized fiber reinforced fly ash-soil mixtures, J Geotech. Eizg. Div., ASCE, Accecpted for publication. Maher, M.H. (1988). Static and dynamic response of sands reinforced with discrete randomly distributed fibers, Ph. D. Thesis. University of Michigan, Ann Arbor, U.S.A. Maher, M.H. & Gray, D.H. (1 990). Static response of sands reinforced with randomly distributed fibers, J Geotech. Eizg. Div., ASCE, I16(11), 1661-1677. Maher, M.H. & Ho. Y.C. (1994). Mechanical properties of kaolinitelfiber soil composite, J Geoteclz. Eizg. Div., ASCE, 120(8), 138 1 -1 393. Michalowski, R.L. & Zhao, A. (1996). Failure of fiberreinforced granular soils, J Geotech. Eizg. Div., ASCE, 122(3). 226-234. Nataraj, M.S., and McManis. K.L.( 1996). Strength and deformation properties of soils reinforced with fibrillated fibers. Geos.ynthetics International, 4 (l), 65-79. Shewbridge, S.E. & Sitar, N.(1 989). Deformation characteristics of reinforced soil in direct shear, J Geotech. Eng. Div., ASCE. 115(8), 1134-1 147. Wu, T.H., Beal. P.E. & Lan, C. (1988). In-situ shear test of soil-root system, J Geotech. Eng. Div., ASCE, 1 14(12), 1376-1394. Wu, T.H. & Erb, R.T. (1988). Study of soil-root interaction, J Geotech. Eizg. Div., ASCE, 114( 12), 135 1-1 375.
Figure 4. p’-q’ plots for consolidated drained tests.
7 ACKNOWLEDGEMENTS The authors are thankful to Mr. P. Ratna Rao of R.R. International. Gurgaon, India for providing the polyester fibers. REFERENCES ASTM C-6 18 (1993). Specification for fly ash and raw or calcined natural pozzolana for use as a mineral admixture in portland cement cocrete, Annual Book of ASTM Standards, 4.02 (4), 3 10-3 12. Brown, C.B. & Sheu, M.S. (1975). Effect of deforestation on slopes, J Geotech. Etzg. Div., ASCE, 101, GTI, 147-165. Chakraborty, D.K. & Dasgupta, S.P. (1996). Randomly reinforced fly ash foundation material, Indian Geotechnical Conference, Madras, India, 1, 23 1-235. Consoli, N.C., Prietto, P.D., and Ulbrich, L.A. (1998). Influence of fiber and cement addition on behavior of sandy soil, J. Geotech. and Geoeav. Eiig. Div., ASCE, 124 (1 2). 121 1 1214. Gopal Ranjan, Vasan, R.M., & Charan, H.D. (1996). Probabilistic analysis of randomly distributed fiber-reinforced soil, J Geotech. Eng. Div., ASCE, 122(6), 419-426.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Partial factors for geosynthetics specific to lirnit state approach A. J. Khan Bangladesh University of Engineering and Technology, Dhaka, Bangladesh
ABSTRACT: In limit state approach, Partial Factors are applied to geosynthetics both for Ultimate Limit State [ULS] and Serviceability Limit State [SLS] analyses. Most commonly, Partial Factors are determined by comparing the ‘before’ and ‘after’ strengths of geosynthetics obtained from short-term Constant Rate of Strain [CRS] test. The Partial Factors, so obtained, are then applied to Reference Strength obtained from long-term sustained load creep test in order to obtain long-term Design Strength of geosyntheics. It is indicated in this paper that totally different amount and nature of strain components are measured in CRS and creep tests. Therefore, Partial Factors obtained from CRS test should not be applied to obtain long-term Design Strength. Further, it is identified for a range of geosynthetics that Partial Factors are both time and strain level dependent. This suggests that different values of Partial Factors should be used for the limit state being considered. BS8006 (1995) and Jewel1 (1996) defined the Reference Strength as the load to cause creep mpture of ‘Ex-works’ specimens at the end of design life. Typically, geosynthetics exhibit a wide range of scatter of creep rupture strains for different sustained [creep] load levels. Hence, the Reference Strength, defined on the basis of load at creep rupture for a specific design life, can be very difficult to select with any certainty related to the strain level developed at creep rupture, McGown et al(1998). Some design methods define the Reference Strength as the load obtained from the Isochronous Load-Strain curves, corresponding to a Performance Limit Strain, for example the TBW Method (1998) and the HA 68/94 Design Method (1997). Performance Limit Strain is always less than Instability Strain Limit and hence Reference Strength for Ultimate Limit State [ULS] design based on Performance Limit Strain can be seen to be a conservative choice, McGown et a1 (1998). To summarize, currently the Reference Strength for geosyntheics is defined in two different ways, (i) on the basis of long-term creep rupture strength and (ii) on the basis of long-term creep at a limiting strain.
1 INTRODUCTION Important input parameters for geosynthetics in designs are their Reference Strength, Partial Factors and Design Strength. Reference Strength is the strength of Ex-works materials. Partial Factors are the factors that allow for any change of geosynthetics due to construction damage and environmental degradation. Design Strength is the strength obtained by dividing the Reference Strength by the Partial Factors. 1.1 Reference strength Various design codes and methods define Reference Strength in different ways for geosynthetics. DIBt (1998) and AASHTO (1997) use the maximum load at rupture of the ‘Ex-works’ materials under Constant Rate of Strain [CRS] tensile testing as the basis of defining the Reference Strength of geosynthetics. The Reference Strength, to avoid longterm creep rupture, is then obtained by dividing the CRS rupture strength by a Reduction Factor. The DIBt and the AASHTO design methods specify 33% per minute and 10% per minute strain rates respectively, for the CRS tests employed. Kabir (1984) have shown that the CRS test can give different strengths at different test strain rates and temperatures. Therefore, even for a particular type of geosynthetic the same Reduction Factor may not be applicable to all CRS data in order to obtain the longterm creep rupture strength, i.e. the Reduction Factor is dependent on the strain rate used in the CRS test.
1.2 Partial factors The strength of Ex-works geosynthetics requires to be modified by applying Partial Factors in order to obtain their Design Strength. Four Partial Factors of major concern have been identified by Voskamp and
67
tained from long-term creep test does not seem to be compatible. This is explained more in detail in the following section.
Risseeuw (1987) and Jewel1 and Greenwood (1988) for geosynthetics. These are: a) The Damage Factor; to allow for mechanical damage during handling, transportation and construction. b) The Environmental Factor; to allow for the chemical environment, UV radiation and microbiological exposure in the ground. c) The Material Factor; to allow for the uncertainty inherent in the extrapolation of test data, and d) The Overall Factor; to allow for the properties of materials not meeting the manufacturer’s specification. Partial Factors are determined by comparing the strength of geosynthetics ‘before’ and ‘after’ the construction damage or environmental degradation. Different design codedmethods and different researchers suggest different methods of determining Partial Factors for geosynthetics. Bush (1988), Watts and Brady (1990) and KOerner and Koerner (1990) used CRS tests to identify damage effects on geosynthetics. They determined the Damage Factors for various geosynthetics by comparing the CRS loads for rupture ‘before’ and ‘after’ damage effects. BS8006 (1995), DIBt (1998)’ AASHTO (1997), TBW (1998) and HA 68/94 (1997) adopt the same method for the determination of Partial Factors for geosynthetics. Partial Factors, determined by comparing the CRS strengths ‘before’ and ‘after’ an event, are considered to be constant in these codes over the whole design life of a Geosynthetic Reinforced Soil Structure [GRSS]. Therefore, most of the design codes/ methods specify a single value of Partial Factor for geosyntheics to obtain their Design Strength, regardless of design life and operational strain level. The underlying assumption of such specification is that the effect of an event, i.e. construction damage or environmental degradation, on a particular geosynthetic ‘today’ will cause no further deterioration of its properties after ‘100 years’. Due to the elastovisco-plastic nature of geosynthetics, this assumption is very unlikely to be true always.
2 BEHAVIOUR OF GEOSYNTHETICS UNDER DIFFERENT LOADING REGIMES Geosynthetics exhibit elasto-visco-plastic behaviour. This means that their mechanical behaviour is in part similar to that of elastic solid, in part similar to that of a viscous liquid and in part similar to that of a plastic, with all these parts being temperature dependent. Therefore, when subjected to an externally applied load as shown in Figure la, they respond by exhibiting a combination of elastic displacement, viscous flow and irrecoverable plastic deformation, Figure lb. The elastic deformation may be termed the Immediately Recoverable Strain and is denoted by ER. The viscous and plastic component together may be termed the ‘Locked-in’ Strain and is denoted by EL. The ‘Locked-in’ Strain, EL, comprises of one time dependent recoverable part, i.e. viscous strain and the other never recoverable part, i.e. plastic strain. It may be appreciated that for a certain loading regime a particular limiting strain can be reached by a geosynthetic with a number of combinations of Immedi-
1.3 Design strength It was mentioned earlier that most of the design codes/methods define Reference Strength in two different ways, i.e. on the basis of long-term creep rupture strength or on the basis of long-term creep at a limiting strain. Partial Factors, on the other hand, are specified on the basis of short-term CRS test results. Design Strength is then obtained by dividing the long-term Reference Strengths by the Partial Factors obtained from short-term CRS test. It may be appreciated that totally different amount and nature of strain components are measured in CRS and creep tests. Therefore, applying partial factors obtained from short-term CRS test to Reference Strength ob-
Figure 1. Typical behaviour of geosynthetics.
68
ately Recoverable Strain, ER and ‘Locked-in7 Strain, means that the same limiting strain ~1 can be reached with a high value of ER and a low value of EL (point A) or a high value of EL and a low value of ER (point B). The probable strain envelop for a limiting strain ~1 is shown in Figure 2. If a CRS test is carried out on a geosynthetic to reach a limiting strain of ~1 at a very fast rate of strain, more Immediately Recoverable Strain, ER will develop than ‘Locked-in7 Strain, EL (point C). The ‘Locked-in’ Strain, EL will not have sufficient time to develop and mostly the contribution to total strain ~1 will be due to the Immediately Recoverable Strain, ER. If, however, a long-term creep test is carried out on the same geosynthetic to reach the same limiting strain of ~1 in say 1000 hours, more ‘Locked-in7 Strain, EL will develop than Immediately Recoverable Strain, ER (point D). In this case, there will be sufficient time for the ‘Locked-in’ Strain, EL to develop and mostly the contribution to total strain ~1 will be due to the ‘Locked-in’ Strain, EL. This indicates that a geosynthetic is likely to respond differently in terms of its strain components in different loading regimes and it may not be appropriate to apply data obtained from one test methodology to the other. Therefore, it may be suggested that for long-term applications, both the Reference Strength and Partial Factors should be determined on the basis long-term sustained load creep test results. EL. This
considered, i.e. different strain level should be identified for ULS and SLS analyses. According to this approach, first, the Isochronous Load-Strain curves for a geosynthetic should be obtained ‘before’ and ‘after’ an event, Figures 3a, b. The ratio of the load carrying capacity of the geosynthetic is then obtained at different strain levels to obtain the Partial Factors at that strain level. Partial Factors, so obtained, are then plotted against log-Time for different strain levels, Figure 3c.
3 SUGGESTED PROCEDURE FOR DETERMINING PARTIAL FACTORS In this approach, the Partial Factors may be defined as the ratio of the load carrying capacity of a geosynthetic at a particular strain ‘before’ and ‘after’ an event, e.g. ‘before’ and ‘after’ the construction damage or environmental degradation. The strain level should be chosen on the basis of Limit State being
Figure 3. Proposed approach for determining partial factors.
Figure 2. Possible variation of strain components.
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It should be noted that a value of Partial Factor greater than unity indicates a reduction of the original intrinsic property of a geosynthetic and a value less than unity an increase in the original intrinsic property. Further, Partial Factors obtained according to this approach is likely to exhibit their time and strain level dependency.
4 DETERMINING PARTIAL, FACTORS FROM THE EXPERIMENTAL DATA Al-Mudhaf (1993) and Esteves (1996) studied the effects of construction damage and environmental degradation for a range of geosynthetics. A brief description of the test sites, the type of soil, construction equipment and conditions are given in the following sections. The data obtained from these tests are then analysed and Partial Factors are determined according to the approach proposed in Section 2 of this paper.
Figure 4(a). Isochronous load-strain curves for ex-works uniaxial geogrid at 20°c (after Esteves, 1996).
4.1 Damage factor For the evaluation of the construction damage effects, a site damage trial was carried out in Kuwait on a uniaxial geogrid and a woven geotextile. The compaction was performed during the month of May, when the maximum temperature reached 45°C and the humidity ranged between 40% and 80%. The storage room in the laboratory was kept at a constant temperature of 25°C. The soil at the site was a fine medium sandy soil without any traces of water. The soil from the site was used for fill and placed over the geosynthetics in a 0.3m thick layer. It was compacted to the maximum dry density at the optimum moisture content- The compaction was performed using hand operated compaction equipment of 3.0 kN of weight, vibrating at a rate of 3.33 Hz. The compaction plate dimensions were of 0.3m x 0.5m. For the trial, the materials were laid out one beside the other, without any overlapping, in a test bay of 12m x 6m. They were covered with the fill and then subjected to 12 compaction passes over a period of two working days. The materials remained in the soil for one more week. Thereafter, they were carefully removed, stored and returned to the laboratory for testing. CRS and sustained load creep tests were carried out on the ‘Ex-works’ and ‘Damaged’ specimens of the geosynthetics in the laboratory. CRS tests were carried out at 10% strain/minute and the creep tests were carried out up to 1000 hours for each load level. Isochronous Load-Strain curves for the ‘Exworks’ and ‘Damaged’ specimens of the uniaxial geogrid were plotted from these test data, Figures 4 a, b. The Damage Factor was then calculated according to the method described in Section 2. The varia-
Figure 4(b). Isochronous load-strain curves for damaged uniaxial geogrid at 2OoC (after Esteves, 1996).
Figure 5. Damage factor-time relationship for uniaxial geogrid at 200c.
70
tions of Damage Factor for the uniaxial geogrid with time and limiting strains of 2, 5 and 10 per cent are shown in Figure 5 . Similar test data were analysed for the woven geotextile and the variations of Damage Factor with time and strain levels are shown in Figure 6.
Isochronous Load-Strain curves for the ‘Exworks’ and ‘Exposed’ specimens of uniaxial geogrid were plotted from these test data, Figures 7a, b. The Environmental Factor was then calculated according to the method described in Section 2. The variations of Environmental Factor for the uniaxial geogrid with time and limiting strains of 2, 5 and 10 per cent are shown in Figure 8. Similar test data were analysed for the biaxial geogrid and the variations of Environmental Factor with time and strain levels are shown in Figure 9.
4.2 Environmental factor Another study was undertaken in Kuwait for the evaluation of the environmental degradation on the strength of a uniaxial and a biaxial geogrid. Kuwait experiences air temperatures ranging from 45°C in summer to 0°C in winter and has an extremely high UV radiation level with long hours of uninterrupted sunshine most of the days of the year. The test set-up for air temperature measurements comprised one temperature sensor located in a shaded housing above ground. A Pyranometer was used to measure the solar radiation at the test site. The recorded variations in air temperatures indicated a maximum of 49°C in summer and 12°C in winter with daily variations of 1 to 3°C. The cumulative UV radiation over the 12 months of the study was measured as 1,800,000 Wh/m2. The rate of radiation in the summer was twice that in the winter. Each type of geosynthetics was set out vertically on the test site and left open to all weathering conditions including direct sunlight. These were removed for testing after 12 months of exposure and severe temperature cycling on both a daily and seasonal basis. Test specimens were then cut from these samples and prepared for testing in the laboratory. Thereafter, laboratory CRS and sustained load creep tests were carried out on the ‘Ex-works’ and ‘Exposed’ specimens of the geosyntheics. CRS tests were carried out at 10% straidminute and the creep tests were carried out up to 1000 hours for each load level.
5 CONCLUDING REMARKS Total strain of a geosynthetic comprises of an Immediately Recoverable Strain part and a ‘Locked-in’ Strain part. A particular limiting strain can be reached by a geosynthetic with different combinations of these strain components.
Figure 7(a). Isochronous load-strain curves for ex-works uniaxial geogrid at 20°C (after Mudhaf, 1993).
Figure 7(b). Isochronous load-strain curves for exposed uniaxial geogrid at 20’C (after Mudhaf, 1993).
Figure 6. Damage factor-time relationship for woven geotextile in machine direction at 20’C.
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applied to long-term Reference Strength of a geosynthetic in order to obtain its long-term Design Strength. Partial Factors vary with time and strain level. Therefore, different values of Partial Factors should be applied over the design life of a GRSS for the limit state being considered, i.e while for ULS analysis Partial Factors associated with large limiting strain require to be considered, for SLS analysis Partial Factors associated with small limiting strain should be used. REFERENCES AASHTO. 1997. Standard specifications for highway bridges. Division Ii, section 7: earth retaining systems. AI-Mudhaf, H.F. 1993. The behaviour of geotextiles and geogrids following environmental conditioning in Kuwait. Ph.D. thesis, University of Strathclyde, Glasgow, UK. BS8006.1995. Code of practice for strengthenedreinforced soils and other fills. BSI, UK. Bush, D.I. 1988. Evaluation of the effects of construction activities on the physical properties of polymeric soil reinforcing elements. Intl. geotechnical symp. on theory and practice of earth reinforcement, Japan, 5-7 Oct., p. 63-68. DIBt. 1998. Theory used in the Deutches Institut fur Bautechnik Design Method. Esteves, S.A.C. 1996. The strength of geosynthetics and construction effects. M.Phil. thesis, University of Strathclyde, Glasgow, UK. HA 68/94. 1997. Design methods for the reinforcement of highway slopes by reinforced soil and soil nailing. Design Manual for Roads and Bridges, v. 4, section 1, part 4, HMSO. Jewell, R.A. & Greenwood, V.H. 1988. Long term strength and safety in steep soil slopes reinforced by polymer mat. Geotextiles and geomembranes, v. 7, p. 81-1 18. Jewell, R.A. 1996. Soil reinforcement with geotextiles. Special publication 123, CIRIA, UK. Kabir, M.H. 1984. In-isolation and in-soil behaviour of geotextiles. Ph.D. thesis, University of Strathclyde, Glasgow, UK. Koemer, G.R. & Koemer, R.M. 1990. The installation survivdbility of geotextiles and geogrids. 4th intl. conf. on geotextiles, geomembranes and related products, Hague, p. 597 -602. McGown, A. et al. 1998. Limit state design of geosynthetic reinforced soil structures. Keynote lecture, 61h int. conf. on geosynthetics, Atlanta, USA, v. 1, 25-29 Mar., p. 143-179. TBW. 1998. Theory used in the Tensar Tie-back Wedge Walls Design Method. Voskamp, W. & Risseeuw, P. 1987. Method to establish the maximum allowable load under working conditions of polyester reinforcing fabrics. Geotextiles and geomembranes, v. 6, p. 173-184. Watts, G.R.A. & Brady, K.C. 1990. Site damage trials on geotextiles. qL''intl. conf. on geotextiles, geomembranes and related products, Hague.
Figure 8. Environmental factor-time relationship for uniaxial geogrid at 20°C.
Figure 9. Environmental factor-time relationship for biaxial geogrid in longitudinal direction at 20°C.
Different amount and nature of strain components are measured in short-term CRS and long-term creep tests. These two testing are, therefore, unlikely to provide the same value of a Partial Factor even for a particular geosynthetic. Indeed, it has been shown for a range of geosynthetics in this paper that Partial Factors obtained from short-term CRS test and longterm creep test are different from each other both in terms of their value and nature. It is suggested that for long-term applications Partial Factors obtained from long-term sustained load creep test should be
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Landmarks in Earth Reinforcement, Ochiai et a/ (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 852 8
Mechanical properties in short-fibers mixture stabilized volcanic cohesive soil M. Kudo Department of Civil Engineering, Oita National College of Technology, Oita, Japan
H. Ochiai & K. Omine Department of Civil Engineering, Kyushu University, Fukuoku, Japan
ABSTRACT: Short fiber-reinforced soil, which is made up of the mixture of low quality soil materials with short fibers of polyester and polypropylene, can improve the mechanical properties and erosive resistance of materials. The stabilized soil has been known to increase strength properties, but not to reach satisfactory expectation. Therefore, the above soil is considered to mix with the stabilized soil. In the present study, short fibers are mixed with the stabilized soil of volcanic-cohesive soil in order to examine the influence on the strength-deformation properties.
1 INTRODUCTION
Table]. Physical properties of Kuroboku soil.
The short fiber-reinforced soil is the material mixed with short fibers such as polyester and polypropylene with the lengths from dozens of to hundreds of millimeters, and with thickness of about 1-1OOD. This soil can enhance such the advantageous quality as the mechanical properties and erosion-resistance (Miki,H., Fujii,K. and Obata,T, 1997j, especially, it is effective on the soil with low quality materials to be employed. By mixing short fibers with the stabilized soil of the volcanic-cohesive soil, the effect of the kinds of the stabilizer, added amount of the stabilizer, the length of the short fibers, the mixed amount of short fibers etc. upon the strength-deformation properties was examined in the present study.
density of soil particles (Mg/m3) liquid limit (%) plastic limit (%) plasticity index
2.367 152.4 93.95 58.45
~
maximum dry density (Mg/m3)
0.638
200kg/m3) per dry weight of the soil. The test specimens were prepared by mixing about 300gram of Kuroboku soil containing natural water content and 2%, 5%, 15.6% and 31.2% of stabilizer, respectively. Each stabilizer mixed with the above fixed quantity of short fibers was disentangled with thumbs and fingers and carefully mixed uniformly. The test specimens were made by the practice of “Making and Curing Statically Compacted Stabilized Soil Specimen (JGS 08 12-2000j”. The target of the compaction-degree was for the dry density to become more than 90% of the maximum density. The standard dimension of a specimen was 5cm in diameter, and 12cm in height. A mold was made of a cylinder of vinyl chloride divided into three equal parts in the vertical direction and wrapped from outside by another cylinder. And the specimen in the mold was cured for seven days. After that, unconfined compression tests were carried out with the loading rate of 1% /min. of compression strain. In addition, the direct shear test was carried out to check the cohesion and the angle of shear resistance. The specimen for direct shear test was made in the similar way as the practice of “Test Method for Onedimensional Consolidation Properties of Soils (JGS 04 11-2000)’’. The standard dimension of the specimen was 6cm in diameter, and 2cm in height. After
2 SAMPLE SOIL AND EXPERIMENTAL METHOD The specimen used in this test is “Kuroboku” soil, which is the volcanic cohesive soil in Kyushu District, and collected near Taketa City in the southwestern part of Oita Prefecture. Table-1 shows the physical properties of the soil. The specimen was collected in the disturbed state, and the natural water content was 110%. The short fibers used were made of polyester with thickness 6de(diameter 25pm) and the length 15mm and 30mm. The amount of fibermixture was set as 0.25%, 0.5% and 1.0% per dry weight of the soil. The stabilizers employed were the cement stabilizer (cement) and the lime stabilizer (lime), and the added amount was 2%, 5%, 15.6% (equivalent to 100kg/m’), and 3 1.2% (equivalent to 73
the effect of the stabilization at the level of small strain, and, as the strain becomes larger, the effect by short fiber mixture appears. Fig-3 and Fig-4 show the relationship of the peak strength between the mixed and not mixed. The vertical axis in Fig-3 shows the value qu /quso. Here quis the peak strength of the stabilized soil mixed with short fibers, and quso is the peak strength of the stabilized soil in which only stabilizer mixed. The horizontal axis is index value (= nxL / D n: amount of addition, L: length of short fiber, D: thickness ). (Miki,H., Fujii,K. and Obata,T, 1997) Index value is the one which judges the effect on reinforcement quantitatively by the difference in the length and thickness, and amount of mixture. The vertical axis in Fig-4 shows the value qu /quo. Here quis the peak strength of the stabilized soil mixed with short fibers, and quo is the peak strength of usual soil. Fig-3 shows that the increaserate of strength rises uniformly in proportion to the index value, irrespective of the existence of stabilizer, the kinds, and the added amount. Fig.4 shows that the peak strength of specimen with lime 15.6%(equivalent to 100kg/m3) added expresses the same peak strength of the specimen
cured for seven days, the direct shear test was carried out with the loading rate of 1% /min. of shear strain. 3 RESULTS AND DISCUSSION Typical stress-strain curve is shown in Fig-1 and Fig-2. While the usual soil generates the brittle failure abruptly after the peak strength, the soil mixed with short fibers shows the increase of strength, the increase failure strain and keeps the strength in spite of the progress of the strain. Fig-2 is the stress-strain curve when a large amount of stabilizer is added. In Fig-2, though the strain at peak strength of short fiber mixed soil is smaller than that of the usual one, each stress-strain curve of stabilized soil shows almost the same gradient near the part of the starting point of the curve. From this, it can be considered that the influence of the increase in the modulus of deformation due to the stabilization is bigger than that of the increase in the strain due to the fiber mixture. The stabilized soil mixed with short fibers shows
Figure 1. Stress-strain curve (quantity of cement addition 2%).
Figure 3. Rate of strength increase (qu/qUs&
Figure 2. Stress-strain curve (quantity of cement addition 15.6%).
Figure 4. Rate of strength increase (qu/quo).
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Relationship between the change of the rate of modulus deformation and the index value is shown in Fig-6. Vertical axis of the Fig-6 is the value (E50 / E50S0); the modulus deformation E50 of fiber mixture stabilized soil was divided by the modulus deformation E50~0ofstabilized soil. Modulus deformation show a tendency of increasing when index value grows big, but, as the modulus deformation can be thought to be greatly influenced by stabilization, a new value (E50 / E50.0) which is obtained from the division of the modulus deformation E50 of fiber mixture stabilized soil by the modulus deformation E50-0 of usual soil is expressed on the vertical axis in Fig.7. From this figure, the increase of modulus of deformation is thought to be due to the stabilization. Hence, the influence to the modulus of deformation by short fiber mixture is small at the level of a large amount of stabilizer addition, while the increase, though very little, can be seen by the influence of the short fiber mixture at the level of small amount of stabilizer addition. Relationship between the cohesion and the index value is shown in Fig-8. Although the cohesion is considered to be greatly influenced by stabilization, Fig. 8 shows that the cohesion also tends to increase
mixed with 5% lime and the short fibers with index value 6. The specimen added 31.2% of lime(200kg/ m3) also shows the same phenomenon when the specimen with 15.6% lime is mixed with short fibers with index value 6. The above shows that the specimen with combination of the stabilizer with smaller amount and the short fibers can produce the same strength as the strength of the soil processed only by the usual stabilization. Among the mechanical properties, it is well known that the mixture of short fibers can improve toughness of the soil. The relationship between the residual rate of the strength and index value is shown in Fig.5., and from this figure the increase of toughness can be confirmed. The vertical axis in Fig.5 is the value, when the unconfined compression strength qu15% in the compression strain 15% of each fiber mixture soil was divided by the peak strength qu. From Fig.5, according to the growth of the index value, the residual rate of strength approaches 1 and the residual strength can be expected even if big strain occurs. But, when the index value is 6, the residual rate of strength becomes 1, so that the effect cannot be expected even if the index value is enlarged any farther.
Figure 5. Rate of strength residual (qu/qu). Figure 7. Rate of modulus of deformation change (E5dE5,,.,,).
Figure 8. Change of cohesion.
Figure 6. Rate of modulus of deformation change (E5&soso).
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4 CONCLUSION The results obtained can be summarized as follows. 1 The strength is increased by mixing fibers, regardless of the existence of the stabilizer, and the added sort and amount. 2 The strength of the stabilizer mixed with short fiber efficiently is about the same as the strength of the stabilizer without any short fibers added with a large quantity of stabilizer. 3 The short fiber-reinforced soil of stabilized soil has shown the effect of stabilization at small strain level, and, as the strain becomes larger, the effect of fiber-mixing appears more and more and the toughness is improved. 4 The deformation-coefficient of the short fiberreinforced soil is greatly influenced by the increase of the coefficient by stabilization and the influence of increase of the deformation coefficient by fiber mixing is not so large.
Figure 9. Change of angle of shear resistance.
according to the increase of the index value. The relationship between the angle of shear resistance and the index value is shown in Fig-9, which shows that the angle of shear resistance tends to increase according to the increase of the index value, regardless of the addition of stabilizer. By Fig. 8 and 9, the improvement of strength properties of the soil is confirmed by the mixture of the short fibers.
REFERENCE Miki,H., Fujii,K. and Obata,T, 1997. Joint Research on Development of High Grade Soil - Design and Construction Manual of Fiber Mixed Soil Method - (in Japanese) Cooperative Research of PWRI, 1997.
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Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 02001 Swets & Zeitlinger, ISBN 90 265 1852 8
Progressive pullout failure of geosynthetic reinforcement J. Mak Road and Traffic Authority, New South Wales, Australia
S-C.R. LO University College, University of New South Wales
ABSTRACT: Progressive pullout failure of geosynthetic reinforcement in pullout testing was studied by modelling the interface strain softening with a family stress transfer curves and finite difference formulation. The numerical adequacy of such analyses was checked. A parametric study was conducted to study the influence of various factors on the simulated mobilisation curves in pullout testing.
1 INTRODUCTION
A finite difference formulation was adopted in the analytical study reported in this paper. Such an approach enable easier modelling of interface strain softening. The broad objective of this study is to examine the progressive pullout failure of geosynthtic reinforcement in pullout testing under a wide range of conditions.
Due to the elongation of a geosynthetic reinforcement under applied load, its displacement relative to the surrounding soil varies along the bonded length. When the relative displacement at the front zone is at a strain hardening state, the relative displacement at the rear zone is so minute that essentially nil interface shear stress is mobilised. When the interface shear stress at the rear zone reaches a significant value, the front zone is in a post-peak (hence strain softening) state. This leads to progressive pullout failure. Large scale pullout testing is considered to be a suitable method for studying the pullout behaviour of geosynthetic reinforcement, and a number of researchers have designed and commissioned large scale pullout testing equipment (Palmeira and Milligan, 1989; Juran et al., 1988; Fannin and Raju, 1993; and Farrag, et al., 1993, among others). Internal measurements, in the form of strain gauges or tell-tales mounted on the reinforcement, have been used by researchers to gain better interpretation of pullout test results and to examine the influence of box design on the measured progressive failure (e.g., Fannin and Raju 1993). Experimental studies are difficult because of the large number of factors that may influence the results and that large scale pullout testing is extremely resource intensive. Analytical studies have been performed to supplement such researches. These studies (e.g., Bergado & Chai 1994) were largely based on t-z curve analysis with strain softening soil springs of constant properties along the embedded length; and do not consider the evolution of stress non-uniformity of the surrounding soil induced by the pullout force (Raju et a1 1998). In theory, finite element analysis can capture the complete stress distribution in a pullout box. But modelling of strain softening plus the need to have the analysis proceeded to pullout failure, is problematic.
2 ANALYSIS MODEL The layout of the modeled pullout box is shown in Fig. 1. The length of the pullout box was either 2m or Im. However, the height of the box was 0.6 m irrespective of box length. A flexible sleeve with a length equal to 10% of the box length was incorporated in the model. A flexible sleeve is defined as one with nil shear stress transfer between the reinforcement and the surrounding soil, but allows the reinforcement to deflect freely (and compatibly) with the soil in the vertical direction. The design of a flexible sleeve is reported in Lo (1998). It is different from a rigid sleeve which constrained the vertical displacement of the reinforcement to zero by a rigid device. As
Figure 1. Pullout box layout
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discussed in Raju et a1 (1998), a flexible sleeve arrangement gives a more uniform stress field. The front wall of the pullout box (i.e., the wall near exit sleeve) was taken either as a frictionless boundary or as a frictional boundary with an interface friction angle equal to 75% of the peak friction angle of the soil. The rear wall was considered as a roller boundary. To avoid the complication of investigating the boundary condition of the rear wall, the rear end of the reinforcement is at (O.lL+O.lm) from the rear wall. The bottom boundary was fixed whereas the top boundary was subject to a prescribed uniform test pressure. In this paper, only results corresponding to a test pressure of 50 kPa is presented. The soil was modelled as a Mohr-Coulomb elasticplastic material with a non-associated flow rule. Unless stated to the contrary, the soil parameters used in the analyses were: unit weight, y = 20 kN/m3; Young's modulus, E = 20 Mpa; Poisson's ratio, v = 0.3; friction angle, (I = 40"; and dilation angle, y = 10". The reinforcement was modelled as a linear elastic one-dimensional element. Four sets of reinforcement axial stiffness, J, of 300 kN/m, 500 kN/m, 800 kN/m and 1200 kN/m, were considered in this study. Shear springs at the reinforcement nodal points were used to model the interaction between the reinforcement and the surrounding soil. These shear springs had stiffness-strength characteristics described by a family of piecewise linear shear stress transfer functions as shown in Fig. 2a. The maximum shear stress was defined by the peak interface friction angle, 8peak, which was assigned to be 30". The fully strain softened shear stress was defined by the residual interface friction angle, 8re5id, which was assigned to be 21". It is important to emphasise that &peak and 8resid were element material properties, not scale dependent average properties as assumed in some simplified equation(s) for pullout calculating pullout resistance. The algorithm will select the appropriate piecewise linear function depending on the local interface normal stress, on. Normal coupling springs of adequately high stiffness were also included in the numerical model. The analyses were conducted using a finite difference stress analysis program known as FLAC, Fast Lagrangian Analysis of Continua (FLAC 1996). In FLAC, the equations of equilibrium and stress strain behaviour are expressed in finite difference form and are solved by an explicit iterative scheme. The analysis is inherently incremental and can model material strain softening without particular numerical difficulties. The program includes looping control commands and an internal programming language for implementing user-defined procedures. These features were used in the present study to incorporate interface strain softening, and to implement a displacement controlled mode of pullout. The latter ensured that the analysis can proceed beyond peak pullout failure. The smallest element size of 78
16mm wide by 20mm height, and each analysis had at least 600 displacement increments. For the sake of clarity, not all data points for every increment were plotted with a symbol. The analysis required the prescription of horizontal stress, o h , at the initial state (ie, prior to the application of any pullout force). For a normally consolidated deposit, 01, is commonly assessed by the equation o h = (l-sin(I)ov, where Gv is the vertical stress. In a retaining wall, the effect of compaction at shallow depth leads to a horizontal stress considerably higher than that given by the above equation, and may even approach the passive pressure value (Ingold, 1979). Lo (1998) conducted pullout tests with a pre-loading pressure applied prior to the application of test pressure so that the effects of vertical compaction stress could be at least partly simulated. In this study, the initial earth pressure coefficient was taken as 1.0, unless stated otherwise to the contrary. The stability of the analysis was verified by simulating a pullout test with J=5O0x1O3 kN/m. With
Figure 2. Interface shear stress transfer
such a high reinforcement stiffness, reinforcement elongation is negligible and the peak pullout force should be the integration of peak interface shear stresses over the bonded length of the reinforcement. The pullout force displacement curve should also be similar to that of the shear spring characteristics (Fig. 2a). The simulated force displacement curve presented in Fig. 2b evidently conformed with these “known” results. It is important to emphasise that the material models and parameters adopted in this study were not specific to any particular soil or reinforcement. It is not the intention of this paper to quantitatively predict the response in a pullout testing. These were reasonable assumptions so that the complicated behaviour pattern, in particular the influence of test conditions, can be better understood. 3 RESULTS OF REFERENCE CASE
Figure 3. Pullout force displacement responses
of pullout force to 64 kN (MF=0.646), which corresponded to that given by a fully softened interface strength. The force displacement curve so predicted conformed to known trend obtained from large scale pullout testing, with the exception of the rapid reduction of the peak pullout force to a residual value. However, most pullout tests were terminated at less than 130mm displacement, and therefore may not have captured the rapid reduction of peak force to a residual value. This rapid drop in pullout force from peak value to fully softened value appears to be related to the decrease in the reinforcement tension and elongation after peak pullout force. When the reinforcement was pulled beyond the maximum pullout force, the reduction in reinforcement tension led to a reduction in the elongation of the reinforcement. Since the displacement at the reinforcement front end was prescribed as increasing, other points along the reinforcement had to move forward further and thus inducing further strain softening of the interface shear stress.
The reference case is defined as a pullout box of ength 2m with a smooth front wall, under 50 kPa test pressure and J = 500 kN/m stiffness. These conditions were considered as reasonable and probably induce slightly higher progressive failure relative to high strengthhtiffness reinforcement used in reinforced soil structures. Hence the analysis results of the reference case were examined in detail. 3.1 Force displacement behaviour The reinforcement pullout force displacement curve was presented in Fig. 3 (plot with filled-box symbol) using mobilization factor (MF) defined as: Lb tan EjPeak) M F = P / ( 2 ovO where P = pullout force (at exit location), ovO = pressure applied at the top boundary plus vertical stress due to the self weight of 0.3m height of soil, LI, = initial bonded length of reinforcement. The upIn general 1 2 MF 2 tan(Ejres,d)ltan(~peak). per bound value of unity applies when the reinforcement can be considered as inextensible, whereas the lower bound value was for the residual pullout resistance. To eliminate any aberration of the mobilisation curve due to elongation of the unbonded reinforcement (within the sleeve), the reinforcement displacement at the beginning of the bond length (i.e., at 0.1L from the exit) is plotted in this figure. A maximum pullout force of 71 kN (MF=0.717) was mobilized at 130 mm pullout displacement. The peak pullout force was lower than that given by the peak interface strength, as expected, but higher than that given by the fully softened interface strength. Once the peak pullout force was achieved, further application of displacement at the reinforcement front end led to a rapid reduction
3.2 Stress distribution The distribution of vertical stress changed with application of pullout force as illustrated in Fig. 4a. At P = 35 kN (ie, 50% of maximum pullout force), the normal stresses acting on the reinforcement (Gn) were smaller than the average overburden stress (56 kPa) within the front region defined by a distance of less than 0.5 m of the pullout box. In the mid region, defined by a distance of about 0.5 m to 1.4 m from the front wall, O n was higher than the average overburden stress and had a maximum value of 61 kPa. In the rear region, defined by a distance greater than 1.4 m behind the front wall, On was essentially equal to the average overburden stress.
-
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Figure 4. Stress distribution at mid-height of pullout box
Figure 5. Shear displacement and interface shear stress of reinforceinent
At maximum pullout force, On within the front region of the pullout box remained essentially the same as above. In the mid region, On was close to the average value. ol, in the rear zone was highly nonuniform. q1increased rapidly between the distance of I .4 m to 1.6 m. It reduced sharply to 30 kPa at the rear end of the reinforcement, but On jumped back to 72 kPa slightly behind the rear end of the reinforcement. When the pullout force was at the residual value, the on distribution along the reinforcement was similar to that at peak pullout force. The rapid reduction of onin the rear zone of the reinforcement can be explained by looking at the distribution of horizontal stress as presented in Fig. 4b. The pullout force led to reduction in horizontal stress in the rear zone of the reinforcement. At maximum pullout force, the horizontal stress was reduced to such a low value that the soil elements approached failure, which in turn limited the value of vertical stress.
tially zero. Peak shear stress occurring at about 0.6m from the exit. This implies significant strain softening had already occurred along the front portion. Therefore, the relevance of modelling interface strain softening was demonstrated. Maximum pullout force occurred with slippage of the rear end of the reinforcement, as indicated by a shear displacement of about 3mm. A significant portion of the embedded length, however, had attained the fully softened state. The maximum interface shear stress at maximum pullout force was slightly higher than that at P=35 kN. At residual pullout force, the interface shear stress along nearly the entire bonded length was at the fully soften state but there was a "small peak" near the rear end of the reinforcement. Both phenomena was due to stress nonuniformity induced by the pullout force.
4 INFLUENCE OF REINFORCEMENT STIFFNESS AND BOX LENGTH
3.3 Relative displacement and interjace shear stress The shear displacement of the reinforcement (ie relative to the surrounding soil) was presented in Fig. 5a, whereas the interface shear stress was presented in Fig. 5b. At P = 35 kN, ie -50% of maximum pullout force, shear stress and displacement along a significant part of the reinforcement were essen-
The influence of reinforcement stiffness was examined by conducting the analysis for different J values of 1200kN/m, 800 kN/m and 300 kN/m. The mobilisation curves so obtained were compared to that of the reference case (J= 500 kN/m). As evident from
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Fig. 3, the reinforcement stiffness has a slight influence on the peak mobilised value but a significant influence on the rate of mobilisation. The effect of box length was studied by repeating the analyses for a l m box. As evident from Fig. 3, the shorter box gave a higher peak mobilised value and a significantly higher rate of mobilisation. This is in agreement with known trend deduced experimentally (Fannin and Raju 1993). The mobilisation curves of the shorter box also appeared to be less sensitive to the reinforcement stiffness, J.
5 INFLUENCE OF SOIL STIFFNESS The analysis for 2m box was repeated for a range of soil stiffness with E varying from a low value of 8.33 MPa to a high value of 100 MPa, but with J kept at 500 kN/m. This means the J/E ratio was varied. It is sometimes assumed that the effects of change in either J or E can be approximated by changes in J/E ratio. As evident from the mobilisation curves presented in Fig. 6, change of J/E caused solely by change in E had minimal influence on the mobilisation of pullout resistance. This was very different from change in J/E caused by change in J but with E kept constant (see Fig. 3).
Figure 7. Influence of sleeve and front wall 0 0
no-sleeve and a smooth rough front wall, no-sleeve and a rough front wall.
The mobilisation curves for these six cases were presented in Fig. 7. The curves obtained from the first three conditions were essentially the same as the reference case of flexible sleeve and smooth front wall. The curves for the two cases for nosleeve, however, manifested slightly higher pullout displacement. This is due to the additional 0.2m in reinforcement length due to the elimination of sleeve. This finding conformed to that obtained by Raju et a1 (1998) based on a synthesis of a limited experimental results.
6 INFLUENCE OF FRONT WALL AND SLEEVE DESIGN The influence of the front wall roughness and the provision of sleeve were also examined by conducting the following additional analyses for the 2m box. 0 flexible sleeve and a rough front wall, rigid sleeve and a smooth front wall, 9 rigid sleeve and a rough front wall,
7 INFLUENCE OF NUMERICAL MODELLING The assumption of an initial horizontal to vertical stress ratio, Ki, of unity is somewhat arbitrary although tenable. Therefore, the reference case was reanalysed with K, = 0.75 and Kiz0.5. As evident from Fig. 8, The resultant mobilization curves are essentially identical to that of reference case. It is also recognised that friction angle may increase at lower confining stress. To model this effect the variation of friction angle with confining stress in triaxial testing was predicted by Lade's failure criterion (Lade 1977) reproduced as Eqn (2).
where 11 and 13 are the first and third stress invariance, pa is the standard atmospheric pressure, and q and m are soil parameters. q = 65 and m = 0.35 was adopted which then gave 4 = 40" at pa. An increase in friction angle will lead to a corresponding increase in dilatancy angle, y~ and this was modelled using the relationship proposed by Bolton (1985)
Figure 6. Influence of soil stiffness
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minimal effect on the simulated mobilisation curves. The front wall roughness and sleeve design were also found to have only slight influence on the simulated mobilisation curves. Two modelling factors, (i) the value of Ki, and (ii) changes in @ and y~ with 0 3 , were also studied. Both soil modelling factors had minimal influence on the simulated mobilisation curves.
9 ACKNOWLEDGEMENT The authors wish to thank the Roads and Traffic Authority, New South Wales, Australia, for allowing the first author in participating in the publication of this paper. The opinions and findings expressed in the paper are, however, solely those of the authors.
REFERENCES
Figure 8. Influence of numerical modelling
@ = @cv + 0 . 8 ~
(3) where @cv is the friction angle at critical state and a value of 32’ was adopted in the analysis. Despite all these changes, the mobilisation curves for these three analyses were essentially identical to that of the reference case as evident from Fig. 8.
Bergado, D.T. and Chai, J. (1994) “Pullout Force/Displacement Relationship of Extensible Grid Reinforcement”. Geotextiles and Geomembranes, 13, 1994, 294-3 16. Bolton M.D. (1986). The strength and dilatancy of sand. Geotechnique, 36[1], 65-78. Fannin, R. J., and Raju, D. M. (1993). On the pullout resistance of geosynthetics. Canadian Geotechnical Journal, Vol. 30, June 1993, pp. 409 - 417. Farrag, K., Acar, Y.B. and Juran, I. (1993). “Pullout resistance of geogrid reinforcement”. Geotextiles and Geomenzbranes, 11 (2): 133-159. FLAC ( 1 996) “Fast Lagrangian Analysis of Continua”, Version 3.3. Theory and background. Itasca Consulting Group. Ingold (1979). The effects of compaction on retaining walls. Geotechnique, 29: 265-283. Juran I., Knochenmus G., Acar Y.B., and Arman A. (1988). Pullout response of geotextiles and geomembranes. Proc Symposium on Geposynthetics for Soil Improvement, ASCE, Tennessee, 92-1 1 1. Lade P.V. (1977) Elastoplastic stress strain theory for cohesionless soils with curved yield surfaces. J of Structures, Vol. 13, 1019-1035. Lo,S-C.R. (1998). “Pull-out resistance of polyester straps at low overburden stress”. Geosynthetics International, 5 (4): 36 1-382. Palmeira, E.M. and Milligan, G.W.E. (1989). “Scale and other factors affecting the results of pull-out tests of grids buried in sand”. Geotechnique, 39 (3): 5 11-524. Raju D.M., S-C.R. Lo and Gopalan M. (1998) “On large scale pullout testing”. Geotechtzical Eng J of SE Asian Geot. Societ)),29 (4).
8 SUMMARY AND CONCLUSIONS Progressive pullout failure in pullout testing of geosynthetic reinforcement was studied analytically by modelling interface strain softening with finite difference analysis. Despite a uniform stress was prescribed at the upper boundary, the normal stress on the reinforcement became non-uniform due to the application of the pullout force. Therefore, interpretation of internal force and displacement measurements would be difficult unless internal distribution of normal stresses is also measured. A parametric study was conducted to study the effects of different factors on simulated mobilisation curve. Increasing J and/or decreasing pullout box length led to higher peak MF value, considerably higher rate of mobilisation, and an apparently slower rate of strain softening of the pullout force after the peak. Changes in soil stiffness, E, however, had a
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Strain-induced toughness and shearing characteristics of short-fiber reinforced soils Katshuhiko Makiuchi & Kunio Minegishi Department of Transportation Engineering and Socio-Technology, College of Science & Technology, Nihon University, Japan
ABSTRACT: In order to study the quantitative effects of short-fiber reinforced soils on shear behaviors, a series of box shear test and unconfined compression test on non-cohesive and cohesive soils were carried out. Reinforcement materials used in the tests were consisted of synthetic fibers with different sizes and surface friction properties. It was found from the experimental works that length, diameter and surface roughness of fibers had marked effects on mechanical properties of the reinforced soils and the recommended optimum mixing rate of the short fibers was presented. Furthermore test results showed the prominent increase in peak shear resistance of fiber-reinforced soils and their high residual strength within the wide range of large deformation. Its strain-induced toughness can be recognized as effective and well-directed for earthquake-proof geotechnical structures. equipment and the fiber-sand mixture is built up in the field. Another type is a recently developing method using short length fibers (Research Institute of Public Works, 1997). As the results of improvement, the fiber-reinforced soils acquires extremely high ductility and toughness against the deformation of geotechnical structures and contributes to reduce the lateral earth pressure within embankments and back-fill soils. Some interaction effects among soils types, shape, size and surface roughness of reinforcement materials are commented in other reports (Research Institute of Public Works, 1997 and Nakahara,H., 1998). In this experimental works mechanical properties of fiber-reinforced fine sand and sand-clay mixture were investigated using both a laboratory shear box test and an unconfined compression test.
1 INTRODUCTION A short-fiber reinforced-earth construction method is recently developed in attempting to use for unstable embankment slopes and poor ground foundations, and in particular for the backfill soils of wall-type reinforced-earth with the aim of strengthening the shear resistance. However theoretical and experimental reinforcement mechanism and fundamental mechanical properties of fiber-soil mixture are not made fully clear at present. Interface friction among soil particles and fibers and their interlocking or intertwine action are considered as fundamental factors of improvement of fiber-reinforced soil which is called as an internal confining reinforcement in this paper. The fiber-reinforced techniques has some advantageous engineering properties of soft grounds and filling materials. For example, the fiber-reinforced soil construction method can be applied for purposes of supplying an apparent cohesion component for non-cohesive granular sandy soils. Furthermore it can be used for controlling the engineering properties of problematic soils for stabilizing geotechnical structures and for vegetation, erosion control of earth slopes, weak disposal soil materials and so on. There are two major types of earth-reinforcement techniques using synthetic fibers. One is a widely used conventional method in which continuous filament yarns are employed for non-cohesive granular soils, for example, Texsol product (Public Works Research Center, 1992) developed firstly in French. In this type the filaments are mixed with fine sand at the specified moisture content by the jet-mixing
2 MATERIAL CHARACTERIZATION 2.1 Soils Two types of soils; Toyoura fine sand (abbreviation: S) and Kaolin clay (abbreviation: K) were used for this experimental works. Physical characteristics of the fine sand were; particles density ps = 2.64 g / ~ m . ~ 60 % passing gain diameter d60 = 0.2 mm, uniformity coefficient Uc = 2.0, coefficient of curvature Uc’ = 1.45, maximum and minimum void ratios emax= 0.97, emin= 0.59, respectively. Physical properties of the clay (CH: high plasticity clay) were; particles density ps= 2.56 g/cm3, liquid limit W L = 85.1 %, plasticity index Zp = 54.6.
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The fine sand (S) as a non-cohesive material was used in the air dry condition in a direct shear box test. The soil employed as a cohesive material in an unconfined compression test was a mixture (S+K) of the sand and the clay with the mixing rate of 10 % in dry mass. This mixing rate was determined based on the results by pre-compaction test.
3 PROCEDURE 3.1 Specimen preparation In Test I, the air-dried fine sand mixed with the short fibers at the mixing rate 1.0 % were used for a shear box test of 6 cm in diameter and 2 cm in height. The sand specimens mixed with the fibers were compacted under the condition of dry density of 1.43 to 1.45 g/cm3, viz. density index ID (or relative density, Dr) between 45 and 55 %. In Test 11, firstly fine sand and clay were mixed and then the specified amount of reinforcement fibers were blended uniformly with the mixture. The unconfined compression test specimens of 6 cm in diameter and 14 cm in height and the specimen's dry density of 1.74 g/cm' were compacted using a steel mould subjected to the static compression force of 14.7 kN in the condition of the optimum moisture content of 18 %. The specimens used in Test I11 were prepared in the same condition and the similar way to Test.
2.2 Reinforcement materials Reinforcement short fiber materials used in the tests were nylon fibers for Test I and I1 and polypropylene fibers for Test 111. Table 1 and 2 show the sizes of nylon fibers tested in a direct shear box test (Test I) and in an unconfined compression test (Test 11), respectively. One group is L = 1.0 cm in length with diameters of d = 0.175, 0.33 and 0.66 mm, and another group is d = 0.33 mm in diameter with L = 0.5 to 16 cm in length. A polypropylene fiber used in Test I11 is L = 1.0 cm in length and two types of the fiber's surfaces are prepared. One has a non-treated smooth surface and another one is a rough surface fiber to which the fine sand particles are adhered with a chemical bond (silicone-sealant 8060). The mixing ratio of these fibers with the soils was maintained at the constant rate of M = 1.0 % of fibers in dry mass in order to investigate the fiber's size and surface friction effects in these types of soils.
3.2 Direct shear test Direct shearing resistance illustrated in the shear stress-deformation-strength relationships of fiberreinforced sand (S) were investigated using a standard type shear box device. The shear tests were conducted at the deformation rate of 0.25 m d m i n and up to 6 mm in the displacement. 3.3 Unconfined compression test
Table 1. Reinforcement materials for shear box test.
Soil
LengthL (cm) 1.o
Sand: S
0.5 . _ 1.o 2.0
Diameter d (mm) 0.175 0.33 0.66
Stress-strain relationships and peak and residual strength properties of fiber-reinforced soils (S+K) were investigated using an uni-axial compression apparatus. The compressive tests were conducted at the compressive strain rate of 1.0 %/min and up to about 7 to 10 % in axial strain.
Mixing rate M (%)
1
.o
0.33
4 MECHANICAL CHARACTERISTICS OF SHORT-FIBER REINFORCED SOILS
Table 2. Reinforcement materials for unconfined compression test.
Soil
LengthL (cm) 1.o
Mixture: S+K
Diameter d (mm)
1
4.1.1 Effects offiber's length Figure 1 shows the effects of fiber's length on shear resistance of the fiber reinforced fine sand, under conditions of the normal stress 0 = 200 kPa, the fiber's diameter of 0.33 mm and the fiber's mixing rate of 1.0 %, in a direct shear testing. It is found from these shear stress-displacement curves that longer the length of fiber, higher the shear resistance such as peak strength (increase of about 10 to 15 %), residual strength (increase of about 0 to 20%) and modulus of deformation. It is considered that the reinforcement effects of fibers are caused by interface friction, interlocking and intertwining between the fiber's surfaces and the
Mixing rate
M
(%)
0.33 0.66
0.5
1.o
.o
2.0 an 8.0
4.1 Fiber-reinforced sands
0.33
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4.1.2 Effects offiber's diamater Figure 4 demonstrates that the effects of fiber's diameter on shear resistance of the reinforced sands using the fiber's length of 1.0 cm, under the same conditions of mixing rate as indicated in Fig.l. It is found from the figure that smaller the diameter, the higher the peak strength and the modulus of deformation. This increasing shear resistance (increase of about 0 to 25 % in the peak strength) is owing to the increased surface area of fibers that are depending to the numbers of fibers in inverse proportion to the diameter of fibers at the constant fiber's mixing rate. However, No difference can be recognized at the residual strength. Figures 5 shows the internal friction angles affected by the fiber's diameter obtained from the normal stress and shear stress diagrams. It can be seen from the figure that the internal friction angles decrease with the increase in fiber's diameter at the specified mixing rate and will approach finally to almost same value of internal friction angle @= 29.6"of the non-reinforced sand.Therefore, in order
sand particles. In addition, the high peak and residual strength exhibited in the cases of longer fibers are attributable to the phenomenon of expanding shear zone in proportion to the length of fibers. Figures 2 shows the internal friction angles affected by the fiber's length obtained from the normal stress and shear stress diagrams. It can be seen from the figure, however, that no cohesion component may be supplemented to granular soil particles by short-length fiber's reinforcement. Figures 3 illustrates the relationships of the internal friction angles (@)versus the fiber's length (L). It can be seen from Figs. 3 that the internal friction angles of the reinforced sands will increase gradually with the increase in fiber's length. This is probably due to a long continuous friction action of each fiber that will expand effectively the confining zone of sand particles. Therefore, in order to reinforce the fine sands, the optimum size regarding length is supposed to exist at the specified mixing rate and kinds of soil.
Figure 1. Shear curves in direct shear test (Effects of fiber's length). Figure 3. Relation between internal friction angle and fiber's length.
Figure 2. Normal stress vs shear stress curves in direct shear test (Effects of fiber's legth). Figure 4. Shear curves in direct shear test (Effects of fiber's diameter).
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to reinforce the fine sands, the optimum size regarding length and diameter is supposed to exist at the specified mixing rate and kinds of soil.
Especially the residual strength in the cases of specimens having more than 4.0 cm in the fiber's length are maintained for a very wide range of strain, for example beyond more than 5 %. This means that cohesive soils can be extremely improved by the short fiber reinforcement and they have high ductility and toughness for shearing deformation. Concerning the induced compressive strength due to its deformation in an unconfined compression test, the relationships between ~compressivestress and the fiber's length are illustrated in Fig.7. These compressive stresses mean the values mobilized at the arbitrarily specified axial strain, such as the levels of 0.5 to 8.0 %. It is indicated in this figure that the compressive strength mobilized at the strain level of 2 % or more are increased with the increase in fiber's length at the range of 0.5 to 4.0 cm and are maintained constant or decreased at the range of more than 4.0 cm. This tendency is approximately consistent with the empirical value of 10 cm that is supposed to be the optimum length of fibers from the viewpoint of mixing and field construction efficiency. Figures 8 shows the relationships between the strength reduction rate and the axial strain for dis-
4.2 Fiber-reinforced sand-claymixtures 4.2.1 Effects offiber's length Figures 6 shows the effects of fiber's length on the axial stress-strain relations of the fiber-reinforced sand-clay mixtures under conditions of the fiber's diameter of 0.33 mm and the fiber's mixing rate of 1 .O %, in unconfined compression testing. It is found from these unconfined compressive stress-strain curves that longer the length of fiber, higher the shear resistance. This tendency is similar to the cases in Test I with the same reasons of interface friction, interlocking and intertwining action between the fibers and the soils. However, it can be seen that the marked reinforcing effects are obtained in the case of the soils having cohesion component on both the peak strength (increase of up to about 2.5 times) and residual strength (increase of up to about 5 times).
Figure 5. Normal stress vs shear stress curves in direct shear test (Effects of fiber's diameter). Figure 7. Relation between compressive strength and fiber's length.
Figure 6. Stress-strain curves(Effects of fiber's length).
Figure 8. Relation between strength reduction rate and compressive strain (Effects of fiber's length).
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mixing rate. It must be noted that no significant difference of the initial or tangential modulus of deformation can be found among these specimens as shown in Figs.6 and 9. Figure 10 demonstrates that the mobilized compressive strength increases with the decrease in fiber's diameter at this mixing rate. It is found that fabric-reinforced cohesive soils behave prominently high residual strength in the unconfined compression test. Figures 11 shows the relationships between the strength reduction rate and the axial strain for displaying the effects of fiber's diameter, but no difference can be seen among fiber's diameter in this experimental mixing rate condition.
playing the effects of fiber's length. The strength reduction rate that indicates the toughness and ductility is defined as the ratio of the difference between the peak and the residual strength to the peak strength. It can be seen from Fig.8 that the strength reduction rate increases with their strain increases and decreases with the increase in the fiber's length. Namely the soils reinforced using longer fabrics retain high strength.
4.2.2 Effects offiber's diameter Figure 9 demonstrates that the effects of fiber's diameter on the axial stress-strain relations of the fiber-reinforced sand-clay mixtures using the fiber's length of 1.0 cm, under the same conditions of fiber's mixing rate as indicated in Fig.6. It is found from the figure that smaller the diameter, the higher the peak strength (increase of up to about 2 times) and residual strength (increase of up to about 2.5 times at the same strain level). This increasing shear resistance is owing to the increased circumferential area of fibers that are depending to the numbers of fibers in inverse proportion to the diameter of fibers at the constant fiber's
4.3 Efsects offiber's su$ace roughness Figure12 shows that the reinforced soil by rough surface fibers demonstrates higher shear resistance than that by the smooth surface fibers. It must be noted that the smooth surface fiber's reinforcement can not maintain its residual strength and is consistence with that of non-reinforced soil at larger strain range. This effect may be caused by a mobilized friction and an interlocking action between fiber and sand.
Figure 9. Stress-strain curves (Effects of fiber's diameter).
Figure 1 1. Relation between strength reduction rate and cornpressive strain (Effects of fiber's diameter).
Figure 10. Relation between compressive strength and fiber's diameter.
Figure 12. Shear curves in direct shear test (Effects of fiber's surface roughess).
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No cohesion component may be supplemented to granular soil particles by short-length fiber's reinforcement. The marked reinforcing effects are obtained in the case of the soils having cohesion component on both the peak strength and the residual strength. The mobilized compressive strength is increased with the increase in fiber's length at the range of 0.5 to 1.0 cm. This tendency is consistent with the empirical value of about 10 cm that is supposed to be the optimum length of fibers from the viewpoint of mixing and field construction efficiency. The residual strength in the cases of specimens having more than 4.0 cm in the fiber's length are maintained for a very wide range of strain. This means that cohesive soils can be extremely improved by the short fiber reinforcement and they have high ductility and toughness for shearing deformation. The reinforcement effect is influenced by the surface roughness of fibers and the rougher the surface of fiber gives higher the friction angle.
Figure 13. Relation between mixing rate and internal friction angle.
Figure 13 shows that the internal friction angles of both smooth and rough surface fiber's reinforcements increase with the increase in a mixing rate and the angle by rough surface fiber gives a higher value than that by sooth one.
5 CONCLUSIVE REMARKS REFERENCES
Experimental results demonstrated marked effects of the length and diameter of fibers on shear stressdeformational properties of the fiber-reinforced soils. High residual strength in the wide range of large deformation of the reinforced soils was obtained and their toughness and ductility are recognized as beneficial for anti-earthquake geostructures. The principal findings from the results can be summarized as follows; 1) The internal friction angles of the fiber-reinforced non-cohesive sands increase with the increase in fiber's length or with the decrease in fiber's diameter.
Nakahara,H., Watanabe,T., Makiuchi, K. & Minegishi, K. 1998. Effects of Surface Roughness of Fiber Elements on Mechanical Properties Reinforced Soils, Proc. of 13rd Geosynthetics Symposium, Japan Chapter of International Geosynthetics Society: 332-336. (in Japanese) Public Works Research Center 1992. Accreditation Report of the Construction Technique of Texsol:152. (in Japanese) Research Institute of Public Works, Ministry of Construction 1997. Joint Research Report of Technical Manual on Application of Earth Reinforcement mixed with Short Fiber: 66. (in Japanese).
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Landmarks in Earth Reinforcement, Ochiai et al (eds), 0 2001 Swets & Zeitlinger, ISBN 90 265 1852 8
Modelling the behaviour of geosynthetic reinforcements used to resist combined sustained and shock loading J. Kupec Department of Civil Engineering, University of Strathclyde, U K
A. McGown Department of Civil Engineering, University of Strathclyde, UK
ABSTRACT: The shortcomings of the current design codes / methods related to Geosynthetic Reinforced Soil Structures subject to sustained loading plus seismic loading are identified. A possible solution to these is presented, which is based on the use of the Isochronous Strain Energy [ISE] approach. It is suggested that this provides an effective means of analysing the isothermal load-strain-time behaviour of geosynthetics subject to multi-stage loading. In particular it identifies the "Immediately Recoverable" and "Locked-in" components of loads and strains developed during the various stages of loading. Knowing these components, the ability of geosynthetics to resist Multi-Stage Actions, e.g. a sustained loading (self-weight) combined with shock (seismic) loading, can be predicted. On this basis, a fundamental approach to determining the loadstrain-time-temperature behaviour of geosynthetic reinforcements subject to combined sustained and shock (seismic) loading is set out. Thus in this paper, the shortcomings of the current design codes/methods related to sustained plus shock (seismic) loading of GRSSs are discussed. A possible solution to these shortcomings is then presented which is based on the use of the Isochronous Strain Energy [ISE] Approach. On this basis a new approach is set out for the determination of the loadstrain-time-temperature behaviour of geosynthetics subject to combined sustained and shock (seismic) loading.
1 INTRODUCTION Geosynthetic Reinforced Soil Structures [GRSSs] may be subjected to a wide variety of Actions, (loads or deformations), during their design lifetimes. McGown (2000) suggests that some of these are "Sustained Actions", which persist throughout the life of the structure. Others may be represented as "Equivalent Sustained Actions", which may be accurately represented by a sustained load, or deformations. Some other combinations of loading or deformations cannot be represented by either of these and must be treated separately, "Multi-Stage Actions". Thus appropriate input data for the design of "Sustained Actions" or "Equivalent Sustained Actions" may be obtained from "Single-Stage Loading Tests", but for "Multi-Stage Actions", "Multi-Stage Loading Tests" must be used and these should impose a similar sequence of loading, (or deformation), to that imposed under operational conditions. Greenwood (1 996) recognised that data from "Single-Stage Loading Tests" did not represent the behaviour of geosynthetics subject to Multi-Stage Actions. He suggested that geosynthetics possessed a "Residual Strength" which considerably exceeds the "Design Strength" obtained from sustained loading (creep) tests. The theoretical basis of this statement is not made clear but the point made is correct, in so far as it highlights the fact that sustained loading (creep) tests, being Single-Stage Loading tests cannot represent the behaviour of geosynthetics subject to multi-stage loading.
2 CURRENT DESIGN CODESMETHODS USED TO REPRESENT THE BEHAVIOUR OF GEOSYNTHETICS SUBJECT TO COMBINED SUSTAINED AND SHOCK LOADING To date the development of designs for geosynthetic reinforced soil walls and slopes involving earthquake forces has been empirically based. Fukuda et a1 (1994) reported that until 1993, designs for earthquake loading were based on the procedures for structures subject to sustained loading suggested by Jewel1 et a1 (1984). In this procedure the long-term creep rupture strength of geosynthetics was used as the "Reference Strength". The structures so designed, were reported by Collin et al (1992) to have maintained their stability during the Loma Prieta earthquake in 1989, which had a magnitude of 7.1. Fukuda et a1 (1994) also reported a similar situation following the Kushiro Offshore earthquake in 1993, which had a magnitude of 7.8. These data were 89
taken to indicate that geosynthetics were capable of taking higher loads applied rapidly, than the longterm creep strength used in their design. On this basis, Fukuda et a1 (1994), AASHTO (1994) and Jones (1996) suggested that the Reference Strength of geosynthetics for sustained loading should be increased by 1.5 times when designing for sustained loading plus short-term earthquake loading. In more recent design codes/methods, as even more confidence was gained from the performances of Geosynthetic Reinforced Soil Structures during earthquakes, factored CRS strengths were suggested for use in designs for structures subject to sustained loading plus earthquake loading, e.g. AASHTO (1997), NCMA (1997) and DIBt (1998). These suggested approaches to the choice of the Design Strength for geosynthetics subjected to combined sustained and shock loading are all empirically based and have not been technically justified in detail. In fact, like Greenwood (1996), they are essentially reflecting the widely held judgement that geosynthetics can support a greater combined sustained and shock loading than is presently identified for a sustained load alone in design codes/methods. The principal shortcomings of the current approaches are that they do not take account of a) the timing of the shock loading during the design lifetime b) the possibility of the repetition of the shock loading, and c) the strains induced in the structure before, during and after the shock loading. In order to overcome these shortcomings it is necessary to understand the behaviour of geosynthetics under combined sustained and shock loading and secondly to consider the reaction of the soil-geosynthetic composites to such combined loading.
strokes of the Kushiro Offshore and Northridge earthquakes, as reported by Fujii et a1 (1996) and Frankenberger et a1 (1996), respectively, and is actually more critical than the actual earthquake loadings. A sustained loading [P,] of 25 kN/m was applied over 200 hours, with the shock loading [AP,] applied after 100 hours for 20 sec. Five shock loading levels were applied from 10 to 50 kN/m, in increments of 10 kN/m. (The maximum total load of 75 kN/m was the same value as the strength obtained from CRS testing at 20°C with a strain rate of 25% per minute). The test data are given in Fig.1 and it can be seen that only with the additional shock load of 50 kN/m did the material rupture. For all lower levels of shock loading, the strain induced was partially recovered immediately on unloading. The strain then steadily reduced to an almost constant value within the next 100 hours. Indeed for shock loads of 10 and 20 kN/m, the subsequent strain behaviour rapidly approached that for the sustained load 25 kN/m with no shock load.
3 TESTS DATA FROM A GEOSYNTHETIC SUBJECT TO COMBINED SUSTAINED AND SHOCK LOADING
To illustrate the behaviour of geosynthetics subject to combined sustained and shock loading, Kupec (2000) undertook a series of laboratory tests on a HDPE uniaxial geogrid. The tests were all carried out at 2OoC, at which temperature the geogrid exhibited a wide width strength of 80 kN/m at a constant rate of strain of 30 % per minute. It has a long-term rupture strength under sustained loading of 33.33kN/m according to BS8006 (1995). The shock load used was chosen to represent a single earthquake loading. Usually, earthquakes are cyclic in nature with irregular frequency, however, to avoid the complexities of simulating these loading cycles, an earthquake was represented by a uniform load applied over 20 seconds. This loading period was chosen on the basis of the durations of the main
Figure 1. Loading scheme and test results.
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Thus, these data indicate that at combined loading values much less then the wide width CRS strength, a very high proportion of the induced shock loading strain was recovered over a short period of time. However, it must be made clear that in these tests the sustained load was applied for only 100 hours prior to the shock loading. The strain developed in the geosynthetic prior to the shock loading was therefore relatively small. If the sustained load had been applied over a much longer period, then the strain have been greater and it is likely then that the combined load required to cause failure would have been less. Indeed, if the sustained load had been left long enough, it would have caused strains large enough to cause creep rupture. In such a case no additional shock loading is required to cause rupture. Thus it may be suggested that the strength available to resist shock loading in geosynthetics following different periods of sustained loading will vary. In order to determine just how this varies requires an understanding of the response of elasto-visco-plastic materials to combined loading.
Figure 2. Idealised sustained load (creep) curve at constant temperature.
Figure 3. Strain response from sustained loading at different load levels.
4 THE RESPONSE OF ELASTO-VISCOPLASTIC MATERIALS TO LOADING Geosynthetics exhibit elasto-visco-plastic behaviour, viz. when loaded they develop initial elastic and plastic strains then with time develop primary, secondary and then tertiary creep strains which may lead to rupture, Fig. 2. The possibility of developing creep rupture depends on there being both a sufficient load level and a sufficient period of loading, Fig. 3. If either the time or load is insufficient to cause creep rupture, then when the load is removed in part or in whole, there will be an immediate elastic rebound followed by a time dependent rebound. In most situations, there will always be a permanent irrecoverable strain, Fig. 4. Representation of this complex behaviour can be achieved using rheological models, however, McGown (2000) has proposed a new proach to modelling this behaviour which is termed the “lSoChronoUS Strain Energy” [ISE] approach* This has the advantage Of being to make direct comparisons of data obtained from different testing methodologies and to allow consideration of the effects of combining different load, or deformation sequences, i.e. “Multi-Stage Loading”. Thus it is proposed that this approach should be used to analyse the behaviour of geosynthetics under combined sustained and shock loading.
Figure 4. Idealised strain response from sustained loading and
5 THE ISOCHRONOUS STRAIN ENERGY APPROACH For single-stage loading under isothermal conditions, the external work done per unit width of a geosynthetic at any time (t) may be taken to be equal to the llAbsorbed Strain Energy All single-stage loading test data can be represented by Isochronous Load-Strain curves and the areas under the curves represent for any specific time, i.e. the Isochronous Strain Energy, Fig. 5. It should be noted that a feature of the ISE approach is that data obtained at the same temperature from different load-strain paths may be plotted on the same ISE - Time plot, McGown (2000). I!.
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Figure 5. Calculation of isochronous strain energy.
The unit of Isochronous Strain Energy [ISE] for geosynthetics is: Force per unit width times unit strain = (kN/m) x (m/m) = kN/m. To avoid confusion with existing definitions of strain energy it is suggested that another unit for ISE is used as follows: ISE = (kN/m) x (m/m>= (kNm/m2>= H/m2 Thus at any temperature (T) and time (t) after the application of a particular loading regime, there will be a finite amount of work done per unit width, which can be represented as the "Absorbed ISE" [A],. The amount of ISE to develop a limiting strain or rupture at that temperature for a particular SingleStage Loading regime is termed the "ISE Capacity" [ C ] , of the geosynthetic at the specified time (t), Fig.
6(4.
Additionally, Fig. 4 can be reinterpreted in terms of the ISE approach. Upon application of the load immediate strains occur so that there will be an Absorbed ISE at time to (point 1). With time, the Absorbed ISE will increase to the Absorbed ISE at time tl (point 2). If the load is removed the Absorbed ISE at time t 2 (point 3) will be reduced and then continue to reduce to that at time t3 (point 4). This shows that at time tl there were two components of ISE. One component was "Immediately Recoverable" on unloading whilst the other was "Locked-in" at time tl. This "Locked-in ISE" was partially recoverable with time however some was "Irrecoverable Locked-in ISE". Hence at any time, [tl], the Absorbed ISE comprises two components, which are the "Immediately Recoverable ISE" [RIt and the "Locked-in ISE" [L],. These components vary with time for any limiting strain condition or rupture. The isothermal ISE Components for any singlestage loading, i.e. [RI, and [L],, can be calculated from the areas under the respective "Isochronous Load-Strain'' curves for different times and strains, Fig. 6 (b) and (c), McGown (2000). These can then be used to produce the plots shown in Fig. 7.
Figure
'.
Of
IsE.
Figure 7. Derivation of ISE Capacity [C], and ISE Components [RI, and [L], plot.
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6 INTERPRETATION OF THE BEHAVIOUR OF THE GEOSYNTHETIC SUBJECTED TO COMBINED SUSTAINED AND SHOCK LOADING USING THE ISE APPROACH The ISE approach can be used to interpret the test data from the uniaxial geogrid under the combined sustained-shock loading, using the following procedure: a) The Recoverable ISE [RI,, and the Locked-in ISE [L],1 after lOOhours of sustained loading are calculated in the manner set out in the previous section and as described in detail by McGown (2000). b) The Additional Recoverable ISE [AR], and the Additional Locked-in ISE [AL], due to the shock loading are determined by constructing the Isochronous Load-Total Strain, LoadRecoverable Strain and Load-Locked-in Strain curves. The Additional Recoverable ISE [AR], and Locked-in ISE [AL], in this stage can then be calculated as the areas under the curves of Isochronous Load-Recoverable Strain and Isochronous Load-Locked-in Strain curves. Similarly, in order to calculate the Recoverable ISE [RIt3 and the Locked-in ISE [LI13after the removal of the shock loading, it is necessary to construct the Isochronous Load-Total Strain, Load-Recoverable Strain and Load-Locked-in Strain curves for each level of additional shock loading [APJ. The Recoverable ISE [RIt3and the Locked-in ISE [LIt3in this stage can then be calculated as the areas under the curves of Isochronous Load-Recoverable Strain and Load-Locked-in Strain curves. Using the above procedure the [RI,-EL], plots for the various stages of loading can be plotted. The [RI,-[L], relationship for Multi-Stage Loading test data at a limiting strain of 10% is shown in Fig. 8. This figure shows that only at the additional shock load [APJ of 50 kN/m did the geogrid reached the 10% limiting strain. For additional shock loads less than this, the Locked-in ISE at the end of 200 hours almost reverted back to the Absorbed ISE for the sustained load alone at 100 hours. However, it is important to appreciate that in the longer term, the Absorbed ISE due to the sustained load alone will continue to increase and will eventually exceed this value. Further it is important to note that the uniaxial geogrid reached the 10% limiting strain in 7 seconds only at 50 kN/m of additional shock loading [AP,]. For other levels of shock loading, 10% strain was not reached within 20 seconds. Therefore, it may be thought that this geogrid would be able to withstand at least 40 kN/m of additional shock loading [AP,] at any time during its operationa1 lifetime, however this may not always be true.
Figure 8. [RI, - [L], relationship for the various stage of loading
It is suggested that the amount of additional shock load [AP,] that can be taken by any geosynthetic depends upon when this load is applied. Under the action of the sustained load the geosynthetic will continue to strain (creep) with time, hence the difference between the developed strain and the limiting or rupture strain will diminish with time. Thus the additional shock load [AP,] that the geogrid will be able to take will depend on the “Available Strain”, i.e. the difference between the strain before the shock (earthquake) loading and the limiting or rupture strain. For example, if for a Sustained Load [P,] the strain in a geosynthetic is equal to 8% and the limiting strain is 10%, then the Available Strain is 2%. Thus the geogrid will be able to take only the amount of shock loading to develop 2% strain. 7 MODIFIED MATERIAL PROPERTIES APPROACH
The above indicates that the current practice of using a single value of Design Strength over the entire design lifetime could be unsafe, particularly towards the end of design life of a GRSS. Further, it implies that GRSSs which survived the recent earthquakes will not necessarily survive similar earthquakes in the future. Thus more understanding of the response of the ISE components under sustained plus shock loading is required. In addition, the effects of being confined in soil need to be further assessed. However, for design purposes it is suggested that in order to allow for sustained loading plus shock loading, a “Modified Material Properties Approach” should be adopted. Within this approach, a new Par-
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rupture. This allows an Available Strain for additional shock loading to be developed. Further research is required to provide a better understanding of the behaviour of geosynthetics under sustained loading plus shock (earthquake) loading when confined in-soil. REFERENCES AASHTO 1994. Standard specifications for highway bridges. Washington DC, 50“’ ed. AASHTO 1997. Standard specifications for highway bridges. Division 11, section 7, earth retaining systems, USA. BS8006 1995. Code of Practice for strengthened/ reinforced soils and other fill. British Standard Institution. Figure 9. Modified Material Property Approach. Collin, J.G., Chouery-Curtis, V.E. & Berg, R.R. 1992. Field observation of reinforced soil structures under seismic loadtial Factor, called the “Short Term (or Shock) Load ing. Proc. Int. symposium on earth reinforcement practice, ing Factor”, should be applied to the Limiting orv.1: 223-228. Rupture Strain [EL] in order to obtain a Factored DIBt 1998. Deutches Institut fur the Bautechnik Design MeLimiting Strain [ ~ d ~ , ,Fig ~ ~9.] ,The sustained load at thod. Netlon Ltd, Blackburn, UK. the design lifetime corresponding to this Factored Frankenberger, P.C., Blomfield, R.A. & Anderson, P.L 1996. Reinforced earth walls withstand Northridge earthquake. Proc. Limiting Strain would then be the “Modified Design Int. symposium on earth reinfocement, FukuokdKusushu, JaStrength” [PdesJ. The purpose of applying the Short pan, 12-14 Nov.: 345-350. Term or (Shock) Loading Factor to the Limiting Fujii, T., Fukuda, N. & Najiri, N. 1996. Dynamic response Strain [EL]is to allow sufficient Available Strain to analysis of geogrid reinforced steep embankments. Proc. accommodate the shock load (or possibly several Int. symposium on earth reinforcement, FukuokdKyushu, shock loads) at any time during the design lifetime Japan, 12-14 Nov.: 197-202. Fukuda, N., Yamanouchi, T., Sakai, N. & Shinatani, H. 1994. Apof the GRSS. plicability of seismic design methods to geogrid reinforced Thus in this approach, the GRSSs is first designed embankment. Proc. 5* Int. conference on geotextiles, gefor Single-Stage Actions or Equivalent Sustained omembranes and related products, Singapore: 533-536. Actions using the Modified Design Strength [Pdes,eq] Greenwood 1997. Residual strength: an alternative to stress rupand then the reinforcement layout is checked for an ture for earth reinforcement design. Proc. Int Sym. on Earth Reinforcement, Fukuoka, Kyushu, Japan., v0.2: 1081-1083. Additional Short Term (or Shock) Load [AP,], (or Jewell, R.A., Paine, N. & Woods, R.I. 1984. Design method for several of these as considered appropriate). steep reinforced slopes and embankments. Symposium of polymer grid reinforcement, London, UK. Jones, C.J.F.P. 1996. Earth reinforcement and soil structures. 8 DISCUSSION Thomas Telford Publishing, London, UK. Kupec, J. 2000. Combined Sustained and Short Term Load Testing of Geosynthetics. M.Sc. Thesis, University of It has been shown that the present methods of deStrathclyde, Glasgow, UK. signing for sustained loading plus shock loading are McGown, A. 2000. 4Ih Mercer Lecture: The behaviour of geoempirical. It is suggested that in order to design for synthetic reinforced soil systems in various geotechnical sustained loading plus shock loading, a Modified applications. EuroGeo 2000, Bologna, Italy: 3-26. Material Properties Approach should be used. NCMA 1997. Segmental retaining wall seismic design procedure. Final draft. Within this approach, a Short Term (or Shock)
Loading Factor is applied to the limiting strain or
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Testing related to the introduction of a new geogrid with welded flat bars for use as a soil reinforcement G. Heerten & E. Reuter Naue Fasertechnik GmbH & COKG, Germany
A. McGown & J. Kupec Department of Civil Engineering, University of Strathclyde, UK
ABSTRACT: When introducing a new geosynthetic soil reinforcing product, a range of test data is required for specifying authorities and design codes/methods and there are no general correlations between them. As a result, manufacturers have to undertake a wide range of testing and specifiers/designers then find it difficult to identify which of the data available is critical. In this paper, the recently developed Isochronous Strain Energy Approach for the load-strain-time-temperature behaviour of geosynthetics is presented and it is shown how this can be used to characterise geosynthetic products generally and a new geogrid with welded flat bars in particular this is not the case. Additionally, there are few general correlations between the data obtained from different test methods. This has had the effect of putting a great deal more emphasis on the provision of Index test data than on the provision of the more complex and more expensive Performance test data. Recently a new approach to the characterisation of the load-strain-time-temperature behaviour has been developed, McGown (2000), which allows correlation of data from a wide variety of test methodologies. This is called the “Isochronous Strain Energy” [ISE] approach. In this paper the ISE approach and the test methodologies to be used in relation to the introduction of geosynthetic reinforcing products in general are described and then applied to a new range of geogrid products made from welded flat bars.
1 INTRODUCTION A wide range of predominantly synthetic polymers mixed with a variety of additives are formed into geosynthetics using a number of manufacturing processes. The combination of different polymer mixes and manufacturing processes results in products possessing diverse micro and macrostructures. As a result, these products exhibit a very wide range of physical and engineering properties. Specifiers and Designers of geotechnical structures incorporating these products require to obtain test data from the products for the purposes of quality control and design. So-called “Index” tests are generally appropriate for specification (quality control) purposes but for design more complicated, (sometimes confined in-soil), methods are required. These are termed “Performance” tests, Murray and McGown (1982, 1987 and 1992). Various national and international design codes/ methods and specification authorities have recognised the need for both Index and Performance tests but to date there is a lack of standardisation in these test methodologies and to some extent in the interpretation/presentation of the test data. Thus when introducing a new product range, manufacturers require to obtain a wide range of test data using different test methodologies and data presentation techniques in order to satisfy the various national and international specifications and design codes/ methods. The test methodology and terminology used to identify the basic characteristics are generally agreed internationally, but particularly for design input data,
2 CHARACTERISING THE LOAD-STRAINTIME-TEMPERATURE BEHAVIOUR OF GEOSYNTHETICS 2.1 Loading regimes Whether for specification or design purposes it is always necessary to identify the load-strain behaviour of geosynthetic soil reinforcing products at specified times and temperatures. This usually involves testing the products under Constant Rate of Strain and Sustained Loading (Creep) test conditions at various specified temperature conditions. Tests on ex-works materials provide data which can be used to identify what is termed the “Reference Strength”. Tests must then be carried out on materials that have
95
ferent strengths due to different test conditions. Hence, it can be stated that to avoid long-term creep rupture, a single value of the Reduction Factor will not be applicable to data obtained from different test methodologies. Indeed, it is apparent that Reduction Factors are likely to be specification and design code / method specific. Additionally, Troost and Ploeg ( 1990), BS8006 (1995) and Jewel1 (1996) defined the Reference Strength as the load to cause creep rupture of exworks specimens at the end of their design life [tdl]. Typically, many geosynthetics exhibit a wide range of creep rupture strains for different sustained (creep) load levels. Hence, the Reference Strength, defined on the basis of load at creep rupture for a specific design life time, can be very difficult to select if it has to be related to a specific strain level developed at creep rupture, McGown et a1 (1998). Some other design methods define the Reference Strength as the load obtained from the Isochronous Load-Strain curves, corresponding to a Performance Limit Strain, for example the HA 68/94 Design Method (1997) and the TBW Method (1998). The strength at the “Performance Limit Strain” is always less than the strength at creep rupture, hence it is a conservative choice. Overall it may be suggested that currently the socalled Reference Strength of a geosynthetic should be viewed as a range of values which depend on the application and the test methodology specified in the code/method.
been damaged or environmentally exposed in order to establish what level of degradation of the loadstrain properties of the geosynthetic will occur with time under operational conditions. This then allows Factors of Safety for Limit Equilibrium designs and Partial Factors for Limit State designs to be assessed. “Design Strengths” are then obtained by applying these Factors of Safety or Partial Factors to the Reference Strengths. To date the complex loading regimes applied to Geosynthetic Reinforced Soil Structures [GRSSs] are represented as quasi-static long-term loading. In fact the Actions to which GRSSs are subject may be permanent or variable; accidental free or fixed; static or dynamic, Eurocode 7 (1995). In view of the wide range of Actions that can affect GRSSs, their performance can be very difficult to assess. Thus for simplicity in specifications and designs it is suggested that the various types of Actions should be split into only three general categories, viz. “Sustained Actions”, “Equivalent Sustained Actions” and “Multi-Stage Actions”. The first two may be represented as long-term sustained loads or deformations, i.e. “Single Stage Actions”, however the third must be treated as a series of loads or deformations acting for different periods of time, i.e. Multi-Stage Actions. Test data related to the performance of geosynthetics subject to Single-Stage Actions can be obtained from “Single-Stage Loading” tests. For situations where Multi-Stage Actions are applied, then “Multi-Stage Loading” tests are required, however, these are still at the development stage. Hence in this paper, only Single-Stage Loading test data will be referred to in respect of the determination of Reference Strengths, Factors of Safety/Partial Factors and so Design Strengths.
2.3 Factors of safety and partial factors To date, most specification and design codedmethods, suggest the use of Factors of Safety or Partial Factors obtained from CRS tests. For example, Watts et a1 (1990) and Koerner et a1 (1990) used CRS tests to identify damage effects on geosynthetics. They determined the Damage Factors for various geosynthetics by comparing the CRS loads for rupture “before” and “after” damage effects BS8006 (1995), AASHTO (1997), HA 68/94 (1997), TBW (1998) and DIBt (1998) adopt the same method for the determination of Partial Factors for geosynthetics. The above procedure implicitly assumes that the effects of construction damage or environmental degradation on the properties of geosynthetics are the same in both the short-term and the long-term. Furthermore, it is likely that Factors of Safety and Partial Factors will be strain level and time dependent. Therefore specifying a single value of a Factor of Safety or Partial Factor is likely to be inappropriate.
2.2 Reference strength For Single-Stage Loading the Reference Strength of geosynthetics is defined in existing specifications and design codedmethods in different ways. Some define it as the factored strength determined from short-term Constant Rate of Strain [CRS] testing, whilst others define it as the long-term, sustained load (creep) strength at rupture or at a limiting strain. For example, DIBt (1998) and AASHTO (1997) use the maximum load at rupture of the exworks materials under Constant Rate of Strain [CRS] testing as the basis of defining the Reference Strength of geosynthetics. The Reference Strength to avoid long-term creep rupture, is then obtained by dividing the CRS rupture strength by a Reduction Factor. The DIBt (1998) and the AASHTO (1997) design methods specify 33% per minute and 10% per minute strain rates respectively, for the CRS tests employed. Thus as suggested by Kabir (1984) and Yeo (1985), these tests are likely to provide dif-
2.4 Design strength In view of the differences identified in the definitions of Reference Strength and in the means of determining Factors of Safety or Partial Factors, it is 96
apparent that Design Strengths will be different from specification to specification, code to code and method to method, for any particular geosynthetic. Further, it is likely that the geosynthetic strains corresponding to the Design Strengths for the Ultimate and Serviceability Limit State in Limit State Analyses will differ.
i) Immediately Recoverable ISE [RI, and ii) Locked-in ISE [L],, part of which is recoverable with time. It should be noted that these components vary with time for any limiting strain or rupture condition, as shown in Fig. 3.
4 DETERMINING THE ISE CAPACITIES AND COMPONENTS FROM TEST DATA
3 CHARACTERISING THE LOAD-STRAIN-
TEMPERATURE BEHAVIOUR OF GEOSYNTHETICS SUBJECT TO SINGLE STAGE ACTIONS USING THE ISE APPROACH
4.1 Determining ISE capacities CRS testing needs to be carried out at different rates of strain at a specified temperature, to allow the
3.1 Definition of Isochronous Strain Energy For a Single-Stage Loading test, e.g. CRS or sustained loading (creep) test, under isothermal conditions, the external work done per unit width of a geosynthetic may be taken to be equal to the Absorbed Strain Energy at any time (t). From Isochronous Load-Strain plots developed from these test data, the areas under the curves can be calculated to give the Isochronous Strain Energy [ISE] for the geosynthetics at different times, Fig. 1, McGown (2000). The unit of ISE for geosynthetics is: Force per unit width times unit strain = (kN/m) x (m/m) = kN/m To avoid confusion with existing definitions of strain energy it is suggested that another unit for ISE is used as follows:
ISE = (kN/m) x (m/m>= (kNm/m2>= H/m’ 3.2 The components of ISE Figure 2 represents the load-strain-time behaviour of a geosynthetic in terms of the “Absorbed ISE” [A]t. Upon application of load [PI1 at time [to], as shown in Fig. 2(a), there will be an Absorbed ISE [Alto (point 1) within the geosynthetic. Thereafter, over a period of time between [to] and [tl] more Strain Energy will be absorbed by the geosynthetic, i.e. the Absorbed ISE will increase to [Altl, (point 2), Fig. 2(b). At time [tl] when the load is removed, a part of Absorbed ISE will be recovered and this is termed the “Immediately Recoverable ISE’ [Rltl, (point 2 to 3), Fig. 2(c). At this point of time, i.e. time [tl], the Absorbed Strain Energy remaining in the geosynthetic is termed the “Locked-in ISE’ [Lltl, Fig. 2(c). If no further load is applied to the geosynthetic, then with time part of this Locked-in ISE will be recovered due to viscous rebound, however, part will b e irrecoverable, the so-called “Irrecoverable LOCked-in ISE’ EL],. Clearly, this indicates that at any time the Absorbed ISE comprises two components, which are:
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Figure 2. Load strain-time behaviour in terms of ISE.
Figure 3. Variation of ISE components with time.
Isochronous Load-Strain curves to be drawn as shown in Fig. 4(a). Sustained load (creep) tests carried out at different load levels may similarly be used to plot Isochronous Load-Strain curves, Fig. 4(b). The areas under these Isochronous Load-Strain curves may then be plotted together in an ISE1og.Time plot, as shown in Fig. 4(c). If these data are re-plotted to a log-log scale and the best-fit curves drawn through the data, it will be seen that the data from the CRS and creep tests will lie close to the same best-fit curves. The best-fit curves at each strain level represent the ISE Capacity [C]t for this material. Strictly speaking, for a particular strain at a specific time, the load required to achieve this strain in a CRS test will be different from the load required in a sustained load (creep) test. Therefore theoretically, the ISE Capacity [C], at a particular limiting strain [ E ] L and at a specific time from CRS tests should be always different from data from sustained load (creep) tests. However, for relatively short-term tests, the response of geosynthetics to loading is dominated by their initial elastic and plastic strains and a limited amount of rapidly developed primary creep. The result is that for all practical purposes, the ISE Capacity [C], determined from different shortterm test methods will be very similar and can be represented by the same best-fit curves. Therefore, such curves can be used to compare and correlate the data obtained from CRS and other short-term tests, such as sustained load, sustained strain (stress relaxation) and cyclic loading tests, McGown (2000).
Figure 4. Test data from different load-strain paths and their correlation in the ISE-logtime plot.
4.2 Determining the ISE components Calculation of the ISE components requires the identification of the “Isochronous Load-Locked-in Strains”. To do so the “Isochronous Load-Total Strain” and “Isochronous Load-Immediately Recoverable Strain” curves require to be developed from loadinghnloading tests. The “Total Strain” at any time is equal to the summation of “Immediately Recoverable Strain” and “Locked-in Strain”, therefore the Isochronous Load-Locked-in Strain curves are obtained by the method shown in Fig. 5 . The ISE capacity and its components for any Singlestage Loading regime, i.e. [C],, [RI, and [L],, are then determined from the areas under the respective Isochronous Load-Strain curves for different times and strains, as shown in Fig. 1.
Figure 5. Construction of Isochronous Load-Locked-in Strain curves.
ISE Components at various isothermal conditions, then it is possible to reverse the process. Thus given the availability of such ISE data, the load-straintimetemperature behaviour of geosynthetics may be derived for any specified conditions. Therefore ISE Capacity and ISE Component data provide the basis for Manufacturers, Specifiers and Designers to predict the performance of geosynthetic reinforcements under different test methodologies in a manner not previously possible.
5 DETERMINING TEST DATA FROM THE ISE CAPACITY AND COMPONENTS It is clearly the case that if a product is characterised using CRS tests and sustained load (creep) tests and the data analysed to obtain the ISE Capacities and
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6 CHARACTERISATION OF A NEW GEOGRID
6.3 Derivation of LSE capacity and ISE Components
WITH FLAT WELDED JUNCTIONS
To obtain the ISE capacity, [C],, and ISE Components the short-term and long-term test data were employed in the manner described previously, see Fig. 4. The ISE Capacity [C], - 1og.Time plot was constructed, Fig. 6,and the ISE Components derived as shown in Fig. 7.
6.1 Basic properties of the geogrid To increase the current range of geosynthetic reinforcement, a new range of geogrid reinforcements has been developed in Germany. It is made from prestressed / pre-strained monolithic flat polyester (PET) bars with welded joints. These are intended to provide low elongation at high strength with low creep characteristics. The product range covers uniaxial and biaxial geogrids with tensile strengths up to 600kN/m when tested under CRS conditions at 20°C.
6.4 Use of ISE capacity and ISE Components to derive test data The test data presented in Figs. 6 and 7 were extrapolated by two log cycles. The data were then used to predict the material behaviour for test conditions not included in the test programme, as shown in Fig. 8, and as suggested would be possible in the previous sections. Test data from environmental degradation and construction damage are not yet available so that Factors of Safety and Partial Factors cannot be calculated to obtain the Design Strength at any condition and time but they will be obtained in a similar manner following further testing.
6.2 Choice of representative test specimen The size and shape of representative samples of geogrids are dominated by their macro structure. The gauge length of test specimens is taken as the distance between the centre of the elements and a minimum dimension of lOOmm in any direction of testing is generally considered satisfactory to account for the local variability of most geogrids. However, many test standards also require that at least three complete tensile elements (ribs) and at least one row of nodes or cross-members along the width, (excluding the nodes or cross-members by which the sample is being held in the clamps), should be included. In addition, all ribs must be cut at least lOmm away from any node. The test procedures outlined in I S 0 13431 (1999)state that at least three test specimens for each product should be tested. Recently, Wrigley et a1 (1999)suggested that the number of test specimens required to characterise a product can be reduced using timetemperature superposition, For the new geogrid product the testing recommendations of I S 0 13431 (1999) and BS 6906 (1987) have been employed. A gauge length of 2 1 0 m was selected which included 6 cross members. Specimens were held in specially developed friction clamps as roller clamps proved unsatisfactory. CRS tests were conducted at different rates of deformation and repeated three times. Specimen rupture occurred well within the clamped area and across the specimen. Sustained loading (creep) tests were conducted on 10, 20, 30 and 40% of the 1 O%/min short-time CRS strength. The sustained loads were applied over periods of 1, 10, 100 and 1000 hours and unloaded for a tenth of the loading time or at least 10 hours. The sustained loading (creep) tests were repeated three times to observe the specimen variation. The test data indicate that specimen reproducibility from CRS and sustained loading (creep) tests was 95% at any time.
Figure 6. ISE capacity [Cl,-log. Time and variation of the ISE Components.
Figure 7. ISE Components [RI, - [L], plot.
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BS 6906 1987. Determination of tensile properties using a wide width strip, Part 1, BSI, UK. BS 8006 1995. Code of Practice for strengthenedkeinforced soils and other fill. BSI, UK. DIBt 1998. Theory used in the Deutsches Institut fur Bautechnik Design Method. “Netlon Ltd, Blackburn, UK. Eurocode 7 1995. Geotechnical design. “DD ENV 1997-1 1995. General Rules, (together with the UK application document): 5-7, 12-15,72-86. HA68/94 1997. Design methods for the reinforcement of highway slopes by reinforced soil and soil nailing. Design manual for roads and bridges, v. 4, section. 1, part 4, HMSO. IS0 13431 1999. Determination of the Tensile Creep and Creep Puncture Behaviour. Jewell, R.A. 1996. Soil reinforcement with geotextiles. “Special Publication 123, CIRIA, London, UK’. Kabir, M.H. 1984. In-isolation and in-soil behaviour of geotextiles. Ph.D. thesis, University of Strathclyde, Glasgow, UK. Koerner, R.M. 1994. Designing with Geosynthetics, Englewood Prints, NJ, USA. Koerner, G.R. & Koerner, R.M. 1990. The installation survivability of geotextiles and geogrids. 4th International Conference on Geotextiles, Geomembranes and Related Products, Hague, Netherlands: 597-602. McGown, A., Andrawes, K.Z., Pradhan, S. & Khan A.J. 1998. Limit state design of geosynthetic reinforced soil structures. 6th Int. Conf. On Geosynthetics, Atlanta, USA. McGown, A. 2000.4th Mercer Lecture: The behaviour of geosynthetic reinforced soil systems in various geotechnical applications. EuroGeo 2000, Bologna, Italy: 3-26. Murray, R.T. & McGown, A. 1982. The selection of testing procedures for the specification of geotextiles. Second International Conference on Geotextiles, Las Vegas, USA: 29 1-296. Murray, R.T. & McGown, A. 1987. Geotextile test procedures: background and sustained loading testing. “TRRL Application Guide 5, Transport and Road Research Laboratory, Department of Transport, UK’, 12. Murray, R.T. & McGown, A. 1992. Assessment of Index test methods for geotextiles. “TRRL Application Guide 21, Transport and Road Research Laboratory, Department of Transport, UK.” 12. TBW 1998. Theory used in the Tensar Tie-back Wedge Walls Design Method. Troost, G.H. & Ploeg, N.A. 1990. Influence of weaving structure and coating on the degree of mechanical damage of reinforcing mats and woven geogrids caused by different fills during installation. 4th Int. Conf. on geotextiles, geomembranes and related products, Hague, Netherlands, v.3.: 609614. Watts, G R.A. & Rrady, K.C. 1990. Site damage trials on geotextiles. 4th International Conference on Geotextiles, Geomembranes and Related Products, Hague, Netherlands: 603-607. Wrigley, N., Austin, R.A. & Harrison, P.E. 1999. The longterm strength of geogrid reinforcement. Geosynthetics ‘99, Boston, USA. 2.: 711-721. Yeo, K.C. 1985. The behaviour of polymeric grids used for soil reinforcement. Ph.D. thesis, University of Strathclyde, Glasgow, UK.
Figure 8. Prediction of material behaviour.
7 DISCUSSION The need to have a wide range of test data from different test methodologies in order to comply with various national and international specifications and design codes/methods has caused Manufacturers to focus more on Index testing than on Performance testing. The range of data then available and the lack of correlation between the data has caused confusion to many Specifiers and Designers. In this Paper the use of the Isochronous Strain Energy to correlate test data and provide a fundamental characterisation of the load-strain-time-temperature behaviour of geosynthetics has been identified. The application of this approach to a new geogrid with welded flat bars has been demonstrated. The use of ISE Capacity and ISE Component data to provide data for any specified test methodology, within the range of time and temperature of the ISE data was also indicated. Thus it is suggested that the ISE approach is a fundamental and useful means of characterising geosynthetic reinforcements. In order to comply fully with Specification and Design, other data will be required relating to soilgeosynthetic interaction and perhaps to junction strength in respect of geogrids. Recommendations on the approach to be taken to soil-geosynthetic interaction have been provided previously by McGown et a1 (1998). To date no recommendation can be made with respect to junction strength. Tests have been suggested for this, Koerner (1994), but the applicability of these is still very much open to question. REFERENCES AASHTO 1997. Standard specifications for highway bridges. “Division I1 Section 7: Earth Retaining Systems”.
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Earthquake Engineering Frontiers in the New Millennium, Spencer & Hu(eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Prediction of failure stress of reinforced residual soil by Simplified approach S.A. Mofiz Bangladesh Institute of Technology, Rajshahi, Bangladesh
M.R. Taha Universiti Kebangsaan Malaysia, Selangor, Malaysia
ABSTRACT: A testing program has been carried out to study the stress-strain behaviour of unreinforced and reinforced residual soil. Drained triaxial tests were conducted using computer controlled GDS triaxial apparatus. Triaxial shear test results show that non-woven geotextile reinforced soils exhibit higher failure strains and volume contraction than unreinforced soils. Failure strains and the strength increase with increase in number of layers. A simplified approach for numerical calculations were proposed to predict the shear strength and the coefficient of interface friction of reinforced soils for conventional triaxial compression (CTC) stress paths. Charts were also presented to predict the strength of reinforced soil and to determine the coefficient of the interface friction from triaxial tests. Predictions of failure stress using simplified approach are satisfactory compared to experimental observations.
1 INTRODUCTION Reinforced soil has gained popularity due to its wide application in the construction of geotechnical structures such as retaining walls, foundations, embankments, pavements, etc. The use of reinforcements increases bond in the soil system due to the interlocking of the soil particles with the reinforcement aperture as well as enhancing the bearing resistance of the transverse members of the reinforcement. The effectiveness of the reinforcements in contributing an increase in the shear resistance is highly dependent on the orientation of the reinforcements with respect to the failure plane. The mechanical and hydraulic properties of the non-woven and composite geotextile have been studied in detail Ling et al. (1990 and 1992). The non-linear stress-strain relationship, which may be highly dependent on the confining stress, was formulated and implemented for finite element analysis by Ling and Tatsuoka (1992). Ling and Tatsuoka (1994) conducted a study on silty clay reinforced with three types of geosynthetics, two geotextiles, and a geogrid under plane strain conditions. The importance of plane strain loading was highlighted, and the influence of the geosynthetic mechanical and hydraulic properties as well as the consolidation stress ratio was investigated in their study. Cuzzafi et al. (1994) were conducted large scale triaxial tests on geogrid reinforced gravel and their test results showed that the apparent cohesion were induced due to the reinforcement in soil.
Taha et al. (1999) demonstrated the behaviour of georeinforced residual soil using drained triaxial samples that the reinforced systems increased strength-deformation properties in a significant manner. The results of the failure mechanism indicated strain hardening behaviour and multiple bulging with restraint at the reinforcement layers. Ashmawy et al. (1999) reported that reinforced soils exhibit an improvement in strength and deformation characteristics under monotonic loading conditions, due to the additional “pseudo” confinement caused by the lateral restraint and shear mobilization along the soilinclusion interface. The research works aimed to determine the stress-strain mechanism between the non-woven geotextile and residual soil by triaxial shear test. However, with respect to residual soil, its interaction mechanism and the failure behaviour in the reinforced composites are not well understood due to limited study. Thus, a thorough investigation of the soil reinforcement interaction was conducted. The simplified prediction procedures to determine the strength of reinforced and unreinforced soils for various stress paths were presented. An attempt was undertaken to determine the coefficient of interface friction from triaxial test results. Prediction charts have been presented for different friction angle of unreinforced soil and number of reinforcement layers. Finally, the equivalent angle of internal friction of reinforced residual soil composites has been derived using the soil friction angle, interface friction and number of reinforcement layers.
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2 PROPERTIES OF SOIL AND REINFORCEMENT
4 TESTING PROCEDURE
Residual soils often found in tropical or semitropical area, are formed from intense weathering of rocks under consistently high temperature and rainfall. In this work, the disturbed soil was collected from the area of hospital campus of Universiti Kebangsaan Malaysia (HUKM) located in Cheras, Selangor approximately 8 kilometers south of Kuala Lumpur, Malaysia. The soil is reddish in colour and classified as CH in Unified Classification System (USCS). The soil particle contains about 45 % clay, 19 % silt, 36 % sand and no gravels. The maximum dry density from the standard Proctor test was 14.42 kN/m3 and the optimum moisture content was about 24.6%. In this research work, a non-woven geotextiles (Polyfelt TS 60) was used as the reinforcement material. This group of geotextiles consists of mechanically bonded continuous filaments made from UVstabilized polypropylene. The stress-strain behaviour of five samples of non-woven geotextiles were tested using a universal testing machine. The tensile strength properties of the reinforcement were determined following ASTM D4595- 1992 (ASTM 1992) and can be expressed as
o = -P, ’
W
where 01 is the tensile strength (N/m), P b is the observed maximum tensile force ( N ) and W is the width of the reinforcement specimen in metre. The maximum tensile strength from the tests was obtained 18.68 kN/m and 19.12 kN/m in the longitudinal and transverse direction, and the corresponding elongations were about 74%, 5 1% respectively.
5 TEST RESULTS The shear stress vs. axial strain and volumetric strain vs. axial strain curves for unreinforced soil are shown in Figure l(a) and Figure l(b). The shear stress-axial strain plot indicate that the axial strain corresponding to maximum shear stress increases with confining pressure. It was observed that there were no distinct peak points in the GE curves and the curves levels off at higher strains until failure. The volume change characteristics exhibits contraction behaviour at lower stress levels and expansion at higher stress levels. The cohesion intercept and angle of internal friction of unreinforced soil under gompression loading are c’ = 27.42 kPa, = 28.02 respectively. For non-woven geotextiles reinforced soil, the stress-strain and volume change characteristics of a single layered soil composites are shown in Figure 2(a) and Figure 2(b) respectively. As expected, the reinforced samples exhibit higher shear strength than unreinforced samples and the maximum shear
3 TESTING PROGRAM The testing program was performed by a series of consolidated drained triaxial stress path (CTC) tests on unreinforced and reinforced soil. This stress path is followed using the conventional lO0mm dia and 200mm high cylindrical triaxial samples. For the CTC stress path test, the incremental stress tensor can be expressed as
Ao,
A
0 0
~ 0 ~ 0 O]= 0 0 0
In this investigation, six consolidated drained triaxial tests were performed on the unreinforced residual soil. For reinforced soil, twelve tests were carried out for compression stress path with a single layer and two layer non-woven geotextile reinforced specimens. Based on the unit weight and the volume of the triaxial mold, the total weight of the soil was divided into two equal portions for single layer and three equal portions for double layer, and compacted inside the mold in layers of equal height. For a single layer specimen, circular disc of non-woven geotextile was placed at the mid height and for a two layer the distances are 1/3 height from the top or bottom of the specimens. A rate of 0.15 mm/min for compression on a triaxial press was adopted, and each layer was compacted following the approach by Cui and Delage (1996) to ensure Proctor maximum density with a double piston system. The tests reported in this paper for both unreinforced and reinforced soil were carried out under consolidation pressure 100-600 kPa. A strain rate of 0.00 15 %/min was used that ensured no pore pressure change as required in a drained test. The computer controlled triaxial (GDS) system was adapted to carry out the CTC stress path tests. A microprocessor collects the data from transducers automatically at prescribed intervals. The data were transmitted by the controlling microprocessor for recording, processing and production of results, which could be displayed on the screen, tabulated or plotted by a plotter.
~
where Ao, is the incremental stress tensor and Aolis the incremental deviatoric stress.
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Figure I . Stress-strain characteristics of unreinforced soil: (a) shear stress vs. axial strain (b) volumetric strain vs. axial strain.
Figure 2. Stress-strain curves of a single layered reinforced soil: (a) shear stress vs. axial strain (b) volumetric strain vs. axial strain.
strength were attained at higher axial strains. This increase of shear strength is caused by an increase of the confining pressure in the soil between the reinforcement layers which depends on the interface friction resistance along the reinforcement. The shear stress and volume change behaviour of unreinforced and reinforced samples at the initial shearing were similar since the effect of the reinforcement will only begin to function at some finite axial strain. The result shows different volumetric pattern at higher strain when the soil samples started to dilate. From the test results it is also observed that the volumetric strain vs. axial strain behaviour re-
vealed that expansion is more pronounced especially at lower confining pressure. The results of unreinforced and reinforced soil samples also show that dilatancy is dependent on the confining pressure. The shear strength parameters for a single layered and two layered non-woven geotextile reinforced soil are determined from the MIT stress path method. The cohesion intercept and angle of internal friction for a single layered soil composite are c’=36.99 kPa, 9’=30.76 an$ for a two layered soil are c’=43.85 kPa, @’=32.4 . It is also observed that the reinforced soils exhibit higher failure strain and shows about 20% to 43% higher than that of unreinforced soils.
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The failure strains also increases with the increase of reinforcement layers. However, the increase in failure strains over unreinforced soil is not proportional to the number of layer.
R, = (xdh) A o 3
(51
where h is the height of the sample and d is the diameter of the sample. Combining Equation 4 and Equation 5 the increase in confining pressure can be written as:
6 PREDICTION OF SHEAR STRENGTH BY SIMPLIFIED APPROACH The prediction of the shear strength of reinforced soils depends on different factors such as the number of reinforcement layer, tensile strength of the reinforcement, confining pressure, and interface friction. Yang (1972) reported for the first time the concept of increase in confining pressure due to insertion of reinforcement. Broms (1977) and Chandrasekaran et al. (1989)proposed various expressions for the shear strength of reinforced soil considering Yang’s concept. In this research work, the shear strength expressions for triaxial compression and extension stress paths are derived based on the Yang’s concept. The derivation and calculation procedures are outlined in the following section. According to the Yang’s concept, shear stresses developed in the reinforcement layers is assumed to be transferred to the soil through interface friction and this may cause an increase in confining pressure. In this section, an analysis is presented for a single layer of reinforcement placed at the centre of a triaxial sample. During shearing the deviator stress increases and the reinforcement is subjected to tensile stresses which cause the development of interfacial frictional forces between the soil and the reinforcement. The generation of interfacial friction between the soil and the reinforcement are not constant along the reinforcement. Chandrasekaran et al. (1989) reported that the tensile forces in the reinforcement increases from the periphery of the sample to the centre of the reinforcement. The tensile forces in the reinforcement can be expressed as a function of confining pressure and spacing of the reinforcement. The coefficient of interface friction between the soil and the reinforcement can be written as
From the shear strength theory for unreinforced soil the following equation can be written O,
?r
(4)
where 01 is the axial stress at failure, d is the reinforcement diameter, KO, is the mobilisation factor which is assumed to be equal to (Ku +K0)/2. The total frictional force, RF, is assumed to provide additional confining stress effect. Due to the increase in confining stress, the total friction force can expressed as:
(7)
where N# = tan2 (45 + $72) For reinforced soil, the axial stress, 01, can be expressed as
Substituting the value of do- from Equation 6 into Equation 8, 01 can written as: (9)
For single layer reinforced soil, the axial stress, can be predicted from Equation 9 if the values of .f: d, h, oj and fare known. A general expression for the axial stress in terms of N layers of reinforcement can be written as 01
f = tan6 (31 where 6 i s the angle of interface friction. The total frictional forces, RF, at failure along the triaxial cylindrical sample can be expressed as
R , = -( d CF~2 f K n v ) 4
= 0, N, + 2c’J N p
Equation 10 indicates that the shear strength of reinforced soil is a function of oj,q5:.iN,(s/h) and (d/h). The coefficient of interface friction has been determined using the failure stress at consolidation pressures 0,=200 kPa and 0,=400 kPa for CTC path only. Back predictions of the failure axial stress were then calculated using the average coefficient of friction for the various stress paths. The axial stress of reinforced soil with various types of reinforcement inclusions for CTC stress path is calculated using Equation 10. Comparisons were made with the experimental results. The result shows that the predicted response is a good agreement with the experimental values. 7 PREDICTION OF THE COEFFICIENT OF INTERFACE FRICTION The coefficient of interface friction of the reinforced soil can determined either from the pull out test or
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the modified direct shear test. In the field, the reinforced soil is normally subjected to a confining stress. Thus, conducting triaxial tests at the required confining pressure can simulate such conditions. Therefore, an attempt is made to determine the coefficient of interface friction from the triaxial test results. The coefficient of interface friction, f , can be determined by using any stress path test from the values of axial stress at failure, confining stress, angle of internal friction of unreinforced soil, number of reinforcement layers, (dh) and (&), respectively. It is very convenient to determine the coefficient of interface friction, .f, by conducting the conventional triaxial compression (CTC) stress path tests. In this stress path the value of .f can be determined by rearranging Equation 10.
For the evaluation of the equivalent friction angle, first a relation between the axial stress and confining stress at failure is written down as
(14)
0, = ~3N@O,,l + 2c'JN@,,,,
@Lo,,
where NFolll= tan2(45+ 12) in which @',,,,, is the combined effective friction angle of reinforced soil. Expanding Equation 14 and rearranging, the combined angle of internal friction of reinforced soil can be determined as:
For single layer horizontal reinforcement placing at the centre of the specimen, the coefficient of interface friction can be determined as
Figure 3. Prediction chart for the coefficient of interface friction for a single layer reinforced soil.
The above equation can be rewritten when the reinforcement is placed at the mid height of the specimen (&h = 0.5)
The coefficient of interface friction, j can be determined from the above equation if the values of 01 and 4 3 for reinforced soil and angle of internal friction, for the unreinforced soil are known. Prediction charts for finding,f, values are presented in Figure 3 to Figure 6. From the charts the interface friction can be determined if the values of failure ratio (dl/d~), number of reinforcement layers (N) and angle of internal friction (@') for the soil composites are known. In this analysis, equal spacing of reinforcement layers was assumed. The combined friction angle or equivalent angle of internal friction is one of the important factors which will play a vital role for the improvement of the shear strength of the reinforced soil composites.
Figure 4. Prediction chart for the coefficient of interface friction for a double layer reinforced soil.
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internal friction of soil, numbers of reinforcement layer and interface friction coefficient are known. 3. Prediction charts for obtaining the strength of reinforced soil and the coefficient of interface friction are given. 4. A procedure for estimating the combined or equivalent friction angle from triaxial tests data is also presented. REFERENCES
Figure 5. Prediction chart for the coefficient of interface friction for a double layer reinforced soil.
Figure 6.Prediction chart for the coefficient of interface friction for a double layer reinforced soil.
8 CONCLUSIONS 1. Simplified approaches have been proposed for CTC stress paths to simulate the failure stress of the reinforced soil composites. 2. Predictions of the shear strength of reinforced soils can be determined if the angle of
Ashmawy, A.K., Bourdeau, P.L., Drnevich, V.P. & Dysli, M. 1999. Cyclic response of geotextile- reinforced soil. Soils and Foundations, 39(1):43-52, ASTM 1992.Testing of geotextiles and related products. Annual book of ASTM standards, 04.08.D4595. Broms, B.B. 1977. Triaxial tests with fabric-reinforced soil. Proc. Int. Con$ on the use of Fabrics in Geotechizics, Paris, France, 3: 129-133. Chandrasekaran, B., Broms, B.B. & Wong, K.S. 1989.Strength of fabric reinforced sand under axisyminetric loading. Geotextiles arid Geoniembranes, 8:293-310. Cuzzafi, D.,Picarelli, L., Ricciuti, A. & Rimoldi, P. 1994.Behaviour of geogrid reinforced gravel in large scale triaxial tests. Proceedings of the 13th International Coitfereizce on Soil Mechanics and Foundation Engineering, ICSMFE, New Delhi, India, 1: 271-274. Cui, Y.J. & Delage, P. 1996.Yielding and plastic behaviour of an unsaturated compacted silt. Geotechnique,46(2): 29 1-31 1. Ling, H.I. & Tatsuoka, F. 1992. Nonlinear analysis of reinforced soil structures by modified CANDE (M-CANDE). Geosyrzthetic-Reiizforced Soil Retaining Walls, Jonathan Wu, (ed.), 279-296.Rotterdam: Balkema. Ling, H.I. & Tatsuoka, F. 1994. Performance of anisotropic geosynthetic-reinforced cohesive soil mass. Journal of Geotechnical Engineering, ASCE, 120(7):1167-1184. Ling, H.I., Tatsuoka, F. & Wu, J.T.H. 1990.Measuring inplane hydraulic conductivity of geotextiles. Geosyizthetic Testing ,for Waste Containment Applications, ASTM STP 1081, Koerner. (ed.), 257-272.Philadelphia. Ling, H.I., Wu, J.T.H. & Tatsuoka, F. 1992. Short-term strength and deformation characteristics of geotextiles under typical operational conditions. Geotextiles arid Geomernbranes, 1 l(2): 185-219. Taha, M.R, Mofiz, S.A. & Hossain, M.K. 1999.Behaviour of georeinforced residual soil in triaxial test. Proc. World Engineering Congress 99-Towards Engineering Vision: Global Challenges and Issues, 19th -22nd July, 1999, Kuala Lumpur, 175-180. Yang, H. 1972. Strength deformation characteristics of reinforced sand. PhD Dissertation, University of California, Los Angeles, California.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0200 1 Swets €4 Zeitlinger, ISBN 90 2651 863 3
Effect of connection bar stiffness on failure strength of connected steel grid reinforcements Y. Nabeshima, S.G. Zhou & T. Matsui Osaka University, Osaka, Japan
N. Sakata Geosystern Co. Ltd., Osaka, Japan
ABSTRACT: As the steel grid reinforced earth walls become higher and larger, the required reinforcement length becomes longer, and subsequeutly it is indispensable to connect reinforcements from the economical viewpoint. An effective mechanical connection between steel grid reinforcements has already been proposed by the authors. In this paper, the failure strength and failture mode of mechanical connections between steel grid reinforcements are examined, changing diameters of connection bars. The effect of connection bar stiffness on the failure strength of mechanical connections are discussed based on tension test data. As the results, the connection bar stiffness plays an important role on the failure strength and failure mode of the mechanical connections. The failure strength increases as increasing the diameter of connection bar, however, a suitable diameter of connection bar exists for the mechanical connection from the viewpoint of failure mode. 1 INTRODUCTION The reinforced earth structures have been extensively constructed instead of the gravity retaining walls during the past two decades. Figure 1 shows the transitions of the number of construction sites and the average construction area of steel grid reinforced earth walls constructed by Geosystem Co. Ltd., of which heights are more than 10 m. This figure shows that higher and larger steel grid reinforced earth structures increase and become popular and that their average area is larger than 400 in2. As the result, the required reinforcement length becomes longer. Longer reinforcements are not economical to transport to construction sites and not easy to manufacture. Therefore, it is indispensable to connect
Figure 1. Transitions of number of construction sites and average construction area of the steel grid reinforced earth wall (h2lOm; provided from Geosystem Co. Ltd.).
regular length reinforcements. The mechanical connections between steel grid reinforcements have been developed and improved so far. In this paper, the authors examine the failure strength and failure mode of mechanical connections between steel grid reinforcements, changing diameters of connection bar into four kinds. The tension tests in the soil are carried out to estimate the failure strength of the mechanical connections and investigate their failure mode. The authors discuss the effect of connection bar stiffness on the failure strength and failure mode of mechanical connection. 2 STEEL GRID REINFORCEMENT AND MECHANICAL CONNECTIONS The applied steel grid reinforcement is a grid-type reinforcement, in which longitudinal and transverse members are welded each other as shown in Figure 2(a). The diameters of transverse and longitudinal members are 5.0 and 6.0 mm respectively. Many types of connections have been proposed and the mechanical connection used in this paper was the best among them from the viewpoint of failure strength (Nabeshima et al. 2000). Figure 2(b) shows the schematic diagram of the mechanical connection between steel grid reinforcements. The proposed mechanical connection consists of standard and connection reinforcements and a connection bar. The connection reinforcement is characterized by crank shaped edges. The standard reinforcement is overlapped on the connection one, and the connection bar is used to connect both the standard and connec-
107
Figure 3. Schematic diagram of pullout apparatus and arrangement of mechanical connection in the pullout box. Figure 2. Schematic diagram of steel grid reinforcement and mechanical connection.
tion reinforcements. Therefore, the connection bar seems to play an important role in the failure strength of the mechanical connection. The diameter of connection bar can be varied from 5.0 to 9.0 mm. 3 TEST PROCEDURES
The tension tests of steel grid reinforcement with mechanical connections are performed in the dry sand ground under a surcharge of 98.1 Wa. Figure 3 shows the schematic diagram of the pullout test apparatus. The apparatus and used material are the same as those in the references (Matsui et al. 1997, Nabeshima et al. 2000). The rear end of the connection reinforcement is fixed on the loading flame, to carry out the tension test in the pullout test apparatus. All mechanical connections are pulled at a constant rate of about 1.0 mm/ min. The diameter of connection bar is varied in four kinds of 5.0,6.0,7.5 and 9.0 mm, to examine the effect of connection bar stiffness on the failure strength of mechanical connection. Figure 3 also shows the arrangement of the mechanical connection in the pullout box before tension tests.
4 CONNECTION BAR STIFFNESS AND FAILURE STRENGTH OF MECHANICAL CONNECTIONS Figures 4 to 11 show the variations of tensile force of mechanical connections with displacement and
their failure mode after tension tests for four kinds of diameter of connection bar, respectively. In case of D=S.Omm, the curve shows very brittle behavior and the connection bar visibly slightly deformed after tension test. This means that the stiffness of the connection bar is not enough. On the other hand, in case of D=9.0mm7 the curve shows ductile behavior which is quite different from the other cases. Also, although the deformation of connection bar i s hardly observed, the tension failure of longitudinal member is observed after tension test. This means that the stiffness of connection bar is too big. And in cases of D=6.0 and 7.5mm, their curves show intermediate behaviors between those of 5.0 and 9.0 mm. From the above observations, the connection bar stiffness plays an important role on the failure strength and failure mode of the mechanical connections. The failure mode changes from brittle to ductile as increasing the diameter of connection bar, and the displacement at the maximum tensile force becomes larger. Figure 12 shows the variation of failure strength with the diameter of connection bar. The failure strength of mechanical connection increases as increasing the connection bar diameter. However, the increment of the failure strength gradually decreases in cases where the diameter of connection bar is greater than that of transverse member. The displacement at the maximum tensile force is about 55 mm in case of D=9.0mm, which is the largest in all tension tests. From the viewpoint of failure mode, a suitable diameter of connection bar exists for the mechanical connection, which seems to be 7.5 111111, that is almost 1.5 times the diameter of transverse member. 108
Figure 4. Variation of tensile force during tension test (diameter of connection bar : 5.0 mm).
Figure 5. Failure mode of mechanical connection (diameter of connection bar : 5.0 mm).
Figure 6. Variation of tensile force during tension test (diameter of connection bar : 6.0 mm).
Figure 7. Failure mode of mechanical connection (diameter of connection bar : 6.0 mm).
Figure 8. Variation of tensile force during tension test (diameter of connection bar : 7.5 mm).
Figure 9. Failure mode of mechanical connection (diameter of connection bar : 7.5 mm).
Figure 10. Variation of tensile force during tension test (diameter of connection bar : 9.0 mm).
Figure 1 1. Failure mode of mechanical connection (diameter of connection bar : 9.0 mm).
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2) The failure strength increases as increasing the connection bar diameter. The increment of the failure strength gradually decreases in cases where the diameter of connection bar is greater than that of transverse member. 3) A suitable diameter of connection bar exists for the mechanical connection from the viewpoint of failure mode, which is almost 1.5 times the diameter of transverse member.
6 ACKNOWLEDGEMENTS This paper was summarized when the first author stayed at the Australian Defence Force Academy with a financial support from the Kajima Foundation. The special thanks are due to the Kajima Foundation for giving this support.
Figure 12. Variation of failure strength with diameter of connection bar.
5 CONCLUSIONS The authors examined the failure strength and failure mode of mechanical connections between steel grid reinforcements changing diameters of the connection bar. Main conclusions in this paper are summarized as follows:
REFERENCES Matsui T., Y. Nabeshima, K. Uchihata and J.G. Han : Tensile strength of jointed reinforcements in the steel grid reinforced earth, Proc. of the International CorZference on Ground Improvement Techniques, 355-362, 1997. Nabeshima Y., T. Matsui and S. G. Zhou : Tensile strength of joint systems between reinforcements in steel grid reinforced earth walls, Proc. of 3rd International Conference on Ground Improvement Techniques, 269-276, 2000.
1) The connection bar stiffness plays an important role on the failure strength and failure mode of mechanical connections. The latter changes from ductile to brittle as increasing the diameter of connection bar.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Improvement effect of composite geomaterial by utilization of plastic wastes K. Ornine, H. Ochiai & N.Yasufuku Department of Civil Engineering, Kyushu University, Japan
M. Yamamoto Yamaguchi Prefecture, Japan
Y. Inoue Fukuoka-kitakyushu Urban Expressway Public Corporation, Japan ABSTRACT: Recycle of plastic wastes has not proceeded smoothly, so that it is expected to build up the recycling system of the wastes from geotechnical point of view. As one of the methods in reuse of the wastes, it is considered to mix pieces of plastic waste with soil. This method enables to recycle plastic wastes stably for a long term. In order to clarify improvement effect on mechanical properties of composite geomaterial with plastic pieces, unconfined compression test and tensile splitting test were performed for two kinds of soils, cement-treated soil and air-formed lightweight soil, and the effectiveness is confirmed for various plastic wastes. Furthermore, the application to surface stabilized ground is discussed based on the results of model loading test and deformation analysis. 1 INTRODUCTION The amount of wastes has increased year by year and the disposal becomes a serious problem in Japan. Particularly, recycling ratio of the plastic wastes in life and industry is low and many of them have been reclaimed for the reason of unsuitable ones for incineration. It is necessary to utilize the wastes effectively with technical development in each field. From view point of geotechnical engineering, it is considered that geomaterial mixing small pieces of plastic such as PET bottle is one of the methods (Omine et a1 1996). In this paper, the improvement effects of the strength of cement-treated soil and airformed lightweight soil due to mixing pieces of plastic wastes are investigated based on the results of unconfined compression test and tensile splitting test. Furthermore, an application of composite geomaterial with plastic pieces to surface stabilized ground is verified by the model loading test and the result of deformation analysis on bearing capacity of the surface stabilized ground. 2 MECHANICAL PROPERTY OF COMPOSITE GEOMATERIAL WITH PLASTIC PIECES 2.1 Cement-treated soil (a) Soil sample and testing method First, plastic sheet for the card case (the thickness of 0.4mm, the density of 1.38g/cm3)is used in place of the PET bottle. The plastic sheet is cut in the length of 48mm and the width of 3mm. Kaolin clay
(w~=50.6%,1,,=19.6 and p5=2.70g/cm3)is used as a soil sample. After adding Portland cement to the sample with water content of loo%, the plastic pieces are mixed with it. The cement content C is 50, 100, 200, 300kg/m3 and the plastic content M is 0, 2.5, 5.0, 7.5% in volume fraction. The specimen of unconfined compression test is in the width of lOOxl0Omm and the height of 200mm, and that of the tensile splitting test is in the diameter of 150mm and the height of 75mm. After curing those specimens during 7 days in the thermostatic chamber of 2OoC, each test is performed. Furthermore, other plastic wastes, fishing net (nylon thread) and plastic film for agriculture (soft vinyl sheet), are also used for clarifying the differences in kind of plastic material. (b) Deformation-strength property Figure 1 shows the relationship between compressed stress and axial strain on the cement-treated soil with plastic ieces in the case of cement content 100kg/m , which is obtained from the unconfined compression test. Maximum compressive strength of the cement-treated soil with plastic pieces increases with increase in the plastic content. In addition, after reaching to peak strength, softening of the cementtreated soil becomes small due to mixing plastic pieces and the brittle behavior has been improved. Such improvement effect of the strength has been also confirmed under the triaxial compression stress condition. On the other hand, the result of the tensile splitting test is shown in Fig.2. The vertical and horizontal axes represent a tensile stress and a compression ratio, respectively, where the compression-
9
111
Figure 1 , Relationship between compressive stress and axial strain of cement-treated soil with plasitc pieces
Figure 3. Influence of cement content on Improvement effect for unconfined compressive and tensile strengths
Figure 2. Relationship between tensile stress and compression ratio of cement-treated soil with plasitc pieces
Figure 4. Relationship between compressive stress and axial strain of cement-treated soil with various plasitc wastes
ratio is defined as the value of compressive deformation divided by specimen diameter. The cementtreated soil without plastic occurs a brittle failure suddenly after reaching to the peak strength, whereas the tensile strength in the case of mixing plastic pieces increases remarkably and the high residual strength sustains after reaching to the peak strength. As an index of the improvement effect of the strength due to mixing plastic pieces, the strength ratio of the cement-treated soils with plastic pieces and without it is used. The improvement effects for the unconfined compressive strength and the tensile are shown in Fig.3, where strength, qlJqLt0and ov/ot~, subscript “0” means the case without plastic piece. As shown in the figure, the effect for tensile strength is large in comparison with the unconfined compressive strength. In addition, the maximum value is seen at the cement content of approximately 100kg/m’. It is considered that the improvement effect of strength is not decided meaningfully by only cement content, and it depends on curing period and other factors. On the other hand, when the cement content becomes large, the improvement effect decreases. It should be noted that there is a limitation of the application for the large cement content. It is suggested that this method of mixing plastic pieces is effective for subsurface stabilization of soft
ground, because it enables to improve tensile strength considerably. Next, the figure 4 shows the test result of the unconfined compression test on the cement-treated soil mixed with various type of plastic wastes under the conditions of the cement content of 100kg/m3 and plastic content of 4%. When each plastic waste is mixed, the maximum compressive strength has become large in comparison with it in the case of nonmixture. Softening after peak strength is small in the case of the cement-treated soil with plastic pieces, and it changes to ductile material. The improvement effect is shown in Fig.5 as a relationship between the qlJqI,oand the cement content. The clear improvement effect has appeared in the case of the cement content of 100-200kg/m3. This tendency is seen for each type of plastic material. The effectiveness is in order of the nylon thread, plastic piece and vinyl sheet. The reason is considered that the surface of the nylon thread is rough and the friction is large, and the vinyl sheet has smooth surface and small rigidity.
2.2 Air-formed lightweight soil Not only the cement-treated soil but also air-formed lightweight soil has a brittle behavior. In order toconfirm the improvement on strength of the light
112
Figure 6. Relationship between normalized compressive stress and axial strain of light-weight soil with various reinforcement materials
Figure 5. Improvement effect for unconfined compressive strength of cement-treated soil with various plastic wastes
weight soil, the unconfined compression test is performed. (a) Soil sample and testing method Soil sample is prepared by mixing Kaolin clay and silica sand (p,=2.63glcm3) in dry weight ratio of 7:3. After adding Portland cement into the sample with water content of 56%, plastic pieces and air form are mixed. The specimen is made in a mould for curing of 7 days. For comparing the difference of improvement effect, the unconfined compression test on the lightweight soil lying 4 sheets of Geogrids in horizontal direction is also performed. (b) Deformation-strength property Figure 6 shows the result of the unconfined compression on the air-formed lightweight soil with plastic pieces in a density of approximately 0.8 g/cm3. In consideration of scatter for density of the specimen causing by a difference of mixing condition, the normalized compressive stress by the unconfined compressive strength is represented in the vertical axis of the figure. The lightweight soil without plastic piece shows the maximum strength at small axial strain, and after that, the stress decreases suddenly. It is found that such brittle failure of the lightweight soil is improved by mixing plastic pieces. Next, the improvement effects on peak strength and residual strength of the lightweight soil are discussed. Figure 7 shows the relationship between the unconfined compressive strength and the density of the lightweight soil. As shown in this figure, it is understood that the improvement effect for the peak strength at the same density does not appear clearly in mixing each plastic waste. Because the airformed lightweight soil includes many air bubbles and Poisson’s ratio is approximately 0.1, it is considered that tensile stress does not work in the plastic pieces and the improvement effect does not appear. On the other hand, concerning the improvement effect of the toughness, the relationship between the residual strength and the density of the lightweight soil are shown in Fig.8. Herein, the residual strength is de fined as the compressive stress at the axial
Figure 7. Relationship between unconfined compressive strength and density of light-weight soil with plastic pieces
Figure 8. Relationship between residual strength and density of light-weight soil with plastic pieces
strain of 5%. The residual strength of the lightweight soil with plastic pieces is larger than that in the case of non-mixture at the same density. Namely the improvement effect due to mixing plastic pieces has appeared in residual strength. It may be said that the tension stress of the plastic pieces works when the large deformation of the lightweight soil causes after reaching to peak strength. Thus, when a number of the reinforcement material in a thin and long shape such as plastic piece orny-
113
lon thread are mixed, it is found that the im provement effect appears clearly for residual strength of the air-formed lightweight soil.
lized ground is improved considerably for the increase of toughness due to mixing the plastic pieces.
3 APPLICATIONS TO SURFACE STABILIZED GROUND 3.1 Deformation analysis of su$ace stabilized ground In order to clarify the property of bearing capacity of the surface stabilized ground with plastic pieces, the deformation analysis with two dimensional elastoplasticity is done using finite element method in consideration of the strength property of the improved soil. Usually, crack occurs in the surface stabilized ground when it reaches bending failure with increase of loading stress. Because this crack is a factor to influence the bearing capacity, it is important to consider the occurrence of the crack in the analysis. The model of stress-strain relationship on the element with yield is shown in Fig.9. Type-A supposes the model without the crack for the cement-treated soil with plastic pieces and Type-B supposes the model with the crack for that of nonmixture. Figure 10 shows the failure envelop on the basis of the failure criteria of Mohr-Coulomb. The maximum tensile stress in the element follows the bending strength qYand the maximum shear stress follows the undrained shear strength cl, (=q,/2). The parameters for the analysis are shown in Table 1 and these values were obtained from the unconfined compression test and the bending test of the lightweight soil with density pr =1.0 g/cm3. The mesh used in FEM analysis is shown in Fig.1 1. The cases of the analysis are as follows; Case- 1: Surface stabilized ground without reinforcement (Type-B) Case-2: Reinforcement by laying a sheet of Geogrid in the surface stabilized ground (Type-B+Bar element for Geogrid) Case-3: Mixture of plastic pieces in the surface stabilized ground (Type-A) Figure 12 shows the relationship between loading stress and settlement at the central point based on the analysis. The arrow in the figure represents the elastic limit which means yield stress. In the case-1 without reinforcement, it reaches to brittle failure at a small loading stress. In the case-2 with the bar element, yield stress of the stabilized ground increases somewhat due to tensile stress of the Geogrid, but the brittle behavior is not improved. However, in the case-3 with plastic pieces, yield stress of the stabilized ground increases and the material property changes from the brittle behavior to the ductile behavior. From these results of the analysis, it became clear that the property of bearing capacity of surface stabi114
Figure 10. Failure envelop line used for the cement-treated soil
Figure 1 1. Mesh for deformation analysis Table 1. Parameters used in deformation analysis stabilized ground Plastic piece
soft ground
Non-mixture Mixing
Width (m)
4.5
4.5
Depth (m)
1.0
1 .O
1.8
148000
160000
100
Poisson's ratio v
0.1 67300
0.2 66700
0.3
Shear modulus G (kPa)
Size
Young's E (kPa)
("1
7.0
65
30.8
27.9
45.0
k &Pal
126
145
1000
qu( kw
444
480
$uh
Bending strength CTb, (kPa)
I 43
174
shows the comparison between calculation and test results on the yield stress of the surface stabilized ground with different thickness. The parameters used in the analysis are shown in Table 2. These values of the parameters were decided from the unconfined compression test and the bending test of the cement-treated soils with or without plastic pieces. Predicted yield stress corresponds well with the test results. It is said that the FEM analysis is ef-
Figure 12. Relationship between loading stress and settlement at the center (Result of analysis)
3.2 Verification by model loading test (a) Testing method and test results Figure 13 shows the apparatus of model loading test. Soft ground has been supposed as the Winckler model gathering of the spring. The stabilized ground is prepared using mould by the same method in the section 2.1. Size of the model ground is the length of 1.09m, the depth of 0.3m and the depth of 0.1, 0.15, 0.2m. Soil sample is Kaolin clay and the cement content is 100kg/m3.Loading stress is applied at the center of the model ground using rigid plate in the width of 0. Im under displacement control condition of i m d m i n , and the settlement is measured at 11 points on the one side from the center. The relationship between normalized settlement S/B at the center and loading stress p on the stabilized ground in the thickness of 0.lm is shown in Fig.14, where B is width of the loading. The stabilized ground shows almost proportional relation between the S/B and the loading stress until yield stress independent of plastic content. Concerning the property of bearing capacity, the stabilized ground in case of M=O% shows brittle behavior with clear failure, but that in case of M=5% has a ductile property without clear failure. When the yield stress is defined at the largest curvature point for the case of M=5%, it is clear that the yield stress increases due to mixing plastic pieces. The sketch of the stabilized ground after the test is shown in Fig. 15. When the plastic piece is not mixed, the large crack occurs at the both edges of the loading plate. On the other hand, in the case of M=5%, several small cracks occurs and the excessive crack is restricted by mixing plastic pieces. The same results have been obtained for the stabilized ground in the thickness of 0.15 and 0.2m. Therefore, it became clear from the experiments that the bearing capacity of surface stabilized ground is improved considerably by mixing plastic piece.
Figure 13. Model loading test apparatus for surface stabilized ground
Figure 14. Relationship between loading stress and at the (Test
(b) Comparison between calculation and test results The FEM analysis for the loading model ground is done by the same way of section 3.1. Figure 16
Figure 15. Failure pattern of surface stabilized ground
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Table 2. Parameters used in deformation analysis for model test stabilized ground soft ground Plastic content Width (m) Depth (m)
Size
Young's modulus E (kPa) Poison's ratio v
0.55 0.15
63000
100
0.3
0.3
0.3
26000
40950
65
("1
22.8
21.7
45
(kPa)
27.9
38.5
1000
84
84
37
52
@ub
qu Wa) Bending strength (Thy
M=5%
0.55 0.1, 0.15, 0.2 40000
Shear modulus G (kPa) &b
M=O%
(kPa)
Figure 16. Comparison between calculation and test results on the yield stress
2) Mixing plastic pieces enables to increase residual strength of the air-formed lightweight soil and the brittle behavior is improved. 3) The toughness and yield stress of the stabilized ground are increased by mixing plastic pieces. 4) Composite geomaterial mixing plastic wastes such as PET bottle or fishing net is effective for the utilization of the wastes.
fective for such surface stabilized ground with bending failure. In addition, it is also possible to evaluate settlement behavior of surface stabilized ground by the analysis considering the occurrence of the crack as indicated before section.
4 CONCLUSIONS The main conclusions obtained from this study are as follows: 1) The unconfined compressive strength and tensile strength of the cement-treated soil are increased by mixing the plastic pieces. The improvement effect of the strength depends on the cement content and the optimum condition exists.
REFERENCES Omine, K., Ochiai, H., Yasufuku, N.& Kato, T. 1996. Effect of Plastic Wastes in Improving Cement-Treated Soils. The 2nd International Congress on Environmental Geotechnics: 875-880.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Visualization of interaction behavior between soil and reinforcement using X-ray CT J. Otani, K. Miyamoto & T. Mukunoki Graduate School of Science and Technology, Kumamoto University, Japan
T. Hirai Mitsui Petrochemical Industrial Products, LTD, Japan
ABSTRACT: The objective of this paper is to visualize the behavior of soil-reinforcement interaction using industrial X-ray CT scanner. Here, a new pull-out testing apparatus for the use in the system of X-ray CT scanner was developed and a series of pull-out tests was conducted for geogrid type of reinforcement. Then, the pull-out box was scanned at several steps of pull-out displacement. And a number of cross sectional images for the interaction behavior in each step were obtained. Three-dimensional images were also reconstructed by summing up all the images in each step. Here, the sheet type of the reinforcement was also examined in order to evaluate the effect of the reinforcement geometry on the soil behavior around the reinforcement. Based on those results, the change of the density in the soil around the reinforcement during pull-out displacement was observed and the effect of the shape of reinforcing material on that behavior was also discussed. Finally, the effectiveness of industrial X-ray CT-scanner to geotechnical engineering was confirmed based on those test results. 1 INTRODUCTION Earth reinforcement technique has been widely used around the world and so far, many research projects for both experimental and analytical aspects such as model testing and numerical analysis have been conducted. But the modeling of interaction between soil and reinforcement is still on going issues and its real behavior has not been observed precisely. Recently, an X-ray CT scanner became a useful tool even for geotechnical engineering as a nondestructive testing apparatus. It is known that the result from CT scanning can be realized as the change of the density in the material. This apparatus produces visual images for not only in cross section but also in three-dimension. The objective of this paper is to visualize the behavior of soil-reinforcement interaction using industrial X-ray CT scanner. Here, a new pull-out testing apparatus for the use in the system of X-ray CT scanner is developed and a series of pull-out tests is conducted for grid type of reinforcement. Then, the pull-out box in the apparatus is scanned at several steps of pull-out displacement and a number of cross sectional images for the interaction behavior in each step are obtained. Three dimensional images are also reconstructed by summing up all the images in each step. Here, the sheet type of the reinforcement is also examined in order to evaluate the effect of the reinforcement geometry on the soil behavior around the reinforcement. Based on those results, the change of
the density in the soil around the reinforcement materials during pull-out displacement is observed and the effect of the geometry of reinforcing material on that behavior is also discussed.
2 X-RAY CT SCANNER An X-ray computed tomography(CT) method which
is one of the nondestructive testing method has been used with the name of an apparatus commonly wellknown by the medical diagnostic method which provides the cross sectional images based on the absorption of the x-ray beam through the materials. Recently, an industrial X-ray CT scanner (TOSCANER -23200min:TOSHIBA Corp.) has been installed at the group of Rock Engineering in Kumamoto University, Japan(Sugawara et al. 1997 and Otani et al. 1999). In the system of the X-ray CT scanner, the collimated x-ray is moved around the circumference of the specimen by rotating and translating the specimen table. The detected data are assembled and then, the cross-sectional images are reconstructed using engineering workstation. By using all these cross-sectional images around the circumference of the specimen, three-dimensional (3D) image can be reconstructed. In order to evaluate these nondestructive test results quantitatively, following so-called “CT-value” is used: CT-value = (pt-pL,)KJp,
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(1)
Figure 1. Pull-out test apparatus.
where pt:coefficient of absorption at scanning point; pw: coefficient of absorption for water; and K: constant called Hounsfield value. It is noted that the coefficient of absorption for air is zero. When the value of K is set to be 1000, the CT-value of the air should be -1000. It is known that the CT-value has a linear relationship with the density of the specimen, so that the CT-value is a representative value for the density of the specimen. The CT images are presented with shaded darker or black for lower density region and light gray or white for high density one for all the subsequent black and white images. The total number of the gray level distribution is 256. Precise contents about CT scanner are shown in the reference by Otani et al. 2000. 3 PULL-OUT TEST AND CT SCANNING Figure 1 shows a pull-out testing apparatus which was newly developed by the authors. Figure l(a) shows the total apparatus while Fig. 1(b) shows the soil box part in which the size of the reinforcement and the area of scanning are denoted. Toyoura sand was used and its soil property is shown in Table 1. Two different types of reinforcing materials were
used which are grid type and sheet type. Those are shown in Fig.2. In the test, following process was conducted: 1) After installing the reinforcement in the soil, a series of CT scanning are conducted as a initial condition. 2) Then the pull-out loading is applied at the head of the reinforcement under displacement control(lmm/min.). This loading is stopped at one pull-out displacement level (pre-peak condition j and start CT scanning with every lmm thickness. The total number of scanning is 50. 3) After scanning the soil box, the loading is reapplied and same scanning is conducted at the next strain level (post-peak condition) which is the condition of relatively large displacement for the reinforcement. Figure 3 shows the pull-out force - displacement relationship for the case of geogrid, in which the scanning points are indicated in the figure. It is noted here that stress relaxation was occurred as shown in this figure and this was caused by stopping the pull-out test during CT scanning.
Table 1. Soil property of Toyoura Sand.
Maximum dry density (t/m3)
1.61
Minimum dry density (t/m3)
1.35
Relative density
100
(%)
Figure 2. Types of reinforcement.
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Figure 3.Pull-out force-displacement relationship for the case of geogrid.
4 RESULTS AND DISCUSSION Figure 4 shows one of cross sectional image at the depth of 35mm from the surface of the ground for three different strain levels and the change of the density is shown by gray level distribution. As easily realized from those figures, the reinforcement makes the soil density variable even from the initial condition. Figure 5 shows the vertical cross sectional images for those three cases for total depth of 50mm from 20mm below the surface. These are reconstructed using all the horizontal cross sectional images and the clear shear zones can be observed at the large strain level as shown in Fig.S(c) in which there is not much change between initial and pre-peak strain levels. Figure 6 shows the same images as Fig.5 for the case of sheet type of reinforcement under the same pull-out test. There is no change of the density around the reinforcement in the ground, so that the geometry of the reinforcement makes the in-
Figure 8. 3-D reconstruction images for the shear zone in the soil.
teraction behavior different. Figure 7 shows threedimensional reconstruction images for those three strain levels in which the subtraction image between two images is also shown in this figure. As realized from these figures, the shear banding due to pull-out loadings is clearly occurred around the reinforcement with not single zone but multiple ones in threedimension. Figure 8 shows special three-dimensional images which are reconstructed in order to emphasize the area of shear banding. This is enable to visualizing the inside behavior of the ground. Thus, it is confirmed that the industrial X-ray CT scanner has a possibility of characterizing the effect of the reinforcement in the soil and it may produce the valuable information about modeling of interaction between soil and reinforcement. 5 CONCLUSIONS
Figure 6. Vertical cross sectional images for the case of geomembrane.
A series of pull-out test was conducted for geogrid type of reinforcement. And a number of cross sectional images for the interaction behavior in each loading step were obtained using industrial X-ray CT scanner during pull-out forces. The same pullout test for sheet type of the reinforcement was also examined and the effect of the geometry of the reinforcement on the interaction behavior between soil and reinforcement was discussed. Finally, it is concluded that the progressive failure in the soil could be precisely discussed without any destruction using industrial X-ray CT scanner. It is also confirmed that the possibility of the other applications of this apparatus would be highly expected.
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6 ACKNOWLEDGEMENT
REFERENCES
The author- would like to express their gratitude to M ~&ra . who is a graduate student of K ~ mote University for his assistance for preparing figures and tables in this paper.
Otani, J., Mukunoki,T. and Obara,Y. 2000. Application of X~ ray ~CT Method ~ -for Characterization of Failure in soils, Soils and Foundations Vo1.40, No.2:lll-118. Sugawara,K., Obara,Y., Kaneko,K., Koike,K., Ohmi,M. and Aoi,T. 1997. Visualization of three-dimensional structure of rocks using X-ray CT method, Proc. Asian Regional Conf. On Rock Mechanics, Seoul:769-774.
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Laboratory testing of long-term performance of clay-geogrid interaction A. Pamuk Department of Civil Engineering, Rensselaer Polytechnic Institute, USA
D. Leshchinsky & V.N. Kaliakin Department of Civil and Environmental Engineering, University of Delaware, USA
H.I. Ling Department of Civil Engineering and Engineering Mechanics, Columbia University, USA
ABSTRACT: Pullout tests are often conducted to evaluate soil-to-geogrid interaction properties in design. Most of these tests have used static loads with cohesionless soils to assess stress distribution along the soilgeogrid interface as well as pullout resistance. In this work, laboratory pullout tests were carried out to study the long-term performance of a polymeric geogrid embedded in a cohesive soil. Incremental long-term static or cyclic tensile pulling loads were applied to the embedded geogrid under various confining pressures. Some in-isolation creep tests were conducted to compare the long-term effects of confinement. Under dynamic loading conditions, different frequencies of tensile load were applied. Strain progress and creep development were measured. The testing method and results are presented and discussed in details. 1 INTRODUCTION A limited number of reinforced earth structures constructed with cohesive soils has performed well, showing that they can be used in place of granular soils and thus, reduce the cost of construction. The results from both laboratory and full-scale field tests backfilled with cohesive soil demonstrated that both the short-term and long-term shear strength of cohesive soil might be increased by grid reinforcement (Jewel1 and Jones, 1981). Bergado et al. (1993) reported that cohesive soils compacted to 95% of standard Proctor’s density on the dry side of optimum moisture content could generate pullout capacities comparable to those of the good quality granular soils. Accordingly, there has been great demand on using locally available cohesive soil in mechanically stabilized earth structures, especially in places where good quality granular soil is not readily available. A variety of geogrids has been used for soil reinforcement to provide short-term and long-term stability under static and dynamic loads. Appropriate testing methods are needed to evaluate the long-term stress-strain properties of confined geogrids subjected to pulling loads. Laboratory and field pullout tests are often conducted to evaluate the properties of soil-to-geogrid interaction. Field tests are less common because of the high cost in constructing full-scale structures, but may be more representative. Laboratory pullout tests are often conducted because they offer more controlled testing environment with
less cost. On the other hand, boundary effects (e.g., side wall friction, rigid front face etc.) increase due to various limitations in the size as well as various configurations of pullout apparatus geometries used (Juran et al. 1988; Juran and Christopher, 1989; Abremento and Whittle 1995; Farrag et al. 1993). These variations make it difficult to consistently compare the performance of geosynthetics in different soils (Farrag and Griffin, 1993; Abremento and Whittle 1995). Most of the tests reported in the literature have used monotonically increasing loads with granular backfill materials to assess stress distribution along the soil-geogrid interface as well as pullout resistance. The related testing procedures (e.g., Christopher et. al., 1990) were developed for both short and long-term. This paper presents the results from laboratory pullout tests that were conducted to study the longterm performance of a polymeric geogrid embedded in kaolin clay. The incremental static or cyclic tensile pulling loads were applied to the embedded geogrid under various confining pressures. Strain gages were used to monitor strain distribution and creep development in the geogrid reinforcement subjected to incremental tensile loads. Some in-isolation creep tests were conducted to compare the long-term effects of confinement on the geogrid reinforcement. Under cyclic loading conditions, and in-isolation, different frequencies of tensile load were applied. The testing method and results are presented and discussed in details.
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2 TESTING FACILITY, MATERIALS AND PROGRAM
measured using strain gages bonded at five locations. Figure 2 displays the instrumented geogrid specimen along with configurations of bonded strain gages on the geogrid ribs. The location of the first strain gage was approximately 1.9 cm away from the applied tensile load. The testing plan is given in Table 1. The vertical pressures of 34.5, 69.0 and 103.4 kPa were applied uniformly through an air bag at the top boundary of the box (Fig. 1). The details of the testing facilities and procedures can be found in Pamuk (1997). The long-term tests with the cyclic loads were conducted only with a frequency of 0.1 Hz. A series of laboratory tests was carried out to obtain the index properties and the shear strength of kaolin clay used. The direct shear tests were con-
2.1 In-soil tests Kaolin clay was chosen as backfill material because of its commercial availability as an ideal cohesive soil. Figure 1 displays the in-soil testing setup. The frictional resistance developing between the side walls of the box and the soil was reduced considerably by using high quality silicone grease to coat the walls as well as a 0.4 mm-thick latex membrane separating the wall and the soil. The soil was carefully prepared and then compacted in layers within the testing apparatus with inner dimensions of 60 cm long, 20 cm wide and 30 cm high. The compacted soil was slightly on the wet side of optimum (i.e., 1% above the optimum moisture content) corresponding to about 97 % of standard Proctor density. A normal pressure of 34.5 kPa was applied for 24 hours to settle the soil in the box and to equalize the water content especially around the confined geogrid. A biaxial Polypropylene geogrid (MD 30.5 kN/m and TD 45 W/m) was used during in-soil tests. A load actuator that is capable of generating both static and cyclic loads was used to apply tensile load in the transverse direction. The tensile loads were transmitted through the clamp of the geogrid specimen as shown in Figure 2. The clamp was made by bonding the geogrid specimen between a pair of rigid metal plates using epoxy glue. The epoxy, together with the metal plates, made very stiff clamping so that negligible deformation of the grid occurred within clamp. During the test the clamping assembly was embedded 10 cm into the soil to reduce the boundary effect of the rigid front wall on the applied net normal pressure. The frictional resistance on both faces of the clamp was further reduced by a similar method that was used for reducing side wall friction. Strain distribution along the geogrid specimen was
Table 1. Loading sequence during in-soil tests. Loading Type
Initial Load (kN/m)
Load Increment (kN/m)
Static
1.75
1.75
Cyclic
0.25-1.75
1.75
Figure 2. Geogrid specimen
Figure 1. In-soil testing: (a) Testing setup, (b) Geogrid specimen.
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Frequency
(Hz)
Normal Pressure (kN/m’) 34.5,69, 103.4
o.l
34.5,69, 103.4
ducted on the soil specimens (3.1 cm high, 6.4 cm in diameter) extracted from the testing box at the end of each pullout test as well as on the compacted soil specimens at the desired Proctor’s densities. Consolidated-drained conditions were considered to simulate long-term effects. The tests were carried out under the same confining pressures that were used in pullout tests with a constant shearing rate of 0.23 mdmin. The triaxial tests were conducted on the soil specimens compacted to the desired density, as used in the direct shear tests and than tested under consolidated-drained conditions with an axial displacement rate of 0.13 m d m i n . The index properties and the results from direct and triaxial tests are summarized in Table 2 and Table 3, respectively.
Table 4. Loading sequence during in-isolation tests. Loading Type Static Cyclic
Initial Load
Load Increment
Frequency
iId\J/m)
imm
iHz)
1.75 0.25-1.75
1.75 1.75
0.1 & 0.5
ing (see Table 4). Some results from these tests are and 0.5 Hz. It was observed that the frequency variations of an applied load did not affect the strain-time relationships. The figure indicates that strain increased with the tensile load, and creep strains appeared to be higher under static load than that under cyclic load.
Table 2. Index properties of the soil. Atterberg Limits: -Liquid Limit (LL) -Plastic Limit (PL) Compaction: Standard Proctor: -Maximum dry unit weight -Optimum moisture content (OMC) Specific gravity
59% 32% 14.7 kN/m’ 26% 2.6
Table 3. Direct and triaxial test results of the soil. Test type Direct shear Triaxial
CD
C Water Dry unit (kN/m’) content weight (%) (kN/m3) Peak
27 27
(degree) Peak
dual
14.5
52
19
14.7
14.1
14.5
73
-
15.8
-
Figure 3. Strain-time relationships from in-isolation tests: Effects of frequency variations.
2.2 In-isolation tests The testing apparatus (Fig. I ) that was used in insoil tests was modified to conduct in-isolation tests. The tensile load was applied with different frequencies of 0.1 and 0.5 Hz to measure the effects of frequency variations of the cyclic tensile load on the geogrid reinforcement. The loading program is given in Table 4. 3 TESTING RESULTS AND DISCUSSION In-isolation tests were carried out to study the fundamental behavior of the geogrid without soil confinement so that the effects of the confining soil and the normal pressure on the soil-to-geogrid interaction could be assessed well. Under cyclic loading conditions, the tensile loads were applied at the different frequencies of 0.1 and 0.5 Hz, and with a minimum tensile load of 0.25 kN/m during unload
Figure 4. Strain-time relationship at the front of geogrid: Static load, 0,=34.5 kN/m’.
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the actual load depends on the interface performance). Strain distributions along the length of the geogrid specimen embedded in the clay were recorded by the strain gages bonded at five different locations as shown in Figure 2. The development of strain at various locations of the geogrid is shown in Figure 7 and 8 for static and cyclic loads with a normal pressure of 34.5 Wa. Note that these measurements were recorded 24 hour after the load application. The strain in the geogrid reduced along the length due to interaction between the soil and reinforcement, and then eventually became zero at the end of the geogrid. The zero strain in the grid reinforcement indicated that the applied tensile load did not cause a pullout failure. The tensile load that may cause the failure could not be increased due to limited capacity of the load actuator. Accordingly, for
Three different tests were conducted under various confining pressures with static and cyclic pullout loads. The confining pressures varied from 34.5 through 103.4 kPa, and the pullout loads were increased with a load increment of 1.75 kN/m. The strain-time relationships that were measured approximately 1.9 cm away from the point of load application (i.e., at the front) are shown in Figure 4 for the static loading with a confining pressure of 34.5 kPa. The strain-time relationships that were obtained at the front of the geogrid specimen for each in-soil test are summarized in terms of creep strain rates in Figures 5 and 6 for the static and cyclic loads, respectively. The creep strain rate was higher under static load than that under cyclic load The effect of the confining pressure was negligible at the point of load application where the actual load in the geogrid is known (unlike nodes away from the applied load where only the strain is measured and
Figure 5. Creep strain rate at the front: Static load, 0 ~ ~ 3 4 . 5 , 69.0 and 103.4 kPa.
Figure 6. Creep strain rate at the front: Cyclic load, 0,=34.5, 69.0 and 103.4 kPa.
Of strain at various locations Figure 7. of geogrid during in-soil tests: Static load, 0,=34.5 kPa.
Figure 8. Development of strain at various locations of geogrid during in soil test: Cyclic load, 0,=34.5 Wa.
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most tests no pullout failure as well as breakage occurred in the confined geogrid. The results of the static and cyclic tests as shown in Figures 7 and 8 are compared in Figure 9 in such a way that the applied load required to produce 0.5, 1.0 and 2.0 % strain along the length of geogrid is displaced. It is seen that the difference in strains under the static and cyclic loads diminishes towards to the rear end, and increases in the front end. Strain gage readings in the geogrid provided the measurement of the creep strain that was developed under the constant long-term loading. This helped interpreting the effects of the long-term loading on
the pullout resistance of the geogrid reinforcement in the soil. The distribution of the creep strains developed under the long-term loads for a normal pressure of 69.0 kPa is shown in Figure 10. Figure 11 compares the creep strains developed along the length of the geogrid under various confining pressures with an applied load of 8.75 kN/m. The creep strain increased with an increase in the applied load, and decreased with the confining pressure along the length of the geogrid. However, the creep strain was not considerably influenced by the presence of the soil and normal pressure in the front.
Figure 9. Strain propagation along the geogrid: Static and cyclic loads, 0,=34.5 kPa.
Figure 1 1 . Creep strain along the geogrid specimen: Static load, 8.75 kN/m.
4 CONCLUSION
Figure 10. Creep strain distribution along the geogrid: Static load, 0,,=69.0 kPa.
Some results are given from a long-term laboratory testing on geogrid embedded in clay. The purpose of the testing is to increase the understanding of cohesive soil-to-reinforcement interaction for the future research on utilizing low quality in-situ soils in mechanically stabilized earth structures. The long-term tensile loads were applied statically and repeatedly to the reinforcing geogrid confined in clay. The interaction behavior of the reinforcement was interpreted in different ways; that is, creep strain, creep strain rate, and strain propagation along the geogrid were evaluated under the static and cyclic loads with various confining pressures. The use of strain gages was found an efficient way to investigate the longterm loading effects on a confined reinforcement. The effects of both loadings were compared. It was found that creep strain was higher for static load than for cyclic load, and proportionally increased with the static and cyclic tensile load. Furthermore, for a known tensile load in the reinforcement, the 125
creep strain is practically independent of soil confinement. That is, creep is an intrinsic property of the polymeric material though it could be also dependent on the reinforcement structure (e.g., some nonwoven geotextiles creep and tensile properties may depend on confinement).
5 ACKNOWLEDGEMENTS The funding for this research was obtained from the Federal Highway Administration (FHWA) and the Delaware Department of Transportation (DelDOT) in cooperation with the Delaware Transportation Institute, University of Delaware. REFERENCES Abremento, M. and Whittle, A.J. 1995. Experimental Evaluation of Pullout Analyses for Planar Reinforcements, Journal of Geotechnical Engineering, ASCE, Vol. 121, No. 6: 486-492. Bergado, D.T., Shivashankar, R., Alfaro, M.C., Chai, J.C. and Balasubramaniam, A.S. 1993. Interaction Behavior of steel grid Reinforcements in Clayey Sand, Ceotechnique, Vol. 43, NO. 1: 589-603.
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Christopher, B.R., Gill, S.A., Groud, J.P., Mitchell, J.K., Schlosser, F. and Dunnicliff, J. 1990. Reinforced Soil Structures: Summary of Research and Systems, FHWA RD 89-034, Washington, D.C. Farrag, K.A. and Griffin, P. 1993. Pull-out Testing of Geogrids in Cohesive Soils, Geosynthetic Soil Reinforcement Testing Procedures, ASTM STP 1190, Cheng, S.C.J. (ed.), ASTM, Philadelphia: 76-89. Farrag, K., Acar, Y.B. and Juran, I. 1993. Pull-out Resistance of Geogrid Reinforcements, Geotextiles and Geomeiw brunes, 12: 133-159. Jewell, R.A. and Jones, C.J.F.P. 1981. Reinforcement of Clay Soils and Waste Materials Using Grids, 10”’ International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Sweden, Vol. 3: 701-706. Juran, I., Knochenmus, G., Acar, Y.B and Arman, A. 1988. Pull-out Response of Geotextiles and Geogrids (Synthesis of Available Experimental Data), Geosynthetics for Soil Improvement, Holtz, R.D. (ed.), Geotechnical Special Publication No. 18, ASCE: 92-1 1 1. Juran, I. and Christopher, B.R. 1989. Laboratory Model Study on Geosynthetic Reinforced Soil Walls, Journal of Geotechnical Engineering, ASCE, Vol. 115, No. 7: 905-926. Palmeria, E.M. and Milligan, G.W.E. 1989. Scale and other Factors Affecting the Results of Pullout Tests of Grids Buried in sand, Geotechnique, Vol. 39, No. 3: 5 1 1-524. Pamuk, A. 1997. Long-Term Interaction Behavior of Geogrids Embedded in Clay Subjected to Static and Repeated Loads, MS thesis, Dept. of Civil and Environmental Eng., University of Delaware, Newark, Delaware.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Strength behaviour of lateritic soils randomly reinforced with jute fibre B.R. Phani Kumar & M.V.B. Ramana Sastry Department of Civil Engineering, JNTU College of Engineering, Kakinada, India
ABSTRACT: This paper presents strength characteristics of a lateritic soil randomly reinforced with jute fibre. Jute fibre of different lengths was mixed with the soil. The fibre content was varied from 0 to 0.6% by weight of dry soil. Unconfined compression tests, triaxial compression tests and tensile strength tests were conducted on the soil-fibre mixes. Specimens for strength tests were prepared at the optimum moisture content and dry density of the mix. Results of strength tests indicated that the strength of reinforced soil increased with the % of fibre content and aspect ratio. However, in the case of unconfined compressive strength. there is an optimum fibre content for maximum strength. Jute fibre may be considered for reinforcement in lateritic soils when an open cut is to be made and stability is a problem. Reinforcing with jute fibre of pavements with heavy traffic and on lateritic soils improves tensile strength.
1 INTRODUCTION Lateritic soils, which constitute an important group of residual soils of India, are essentially products of tropical or sub-tropical weathering and are predominantly coarse-grained and ferruginous in character. They form a good foundation material. They pose certain probiems such as high permeability, and decrease of strength at a depth from the surface. Stability of cuts in vesicular laterites-is a problem in the high shear stress level zones near the cuts. A simple flattening of the slope will not increase the stability. Soils are commonly reinforced by continueus oriented inclusions. namely. strips, fabrics and grids (Vidal, 1969). Early civilizations mixed soil with straw or other fibre available for strengthening it. Several soil engineers used material like straw, jutc. saw dust, horse-hair etc. as reinforcement to improve the engineering behaviour of soils. Satyanarayana et al. (1977) studied the behaviour of a cement-treated expansive clay reinforced by fibres of glass and asbestos in compression and tension and found that addition of these fibres increased the tensile and unconfined compressive strengths. An increase of 25 to 55 percent in tensile strength and 5 to 30 percent in unconfined compressive strength is reported. Tensile strength behaviour of lime-stabilized lateritic soil, plain and reinforced with coconut fibre were reported by Bhattacharya et al. (1984). Fibrereinforced mixes were found to have developed increased flexural strengths over plain mixes. Freitag
(1986) observed that lean sand-clay mixes and different types of synthetic fibres showed higher compressive strengths than those unreinforced. The type of the fibre used did not seem to have significant effect on strength. Reinforcement of an expansive soil with 'Garware twine' has resulted in increased shear strength and tensile strength with increase in fibre content, where as the maximum dry density and flexural strength decreased with an increase in fibre content (Setty et al. 1990). It has been reported that reinforcing a clayey soillime mix with coconut fibre has improved the compressive strength as well as tensile strength (Ramana Sastry et al. 1993). Rama Sarma et al. (1997) reported that coconut fibre can be directly mixed with soil far strengthening weak sub-grades. Considerable increase in CBR values has been reported. For the maximum compressive strength, there is an optimum content of fibre reinforcement. Agarwal et al. (1997) studied the influence of fibre properties like weight, fraction and aspect ratio on the strength characteristics of a Kaolin soil mixed with random discrete fibres of polypropylene and jute. Their studies show that the unconfined compressive strength and undrained cohesion increase with fibre content for both the types of fibre. This increment is more with jute fibre as compared to polypropylene fibre. In this paper the authors present results of laboratory tests (Suryanarayana, 1999) carried out to find the efficacy of jute fibre as a reinforcement in improving the strength of lateritic soils.
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Table 1. Engineering properties of the soil.
2 MATERIALS AND THEIR PROPERTIES 2.1 Soil
Value
The lateritic soil used in this investigation has been obtained from a depth of about 1.0m below ground level, freed from organic matter, pulverised and sieved through the set of sieves specified in relevant Indian Standards for different tests and stored in airtight containers.
17.10 20 Table 2. Results of compaction tests.
2.2 Jutefibre Air-dried jute fibre, generally used for packing purposes, has been obtained from local market. The fibre is cut into pieces of two different sizes. 20 mm and 30 mm and kept separately in air-tight containers for use in the investigation. Diameter of the jute fibre is 1 rnm.Jute fibre is a bio-degradable material. In order to improve its resistance against degradation the fibre was dipped in hot molten asphalt so that a uniform thin film of asphalt is obtained on the fibres.
0.2 0.4 0.6
17.00 16.95 16.90
16.95 16.90 16.85
18.6 20.2 21.0
19.6 20.4 21.2
creased the optimum moisture content marginally. Fibre being very light material (specific gravity 0.5 ) compared to soil (specific gravity 2.62) even small quantities have a larger bulk. and addition of it to soil has replaced larger quantities of heavier material, resulting in marginal changes in compaction characteristics of the soil-fibre mixes. This is in accordance with the observation of earlier workers (Bhattacharya, et al. 1984, Ramana Sastry et al. 1993, Rama Sarma et al. 1997).
3 TESTS CARRIED OUT The tests have been carried out far finding the index properties, compaction characteristics, unconfined compressive strength, tensile strength behaviour and shear parameters as determined from triaxial compression (quick) tests. Tensile strength of soil-fibre mixes has been determined by ,conducting indirect tensile strength test (Cylinder split test or Brazilian Test). Specimens used for the strength tests have been moulded at the maximum dry density and optimum moisture conter.t of the particular reinforced soiljute mix as obtained from the compaction tests.
4.1 Unconfined compressive strength Figure 1 presents the influence of fibre content and aspect ratio on the unconfined compressive strength (UCS) of the jute-soil mixes. For any aspect ratio the unconfined conipressive strength increased with fibre content up to an optimum fibre content (0.4 percent in this study). With further increase in fibre content, the strength decreased. For any fibre content, mixes with 30 aspect ratio have resulted in higher strengths compared to mixes with aspect ratio of 20. Aspect ratio of the fibre appears not to have significant effect on the optimum fibre content for strength. Jute fibre reinforcement has contributed to
3.1 Variables considered in the investigation The following variables have been studied. Fibre length (mm): 20 and 30 Fibre content (%): 0,0.2,0.4 and 0.6 Aspect ratio (length/diameter of fibre): 20 and 30 The length of fibres and fibre content (%) have been fixed taking into consideration factors like ease of mixing and compacting of the jute-reinforced soil mixes.
4 TEST RESULTS AND DISCUSSION Tables 1 and 2 show the engineering properties of unreinforced soil and the compaction results of the reinforced soil respectively. A perusal of the results presented in Table 2 shows that addition of fibre to the lateritic soil in general decreased the maximum dry density and inFigure 1 . Effect of fibre content on UCS.
128
a significant increase of cohesion and reduction in the value of angle of internal friction and may be responsible for the increase in unconfined compressive strength with the increase in fibre content upto the optimum value. Beyond the optimum fibre content. the increase in compressive strength is less and the angle of internal friction remained practically constant with increase in fibre content. This may be responsible for the decrease in unconfined compressive strength. Studies carried out by earlier researchers (Satyanarayana et al. 1977, Ramana Sastry et al. 1993 and Bhattacharya et al. 1984) on stabilised soil mixes reinforced with various randomly oriented fibres also show that fibre-reinforcement improves the compressive strength of the reinforced soil and that there is an optimum fibre content for maximum compressive strength.
further increase in fibre content. though there is fairly good increase in cohesion, a marginal decrease in the angle of internal friction is noticed. For any fibre content, reinforced mixes of aspect ratio of 30 have shown higher values of cohesion than those of aspect ratio of 20. However, with respect to angle of internal friction, reinforced mixes of aspect ratio 30 have shown slightly lower values of angle of internal friction than those of aspect ratio 20. Shear strength of the reinforced soil mixes increased with the jute fibre content as well as with the aspect ratio (Fig 4). These findings are in accordance with the observations of Agarwal et al. (1997). Open cuts in lateritic soils have posed stability problems to field engineers. In the light of improvement of shear strength of lateritic soil when reinforced with jute fibre; this technique may offer a solution to the problem.
4.2 Results of triaxinl compression tests
4.3 Results of tensile strength tests
Figures 2 and 3 show the variation of cohesion and angle of internal friction with % fibre content as obtained from triaxial compression tests. Irrespective of the aspect ratio, addition of jute fibre (0.2 percent) has resulted in a considerable increase in cohesion and large decrease in angle internal friction. With
The study of tensile strength of soils and stabilized materials has received very little attention of the engineers. Tensile stresses are set up in soils due to movement of traffic on pavements, shrinkage of soils due to seasonal variation in temperature and al-
Figure 2. Effect on cohesion.
Figure 4. Effect on shear stress
Figure 3. Effect of the fibre content on $.
Figure 5. Effect on tensile strength.
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ternate drying and wetting of soils. A knowledge of the tensile strength is needed in the study of earth dams, airport pavements etc. Reinforcement of a lateritic soil with jute fibre improves its tensile strength. For any fibre content. tensile strength increases with the aspect ratio. For any aspect ratio, randomly distributed fibrous material throughout the matrix of the soil contributes to the increase in tensile strength. Figure 5 shows the variation of indirect tensile strength (ITS) with fibre content. Studies carried by Bhattacharya et al. (1984 ), Ramana Sastry et al. (1993 ) and Datir et al. (1997) on lime-stabilized lateritic soils reinforced with fibres like coir, jute, and polypropylene fibres reported remarked increase in tensile strength (flexural and direct) over those stabilized with lime alone.
struction of high density and heavy traffic pavements. Further, reinforcement of lateritic soil with jute may be used in zones of high shear stresses to improve the stability of open cuts in the regions of lateritic soils.
5 CONCLUSIONS Studies carried out on a lateritic soil randomly reinforced with jute fibre show that this natural fibre, available in plenty as a by-product of jute industry, is quite an efficacious and an economic material for reinforcement. The size (length) of the jute fibre and the amount to be used for reinforcement has to be fixed based on ease with which a uniform mix can be made and compacted. I t is observed that the unconfined compressive strength, shear strength and tensile strength of the reinforced mixes increase with the fibre content and aspect ratio. However, the unconfined compressive strength of the randomly reinforced mixes increased up to an optimum fibre content (0.4 percent in the present study). Further increase in fibre content resulted in a loss of strength. Increase in tensile strength of reinforced soil mixes may be taken as an advantage in the con-
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REFERENCES Agarwal K.B. & Subash Chandra 1997. Improvement in strength and bearing capacity of clay due to fibre reinforcement. Proceedings of the Indian Geotechnical Conference. Bhattacharya P.G. & Pandey B.B. 1984. Effect of density on strength and modulus of plain and fibre reinforced lime lateritic soil mixtures under static and repeated load. Proceedings of Indian Geotechnical Conference. Datir D.D., Pandya H.P., Mankad V.M. & Panehal D.A. 1997. Effect of Polypropylene Fibre on tensile strength of stabilized soil. Proceedings of Indian Geotechnical Conference. Freitag R. 1986. Soil randomly reinforced with fibres. Journal of Geotechnicul Engineering. ASCE. Vol. 112. Rama Sarma, K & Rajesh.G. 1997. Improvement of CBR values using coconut fibre. Proceedings of Indian Geotechnical Conference. Ramana Sastry. M.V.B. & Satyakumari T. 1993. Strength characteristics of a plain and fibre reinforced lime treated expansive soil. Indian Roads Congress Highway Research Bulletin No 49. Satyanarayana B. & Srivasthava K.M. 1977. Tensile and compressive strength of soil-cement-glass and soil-cementasbestos fibre mixes. 1 " National Symposium on expansive soils held at H.B.TI. Kanpur. Setty K.R. Narayana Swamy & Anantha Krishna Murthy. 1990. Behavious of fibre reinforced black cotton soil. Proceedings of Indian Geotechnical Conference. Suryanarayana N. 1999. Strength characteristics of jute - fibre reinforced lateritic soil. Thesis submitted for award of M. Tech Degree. J.N. Technological University. Aiidhra Pradesh. India Vidal H. 1969. The principle of reinforced earth. Highwen Research Record No. 282. H. R. B. Washington.
Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 0 2001 Swets & Zeitlinger, ISBN 90 265 1852 8
Studies on geotextile/ soil interface shear behavior M. Salehi Geotechnical Engineers, Khak Azma Co., Ministry of Energy, Tehran, Iran
ABSTRACT: Shear frictional behavior of soil/geosynthetic interfaces plays a pivotal role in the overall performance of geotextile-reinforced roads. Since a substantial proportion of the total land areas in many Southern parts of Middle east countries is composed of organic soils, it was seen of particular importance to investigate the shear frictional behavior of such soils when subjected to loading with geotextiles used as reinforcement. Two types of soils were used; namely organic silty clay and a fill material, which is a sandy type of soil. From the results of the shear box tests performed it appears that there is a relationship between the tensile strength of the geotextile used and the shear strength of its interface with the organic clay, with the shear strength of the interface increases with the increasing tensile strength of the geotextile. The shear strength of geotextile/fill interfaces did not show a consistent relationship with the geotextile tensile strength. mechanism of geotextile/ soil interfaces and the design of reinforced unpaved roads.
1 INTRODUCTION Organic soils are considered some of the most problematic types of soils for their compressibility and high moisture contents. However because they constitute a considerable proportion of the total land areas in many parts of the world in general and in Northern part of Iran in particular it is necessary to consider these soils as potential subgrades for the construction of reinforced unpaved roads, Hobbs, (1986). One of the major factors that controls the performance of reinforced soil structures is the interaction between the soil and the reinforcement. It is necessary to obtain accurate bond parameters for the design of these structures. It was desired to study the behavior of geotextiles as soil reinforcement materials for their availability in the local market and their wide-spread use all over the world for soil reinforcement applications. Accordingly a test program was carried out to investigate the shear frictional behavior of geotextile/soil interfaces. A series of shear box tests were carried out in the laboratory for this purpose. The experimental results will provide a better understanding of the shear behavior of reinforced unpaved roads. Two types of soils were used in these tests; namely sandy soil and organic clay along with nonwoven needle punched geotextile with four different tensile strengths. It was desired to study the effect of the variation of the geotextile tensile strength on the behavior of this system. Such knowledge would provide a better understanding of the shear frictional
2 DESCRIPTION OF MATERIAL A brief description is given of the materials used in this experimental study. The fill and the organic clay used were a scaled down version of the original site materials. However the type of scaling down of these two materials differs in that the fill was scaled in terms of its particle size distribution whereas the clay was scaled down in terms of its undrained shear strength. The fill material was scaled down using a scale of 1:4 in order to account for the modeling requirements. The Particle Size Distribution curve revealed that more than 50% of the organic clay is in the clay fraction with the rest of the soil being in the silt fraction. Liquid and plastic limits of this organic clay were found to be 83.5% and 48.1% respectively. This gives the Plasticity index of the soil as being equal to 35.4%. It has been found also that the soil had an organic matter content of 1 1.1% with an average specific gravity of 2.54. Since our interest lie in subgrade shear strengths in the range of 20-60 kN/m’ in the field, a range of subgrade undrained strengths of 5-15 kN/m2 was used in this study. To satisfy this modeling requirement adding measured amounts of water and by mixing it to reduce its density until the undrained strength of the soil being used was ‘/4 that of the same soil in the field reduced strength of the soil. 131
The geotextiles used are non-woven needlepunched geotextiles. These types of geotextiles were found to be most effective when filling over very soft cohesive soils. It was proven to be able in reducing settlements and providing a platform for sewing and rolling for site applications. A description of their properties and specifications are shown in Table 1. A square shear box lOOmm by l o o m , split horizontally at mid-height was used for direct shear testing. For testing soil only, whether organic clay or fill, two porous plates were used one at the bottom of the sample and the other on the top. All tests were strain controlled under the same constant rate of shear loading. The range of normal stresses applied was (2998.7) kN/m2, which was similar for all tests conducted in this study with and without a geotextile to simulate the site stresses. It was found earlier that using lower normal stresses would render the test results for these types of soils inconclusive. High normal stresses were used by other investigators when performing direct shear tests, Fishman and Pal, (1994). The geotextiles were cut to square pieces of 100 by lOOmm and then each piece was glued using epoxy glue to the top of a piece of hard wood having the same dimensions (100 by 100mm). This procedure was used previously by other investigators when conducting their soil/geosynthetic Friction tests, Fishman and Pal (1994). After each shear box test the geotextile piece was removed and replaced with another one with the same dimensions to account for the damages in the geotextile texture that might have occurred as a result of the previous test. Table 2 shows the index properties of the tested organic clay and fill specimens. The organic clay
specimens were chosen from all parts of the subgrade material. The organic clay specimen was placed inside the upper half of the shear box with the geotextilewooden block assembly occupying the lower half. Care was taken in excavating the organic clay and any organic components that were to coincide with this soil specimen where cut using a sharp knife to the dimensions of the sampler. After placing the organic clay specimen inside the upper half of the shear box, a cheese wire was used to cut it to proper dimensions to fit inside the shear box. Then normal load was applied and the test was carried out. Fill material was used as the second type of soil in contact with the geotextile. For testing fiWgeotextile assembly, the geotextile-wooden block assembly was fitted inside the lower half of the shear box occupying it first, then the upper half would be screwed on top of the lower and the fill material compacted above the geotextile-wooden block assembly. The rate of shearing applied was 1.27 &min., which was chosen to ensure undrained conditions when testing the organic clay. It was proven through previous studies that this rate of loading was fast enough to cause undrained conditions in the clay, Fishman and Pal, (1994). The same rate of loading was applied to all tests using the two types of soils. 3 RESULTS OF TESTS Figure 1 shows failure envelopes for soil/geotextile interfaces and soil-alone interfaces. Shear stress is plotted against the normal stress both expressed in kN/m2. Figure 2 details the shear strength parameters of the interfaces versus the tensile strength of the geotextiles used. The parameters are expressed in kN/m2 and the tensile strength is in kN/m.
Table 1. The geotextile specifications. ~
Unit A Characteristic Test Standard 180 Nominal ASTM glm’ D 3776 Mass/m2 B.S 6906/1:+ kN/m 12.5 Tensile strength 45 Elongation at B.S 6906/1“ % break Thickness under Pressure ASTM mm 1.80 2m/m’ D 1777 mm 0.80 200 kN/m’
0
B
C
Table 2. Index properties of the sshear box tests performed.
--
200
I Density
280 Type of Test Peaty clay-Only
--
13.9 19.3
(g/cm3) 1.36
--
45
I Moisture
I Organic
Content % 115.4 110.6
Content % 14.7
11 1.9 112.5
15.4 14.7
45
2.00 2.50 3.90 1.20
British Standards
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Fill-Onlv Fill-A Fill-B Fill-C
1.61
Fill-D
1.61
0.8
I
Figure 2. Shear strength versus tensile strength for the soil/ geotextile interface.
Figure 1. Failure envelopes of the soil/geotextile interface
4 DISCUSSION AND CONCLUSION As a result of shear box tests, the following is an attempt to discuss and interpret the results of these tests. Examining the organic clay failure envelopes in Figure 1 reveals a behavior that differs according to the tensile strength of the geotextile used in contact with the soil. By examining the organic clay-geotextile curves a trend appears to dominate these curves associated with the increase of the geotextile strength. It can be seen clearly from Figure 2 that increasing the tensile strength of the geoetextile in contact with soil increases, in return, the shear strength tolerated by the soil-geotextile interface. This can be seen clearly by observing the increase in the shear strength angle, (@),and the cohesion or adhesion, c, for the range of tensile strengths used, Figure 2. There is a significant increase of shear strength associated with the increase of geotextile strength. It can be seen that the highest shear strength in terms of both @ and c was gained by the C-organic clay interface.
Using a geotextile with a lower strength resulted in lower shear strength for the B-soil interface followed by lower shear strength for the lightest geotextile used (A). Organic clay-only test gained a shear strength lower than all of its interfaces with the geotextiles. The only abnormal behavior was exercised by D, which, although having the highest strength, achieved an interface shear strength lower than all other geotextiles. This is thought to be due to its complete interlocking with the organic soil. A result of which failure might have occurred inside the organic clay near the joint soil-geotextile interface. Another reason might be the high pore water pressures on the soil-geotextile interface reducing the active stresses in that region causing a reduction in the shear strength of the interface. This observation is consistent with that of Koerner et al., (1986) who conducted tests on a variety of cohesive soils in contact with various geomembranes. He noticed lower friction coefficients for the harder PVC and HDPE geomembranes. Examining fill-geotextile interfaces reveals a behavior that differs completely from that of the organic clay-geotextile interfaces discussed above. Fill-geotextile interfaces seem to be relatively close in terms of both Angle of Shear Friction (@) and Cohesion (c). Although on average they exhibited angles of shear strength much higher than those of the clay-reinforcement interfaces, it seems that the effect of the variation of the geotextile strength has no significant effect on the magnitude of their angle of shear friction. The only-fill envelope had shear strength consistently lower than all its interfaces with the other geotextiles.
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As for the rest of the shear envelopes that represents the behavior of the interfaces of the four geotextiles in contact with the sandy fill material, all envelopes seem to share approximately the same angle of friction and the same cohesion. By examining Figure 1 that shows failure envelopes of the soil-geotextile interfaces and soil-only tests, it can be observed, in general, that fill geotextile interfaces gained higher angles of shear strength with lower cohesion. This in comparison with the organic claygeotextile interfaces that on the contrary had higher cohesion values with significantly lower angles of shear friction. This increase in cohesion of the organic claygeotextile interfaces associated with the increase of the geotextile strength can be attributed to the fact that as the thickness of the geotextile increases, its capacity for performing drained cohesion increases. In other words when the geotextile thickness increases, a corresponding increase in its capacity as a drainage media occurs. This in turn gives space for more water to be drained from the interface with the soil. The result of which is higher interlocking with the organic soil.
It should be pointed out that high cohesion values were obtained for the organic soil in contact with the geotextiles and that these values are plotted in Figure 2 that shows their magnitude versus the tensile strength of the geotextiles used. It can be observed that cohesion plays a very important part of the bond resistance in organic clays. It contributes to the overall shear resistance of the organic soil more than its angle of shear friction. Therefore cohesion in organic soils is an important part of the bond resistance that should be taken into account when designing reinforced unpaved roads. REFERENCES Fishman, K.L. & Pal, S. (1994), Further study of geomembranelcohesive soil interface shear behavior, Geotextiles & Geonienzbranes, 13, pp. 57 1-590. Hobbs, N.B. (1986),Mire morphology & the properties & behaviour of some British & foreign peats, Quarterly Journal of Engineering Geology, London, Vol. 19, pp. 7-80. Martin, J.P., Koerner, R.M. & Whitty, J.E. (1984). Experimental friction evaluation of slippage between geomembranes, geotextiies & soils. Proc. Int. Cant on Geoniembraries,Industrial Fubr-ics Association International, pp. 191-196.
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Earthquake Engineering Frontiers in the New Millennium, Spencer & Hu(eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Discussion of safety from a study of the creep rupture of polyester P. Segrestin Freyssinet International, Soil Technical and Marketing Department, Nozay, France
P. Orsat Freyssinet International, Mknard Division, Nozay, France
ABSTRACT: The extensive creep rupture testing by Freyssinet of about 160 polyester tendons used in Paraweb strip soil reinforcement led to interesting results and discussion. The paper first summarises the test program and outcomes. The tests demonstrated that the tendons retain a significant so-called "residual strength" before rupture. This gave rise to the feeling that the combination of reduction factors and safety factors generally used for design might be overly conservative. The paper demonstrates that this idea is misleading and deceptive. The paper however acknowledges that the allowable tensile load could be noticeably larger (in the context of a "Load and Resistance Factor Design") if the resistance factor of safety were applied to the loss of strength rather than to the remaining strength. The paper finally recognises the value of the residual strength with regard to short duration loading, such as seismic effects. 1 INTRODUCTION Between 1995 and 1998, the Technical Research Centre of the Highways Department in Normandy, France undertook, at the request of the Freyssinet Company, a series of trials on the creep behaviour of Parafil high tenacity polyester tendons. These tendons are of the same type as those which form Paraweb soil reinforcements. The series was carried out on a large number of samples, 157 in total. For half these samples, the trials consisted of obtaining creep rupture of tendons under constant load, at ambient temperature (20°C). The duration of loading ranged between a few hours and 11 months, i.e. nearly 8,000 hours. For the other half, loading was interrupted after periods of 10 days to 15 months and the residual strength of the tendons was measured. The results produced by the series of experiments have renewed consideration of numerous subjects, beginning with the admissible strength to use for design, i.e. the definition of a pertinent and reasonable safety factor. We summarize here some aspects of this reflection in the imaginary form of a discussion which could have occurred between a "salesman" whom we will call Mark, and a "user", whom we will call Bob.
usual way, with, as abscissa, a logarithmic scale of time at the end of which rupture occurred, and as ordinal, the applied load, expressed as a percentage of the reference nominal strength. I should emphasise that this reference strength has been verified for each trial series on around twenty samples (in particular to detect possible problems with the jaws). From the experimental results a minimal value of the logarithm of rupture time has been calculated for each load level at a level of confidence of 95% (Figure 1). Then a regression line was drawn for these values and extrapolated to the abscissa which corresponds to a service life of 100 years. It shows that there is a 95% chance that rupture will not occur before the end of the service life if the load is equal to 66% of the nominal strength (Figure 2).
2 DISCUSSION Mark First of all I will show you a graph summarizing all the creep rupture results, presented in the usual way, with, as abscissa, a logarithmic
Figure 1. Creep rupture tests. 95% confidence regression line.
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the "material" safety factor and we will take it as equal to 1.5 (this is in fact the order of magnitude that everybody has in mind). We will apply it to the creep rupture strength which has been determined for a 100 years service life, i.e. 66% of the nominal strength. The load is brought down to 44%. It can be seen from the graph that it corresponds to a period of about 15 billion years, anyhow!
Bob
What you are in the process of suggesting is that safety should be considered over time, on the length of service life, rather than on strength? Don't forget that we are on a logarithmic scale. If you take a safety factor of say 2 on the length of service life (let's be generous!), aiming at 200 years instead of 100, you deduce from the graph that we can work at a level of 65%, instead of 66% (Figure 3). The difference doesn't seem to me to be sufficiently great to make me feel that we have really put ourselves on the side of safety.. .
Figure 2. Creep rupture tests. Extrapolation to 100 years.
Bob
That's quite clear. It seems to me that these results are very close to those already found in the literature for other high tenacity polyester fibres? Mark Exactly. But if you examine this graph while remembering that polyester soil reinforcements are generally employed at a level of about 30% of their nominal strength, you will see that under this load the reinforcements have no chance of rupturing before 3 million billion years! The safety is enormous! Hold on a minute ... In your 30% factor you are somewhat mixing everything up. First of all you are including all losses of strength: losses which result from creep certainly, but also those dues to installation damages at the time of construction and those produced by chemical alteration such as hydrolysis. That is not safety - they are real losses. Then you effectively include actual safety, but it applies as much to calculated forces as to the long-term resistance of the material. Let's try to estimate what safety there is on the material itself. I calculate that, taken overall, in general 95% of the strength remains after construction, 95% after hydrolysis and 66% after creep, that makes 60%, all losses considered together. Safety on the calculated forces depends on the local Codes. It is about 1.35 in France; it will be about the same in the USA under the AASHTO "LRFD" specifications. Thus we arrive at about 45%. In effect there only remains a safety factor in the order of 1.5 on the material strength, if it is employed as you say at 30% of its nominal strength.
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Mark I believe that it is now time to tell you about the other phase of the trials, those concerning residual strength. As I have already told you, on half of the samples subjected to the creep tests, they were not taken right up to rupture under constant load. The trial was interrupted at different dates, getting closer to the estimated date of rupture. The sample was released from load and its remaining strength was measured, using the same operating method at controlled strain, used to define the nominal resistance of the product. Half a dozen samples were even not released from their load but the load was progressively increased, with lead balls added to the weight. Whatever the method the result was essentially the same: the residual strength before reaching creep rupture is very high, very close to the nominal reference strength. The shape of the curves which represent the variation in mean residual strength (although some look a bit bumpy..) suggests that this does not really begin to decrease until the last mo ment, if you can say that, in any case shortly before rupture occurs (Figure 4).
Mark OK. But, please, for the moment let us forget installation damages and hydrolysis. Let us suppose that we have a case where the losses caused by them are negligible. Let's only concern ourselves with creep rupture and safety. We will set aside the safety on the calculated forces, because that comes from general standards and has nothing to do with the reinforcing material used. We will consider only
Figure 3. Hypothesis of a safety of 2 over the service life.
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Figure 4. Residual strength.
Bob Don't forget that we are on a logarithmic scale and what appears to you to be an acceleration of the loss of strength shortly before rupture might be a reasonably regular decrease, perhaps even linear, on a natural scale.. . But what is your point? Mark This. Suppose that the tendon is loaded to a level corresponding to its creep rupture strength, not even at 200 years as was envisaged just now, let's be more generous, at, say, a thousand years, i.e. at about 63% of its nominal strength. From the shape of the experimental curves of residual strength it can certainly be estimated that at the end of 100 years, i.e. at the end of the service life, the tendon will still have a resistance of at least 95% (Figure 5). In other words, without any problem you will have the safety that you require, because 95 divided by 63 makes precisely 1.5!
Bob
Let's think about this, there is certainly a paradox somewhere!... We have noted that the reinforcement breaks, or risks breaking, at the end of 100 years if loaded at 66%. You are suggesting that by loading it at only 63 it becomes capable of resisting to 95? In other
words, you would only need to apply a safety factor of 1.05 to automatically obtain one of 1.50? The error is that we have forgotten that this residual strength of 95% is not permanently available. Like the reference nominal strength it is ephemeral. At any time a load equal to 95% of the nominal strength will cause rupture in a fraction of a second. It is just that which occurs in your measurements of residual strength: you start from a constant load level then at a given moment you increase the load to go as far as rupture. Once rupture has occurred, its over, the reinforcement is out of service.. . What interests me more is to know just how far I can permanently load the reinforcement without risk of it breaking before 100 years, and that in the mostunfavourable hypotheses. You have already given me the answer: there is no question of exceeding 66% (allowing that we are only talking about creep). Mark What do you mean by unfavourable hypotheses?
Bob
Let's return to the general question of safety factors. I think that the graph below illustrates it well enough (Figure 6). It represents on the left the probability of the value of the calculated tensile force, on the right, that of the long term strength of the reinforcement . The problem is to ensure that the minimum longterm strength is greater than the possible maximum force (accepting a priori a very small risk that this is not the case). In what is called the maximum force is included the load factor, i.e. the partial safety factor which takes into account all the uncertainties concerning the real value of the loads applied, the validity of the calculation methods etc. This has already been talked about, essentially returning us to the common rules for all types of construction.
Figure 5. Creep rupture and residual strength.
Figure 6. Safety illustration.
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which indefinitely retains 100% strength. Why should we only allow two thirds, as if we estimate the uncertainty on the loss as 33%? Imagine on the other hand a perishable material which, from the extrapolations, may lose 70% of its initial resistance over one hundred years and retain only 30%. Why should we have sufficient confidence in it however to count on 20%, as if the uncertainty this time was no greater than 10%, i.e. less than 15% of the probable loss?
For what is known as minimal strength we start from the most likely strength, which is determined from the anticipated losses, i.e. by bringing into play the reduction factors due to installation damages, chemical alteration and creep. Then, safety must be taken into account, i.e. allowance must be made for all the uncertainties which remain. Above all three things must be considered: - the validity of the extrapolations - possible synergy between the different types of loss of strength - and then any unknowns in changes in the material with time, in the environment in which it is placed. Mark Let's consider them one by one. Concerning the validity of the creep effect extrapolations, it seems to me that there isn't a problem: we have analysed the experimental results statistically and have determined a minimum envelope curve at a 95% confidence level.
Mark The idea of applying the safety factor to the losses seems to be completely justified. But we must again agree on the losses, or the part of the losses to which the safety factor should be applied. In fact there are losses where there is no uncertainty. This is above all the case for those resulting from installation damages, because they are measured and are not extrapolated. In the case of creep and chemical damage, this should also be the case of what was seen and measured during direct experiments. If we again look at the creep rupture graphs of our polyester tendon (Figures 1 to 3), we can see that the trials carried out over a year cover a loss of about 25% relative to the reference strength (from 100% to 75%). There is no uncertainty on that 25% and there is no reason for it to be amplified by the safety factor.
That effectively gives us a satisfactory value for the reduction factor. But that does not really include safety. It must be indeed recognised that the loading range is narrow, that there is a lot of dispersal, that the trials lasted for less than a year while they are extrapolated over one hundred, etc. It is doubtful whether with other trial series exactly the same regression line would have been found and particularly the same slope. Thus it cannot be excluded that this line may be inclined a little more. Mark
Bob I agree. Also I propose to illustrate the way in which the safety factor should be applied with the following graph (Figure 7). On the time axis, still a logarithmic scale, can be seen two periods: the period te of the direct experimentation, or more exactly that from which the extrapolations are made, and the required service life ts. Two strength values correspond to it: F, to time te, and F, to time ts, as results from the extrapolations. If you wish to express these values as a function of reduction factors as defined in most codes (RFI for installation damages, RFd for deterioration due to the environment, RF, for creep), the following can be written adding a suffix relative to the date:
You also mention synergy.
Bob I am thinking for example about doubts that may exist concerning possible acceleration of creep in a damaged or deteriorated material. But that is in fact already posing the third question that I mentioned concerning everything that is not suspected, everything not yet known about other causes which may affect the long-term strength of reinforcement in the fill. Mark Finally, how do you propose to take into account all these uncertainties?
Bob Because it essentially involves taking into account unknowns, it is of necessity a bit arbitrary. Inevitably we are again going to think of a safety factor in the order of 1.5. On the other hand, it seems completely logical to suggest applying this safety factor not to the remaining strength, as is habitually done, but rather to the strength which is lost. It is after all the loss of strength that we are trying to estimate, and it is on this loss that there are uncertainties. It is this which should be increased for safety, a priori by 50%. On the other hand it is not satisfactory, as is generally done, to divide the remaining strength by 1.5, i.e. to reduce it by one third. Imagine an extremely stable, extremely durable material,
Figure 7. Principle for application of the "material" safety factor.
138
Fe = F, /(RFi, * RF,,
* RF,, )
F, = F, /(RFi, * RF,,
* RF,,)
pending whether the product has been already widely used and documented or not. Some authors have suggested giving the "material" safety factor the following value (cf. Segrestin, Boyd & Jailloux, 1999):
where Fo is the reference strength. If Tmis our safety factor for the material strength, the allowable strength for design, Fa, will be given by: ~a
= Fe - r n i (Fe - F s
1
(4)
(3)
Mark I am curious to know what we arrive at in the case of our polyester reinforcements, if we consider the same situation as before: 95% of strength after construction, 95% after hydrolysis and 66% after creep.
where r k varies between 1.35 and 1.05 following experience acquired with the product. Because it already depends on the range of the extrapolations, this factor r,, should however likely apply to all of the losses from construction until the end of the service life, rather than to those produced after te.
Bob
That corresponds to the following values of w i s , RFds and w c s :
Mark I return to residual strength. If we adopt, in a routine case, a working level of the order of 54%, including the load factor, we must sometimes however be able to count on the reserve of strength which the trials brought to light, particularly if this is for a short period and if it does not reach an excessive load level.
RFis = 1/ 0.95 = 1.05 RF,, = 1/ 0.95 = 1.05 RF,, = 1/ 0.66 = 1.52 from which F, = 100%/(1.05 '!: 1.05 * 1.52) = 60% At time b, which corresponds to creep rupture tests at 75% of the reference load, the damage is the same, and there is still no hydrolysis, thus: RFie = 1/ 0.95 = 1.05 RF,, = 1/ 1.0 = 1.00 RF,, =1/0.75=1.33 from which F, = 100%/(1.05*1.00*1.33)= 71.2% With Tn,= 1.5, we arrive at: Fa = 7 1.2 - 1.5(71.2 - 60) = 54.4%
Bob
Certainly, you are right. Imagine a structure where the reinforcements are designed for a permanent tensile force of about 54% of their nominal strength. We have shown that the risks that they will break before the end of a service life of 100 years is extremely low. If, during this period of service, the load level rises accidentally for only a few minutes to say 70%, that will have practically no consequence (Figure 8). In effect we know from the creep rupture tests that such a load must be applied in a constant fashion for more than two years for a rupture to ensue. A period of application of a few moments will thus not appreciably affect the length of serviceof a reinforcement which can be permanently loaded at 54% (but more probably only at about 40%). The benefit arising from this statement can be
Mark Should this value be compared with 63% which we mentioned earlier, when we were wrongly considering residual strength?
Bob
Exactly. And if you want to know how that compares with the "all in" 30% that you talked about at the start, it should be divided by the load factor, which has been estimated as 1.35. We arrive at a little over 40%. You can see we have come some way along the road by looking at things more closely! You should take note that we can certainly do a bit better if we refer to tests on similar materials running over a longer period, i.e. the period of extrapolation is shorter. We could also discuss the safety factor and not necessarily use 1.5. This value could be adjusted de-
Figure 8. Effect of momentary accidental overload.
139
appreciated when considering the effects earthquakes or other exceptional overloads.
of
Mark Does that also apply to unexpected leaps in temperature?
No, because we cannot see any reason why that should happen accidentally, only shortly, once or twice in the life of a structure. It is true that we have not yet talked of temperature, and all the examples that we have quoted suppose that we are at around an ambient temperature of 20°C. Of course, the values of the reduction factors RFd and RF, should be carefully revised when the mean temperature is noticeably higher!
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3 CONCLUSIONS This (fictitious) discussion has mainly demonstrated three things: - the safety factor on the long-term strength of polyester (as for all other materials used for
-
fill reinforcement) should apply to the loss of strength which will occur during the period from which the data is extrapolated an adequate combination of this safety factor and classic reduction factors a priori justifies a small increase in the usually admissible tensile load the so-called "residual" strength only plays a role where accidental overloads of short duration are concerned.
REFERENCES Orsat, P., Khay, M., Mc Creath, M. 1998. Study on creep rupture of jmlyester tendons: full scale tests. Sixth International Conference on Geosynthetics, Atlanta, GA USA. Segrestin, P., Boyd, M.S., Jailloux, J.-M. 1999. Soil reinforcements in MSE structures: a rationale for the determination of long-term admissible tensile loads. TRB 78th annual meeting. January 10-14, 1999, Washington D.C. USA. Orsat, P., Khay, M., Mc Creath, M. 1999. High tenacitypolyester fibre reinforcing strips : study of creep, creep rupture and residual strength. Rencontres GCosynthCtiques 99. October 12-13, 1999, Bordeaux - France.
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Modelling of reinforced unpaved composite material using HISS model K.G. Sharrna & K.K. Gupta Department of Civil Engineering, IIT Delhi, India
Praveen Aggarwal Department of Civil Engineering, REC Kurukshetra, India
ABSTRACT: With the advent of high strength geogrids, the interest of civil engineers in using geosynthetics as reinforcement material in pavement construction has increased. An experimental study is carried out at IIT Delhi to understand the effect of geogrid in unpaved roads. Behavior of composite material, which comprises of Yamuna sand as subgrade and Water Bound Macadam (WBM) as base course is studied with and without geogrid reinforcement under drained conditions. The geogrid manufactured by Netlon India product code CE20 l is used as reinforcing material. Drained triaxial tests were performed at three different confining pressures of 50, 100 and 200 kPa on both unreinforced composite and reinforced composite materials of specimen size lOOmm diameter and 200mm height. Hierarchical single surface (HISS) model developed by Desai and coworkers is used to model the unreinforced and reinforced composite material. A computer program PARA6 is used to calculate the various parameters and to back predict the stress-strain-volume change behavior of unreinforced and reinforced composite materials. The predicted results match closely with the observed results.
dent on many additional factors such as the quantity, type, spacing, interface properties and tensile strength of the reinforcement. Stress-strain relationship of reinforced soil is thus function of these factors in addition to the other factors of unreinforced soil. A number of experimental and analytical studies have been undertaken by researchers to understand the behavior of the reinforced soils. For interface friction behavior, modified direct shear test or pullout test has been used. Stress-strain and strength characteristics have been studied mostly by conducting triaxial tests. Triaxial tests were conducted on cylindrical specimen of reinforced sand using woven fabric glass netting as reinforcement (Yang 1972). It was observed that the axial stress increased with the number of reinforcement layers. Axial strain at failure also inereased due to insertion of the reinforcement. Strength of the reinforced soil was observed to be increasing with the number of reinforcement layers. Constitutive model is a mathematical relation, which reproduces the observed response of a continuous medium. Constitutive models can be broadly classified into following three categories. 1. Empirical models 2. Elasto- plastic models 3. Elasto-viscoplastic models All geological materials show plasticity almost from the beginning. Therefore, stress-strain and volume change response of many geologic media includ-
1 INTRODUCTION Pavement is a structure made in between the wheel and the natural ground. The basic aim of pavement is to provide a hard surface for the movement of wheels without significant deformation and to distribute the wheel load effectively to the larger area of natural ground so that the stresses are within bearing capaity. Temporary or unpaved roads with low volume of traffic are required for construction and access roads, contractors’ haul roads, short-term detour around, for example, bridge replacement construction etc. Further such roads are also frequently constructed world wide to support resource industry viz. forestry, mining, oil and tarsand extraction, agriculture and others. Considering the economic significance of unpaved road, attempts have been made to understand their behavior, so that the benefit of geosynthetic can be quantified. Initially low strength geotextiles were used as a separation layer only thereby maintaining the effective thickness. Reinforcement function of geosynthetics was realized with the emergence of improved material in the form of strong woven geotextile and geogrids. Various factors, which affect the behavior of soils, are density of the soil, the confining pressure, the drainage conditions and the stress path followed. Stress-strain relationships of the soils are important to be studied to analyze load-deformation problems. Reinforced soils have further complex behavior due to insertion of the reinforcement. The behavior is depen141
ing water bound macadam, bituminous conerete can only be predicted by plasticity models. Some of the yield criteria are 1. Mohr-Coulomb Criterion 2. Drucker-Prager Yield Criterion 3. Critical State Model 4. HISS model Two models, the SIGMA-model and TAU-model were postulated using Yang’s results for describing the strength of soil mass reinforced with horizontal layers (Hausmann 1976). Broms (1977) suggested a semi-empirical formula to evaluate the strength of reinforced soil in a triaxial test. He further observed that placement of reinforcement plays an important role in the behavior of reinforced soils. The empirical relation was however very complex. Rao et al. (1987) attempted to model the results with hyperbolic model. All these studies were conducted using conventional triaxial compression stress path. Baykal et al. (1992) conducted triaxial tests using two stress paths and observed that stress-strain behavior of reinforced soil is stress path dependent. Soni (1995) studied the behavior of reinforced soil using six stress paths and modelled the behavior using HISS model.
2 ROLE OF GEOSYNTHETICS IN UNPAVED ROADS The use of geosynthetics in the pavement construction started to serve as a separation layer. An example of use of geosynthetics as separation layer is in the construction of airfield in Switzerland (Hausman 1987). With the development of new stronger geosynthetics reinforcement funetion of this new construction material became apparent. Today geosynthetics are used in the pavement construction primarily to serve the following four functions: I . Separation 2. Filtration 3. Drainage 4. Reinforcement
2.1
Later on Webster & Watkins (1977), Kinney (1979) indicated that the performance of such systems improved with modulus of geotextile and led to the inclusion of membrane effect in theoretical analysis and design procedures proposed by Barenberg (1980). Giroud & Noriray (198 1). Raumann (1982). Haliburton & Barron (1983 ) proposed the restraint action of geotextiles on the aggregate layer. Limited stu dies have been done on reinforced pavement system under repeated loading. From the above literature it is very clear that a geotextile effectively reduces the permanent deformation of the reinforced pavement system. The beneficial effect of the geotextile in controlling the evolution of permanent deformation appears to be due to the following reasons (Barksdala et al. 1982 & Kinney 1982). 1. Geotextile at the interface improves the load spreading capacity of the aggregate consequently the stress-strain state in the subgrade changes leading to less plastic strain with number of repetitions. 2. Lateral restraint effect of geotextile restricts the lateral movement of the subgrade soil away from the directly loaded zone. Kinney (1982) observed that as the thickness of the aggregate increases the beneficial effect decreases and also some lateral extensions beyond the directly loaded area is essential for better performance. On the other hand, Dougles & Kelly (1986) found that anchorage details of geotextile do not effect the behavior of unpaved models. Omoto et al. (1992) observed the cumulative permanent and elastic deformations at the surface under repetitive load tests on aggregate-soil system with and without reinforcement at the interface. From the results it is concluded that under similar test conditions geosynthetics can reduce the elastic deformation on such systems thereby improving the resilient modulus of the system.
3 HISSMODEL In the present study HISS model is used. The brief description of the model is as follows. Desai (1980), Desai et al. (1986) developed hierarchical single surface (HISS) 60 and 61 models. In these models, a unique and continuous yield function is used that leads to the failure when an ultimate condition is reached. The 60 model is based on associative plasticity and isotropic hardening. The constitutive equation for elasto-plasticity can be written as
Behavior of reinforcedjexible pavement
The strengthening effect of geosynthetics in flexible pavement has been studied through plate bearing tests on laboratory models and through triaxial tests under static and repeated loading. Barenberg et al. (1975), Binquet & Lee (1975) stressed on the mode of failure and the bearing capacity of the systems. According to Barenberg et al. (1975) non-woven geotextile inhibits the punching type failure characteristics of soft soil by restraining the soft subgrade and they proposed to use higher bearing capacity factors in reinforced system irrespective of the mechanical properties of geotextiles. Similar observation was made by Steward et al. (1977) in a test road with non-woven geotextile.
where C kq : is the constitutive matrix for elastoplastic approach. 142
The yield function for 60 model is given as
1/27 J 3 D
S,. = Stress ratio = -1.5
(3)
J2D
where: J1 first invariant of stress tensor J ~ second D invariant of deviatoric stress tensor J ~ third D invariant of deviatoric stress tensor Pu atmospheric pressure a,/3, n and y material constants rn -0.5 for geological materials (Desai et al. 1986). For non-associative model 61, plastic potential function Q is defined as a modification of F with a replaced by a ~i.e.,,
I. Elastic constants ( E , v) 2. Ultimate parameters ( y , /3, m ) . 3. Phase change parameter (n). 4. Hardening parameters 01, q 1. 5. Non-associative parameter ( K ) . The procedure for the determination of material parameters has been described in detail in various references (Varadarajan & Desai 1993, Desai 1994). It is briefly presented herein. 1. Elastic constants ( E , v) The two elastic constants for an isotropic material. Young's modulus, E and Poisson~sRatio, U are determined from the average slopes of the initial part of the stress-strain curves and the ratio of lateral and axial strains respectively. Janbu's relation is used to correlate Young's modulus with confining pressure. E = kPu[:lN where k and N are constants.
where:
a~ = a
+ ~ ( a -o a ) ( l - r,)
(51
in which K is non-associative parameter, a0 is a at the beginning of shear loading and r,, = ,$,, is volumetric part of ,$ (plastic strain trajectory).
p,
3.1
Properties of the HISS yield function
Some of the features of HISS model are as follows: 1. The model involves only one continous surface which describe yield or loading surfaces by a single function and also describe the ultimate behaviour. In the model only two parameters y and B at ultimate are used to define the traditional failure. 2. Entire hardening and ultimate behaviour is defined by only one function. 3. The plot of yield function F is continous and convex in the stress space for geological material. However it intersects the J1 axis at right angles, as a result it can be implemented in the context of the classical theory of plasticity. 4. As the intersection of two or more surfaces and corner in I1 plane are avoided, the model is easier to implement in numerical analysis. 5. A single parameter growth function a can simulate hardening and include the effect of stress path, volume change and coupling of shear and volumetric responses. 4
DETERMINATION OF MATERIAL CONSTANTS
The HISS model requires ni ne parameters for the constitutive modelling of any material, which can be classified into five categories.
2. Ultimate parameters ( y , ,L3, m ) For many geological materials m is found to be -0.5 (Desai et al. 1986). Therefore. in the present The procedure work, m is considered as -0.5. adopted for the calculation of y and ,8 from the laboratory results is described below. At the ultimate state, the value of a tends to zero thus, the yield surface degenerates to an open surfaceintersecting J1 axis at infinity. Applying the condition to the yield function, Equation 1, the slope of the ultimate line is derived as (7) where: S,. = 1 for compression path and S, = - 1 for extension path. The ultimate parameters can be found out by conducting least square fitting procedure on Equation 7 for at least two triaxial tests on J1 - & plane. 3. Phase change parameter ( n ) The phase change parameter, n , is calculated using the zero plastic volume change condition, $ = 0. An average of n values for different tests is taken as an overall value of n for the material.
4. Hardening parameters (a1 and 1) In the present study, growth function a is assumed as the function of ,$ as
Taking natural log on both sides of Equation 8 gives, W a ) = Wa1) - r71 In(,$>
(9)
a1 and v l are determined from the least square fitting procedure for each test. The average value of a1 and 143
ql from various tests are taken as overall values of the hardening parameters.
5. Non-associative parameter
Table 1. Characteristics of Yamuna Sand. Prouei-tv % Sand % Silt Specific Gravity (SG) Coefficient of Uniformity C, Coefficient of Curvature CA Maximum Dry Density (yd max.) Minimum Dry Density (yd min,)
(K)
Non-associative parameter, K in the plastic potential formation, Q is assumed as constant and is determined from the conditions near the ultimate. Basic steps in evaluating, K are given below. From the flow rule,
Value 94 6 2.67 2.24 1. 14 16 kN/m’ 13 kN/m2
-
we get
-
Coarse Aggregate(Grade A)
4-
- Screening (GradeB)
r
d&,P=A or
0. I
I
10
I00
Particle Size inm Figure 1. Gradation of Water Bound Macadam
Then from Equation 5,
state to obtain the required gradation for the WBM mix (IRC: 19-1977).
Even though, K could be calculated for any stress point, t he portion of - E , , curve near the ultimate coudition is used since the deviation ( a -~a ) is the greatest in the ultimate zone. The value of K has been calculated using the program PARA6.
Water Bound Macadam (WBM)mix design Water Bound Macadam mix was designed as per (IRC: 19-1977) specification, for possible use as surfacing course. Delhi Silt (P.1 = 6%) was used as a binding material. Required quantity of both the grades A and B (Coarse aggregate and Screening) was mixed with binding material in the ratio of 1.0 : 0.16 : 0.15 by volume in loose state in dry condition. Then Proctor Compaction Test was carried out to find out the Optimum Moisture Content (OMC) and Maximum Dry Density (MDD) for further use. OMC for WBM is 6.8 % and MDD achieved at this moisture content is 22.30 kN/m3.
5 TEST PROGRAM In order to model the unreinforced and reinforced unpaved composite materials, a series of drained triaxial test is carried out on unreinforced and reinforced unpaved composite materials at three contining pressures of 50, 100 and 200 kPa. 5.1 Materials
Subgrade Soil The subgrade soil used in the present study is Yamuna Sand. Yamuna Sand is locally available sand obtained from the bed of river Yamuna. The characteristics of Yamuna sand are summarized in Table 1. Aggregates Two grades of crushed stone aggregates were used to prepare the Water Bound Macadam (WBM) mix. The particle size distribution of the two grades of aggregates is shown in Figure 1. The grade A is used as coarse aggregates and grade B as screening. The two grades of crushed stone aggregates A and B were mixed with binder having grade C (P.I. = 6), in the ratio of 1.0 : 0.16 : 0.15 by volume in loose
Geogrid The geogrid used was extruded mesh CE 201, manufactured by M / s Netlon India Ltd., Vadodara. Aperture size is 7.1 x 7.1 mm. thickness 1.85 mm and at joint is 3.25 mm. The wide width tension tests were carried out as per (ASTM:D-4595-86 1988) to determine the load deformation behavior in machine and cross directions of geogrid. Tensile strength is 7.1 1 and 6.43kN/m in machine and cross directions respectively at 40 mm elongation. 5.2 Testing
Triaxial test Conventional triaxial apparatus (Bishop & Henkel 1962) was used for the triaxial tests. A perspex triaxial cell capable of withstanding more than 1 MPa 144
and with the facility of 100 mm diameter and 200 mm height was used. Thickness of subgrade (Yamuna sand) and base course layer (WBM) is lOOmm each in both the cases. Axial strains, deviator stresses and volumetric strains were observed during the tests. It is observed that mode of failure of unreinforced composite material is by bulging of the subgrade layer, hence reinforcement is provided at the center of the subgrade layer in reinforced composite material specimen. The experimentally observed behavior is presented in Figures 2-5.
response. The equation is integrated starting from the initial hydrostatic state. The prediction is made using the nine parameters calculated for unreinforced and reinforced composite material under strain control condition. Both predicted and experimentally observed variation of octahedral stress and volumetric strain with axial strain are presented in Figures 2-3 for unreinforced case and in Figures 4-5 for reinforced case. The observed and predicted behavior matches closely. Table 2. The material parameters for unreinforced and reinforced composite materials.
6 MODELLING Constant k N
All the nine constants for HISS model are calculated for both unreinforced and reinforced cases and are presented in Table 2.
Y
B in n
7
PREDICTION
a1 rll
The incremental constitutive relation (Equation 1) has been used to predict the stress-strain-volume change
K
145
Unreinforced 141.97 1.1280 0.0798 0.74 -0.5 2.655
0.001150 0.420 0.15
Reinforced 171.58 1.0996 0.0840 0.74
-0.5 2.820 0.0002 I5 0.686 0.22
8 CONCLUSIONS From the results it is observed that inclusion of geogrid improves the performance of composite material by more than 15%. Dilation starts at a higher axial strain in case of reinforced composite material as compared to unreinforced composite material. Modelled behavior matches closely with the observed results.
REFERENCES ASTM: D4595-86 1988. Test method for tensile properties of geotextiles by wide width strip method. ASTM standard of geotextiles. 63-66.ASTM, Philadelphia, PA, USA. Barenberg, E. J. 1980. Design procedures for soil-fabric aggregate systems with mirafi 500x fabrics University of Illinois at Urbana: Champaign. Barenberg, E. J., Dowland, J. H. Jr. & Hales, J.11 1975. Evaluation of soil-aggregate systems with mirafi fabric. Department of Civil Engineering. University of Illinois. Barksdale, R., Robnett, Q., Lai, J. & Zeevaert-Wolff, A. 1982. Experimental and theoretical behavior of geotextile reinforced aggregate soil systems. Proc. 2nd. h t . Conf oiz Geotextiles, Las Vegas, Vol. 2, 375-380, USA. Baykal, G., Guler, E. & Akkol, 0. 1992. Comparison of woven and non-woven geotextile reinforcement using stress path tests. Proc. Int. syn. on Earth Reinforcement, Kyushu, 23-27, Japan. Binquet , J. & Lee, K. L. 1975. Bearing capacity tests on reinforced earth slabs. Proc. ASCE Jra. of Geotechnical Eiagineering Div., Vol. 101, No. GT 12, 1241-1255. Bishop, A. W. & Henkel, D. J. 1962. The measurement of soil properties in the triaxial test. Edward Arnold Publishers Itd., London. Broms, B. B. 1977. Triaxial tests with fabric reinforced soil. Proc. Int. Conf: on the use of fabric in geotechnics, Paris, 129- 133. Desai, C. S. 1980. A general basis for yield, failure and potential functions in plasticity. Int. Journ. Nurn. Anal. Meth. Ceornech., 15(9), 649-680. Desai, C. S., Somasundaram, S. & Frantziskonis, G. 1986. A hierarchical approach for constitutive modelling of geologic materials. Int. J. Nuin. and Analytical Methods in Geomech., 10,225-257. Desai, C. S. 1994. Hierarchical single surface and the disturbed state constitutive models with emphasis on geotechnical applications. Geotech. Eng. Emerging Trends in Design and
Practice, Chap. 5 , K. R. Saxena (Editor), New Delhi, India. Oxford & IBH Pub. Co. P t. Ltd. Douglas, R. A. & Kelly, M. A. 1986. Geotextile reinforced unpaved logging roads: the effect of anchorage. Geotextiles and Geomernhranes, Vol. 4, 93-106. Giroud, J. P. & Noiray, L. 198 1. Geotextile reinforced unpaved road design. Proc. ASCE Jn. Geotechnical Engineering Div. Vol. 107, NO. GT9, 1233-1254. Haliburton, T. A. & Barron, J. V. 1983. Optimum depth method for design of fabric reinforced unsurfaced roads. Presented at the Annual Meeting, Transportation Research Record Board. Hausmann, M. R. 1987. Geotextiles for unpaved roads - A review of design procedures. Geolextiles and Geomembranes, Vol. 5 , 201-223. Hausmann, M. R. 1976. Strength of reinforced soil. Proc. 8th Australian Road Board Con$, 8(13): 1-8. IRC: 19-1977 1982. Standard specification and code of practice for water bound macadam. The Indian Road Cotzgress, New Delhi, India. Kinney, T. S. 1979. Fabric induced changes in high defonnation soil-fabric-aggregate systems. Ph.D. Thesis. Graduate College. University of Illinois, Urbana, USA. Kinney, T. C. 1982. Small scale load test on a soilgeotextileaggregate system. Proc. 2nd Int. Coi$ on Geotextiles, Vol. 2,405-409, Las Vegas, USA. Omoto, S., Kawabata, K. & Mizobuchi, M. 199?. Reinforcement effect of geotextiles on pavement with weak subgrade. Earth Reinforcement Practice: 67 1-676. Eds. Ochiai, Hayashi and Otani. Rotterdam: Balkema. Rao, G. V., Gupta, K. K. & Kachhawah, R. 1987. Triaxial behavior of geotextile reinforced sand. Proc. Indian Geot. Con$ Bangalore, 1 : 323-328, India. Raumann, G. 1982. Geotextiles in unpaved roads: desi-n considerations. Proc. 2nd Int. Con$ on Geotextiles, Vol. 2. 417-422. Las Vegas, USA. Soni, K. M. 1995. Constitutive modelling of reinforced soil. Ph.D. Thesis, Indian Institute of Technology, Delhi, India. Steward, J., Williamson, R. & Moheny, J. 1977. Guidelines for use of fabric in construction of low-volume roads. USDA. Forest Service, Portland. Oregon. Varadarajan, A. & Desai, C. S. 1993. Material constants of a constitutive model determination and use. Indian Geotech. J., 23(3), 291-313. Webester, S. L. & Watkins, E. J. 1977. Investigation of construction techniques for tactical bridge approach roads across soft ground. Technical Report 5-77-1, US Army Engineers, Waterways Experiment Station, Vicksburg, Miss. Yang, Z. 1972. Strength and deformation characteristics of reinforced sand. Ph.D thesis, University of California, Los Angeles, USA.
146
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lS5N 90 2651 863 3
Construction and monitoring of geotubes E.C. Shin & Y.I. Oh University of Inchon, South Korea
B.M. Das California State University, Sacramento, USA
E.S. Lee E & S Engineering Company, Seoul, South Korea
ABSTRACT: Pilot scale field tests were conducted to evaluate the performance of a geotube. The geotube was made of a woven geotextile which conformed to the minimum specifications provided by the U.S. Army Corps of Engineers. The results of a number of laboratory tests are presented: direct shear tests to determine the geotextile-sand interface friction angle, cyclic fluctuation tests to estimate the soil particle loss from the geotube, and long- and short-term environmental tests to evaluate the environmental effects of dissipated water from the geotube. Based on the present study it appears that the geotube is a fast, efficient, and environmentally sound dredging fill technique. 1 INTRODUCTION Geotubes are made of permeable, soil-tight geotextile. They are hydraulically filled with dredged soil. Attempts are now being made to use geotubes in coastal engineering projects such as shore protection and breakwaters. Geotubes also help store and isolate contaminated materials obtained from dredging. The diameter and length of the geotubes vary, depending on the field conditions. The typical length and width are 150-180 m and 4-5 m, respectively. Initial studies regarding geotextile containers are found in the work of Koerner and Welsh (1980). Botzan et al. (1982) and Harris (1987, 1989, 1994) reported the use of geotextile containers in erosion control. Bogossian et al. (1982) and deBruin and Loos (1995) evaluated the effectiveness of geotubes for erosion control. Environmental dredging and backfill technology using geotubes was reported by Fowler et al. (1994) and Pilarczyk (1996). In most cases, a single layer of woven geotextile is used to construct the geotube. According to the U.S. Army Corps of Engineers, the minimum physical properties of woven geotextile to be used for constructing geotubes should be as follows: Tensile strength: 175 kN/m Elongation: 10% Trapezoidal tearing strength: 140-160 kN/m Seaming strength: 105 kN This paper summarizes the performance of a pilot scale field test using a geotube in Inchon, South Korea.
sand is a standard poorly graded silica sand used in Korea. The dredged sand was obtained from the Songdo Land Reclamation area in South Korea. The dredged organic soil was collected from the detention basin located on the west coast of Inchon, South Korea. The detention basin is used for temporarily holding the water before discharging it to the sea. The physical properties of these soils are given in Tables 1 and 2, and the grain-size distributions for the three soils are given in Figure 1. Table 1. Physical properties of the sand.
Item Effective size, Dl0 (mm) Uniformity coefficient, C, Coefficientof gradation, C,, Maximum dry unit weight, l'd(rnax) ( m / m 3 ) Optimum moisture content, bVop1
("/.I
Specific gravity, G, Unified soil classification
Quantity Jumoon Jin Dredged sand sand 0.37 0.09 1S 3 4.67 1.10 1.06 16.09 15.3
15.2
16.2
2.65 SP
2.65 SP
Table 2. Physical properties of the organic soil. Item Specific gravity, G,y Liquid limit, LL (%) Plastic limit, PL (%) Plasticity index, PI Passing 0.075 mm sieve (%) Organ; content (%) Unified soil classification system
2 PROPERTIES OF SOIL AND GEOTEXTILE The soils used for this study were Jumoon Jin sand, dredged sand, and dredged organic soil. Jumoon Jin 147
Quantity 2.29 39.00 30.25 8.75 21.94 15.33 OL
The geotextile generally used to construct geotubes are either woven geotextile or composite geotextile (i.e., an external layer of woven geotextile and an internal layer of non-woven geotextile). For the present study two woven geotextile, designated in this paper as K-1 and K-2, were used. The physical properties of these two geotextile are given in Table 3. 3 LABORATORY TESTS Before carrying out the pilot test in the field, several laboratory tests were conducted to determine the compactibility of the soils and geotextile. These tests are briefly described in the following sections. 3.1 Large scale direct shear tests Large scale direct shear tests were conducted to determine the interface friction angle between the geotextile and the two types of sand described in Table 1. The interface friction angle is an important param-eter in determining the stability of geotubes when they are placed on sloping ground for shore protection and projects such as the construction of breakwaters. Figure 2 is a schematic diagram of the direct shear test box used for the tests. The size of the geotextile used for the tests was 0.3 m x 0.3 m.
Tests were conducted with normal stress varying up to 700 kN/m2 (ASTM D-5321 test method). The sands were compacted to 90% of maximum dry unit weight [yd(max)] as given in Table 1. The interface friction angles thus determined are summarized in Table 4. It should be noted that the tests showed apparent cohesion (c, ) in the range of 36-38 kN/m2. The interface friction angles reported in Table 4 are fairly large and are workable in the stability analysis of geotubes in the field. 3.2 Cyclic fluctuation tests One important function of a geotube is its capability to hold the backfill soil inside the tube with minimum loss. A quantitative evaluation of this factor can be done using a cyclic fluctuation device. The device used for the present test had a water basin measuring 1.3 m (length) x 1.3 m (width) x 1.7 m (height). The geobag holding box inside the waterbasin was made of steel wire and fixed to a vertical pole which moved up and down at a desired rate. To conduct the tests, dredged sand (see Table 1) was placed in the geobags up to 80% of its total volume. The geobags were subjected to 10,000 cycles of vertical fluctuation in water at the rate of 70 cycledmin. At the end of the test, the loss of soil particles was determined by measuring the grain-size distribution. The total loss of soil for the two geotextiles under consideration is shown in Table 5. The soil particle loss rate for various grain sizes is shown in Fig. 3, which shows that silt size particles have the highest loss rate.
Table 3. Properties of geotextile used for the present study. Unit
Property
Test Method (ASTM)
Mass per unit area Tensile strength Elongation Coefficient of permeability Material
D-526 1 D-4632, D-4595 D-4632, D-4595 D-499 1
g/mL kN/m
__
__
%
cdsec
148
Geotextile K- 1 600 196 10-50 10.' - 1 0 - ~ Polyester (PET)
K-2 700 245 10-50 1 0 - ~- 1 0 . ~ Polyester (PET)
Table 4. Direct shear test results. Interface friction angle " (dep) Dredged sand Jumoon Jin sand 33.6 37.2 34.8 38.3 \
Geotextile K- 1 K-2
"1
Table 5. Cyclic fluctuation test with dredged sand. Geotextile K- 1 K-2
Soil lost (%) 5.05 6.09
ss Geotextile a I\
Time (min) Figure 4. Variation of SS and COD.
-11
I0
I
I
1 0.1 Grain size (mm)
I
I
L
I
I
I
0.0 1
Figure 3 . Cyclic fluctuation test results.
3.3 Environmental tests Two types of environmental tests were conductedshort-term self-weight filtration tests and long-term diffusion tests. The short-term self-weight filtration test was performed in the field at the water detention basin in Inchon. Since the K-1 and K-2 geotextile essentially gave similar results, it was decided to conduct further tests only with the K-1 geotextile. Thus for the environmental tests the geo-textile tube was made from the K-1 geotextile. The geotube was 2 m long and 1 m wide, and it was filled with 3.5 m3 of dredged organic soil. Dissipated water samples of 1000 ml each were collected from the geotube at time intervals of 0, 0.33, 0.5, 1, 2, and 3 hours. These water samples were used to determine the suspended solids (SS) and the chemical oxygen demand (COD). The variations of SS and COD with time are shown in Fig. 4. From this figure it may be seen that the magnitudes of SS and COD rapidly decreased within 20 minutes after filling the dredged organic soil in the geotextile tube. After the first 20 minutes, the rate of decrease of SS and COD with time dropped rapidly. The results of SS= 110 ppm and COD = 45 ppm at time = 20 minutes after the beginning of the tests are below the Korean EPA standards of SS = 120 ppm and COD= 130 ppm.
I
0
1000
2000 3000 Time (hour)
1
4000
Figure 5. Variation of SS and COD with time - diffusion test.
A long-term diffusion test, which lasted for six months, was also performed on a geotube filled with dredged organic soil obtained from the detention basin. The geotube had a circumference of 250 mm and a length of 500 mm. The variations of SS and COD with time obtained from this test are shown in Fig. 5. The magnitudes of SS and COD increased with time and reached maximum values in about 100 days. These maximum values of SS and COD are substantially lower than those specified by the Korean EPA. 4 FIELD PILOT TEST A field pilot test was conducted in the dike construction work of the Songdo land reclamation project in the Bay of Inchon. A single layer of K-1 geotextile (Table 3) was used for fabricating the geotube. The geotube had a circumference of 8 m and a length 149
Figure 6. Schematic diagram of the field pilot test.
of 25 m and it was filled by pumping in dredged sand from the site. The soil-water mixing ratio, pumping pressure, and pumping speed were varied to determine their optimum values. For this case, a soil-water mixing ratio of 4:6 was found to be more workable. The unit weight of the filled soil, elongation of geotextile, and vertical pressure were measured for about 3 months. The schematic diagram of the geotube field pilot test is shown in Fig. 6 along with the placement of the pressure cells. The pressure cells were installed at 4-m intervals right below the non-woven geotextile layer. The non-woven geotex-ile was used to prevent the erosion of soil around the geotube due to dissipation of water from the geotube. The distance between the inlet of the dredged soil and the outlet of overflow was about 19 m. The construction sequence of the geotube was as follows: (1) preparation of subgrade (foundation soil), (2) installation of pressure cell, (3) covering the area of subgrade with non-woven geotextile, (4) placing the geotube over the geotextile, ( 5 ) mixing and pumping the slurry into the geotube; and (6) completing the slurry injection into the geotube. Starting with the beginning of slurry pumping, the effective height, the unit weight of soil, and vertical pressure were monitored by the pressure cells at various time intervals up to 3 months. The variation of the vertical pressure with time is shown in Fig. 7. This figure indicates that the vertical pressure increased up to 100 minutes and then decreased. The pressure stabilized after about 2 days. The observation of pressure with the pressure cells showed that the magnitude of vertical pressure is greater near the inlet compared to that near the outlet. The effective height of the geotube and the unit weight of soil that was monitored with time are plotted in Fig. 8. The effective height of the geotube reached 1.2 m at a time of 100 minutes and stabilized at 0.7 m. This trend is similar to that observed from the pressure cells; however, the unit weight of the pumped soil in the geotube increased continuously with time due to
Figure 8. Variation of effective height of geotube and unit weight of soil with time.
150
4. The optimum ratio of the water-soil mixture to be pumped into the geotube is about 6:4. 5. To avoid rupture, the soil-water slurry pumped
the dissipation of water from the tube. The effective height of 1.2 m reached during pumping is about 80% of the maximum possible height of the geotube. During the test it appeared that any further increase in effective height might cause failure by rupture.
into a geotube should not exceed that required to reach 80% of its effective height. REFERENCES
5 CONCLUSIONS The results of a pilot scale field test to determine the functionality and performance of a geotube are presented. A number of laboratory tests were conducted, such as direct shear test to determine the geotextile sand interface friction angle, cyclic fluctuation tests to estimate the soil particle loss from the geotube, and long- and short-term environmental tests to evalu-ate the environmental effects of dissipated water from the geotube, and their results are presented herein. Based on the various laboratory tests and the field pilot test reported, the following conclusions can be drawn:
A geotextile for fabricating geotubes should have a minimum permeability of 104c d s e c . This will keep the particle loss ratio to less than 10%, which is desirable. For the short-term self-weight filtration test, the magnitudes of SS and COD rapidly decreased within the first 20 minutes, after which the rate of decrease of SS and COD dropped. The environmental test results indicate that the quality of dissipated water from the geotube are far below the minimum Korean EPA standards.
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Bogossian, F., R.T. Smith, J. C. Vertimatti & 0. Yazbek 1982. Continuous retaining dikes by means of geotextile. Proc. I1 Itit.Cot$ Geotextiles, L m Vegas:2 1 1-2 16. Botzan, D., L. Kellner & C. Moisa 1982. Construction elements for river bank defense structures using woven geotextiles. Proc. II Int. Cot$ Geotextiles, Las Vegas:223-227. deBruin, P. & C. Loos 1995. The use of geotubes as an essential part of an 8.8-m-high North Sea dike and embankment. Geosyizthetics world,AprillMay:7- 10. Fowler, J., C. J. Sprague & D. Toups 1994. Dredged inaterialfilled geote.de containers, U S . Army Corps of Engineers, Environmental Effects of Dredging Technical Note, EEDP05-01. Harris, L.E. 1987. Evaluation of sand-filled containers for beach erosion control, an update of the technology. Proc., Coastal Zone ‘87, ASCE:2479-2487. Harris, L.E. 1989. Developments in sand-filled container systems for coastal erosion in Florida. Proc., Coastal Zone ‘89, ASCE:2225-2233. Harris, L.E. 1994. Dredged material used in sand-filled containers for scour and erosion control. Proc, Dredging ‘94, ASCE1537-546. Koerner, R.M. & J.P. Welch 1980. Construction and Geotechnical Engineering Using Synthetic Fabrics, Wiley, New York: 160-229. Pilarczyk, K. W. 1995. Geosystems in hydraulic and coastal engineering-an overview. Proc, 1‘‘ European Geosynthetics Conf (EuroGeol), Maastricht, The Netherlands.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Modelling and instrumentation considerations of a geogrid B.V.S. Viswanadham Indian Institute of Technology, Bombay, India
ABSTRACT: In this paper, similitude conditions concerning modeling of geosynthetic materials in I g and Ng conditions are discussed with an emphasis on geogrids. Secondly, instrumentation and calibration aspects of model geogrid with special type of tiny strain gauges for measuring tensile strength are examined. The geogrid layer was used as a reinforcement inclusion within the model clay liner subjected to non-uniform settlements in geotechnical centrifuge. The measured tensile strength and observations made during the centrifuge test were reported.
1 INTRODUCTION A wide range of geotechnical problems can be investigated using physical modelling techniques and the evaluation of the behaviour of soil structures with geosynthetic inclusions is no exception. However, with the model tests at normal gravity the behaviour of soil mass could not be simulated properly because of the dominance of self-weight forces in geotechnical engineering. In order to replicate the gravity induced stresses of a prototype structure in a I/N reduced model, it is necessary to test the model in a gravitational field N times larger than that of prototype structure (Schofield, 1980). Geosynthetic materials, such as geotextiles and geogrid reinforcement elements have been successfully used in soil structures such as walls, slopes, embankments on soft soil and most recently in landfill liners. The reinforcement elements are mainly planar layers like geotextiles or uniaxial geogrids in the case of slopes and walls or biaxial geogrids for embankments on soft ground, pavements on weak subgrade, and landfill liners. Design theories have been developed alongside an increasing database of prototype, smallscale studies under I g and Ng conditions. A major difficulty encountered in model studies involving geosynthetic materials like geogrid is selection, modelling and instrumentation of ideal materials. In this paper, the modelling considerations of geosynthetic materials are presented for I g and Ng tests. The consequent effects which may arise for not using scaled-down geosynthetic materials, especially under I g conditions and necessity of standardization of miniature version of geosynthetic materials representing the bandwidth of commercially available prototype geosynthetics is brought out. Further, instrumentation and calibration aspects of model
geogrid with strain gauges for measuring tensile strength were discussed. The measured tensile strength and observations made during the centrifuge test were presented.
2 MODELLING OF GEOSYNTHETIC MATERIALS In physical modelling (both I g and Ng), it is assumed that the average particle size of soil used in the model and prototype structure is identical. In the case of studies pertaining to reinforced soil structures, contrary to soils the similitude condition does not allow the use of identical products in model and prototype studies. The scaling down of geosynthetic materials is essential in a small-scale physical modelling studies (both I g and Ng studies) in order to infer the correct response of prototype soil structure. Many investigators have attempted to model the geosynthetic reinforcement for understanding the behaviour of reinforced soil structures either at normal gravity conditions or at accelerated gravity conditions with sandclay as interface. The characteristics of different modelled geosynthetic materials by various investigators for both I g and Ng studies are summarised in Table 1. The range of selected model materials includes weakened medical gauge to miniature version of commercially available prototype geosynthetic materials. As can be seen from Table 1 , very few have attempted to consider modelling of geosynthetic materials in I g model tests. The reason could be attributed to the initial interest in understanding the behaviour of geosynthetic reinforced soil structures qualitatively. Geosynthetic materials to be scaled-down are primarily of two types (i) geotextile and (ii) geogrid. In the case of geotex-
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tile, typically two requirements were seen to be fulfilled (a) simulation of tensile strength-strain behaviour of model material and (b) thickness of model material [for e.g. Ovesen (1984), Taniguchi et al. (1988), Porbaha and Goodings (1994), and Zornberg et al. (1997)l. The model geotextile material was selected such a way that it is as thin as possible and exhibits low tensile strength for the defined strain limit. While modelling geotextile either for I g or Ng studies, many investigators have also considered mass per unit area, With this, it should be possible to ensure the model geotextile material with low tensile strength and thickness. In the second type, at least ten to fifteen varieties of synthetic (many having different styles) geogrids are available commercially. They differ considerably in geometry and mechanical properties. The manufacturers attempt to vary the typical geometrical characteristics and tensile strength characteristics (as manufacturing process variable) in order to achieve a desired geogrid. As tensile strength characteristic of synthetic material is mainly dependent on the composition and type of raw material, it has become one of the manufacturing process variables. In addition, depending upon the type of geosynthetic reinforced soil structure being considered there are two more types, namely (i) uniaxial and (ii) biaxial geogrids. This complicates further to scale geogrids to the desired scale factor altogether. The properties of these geogrids are specified based on (i) rib cross-sectional area, (ii) grid opening size, (iii) tensile strength, and (iv) type of material composition (Fig. 1). Mitchell et al. (1988), and Springman et al. (1992) have modelled geogrids for Ng tests. While studying the deformation behaviour of reinforced clay liners subjected to non-uniform settlements in geotechnical centrifuge, Viswanadham (1996) have dealt in detail about the modelling aspects of geogrid. Perfect scaling-down of the prototype geogrid to the desired scale factor may not be feasible. However, the ideal geogrid for model studies needs to be selected by considering suitable model geogrid representing the bandwidth of prototype geogrid characteristics. For tensile reinforcement (like geogrid) to be functional, not only the tensile strength-strain behaviour but also the reproduction of identical frictional bond behaviour between model and prototype must be achieved. By scaling down the prototype geogrid opening sizes to the desired scale factor in the model studies, there is a danger of loosing interlock between soil and geogrid (Fig. 1). In order to overcome this difficulty, it is required to consider modeling of frictional bond behaviour by scaling down the ratio of rib cross-sectional area and grid opening size (measured from centre to centre of ribs) by factor N (i.e., for example in longitudinal direction, [brtd(ar+br)],n = [W(ai+bi)lp”; A’,,, = A’pIN) and by
~ m g Dlrocfion C . I
b. Tensile strength T, [kN/ml
m.( r $ p
Figure 1. Idealised model geogrid characteristics.
maintaining identical percentage open areas of model and prototype geogrids o”;, =&,). Whereas, f = [(alai)/(al+bl)(n,+br>l. Assuming that the tensile stresses in the geogrid with identical material characteristics in model and prototype (i.e. o, = CJ,, = 0; suffix: m: model; p : prototype), the constitutive law of the geogrid o = EE (where, E = modulus of elasticity of geogrid material), and identical geogrid = (q& = E] are strains in model and prototype [(q?),?! valid, the scale factor for tensile strength of model geogrid Tg can be deduced for Ng modelling as follows: Considering validity of Tg = o A’ and using the above,
Correspondingly, the secant stiffness of the model geogrid shall be 1/N times the prototype geogrid stiffness (by using the validity of Tg= Jg&).Based on the above mentioned modelling considerations, four model geogrids have been selected and their characteristics are summarised in Table 1 (see under No. 8). By considering above scaling laws for I g studies, the tensile strength and secant stiffness of the model geogrid shall be 1/N2times prototype geogrid tensile strength and secant stiffness respectively. Although, the scale factor will be low for I g studies, it is required to consider scaled-down geogrid. As it otherwise ends up in over estimating the effect of reinforcement in interpretation of test results. However, small-scale physical modelling of geosynthetic soil structures tested at normal gravity has been used in the past to provide the insight into failure mechanisms qualitatively and very few have attempted to model the geogrid [for e.g. Love et al. (1987), Moghaddas-Nejad and Small (1997) and Pinto and
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Cousens (1999)l. Love et al. (1987) has reported about the use of miniature version of geogrid for the first time as reinforcement inclusion in their I g model studies on granular fills on a soft clay subgrade. Similarly, Moghaddas-Nejad and Small (1997) have reported about their model track tests on the effect of geogrid reinforcement of pavements ( N = 4). Further, Maghaddas-Nejad and Small (1997) have opined that the geogrid product used had a relatively small opening size and a small rib size (Table 1) and that in the prototype it would correspond to reinforcement with a larger grid size and a higher stiffness. However, as shown in Table 1, at 5 % strain the selected model geogrid is found to have a secant stiffness of 280 kN/m in longitudinal direction. By considering the scaling laws, it should have been 17.5 kN/m (i.e. 280/4’). This suggests the essentiality of scaling-down of geosynthetics even for I g model studies. 3 INSTRUMENTATION OF GEOGRIDS
Very few have reported on the use of strain gauges in centrifuge model testing glued to synthetic materials, such as, geogrids, soft fabrics. The use of high performance Cu-Ni strain gauges were earlier reported by Springman et al (1992) to measure the load and strain independently in the centrifuge tests.
In the present study, a model geogrid layer was used as a reinforcement layer within the model kaolin clay liner subjected to non-uniform settlements using trap-door arrangement in centrifuge. As the bending stiffness of the reinforcement layer is negligible, it is assumed that the reinforcement layer follows the deformation profile attained by clay liner. At the onset of non-uniform settlements, the model geogrid layer is subjected to strain due to (i) change in length and (ii) curvature. The details concerning model test package are discussed in detail by Viswanadham (1996). In this paper, the results and observations made on the deformation behaviour of kaolin clay liner reinforced with instrumented model geogrid are reported. The model geogrid MGGl was selected for instrumentation and its characteristics can be seen from Table 1 (No. 8). A special gauge type GFLA-3-70 (with dimensions 3 mm in length, 2.3 mm in width and with a backing of 9.5 x 4 mm) supplied by M/s Europavia Gmbh, Germany was used. The procured strain gauges are having a nominal resistance of 120 s2, gauge factor [k-factor] of 2.13 and with strain limit of 3%. The calibration of strain gauges fixed to the model geogrids is found to be complex. This is because of the non-linear behaviour of the model geogrid. It is interesting here to note that in addition to the non-linear behaviour of the geogrid material at large strains, the stress-strain
Table 1. Summary of characteristics of different modelled geosynthetic materials. No.
Model, Material type & Composition
N
Material characteristics a1
lmml
a, lmml
bl lmml
bt [mml
t [mml
f
[%I
#(Tg)u
(%)It
1%l 1 a I Ng Geotextile -Cotton 50 __a __a --a 4.8 il a NE Geotextile -70% Nylon+ 30% 50- --a --a 0.217 --‘I 38-29 PET 55 d 1 3. Ng Geotextile 40 ---I --a --* --a 10 1 0.15 0.15 0.15 85 18 Geogrid-Plastic 40 1 0.7 --‘I 68 16 40 3.3 3.4 0.7 4. N g Geogrid -PET 0.4 --‘l 88 23 40 5.7 6.3 0.4 Geogrid-PP Geoirid-PP 40 5.5 7.1 0.2 0.2 --‘l 94 12 il 5. I g Geotextile-PET 10 --a --a _-a __a 79 40 --a a --a __‘l 0.0006 --I 20 6. NE Geotextile 60% PET+40 % Rayon 0.9 82 14 at 5% strain 28 3 3 4 40 I g Geogrid-PP 7. 0.16 71 13.2 27 3.7 0.46 0.91 50 3.4 8. Ng Geogrid -PET (MGG1) Geogrid-FG+PVC(C) 50 1.4 1.3 0.32 0.28 0.13 65 9.5 11 Geogrid-PP 50 3.4 3.8 0.1 0.11 0.02 89 1.2 30 Geogrid-PP 50 5 5 0.16 0.25 0.16 93 2 55 d 1 3 d 100 --“ 0.063 18 9. N g Geotextile-100% PET 1 --d d --* 0.119 29 Geotextile 100 --* 60% PET+40%Rayon 1 a 10. 1g Geotextile-PET 5 --a --a 0.3 --‘ 2.7-3.5 15 1) Ovesen (1984), 2) Taniguchi et al. (1 988), 3) Mitchell et al. (1988), 4) Springman et al. (1992), 5) Palmeria & Cunha (1 993), 6) Porbaha & Goodings (1994), 7) Moghaddas-Nejad and Small (1996), 8) Viswanadham (1996), 9) Zornberg et al. (1997), and 10) ) Pinto and Cousens (1999). PP = Polypropylene; PET = Polyester; FG = Fiber Glass; PVC (C) = PVC coated; FRP = Fiber Reinforced Plastic; Data not available/Not applicable; # Longitudinal direction only.
1. 2.
rl
--It
155
IkN/ml 0.2 13.213.6 0.066 0.53 18 3 1 55 0.18
relationship of the material subjected to loading and un-loading cycles is also significant. The strain gauges are to be fixed to the uniform hardened basematerial. The addition of the base material to the geogrid will influence the stiffness of the geogrid slightly. As a first step, the influence of uniformly hardened epoxy adhesive on the load-strain behaviour of the model geogrid was studied. Figure 2 presents the tensile strength - strain behaviour of the model geogrid with and without epoxy base strips. Three no's of epoxy base strips (of width approx. 18 mm and 2 mm in thickness) were spaced equally within the gauge length. It can be seen from the presented results that the presence of epoxy base strips has little influence at lower strain levels. However a marked influence of epoxy base strips (more than one) was noticed at higher strain levels (i.e. > 4 %). It is necessary to carry out calibration tests in order to study the strain gauge response to the loading and reloading cycles of the geogrid. So it is required to calibrate every new model geogrid with strain gauges meant for the centrifuge tests. It is necessary to investigate the strain gauge response to the defined loading cycles in order to calibrate the model geogrid. This enables to arrive at calibration factors, which can be used to analyse the strain behaviour of the model geogrid in centrifuge test results. As shown in Fig. 3, the calibration tests were carried out with strain gauges glued on to the model geogrid MGG1. The selected model geogrid was cut to desired size such a way that it is convenient for calibration and subsequent use in centrifuge model test. The epoxy base material was spreaded uniformly by hand over = 6 apertures wide (18 mm). After spreading the half portion, a very thin Teflon foil strips were used to reinforce the matrix in the tensile strain direction and the Teflon foil strips are overlaid again with epoxy matrix and left for curing at room temperature for about 12 hours (Fig. 4). To prevent the flow of epoxy base material be
Figure 2. Tensile strength vs. strain behaviour of MGGl with and without epoxy strips.
Figure 3. Instrumented model geogrid.
Figure 4. Details of strain gauge on model geogrid.
fore curing, a temporary restraint was provided. The edges of epoxy strips were smoothened (Fig. 4b) to prevent any influence on the soil-geogrid frictional interaction. After smoothening the surface, the strain gauges (in pairs) were glued centrally to the upper and lower surfaces of the epoxy strips using a Cyanoacrylate adhesive and the electrical connections were given in such a way that the bending effects are eliminated. Each strain gauge is supplemented by the dummy strain gauges (fixed on both the upper and lower surfaces of the Perspex glass piece), which are located in the vicinity of the strain gauges fixed on to the geogrid. The dummy gauges are to nullify the effects due to temperature. The prepared model geogrid sample is fitted within the grips of the specially fabricated roller grips. With this arrangement, it is possible to prevent any damage to the model geogrid. The model geogrid was calibrated through a load-controlled approach and subjected to five loading and unloading cycles. During the calibration, the model geogird was subjected to a maximum loading of 1.25 kN/m (i.e. 14 % of ultimate tensile strength of model geogrid with three epoxy strips spaced at 100 mm c/c) in small increments. In each loading step, a waiting period of 10 minutes approximately was maintained to observe the influence of time. The response of strain gauges is observed to stabilise after completion of two load
156
Figure 5. Calibration curves for strain gauge on the model geogrid.
ing-unloading cycles. This could be attributed to the non-linear characteristic of model geogrid material. Hereafter, the response of strain gauges for the last three cycles was considered for the analysis. Typically, Strain gauge response (SGl) versus tensile strength is shown in Fig. 5 for the last three loading-unloading cycles. The reproducibility of identical response can be noticed. The initial curvature during each loading cycle suggests the nonlinear variation of the model geogrid material. This is observed to be predominant at lower tensile strength levels and once after subjected to adequate loading variation of strain gauge response is found to be linear. Identical behaviour is found to observe for other strain gauges also, except with different slopes of tensile strength-strain gauge response curves. Considering the above aspects, for interpreting the geogrid tensile strength through centrifuge test, the calibration factors obtained within the linear range are used. Based on the calibration tests carried-out with the defined strain levels and limited studies made, the measuring accuracy of the strain gauge response with respect of tensile load is of the order of t- 0.1-0.15 kN/m. However, by considering the reproducibility of calibration tests in this study for further investigations the strain gauges can be used to measure in-plane tensile strength characteristics of the embedded geogrid. It is required to calibrate the model geogrid by loading and unloading cycles till the reproducible response is achieved.
4 RESULTS AND DISCUSSION Figure 6 presents a portion of deformed 25 mm thick model clay liner at the end of the centrifuge test. The reinforced clay liner was subjected to a maximum central settlement a of 25 rnm in steps of 5 mm with a settlement rate of 1 mm/min. Before initiation of non-uniform settlements in flight, the clay liner is a1 lowed to consolidate for about 4-5 hours. Between
Figure 6. Deformed model geogrid at the end of centrifuge test.
Figure 7. Measured variation of tensile strength in model geogrid with time [Model dimensions].
each settlement increment of 5 mm a waiting period of 30 minutes is maintained. The schematic deformation profile of model geogrid is shown in Fig. 6b with an air gap in the central region. The formation of air gap is observed to take place after attaining a central settlement of 15-20 mm. This is attributed to participation of tension membrane effect at large settlement differences (Viswanadham, 1996). The measured variation of tensile strength in the model geogrid with time and for each settlement increment is given in Fig. 7. Typically, the response of SGl, SG4 and SG7 (refer Fig. 3) are presented. With an increase in central settlement a , marked increase in tensile strength can be noted. At a = 0 mm the model geogrid seems to have a tensile strength of the order of 0.25 kN/m. This could be due to adoption of calibration factors obtained from slope of linear portion of tensile strength-strain gauge response curves. The strain gauge SG4 is located in maximum curvature zone (i.e. at S2 in Fig. 6b) and it is found to experience high tensile strength till it attains a central settlement of 15 mm and thereafter the strain
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gauge SG7 was found to mobilise high tensile strength (Fig. 7). This establishes the participation of tension membrane effect resulting due to an increase in stiffness of the instrumented model geogrid after attaining a limiting strain.
5 CONCLUSIONS Based on the above considerations, an attempt has been made to bring out the significance of scalingdown of geosynthetic materials for l g and Ng tests. However, in order to model soil-geosynthetic behaviour satisfactorily, there is a substantial requirement of standardisation of model geosynthetic materials for studies pertaining to the beahviour of reinforced soil structures. The instrumentation and calibration aspects of model geogrid reveals the following; (i) It is required to calibrate the model geogrid by loading and un-loading cycles till the reproducible response is achieved and (ii) Strain gauges can be used to measure in-plane tensile strength characteristics of embedded geogrid. However, the influence of base material on the tensile strength-strain behaviour of the geogrid cannot be ruled out.
Mitchell, J.K., Jaber, M., Shen C.K., and Hua, Z.K. (1988). "Behaviour of reinforced soil walls in centrifuge model tests, " Proc. Centrifuge '88, Paris, 259-271. Moghaddas-Nejad and Small (1997). "Effect of geogrid reinforcement in model track tests on pavements, " J. Tramp. Engrg. Div., ASCE, 122(6), 468 - 474. Ovesen, N.K. (1984). "Centrifuge tests of embankments reinforced with geotextiles on soft clay, " Proc. Int. Symp. on Geotechnical Centrifuge model testing, April, Tokyo, 1423. Palmeria, E.M., and Cunha, M.G. (1993). "A study on the mechanics of unpaved roads with reference to the effects of surface maintenance, " Geotextand Geornemb., 12, 109131. Pinto, M.I.M., and Cousens, T.W. (1999). "Modelling a geotextile-reinforced brick-faced soil retaining wall, "Geosyntheric Znt., 6(5),417-447. Porbaha, A., and Goodings, D, J. (1994). "Geotextile reinforced cohesive slopes on weak foundations, "Proc. Centrifuge '94, Singapore, h u n g , C.F.. Lee, F.H., and Tan, T.S., eds., 623-628. Schofield, A. (1980). "Cambridge geotechnical centrifuge operations, " Geotechnique, London, England, 30(3), 227-268. Springman$., Bolton, M., Sharma, J., and Balachandran, S. (1992). "Modeling and instrumentation of a geotextile in the geotechnical centrifuge, " Proc. of the Earth Reinforcement practice, Ochiai, Hayashi and Otani (eds.), 167-172. Taniguchi, E., Koga, Y., MarimotoJ., and Yasuda, S. (1988). "Centrifuge model tests on reinforced embankments by non-woven fabric, "Proc. Centrifuge '88, Paris, 253-258. Viswanadham, B.V.S. (1996). "Geosynthetic Reinforced Mineral sealing layers of Landfills, PhD dissertation, 28, Institute for Soil Mech. and Found. Engrg, Dept. of Civil Engrg., Ruhr- University Bochum, Germany. Zornberg, J. G., Mitchell, J. K., and Sitar, N.(1997). " Testing of reinforced slopes in a geotechnical centrifuge, " ASTM Geotech. Testing J., 20(4), 470 - 480. 'I
REFERENCES Love, J. P., Burd, H.J., Milligan, G.W.E., and Houlsby, G. T.( 1987). "Analytical and model studies of reinforcement of layer of granular fill on a soft clay subgrade," Can. Geotech. J.,Ottawa, 24, 611-622.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Variation in creep rate at constant loading of PET geogrid strapping W. Voskamp Colbond Geosynthetics, The Netherlands
IF. van Vliet Colbond Geosynthetics, The Netherlands
ABSTRACT: Creep tests have been executed on PET strapping material. The measurements were very precise which resulted in the recognition of different creep rate patterns at various loadings. This data allowed to further verify the molecular chain transformation model which has been established for PET yarns. Creep data from other tests were used to validate the model. A simple method has been proposed to calculate the creep during the service life of a PET reinforcing material using elastic strain, strain during the first 1,000 sec. and the constant creep rate in the rest of the service life. This paper gives additional support to the model of creep deformation at molecular level which is developed in our other paper at this conference (Residual strength of PET after more than 12 years creep loading) and which is based on the stress-strain model developed by Van den Heuvel. 1 INTRODUCTION The creep of polymer material has been investigated by many researchers in the past 25 years. A lot of effort has been put in the transfer of the creep data into design rules. Colbond Geosynthetics has been active in the research on creep on PET material since 1975. Many papers have been published on creep programs which we have executed, at that time under our previous name Akzo Nobel Geosynthetics (Den Hoedt, G., 1986; Viezee D.J. c.s., 1990). This all has let to much better insight in the magnitude of creep and the use of these phenomena in design. Creep ends in rupture and the use of a stress-rupture line for design for certain service lifetime has become a standard design tool (Voskamp W., 1985). Nowadays it is normal practice to execute prolonged loading tests in accordance with the applicable standards and measure the creep up till 10,000 hrs. The loads are taken at higher levels to allow for rupture during the test period. Preferably the loads have to be at at least 3 different levels to allow a reliable extrapolation of the line drawn through all rupture points: the stress-rupture line. Extrapolation is done over 1 or 2 decades. Because creep testing executed in the conventional way takes a very long time, methods have been developed to speed up the process. First Time Temperature Superposition (TTS) became in use. This method uses the concept that increasing temperatures accelerate the creep rate, thus the time for the creep to develop is reduced. Creep curves are shifted and lead to a master curve which is used for extrapolation. In the last years a new method has been published: Stepped
Isothermal Method (SIM) (Thornton, J.S. c.s., 1997). Since 1987 an extensive creep measuring program is being executed at the Colbond Geosynthetics laboratory which has given interesting results. In this paper the results of our creep program on PET straps are published. These very detailed readings gave insight in the variation of creep rates at various levels and showed some surprising results.
2 DESCRIPTION OF THE TEST Creep tests have been executed at samples taken from our newly developed Enkagrid material. This geogrid consists of continuous straps which are connected to each other by means of a laser welding technique. The production method has been described by W. Voskamp, 2000. The tests were made at samples, one strap wide, creep was measured with an optical device resulting in accuracy in reading of 0,Ol mm or 0,033%. The tests were executed in accordance with EN-ISO- 13431, sample length between measuring points was approx. 300 mm. Tests were made at load levels of 10, 20, 30, 40, 71, 73 and 75% of the ultimate tensile strength. The higher loadings were selected to give rupture within the testing period. The lower range was used to get detailed creep results. Further Stepped Isothermal Method tests were executed by ERA and published by J. Greenwood, 2000. Based on this program the stress rupture level of 68,5% was established for a service lifetime of 114 years. The tested material has a strength of 104 kN/m at an elongation at break of 159
8%. The modulus of this material is considerably steeper than that of PET yams. The stress-strain curve is given in figure 1. The modulus of the strap varies at different stress levels, therefore the modulus is also shown in this figure. The creep of the material was measured and is shown in figure 2. This is the traditional way of presenting the creep results. If we compress the time scale much more we can see differences in the creep lines. At 10% load level, the strain increases in time as shown by the concave shape of the curve. The levels 20, 30, 40% shown a convex shape. This convex shape, with decreasing strain in time is the normal shape of the creep curve. At the high levels we can see some increase in creep rate which indicates the start of the well known upswing. However, the concave curve at 10% load level was unexpected. This brought us to investigate if this behaviour could be explained by a description of the molecular changes during loading. 3 MOLECULAR CHANGES OF PET YARNS DURING STRETCHING
In our other paper at this conference (W.Voskamp 2001) we have described the changes which take place at molecular level during the loading of PET yarns during stretching. These changes were investigated at Akzo Nobel Research Laboratories by using rheo-optical infra red spectroscopy and other techniques and published in 1993 by Van den Heuvel, C.J.M. It was found that the modulus of the stressstrain curve of PET material varies a lot and has typically 2 maxima. For clearness sake we quote in the following section the description of the molecular changes during loading as we have written in our other paper to this conference. Quote: Clearly 2 maxima can be found in the modulus curve: one around 0,5 - 1% strain and the other at about 7 - 8% strain.These stress-strain curves can be divided in 3 regions using these maxima. The molecular deformations which take place in these 3 regions are clearly different from each other. Region 1: up to the first maximum in the modulus (around 0,5 - 1%). Region 2: between the first and the second maximum in the modulus (between 0,5 - 1% and 7 - 8%) Region 3: after the second maximum in the modulus.
Figure 2. Creep-strain results in compressed scale.
To understand the processes which take place in the 3 regions it is necessary to look at the physical structure of PET, a two phase model with amorphous and crystalline domains (Den Hoedt, 1986, V.d. Heuvel, 1992).
160
Figure 6. Molecular processes in the 3 regions.
domains. The study of Van den Heuvel had as result a description of the uncoiling of processes which take place. In region 1 entanglements (amorphous chain-chain interactions) of the molecular chains contribute substantially to the modulus. This leads to a maximum in modulus. In region 2, after the first maxirrum, the modulus reduces, which is caused by the break down of the entanglement network and the start of the uncoiling by trans transitions. This uncoiling takes of gauche course place in the amorphous domain. The uncoiling effects a lowering of the non-elastic modulus, while in region 2 the elastic modulus increases. The uncoiling in region 2 leads to straining of the tie-molecules. The chain modulus of the taut-tie molecules is relatively high, which results in increase of the tensile modulus of the yarn. This increase continues up to the next maximum. When the modulus reaches its second maximum some of the taut-tie molecules begin to break, this is the start of region 3. The number of molecules that break is limited. (It is measured to be maximally 3%). The increased reduction of the modulus in region 3 is the result of chain scission in the amorphous zone where much more local stress concentrations are generated, which lead to further accumulation of molecular breakdown and which lead to rupture of the filament. We can summarize this process as: Region 1 : entanglement of molecular chains result in high modulus. ---f
Figure 4. Physical structure of PET yarn (V.d. Heuvel).
Figure 5 . Molecular and physical structure of PET (V.d. Heuvel).
During straining the PET molecules will uncoil. The ethylene groups in the amorphous domains of semicrystalline PET occur in 2 conformations, gauche and trans conformation. Molecule chains with a lot of gauche will be coiled strongly, trans conformers in series give rise to extended chains. The crystalline zones consist only of trans conformers. So we have t o concentrate on what happens in the amorphous
&
Region 2: uncoiling of the molecular chains with gauche-trans transformation and straining of the taut-tie molecules.
Region 3: chain scission in amorphous zones leading to rupture.
161
At 30% the modulus is between the value of 20 and 40% load level. This is represented by the values of the creep as indicated in fi ure 1. The higher load levels are close to the 2 maximum and 75% is more or less at the maximum. In the creep curves we observe the start of an upswing which indicates a reduced modulus and entrance of the loadinghtrain condition in region 3. 3. In the other paper on this conference it is shown that as a result of the creep process the elongation at break reduces, the modulus of the curve becomes higher and the location of the 2"d maximum moves forward to lower strain levels. Figure 7 shows the creep results of the SIM test on geogrid type 90. That is the group of measurements at the high load levels and indicated as VD. We see here that the upswing of these curves start between 6 and 7% elongation. This is in line with the described behaviour of the molecular chains. On the other hand we may not draw decisive conclusions from these measurements as they are made at elevated temperatures which influence the modulus as shown in the study by Van den Heuvel(l993).
This process has been measured and verified by means of various techniques as has been described by Van den Heuvel (1992).The process describes what happens during short term loading of a PET sample'. Unquote. The difference in the shape of the creep curves at various load levels can be explained with this method. In figure 1 the stress-strain curve and the modulus of the type 90 material are shown, also the creep during the 1 year testing is indicated for the various load levels. In the modulus curve the various regions can be described: 0-0.8% entanglement of molecular chains result in high modulus. 0.8-2.5% entanglement and start of uncoiling of the molecular chains. 2.5-7% further uncoiling of the molecular chains by gauche -> trans transition and taut-tie molecules carry the load. 7-8% first part breakage of (3%) taut-tie molecules and stress concentrations in the amorphous zone and chain-scission as a result of it.
I3
4 OBSERVATIONS (FIGURE 3) When we study the results of the stress-strain curve of type 90 (figure 1) and the creep curves of figure 2 and 3 following observations can be made. 1. The creep measurement at 10% load is in the region 1.This means that the entanglement of the molecular chains has not yet completed and will further take place during creep loading. The result of this entanglement is an increase of elongation in time, up to the moment that all entanglement has taken place and the modulus has reached its maximum. Reduction of creep rate at the end of the test period may be expected and indications of it can be found at the end of the curve. The concave behaviour of the creep line in figure 3 is typical for region 1 and is also the result of our test. 2. The creep curves for 20, 30 and 40% have a convex shape. Further the amount of creep during the loading time of approx. 10,000 hrs. is maximal at load level 20% and reduces with every load step 30%, 40% and 70% (figure 1 and 3). This is contrary to what one would expect but it is in line with the molecular changes. At 20% load level the sample is in the strain zone between 1,5 and 2,8%. This is the zone of reduced modulus caused by the entanglement of the molecular chains and especially in the zone with the lowest modulus, consequently the elongation will be greater than in the zones with higher modulus.
Figure 8. Creep of Diolen yarns at 10% load level.
162
Figure 9. Creep of Diolen yarns at 20% load level
5 VERIFICATIONS OF THE RESULTS In 1985 a series of very detailed creep tests have been made at the Akzo Nobel Research laboratory on 2 types of Diolen yarns (type 855T ~ I I UI {U). The results of this study showed exactly the same behaviour as mentioned above for PET straps. Figure 8 and 9$how the resulp. I I I Another conclusion of this study was that the creep rate in the first 10,000 seconds (on log scale) was much higher than afterwards and it became steady after this point. These measurements on different PET yarns show the same concave and convex shape of the creep curve for 10 and 20% as was shown with PET straps. In our other paper on this conference we have published supporting data gained from the residual strength and stiffness of other PET yarns. Altogether we consider the presented description of polymer changes during loading as very realistic. This system can also be used for evaluation of long term extrapolations of data.
,
6 QUICK BUT ACCURATE CALCULATION OF TOTAL CREEP The data from the tests on straps as well as those on PET yarns show that after 10,000 seconds or 2,78 hrs. the creep rate becomes steady (at log scale). This observation is important for future creep testing. The total strain curing a certain service life consists o f : C C C2 C
= total strain = elastic strain during loading = creep during 10,000 s = creep during the rest of the service life.
The C e and C 2 can be measured rather easy with a stress-strain tensile test and with the same
equipment but now keeping the load constant for 3 hours. In this way we get very quickly C and C 2. The creep rate C is rather constant during the rest of the service life using a log scale for the time. For example: the measurement at 30% of the PET strapping. C e = 3.35 C 2 = 0.45 The creep rate in the next part of the graph can be calculated as 0,2% over 2 decades which is 0, 1% per decade. If the lifetime is 100 years or 876,000 hrs. then, log t (hrs.) = 5,94. The log t at the end of C 2 period is log (166 hrs.) = 2.22. The expected creep can be calculated as: log t: 5,94 - 2,22 = 3,72 or 3,72 decades. C = 3.72 * 0.1 = 0.37% The total deformation is then C = 3.35 + 0.45 + 0.37 = 4.17%. 7 CONCLUSIONS
1. Creep curves of PET at 10% loading have a concave shape and above 20% a convex shape. 2. The description of the molecular chain behaviour under stretching as published by Van den Heuvel can be used to explain this behaviour. 3. This description is also applicable for description of the creep process of PET material. 4. The elongation under constant loading can be divided in 3 sections: elastic strain, zone of strongly reducing creep rate up till 10,000 s, zone with creep rate constant. These sections are applicable for all loading cases as long as the total strain remains below the 2"d maximum of the modulus of the stress-straincurve. 5 . Using the method of 3 creep zones it is easy and it does not require large creep testing equipment or prolonged creep testing to determine the total strain during the service life of a PET reinforcement material. REFERENCES Den Hoedt, G., 1986, Creep and Relaxation of Geotextile Fabrics, Geotextiles and Geomernbrriiies Vol. 4, page 8392. Viezee, D.J. c.s., 1990. Designing soil Reinforcement with Woven Geotextiles, 4'h International Conference on Geotextiles, Geomembranes and Related Products, The Hague, page 65 1-656. Voskamp, W. How to determine the long-term design strength of reinforcing mats 1985, Yearly conference of the Institution of Civil Engineers in Hong Kong, Hong Kong. Thornton, J.S. c.s., 1997. Approaches for the prediction of Long-Term Viscoelastic Properties of Geosynthetics from Short-Term Tests, Geosyrztlzetics 1997 Conjerence Proceedings, St. Paul, Miizri., page 2 77-291.
163
Voskamp, W. 2000. Index and Performance testing of a new geogrid made of highly oriented straps, Advances in Transportation and Geo-environmental Systems using Geosynthetics ASCE Geotechnical Special Publication no. 103, Denver, USA. Greenwood, J., Voskamp, W., 2000. Predicting the long-term strength of a geogrid using the stepped isothermal method. 2'Id EuroGeo, Bologna Italy, page 329-332. Voskamp, W. c.s., 2000. Residual Strength of PET after more than 12 years creep loading, 3"' IS Kyushu Corgerence on Earth Reinforcenient, Kyushu, Japan.
164
Van den Heuvel, C.J.M. c.s., 1993. Molecular changes of PET yarns during stretching measured with Reo-Optical infrared spectroscopy and other techniques, Journal of Applied Polymer Science. Van der Bleek, I.M.C., 1985. Creep of Diolen Yarns. Internal report Akzo Nobel Research Laboratories.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Residual strength of PET after more than 12 years creep loading W. Voskamp Colbond Geosynthetics, The Netherlands
F. van Vliet Colbond Geosynthetics, The Netherlands
J. Retzlaff Colbond Geosynthetics, Germany
ABSTRACT: The residual strength of PET material after being loaded for more than 12 years at various load levels has been established. It turns out that no significant reduction in strength has taken place, however, the elongation at break has reduced significantly. An earlier published description of the changes at molecular level during stress-strain tests of PET yarns has been used to explain the results of this test and based on the system also the long term effects could be explained. A method has been developed to determine the rest of the available service life of a materiaUstructure after continuous loading. This method has been validated and a practical example is shown. With this method we can determine at any time and with good accuracy the remaining service life of a structure and therefore we can control the safety of any reinforced soil structure using PET reinforcement material perfectly during its service life. Using this method any uncertainty about the lifetime of a soil reinforcing structure can be eliminated.
1 INTRODUCTION At the second Kyushu Conference in 1997, a paper was presented by W. Voskamp entitled “Proposed method to determine the safety capacity of reinforced soil structures”. In this paper it was indicated that the residual strength of PET after having been loaded for some time is much higher than was expected when the stress rupture model is used for design. A typical stress rupture model is shown in figure 1. T,, = the extrapolated creep rupture strength at the end of the selected design life and at a maximum operational temperature. &sign = the selected design life. Tult = the ultimate strength measured during the short-term tensile strength test. In the literature and based on experience it is anticipated that the residual strength of a material is
higher than the TCr according to the stress-rupture line (Schardin-Liedtke, H, 1990). A method to calculate the effect was proposed by J. Greenwood, 1997. At the laboratories of Colbond Geosynthetics, formerly known as Akzo Nobel Geosynthetics a creep testing program is on its way since 1987. Intermediate reports of the results have been made. We have decided to stop this program and to measure the residual strength of these samples. The samples were tested at various load levels and many have reached rupture in the past, however, a number of them have been loaded for about 13 years. The samples were taken of different PET yarn types and from at least two different suppliers. They were taken from spools, from woven fabrics and even a part of the program was on samples taken from fabrics damaged in a controlled way which resulted in the same amount of damage in all samples. In this paper we will report on the residual strength measurements made on various samples. Some test results have already been published by J. Greenwood in the discussion on his paper (Geosynthetics International, 1997). 2 TEST RESULTS
Figure 1. Creep rupture graph.
In 1987 we started with a creep-testing program on yarns. The yarns were loaded with a constant load which varied between 20% and 70% of the ultimate tensile strength (UTS). At regular times the strain in the sample was measured between 2 marks on the sample. The measurement was made by means of an 165
optical device. The temperature in the test area was 20 “C at average. The time to rupture was recorded independently. The samples had a length of at least 80 cm and the strain measurements were made over a length of 60-70 cm. The strain was recorded with an accuracy of 0.01 cm. Five groups of samples were tested: Category 1: Virgin yarn taken from the spool, not woven into a fabric (Diolen 770). Category 2: Yarn taken out of a Stabilenka 150 fabric (Diolen 770). Category 3: Yarn taken from a damaged Stabilenka 150 fabric (Diolen 770); the damage was applied in a controlled way. Category 4: Virgin yarn from the spool, made by another process (Diolen 776). Category 5: Virgin yarn from the spool, made by another company (type A).
In some cases the residual strength is higher than the initial strength. This is caused by the variations in strength between the individual yarns. We have made a statistical evaluation and will use in our analysis for all materials the mean value. In general it can be concluded that, if there is a reduction in strength, it is very limited and it is within the accuracy of the tests. The loading of the samples was checked at the end of the test. The values of the load are given in the table. Uncertainty remains about the initial strength of the samples. We have records that indicate strength levels which are lower than the mentioned ones. We had stored virgin samples of the test material and after completion of the test program we checked the initial strength again. These values are mentioned. During the test program we found strange results: some yarns broke within 1000 hrs. while they actually were loaded at load levels which are related to a service life of 100 years. Detailed investigations have been made and the conclusion was that the transfer mechanism of the load to the yarn and the application of the load to the yarn could be too sensitive for the sample. This means that some uncertainty remains about the correctness of the calculated normalized load (probably the percentages are too low compared to the actual loads). These uncertainties have no influence on the use of the samples for determination of the residual strength. Although the strength of the yarns has not changed so much, the elongation at break has reduced considerably. Table 3 gives an overview of the results. The reduction in strain is significant in all loading cases and types. The standard deviation is small so it may be concluded that although the strength does not vary so much, the strain has reduced considerably. This means also that the modulus of the material has increased.
In 1997 we recorded the following results: Table 1. Test results after 55500 hrs. Yam type
Initial strength
(N)
Loading IN)
Cat. 1 Cat. 2 Cat. 3 Cat. 4 Cat. 5
76.6 77.2 66.7 75.0 74.0
38.61 38.61 41.53 38.77 40.26
Residual Retained strength strength (N) I%) 77.2 75.2 64.9 74.4 76.9
100.7 97.3 97.4 99.3 103.9
P
Yarn type
Time of loading % % .I..”l._l.l__._..._ (hrs) .__.__.l.l......__._._..__._,_.___I__ 5.28 0.86 55575 5.23 0.73 55573 5.79 0.82 55578 5.16 0.71 5507 1 55572 5.73 1.16 Initial strain
I_______._^
Cat. 1 Cat. 2 Cat. 3 Cat.4 Cat. 5
Creep Strain
I
In 2001 we completed the testing and the following results were recorded:
3 MOLECULAR CHANGES OF PET YARNS DURING STRETCHING
Table 2. Test results after 104800 hrs. -%P----,-I_____e--
Yam Type
Initialstrength
Loading
Residual
strength
Retained strength
Cat. 2 Cat. 3 Cat. 4 Cat. 5
74.20 63.54 74.84 76.33
30.89 30.89 31.02 32.21
74.30 67.34 73.36 72.02
100.1 105.9 98 94.4
Yarn Type Cat. 1 Cat. 2 Cat. 3 Cat. 4 Cat. 5
Initial strain %
Creep strain %
Hrs.
4.39 4.32 4.43 4.51 5.20
0.95 0.45 0.40 0.57 0.54
104832 104832 104832 104832 104832
In 1992 a study was made at Akzo Nobel Research Laboratories in Arnhem on the molecular changes of PET yarns during stretching measured with rheooptical infrared spectroscopy and other techniques. This study is reported in the Journal of Applied Polymer Science (V.d. Heuvel C.J.M. C.S.,1993). They developed a method to measure infrared spectra during the mechanical deformation of yarns. With this rheo-optical technique they were able to study the molecular processes during loading of PET yarns. These results were combined with data obtained from Size Exclusion Chromatography and tensile measurements at elevated temperatures. In summary it was found that the stress-strain curve of a PET yarn or other PET material the modulus of the 166
Table 3. Residual strength results.
Cat. 1 spool Spool loaded residual Cat. 2 fabric Fabric loaded residual Cat. 3 fabric damaged Fabric damaged loaded residual DIOLEN 776 Cat. 4 spool Spool loaded (40-50% residual) Spool loaded (20%) residual TYPE A Cat. 5 spool Spool loaded residual Yarns DIOLEN 770 Cat. 1 spool Spool loaded residual Cat. 2 fabric Fabric loaded residual Cat. 3 fabric damaged Fabric damaged loaded residual DIOLEN 776 Cat. 4 spool Spool loaded (40-50% residual) Spool loaded (20%) residual TYPE A Cat. 5 spool Soool loaded residual
78.33 74.86 74.20 74.30 63.54 67.34
1.40 3.97 1.43 3.73 1.37 4.64
11.55 9.98 9.76 9.84 8.20 7.56
74.84 73.36
0.27 10.68 1.66 9.01
74.41
--
76.33 72.02 S.d.
1.18 12.47 3.06 8.95 n
0.78 1.05 0.21 0.97 0.19 0.75
10 6 10 6 10 6
0.39 0.63
10 6
8.34 Figure 3. Typical stress-strain curve with modulus (V.d. Heuvel, 1993). Note: tenacity is factor * load/specific strength
lecular deformations which take place in these 3 regions are clearly different from each other. Region I : up to the first maximum in the modulus (around 0,5 - I %). Region 2: between the first and the second maximum in the modulus (between 0.5 - 1% and 7 - 8%). Region 3: after the second maximum in the modulus. To understand the processes which take place in the 3 regions it is necessary to look at the physical structure of PET, a two phase model with amorphous and crystalline domains (Den Hoedt, 1986, v.d. Heuvel, 1992). During straining the PET molecules will uncoil. The ethylene groups in the amorphous domains of semi-crystalline PET occur in 2 conformations, gauche and trans conformation. Molecule chains with a lot of gauche will be coiled strongly, trans conformers in series give rise to extended chains. The crystalline zones consist only of trans conformers. So we have to concentrate on what happens in the amorphous domains.The study of Van den Heuvel had as result a description of the uncoiling of processes which take place. In region 1 entanglements (amorphous chain-chain interactions) of the molecular chains contribute substantially to the mo-
2 0.63 0.74
10 6
Figure 2. Stress-strain curve before and after testing Diolen 770 spool.
material, which is the first derivative of the stressstrain curve, varies a lot (see figure 3). Clearly 2 maxima can be found in the modulus curve: one around 0.5 - 1% strain and the other at about 7 - 8% strain. These stress-strain curves can be divided in 3 regions using these maxima. The mo-
Figure 4. Physical structure of PET yarn (V.d. Heuvel).
167
R
which lead to further accumulation of molecular breakdown and which lead to rupture of the filament. We can summarize this process as: Region 1: entanglement of molecular chains result in high modulus.
R
Region 2: uncoiling of the molecular chains with gauche-trans transformation and straining of the taut-tie molecules. -JA
fibre of filament
-?
\
I 00 A crystal
Region 3: chain scission in amorphous zones leading to rupture.
=s=
amorphous
Figure 5. Molecular and physical structure of PET (V.d. Heuvel).
dulus. This leads to a maximum in modulus. In region 2, after the first maximum, the modulus reduces, which is caused by the break down of the entanglement network and the start of the uncoiling by gauche -+ trans transitions. This uncoiling takes of course place in the amorphous domain. The uncoiling effects a lowering of the non-elastic modulus, while in region 2 the elastic modulus increases. The uncoiling in region 2 leads to straining of the tie-molecules. The chain modulus of the taut-tie molecules is relatively high, which results in increase of the tensile modulus of the yarn. This increase continues up to the next maximum. When the modulus reaches its second maximum some of the taut-tie molecules begin to break, this is the start of region 3. The number of molecules that break is limited. (It is measured to be maximally 3%). The increased reduction of the modulus in region 3 is the result of chain scission in the amorphous zone where much more local stress concentrations are generated,
Figure 6. Molecular processes in the 3 regions.
This process has been measured and verified by means of various techniques as have been described by Van den Heuvel (1992). The process describes what happens during short term loading of a PET sample. Analysis of the residual strength measurements and especially of the stress-strain curves of the residual strength test indicate that the mentioned process is not only taking place during short term loading, but that the same processes take place during long term, creep loading. Figure 7 describes the loading pattern of the sample Diolen 770, spool. Phase 1, loading of the sample Phase 2, constant loading and creep Phase 3, unloading Phase 4, loading up to rupture. The sample is loaded to a level that it is clearly in region 2: uncoiling of the molecules and straining of the taut-tie molecules. The modulus pattern of the yarn at the begin of the test and at the end is indicated in the figure. The second maximum of the modulus has moved to a lower strain level during the creep loading in 12 years. As can be seen the modulus of the yarn is higher at the end of the test compared to the begin (while the strength remains at the same level). It can be concluded from these results that during the 12 years loading the modulus strain of the yarn has increased which indicates that the molecules have become more stretched during the loading. This is also described as the mechanism in region 2 during the short term test. Further it is logical that, when during constant loading a further increase in stiffness or modulus takes place, it also will not influence the strength of the material. The strength is determined by passing the second maximum in modulus because then tiebreaks and scissions of molecules take place. This mechanism has been verified with the results of the tests on the other type of PET materials and at various load levels. They all show the same pattern, therefore it may be concluded that the molecular 168
Figure 7. Loading during creep test Diolen 770, spool.
change mechanism which was described by Van den Heuvel (1992) is also applicable for long term loading. In a separate paper to this conference the model is verified with creep measurements on various types of PET materials: different yarns, straps and at various loading levels. A different behavior of creep is shown for region 1 and 2.
4 VALIDATION METHOD OF BUILD STRUCTURES The described mechanisms give us a unique method to establish the long term reliability of a build reinforcement structure. We have seen that the second maximum of the modulus moves to lower strain levels during the continuous loading of a structure in region 2. The PET reinforcement material ruptures when the second maximum is passed. This rupture takes place within 3 months after passing the second maximum as we have observed in our rupture measurement tests. It can easily be established (even with a short term test) what the creep rate will be during the further loading of a structure. The practical method is: 1. Take a sample of the PET reinforcement material from the loaded structure. 2. Execute a stress-strain test on this sample. 3. Calculate the modulus curve from the stressstrain curve and indicate the 2 maximums. 4. Determine the creep rate at the loading level for the rest of the service life of the structure. This could even be at a higher load level, when the structure will get an increase in service load. 5. Calculate the available rest service life by dividing the strain at the second modulus minus at the required load level in the residual strength stressstrain line, by the creep rate. tresr
-
2nd mod
Figure 8.2. Diolen 770 - virgin fabric.
Example: The test on the Diolen 770 ‘blanc fabric’ is taken as an example. We have loaded the sample at 30,89N. The corresponding strain in the stress-strain curve (residual strength test after 83664 hrs) is 3,7%. The second maximum of the modulus has moved from the original material (at 7%) to 5,7% at this test . This means that we have 5,7 - 3,7 = 2% to go before the second maximum equals the loading strain. The creep rate is: 0,03% I 2,1836 yr. tr<,,,= ____. 2% 2,1836 years = 145,6years
0,03%
trest= time before rupture Note: the creep rate is decreasing in time and can be better described with a log formula. We took here simply the last measured interval, which also would be done in practice. When we use the log formula we find in this case a creep rate of 0,98% per decade.
2% logt,,,, = ____ = 2,04 0,98% logt,,, = 2,04+10g104832=7,06 t =11.481.536hrs =1.310years
- 1load
creep rate 169
According to the normal stress-rupture line of PET yarns we may expect a service life of 1,000 years (8760000 hrs. or log t(hrs.) 6,94 for 54% and 10,000 year (87600000 hrs, log t(hrs.) = 7,94) for 50% 1oading.The corrected load value for this sample is 43 x 1,2 = 52%, this means a service life according to the stress-rupture line of log t = 7,44. On the log scale measured this is of the same order as has been calculated with the described rest strength calculation method.
5 CONCLUSIONS 1. The residual strength of PET materials after being exposed to loads for more than 12 years has not decreased significantly. 2. The elongation at break of the PET materials after the 12 years loading has reduced significantly leading to the conclusion that the modulus of the material is increased. 3. The earlier developed description of the molecular changes of PET material during stress-strain loading with the 3 regions, clearly divided by the 2 maxima in the modulus of the material can also be applied to the creep process which has been shown by means of the results of various long term tests. 4. With this process description a method is developed to calculate the time period left until rupture of a loaded reinforcement material. 5. This method has been validated with the results of long term test measurements and the results are
comparable to the results of the stress-rupture analysis. 6. The described system of molecular changes during short term loading has been extended to also long term loading which gives us detailed insight in the creep process of PET and allows us to make a good prediction of the safety of executed soil reinforcement systems. REFERENCES Voskamp, W. 1996. Proposed method to determine the safety capacity of reinforced soil structures during the lifetime. Earth Reinforcement proceedings of IS Kyushu 1996 Conference. Page 1075 - 1080. Greenwood, J.H. 1997. Designing to Residual Strength of Geosynthetics Instead of Stress-Rupture, Geosyiitlzetics Interiiational Vol4, no. I page 1-10 Greenwood, J.H. 1997. Discussion on Designing of Residual Strength of Geosynthetics Instead of Stress-Rupture, Geosynthetics International Vol. 4 No. 6 page 673-677. Schardin-Liedtke, H. 1990. Geotextiles for the support of steep slopes: Approval Procedures. Proceedings of the 4Ih Intemational Conference on geotextiles, Geomembranes and Related Products, Balkema, Vol. 1, The Hague, May 1990, 79 - 85. Van den Heuvel, H.M., Faassen, W.A., Veurink, J, Lukas, L.J. Molecular changes of PET yarns during stretching measure with rheo-optical infrared spectroscopy and other techniques, Journal of Applied Polymer Scietzce, 1993. Den Hoedt, G. Creep and Relaxation of Geotextile Fabrics, Geotextiles arid Geomembranes, 1986, Voluirze 4, page 83 - 92. Voskamp, W. & Vliet, F. Variations of Creep Rate at constant loading of PET Geogrid Strapping, 3rdconference on Earth Reinforcement Kyushu, 2001.
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Landmarks in Earth Reinforcement,Ochiai et al. feds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Static and dynamic strength of cement mixed soil reinforced by fibers S. Yasuda Tokyo Denki University, Saitanza, Japan
H. Ishinabe The Institute of Japanese Union of Scientists & Engineers, Tokyo, Japan
Y. Murasawa Mitsui Harbour and Urban Construction Co., Ltd., Tokyo, Japan
ABSTRACT: Static and dynamic strength of cement mixed soil reinforced by fibers were studied by conducting unconfined compression, split and bending tests. Test results showed that compression, tensile and bending strength increased with mixing of fibers. Especially tensile and bending strength increased about two times. Liquefaction strength increased also with mixing of fibers. 1 INTRODUCTION
2 STATIC STRENGTH OF REINFORCED CLAY
Deep cement mixing method is one of the effective method to improve the ground. In Japan, the deep cement mixing method is used to improve the ground under or adjacent of structures. Main purpose of the improvement is to increase bearing capacity or liquefaction strength of the ground. For existing embankments such as river dikes and road embankments, surrounding grounds of the embankments improve by the deep cement mixing method to prevent the settlement of the embankments during earthquakes as shown in Fig. 1, because the foundation ground of the embankments can not improve. For example, the area of 5m in width and 10m in depth improves by the cement mixing method. The improved zone is a kind of selfstanding wall and must resist against bending stress that is caused due to the pressure by self-weight and inertia force of the embankments. Therefore bending strength of the improved ground is important in the design of appropriate improving area. However, in general speaking, bending strength can not improve enough by cement only. Then the authors tried to mix fibers to strengthen the cement mixed soil.
In the first step of this study, static strength of a clay mixed by fibers and cement was studied by conducting several tests. The clay used for the tests was a mix of “Kibushi clay” and bentonite. Mixing rate of the Kibushi clay and bentonite was 9: 1. Figure 2 shows grain-size distribution curves of the mixed Kibushi clay together with the curve for Toyoura sand which is a clean sand. Fiber used for tests was a flat bar of 0.40mm in diameter and 24mm in length as shown in Fig3. Tensile strength and Young’s modulus of the fiber are 900MN/m2 and 30GN/m2, respectively. Powder of the mixed clay was kneaded dough with a predetermined amount of water which produces 100% of water content, by a mixer for ten minutes. After 24 hours, the clay was mixed with cement and the fibers for two minutes by a mixer and for one minute by hands. Then the sample was filled in a mold and cured for seven days in a constant temperature and moisture vessel. Mixing rates
Figure 1. An example of soil improvement for an existing embankment. Figure 2. Grain size distribution curves of tested soils.
171
Figure 3. A fiber used for tests.
of fiber, f, and mixing rate of cement, C, were 0% to 2% in volume and 50kg/m3 to 150kg/m3, respectively. Three kinds of tests, unconfined compression test, split test and bending tests were conducted to know compression strength, tensile strength and bending strength, respectively. Figure 4 shows relationships between axial strain and compression stress of the samples of C=150kg/m3 in the unconfined compression tests. In the case of the sample without fiber, shear stress decreased rapidly after peak stress, while stress did not deceased drastically after peak stress for the samples of f=0.5% or 1%. Stress did not decreased if the sample
contains 1.5% or 2% of fiber. Namely, residual strength increased with the mixing rate of fiber up to about f=1.5%. Peak strength increased also up to f=1.5%. Similar relationships were observed in the spilt and bending tests as shown in Figs. 5 and 6. Tensile and bending strength increased about two times with mixing of fibers. As mentioned before, it is desired that tensile and bending strength increased with the content of fibers because bending strengths must be improved for the design of the measure against the deformation of embankments. Figures 7 to 9 show relationships among mixing rate of cement, mixing rate of fiber and strength obtained by the three types of tests. Compression, ten
172
100
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Mixingrate ofcement, C (kdm 3,
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Figure 1 1. Comparison of stress strain curves of three types of arrangement of fiber.
Figure 9. Relationships among bending strength, mixing rate of cement and mixing rate fiber.
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tension crack
f
(a)Pcrpendicular-arranged specimen
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Mixing rate of cement, C (kdni ')
Figure 12. Comparison of tensile strength among three types of arrangements of fibers. Figure 10. Parallel and perpendicular arrangement of fibers.
sile and bending strength increased with the increase of mixing rate of cement and fiber. However, strength of the samples of f=2% was almost equal or less than the strength of the samples of f=l.5%. This means that strength does not increase or slightly decreases if a sample contain too many fibers. As shown in Figs 8 and 9, tensile and bending strength under C=150kg/m3 were slightly less than the strength under C=100kg/m3 in some mixing rates of fiber. Appropriate combination of mixing rate of cement and fiber for the tested clay was about C=100kg/m3 and f=1.5%. Though all data are not shown in Figs.4 to 6, residual strengths were almost highest values also under the condition of C=100kg/m3 and f=l.5%.
3 EFFECT OF DIRECTION OF FIBERS In the tests mentioned above, the fibers were mixed in random direction. Special tests were carried out to demonstrate the effect of the direction of mixed fibers. Three types of arrangement of fibers: parallel, perpendicular and random directions to tensile force, were selected as shown in Fig.10. Tested soil was
0
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M king rate of cenient, C (kgm ')
Figure 13. Comparison of residual strength among three types of arrangement of fibers.
the mixed Kibushi clay. Mixing rates of fiber and cement were f=1.5% and C=50kg/m3 to 150kg/m'. Split tests were conducted for the three types of specimens. Figure 11 shows test results for the specimens of f=l.5% and C=100kg/m3, together with the tests for the specimens without fiber (f=O%, ~ = 1 0 0 k ~ / Stress-strain ~~). curves for four types of specimens were quite different. Figures 12 and 13 compare the relationship among mixing rate of cement and tensile and residual strengths. Tensile
173
the cement mixed soil, the effect of enlarge parts can not be neglected because the effect of friction between the surface of the fiber and soil is small.
strength of the parallel-arranged specimen was not so different from the specimen without fiber. On the contrary, residual strength of the parallel-arranged specimen was larger than the strength of specimen without fiber. Tensile and residual strengths of the perpendicular and random arranged specimens were larger than the strength of the parallel-arranged specimen. The strongest arrangement was random arrangement. Figure 14 shows schematic diagram of magnified shape of a fiber. The tops of the fiber are enlarged slightly. Then special mended fibers were prepared by cutting the enlarged tops to become straight shape. Figure 15 shows test results for the specimen mixed by the special mended fibers together with test results for the specimens with normal fibers and without fiber. As shown in the figure, strength of the specimen with the top-cut fibers was less than the strength of the specimen with normal fibers. This implies that the enlarged parts strengthen the cement mixed soil. The fiber use in this study has been used for reinforcing concrete members. In the design for the reinforced concrete, effect of the enlarged parts on the strength is neglected because the effect of reinforcement is mainly due to the friction between surface of the fiber and the concrete. However, in
4 STATIC STRENGTH OF REINFORCED SAND Static strength of fiber and cement mixed sand was studied in the same method for Toyoura sand which is a clean sand as shown in Fig.2. Dried sand was mixed with 2% of cement in weight and 1% or 2% of mixing rate of fiber, for two minutes The mixed sand was filled in a mold under the condition of 30% of relative density. Then the specimens were submerged and cured in a container for seven days. Three kinds of tests, unconfined compression tests, split tests and bending tests were conducted same as the tests for clay. Figure 16 compares the relationships between compression strain and compression stress for different mixing rate of fiber. Peak and residual strengths increased with the mixing rate of fiber. Figure 17 compares the relationships between axial strain and tensile stress. Peak tensile strength increased with the mixing rate of fiber. Moreover, fiber-mixed sands had some residual strengths while the sand without fiber had no
'OI-
i
c Figure 14. Schematic diagram of the enlarged top which was cut to become straight shape.
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Figure 16. Stress-strain curves in unconfined compression tests for Toyoura sand.
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Figure 15. Effect of the cut of the top of iibers on stress strain curves.
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Figure 17. Stress-strain curves in split tests for Toyoura sand.
174
residual strength. Though the data are not shown here, results in bending tests were similar as the results in the split tests. 5 LIQUEFACTION STRENGTH OF REINFORCED SAND Undrained cyclic triaxial tests were conducted for the same cement and fiber mixed sand mentioned above, to obtain liquefaction strength. Cyclic axial stress of 0.1Hz was applied under the effective confining pressure of 49kPa. Relative density of the sample was 30% to 50 %. Figure 18 compares the relationships between cyclic stress ratio and number of cycles to cause liquefaction with different rates of fiber and cement. As shown in the figure, cyclic stress ratio to cause liquefaction in a certain number of cycles was small if the sand was not mixed with cement even though the sand has some fibers. Figure 19 shows relationship between mixing rate of fiber and undrained cyclic strength, Rl(N1=20, DA=5%). The undrained cyclic strength increased with the mixing rate of fiber if the soil was mixed with cement also. The undrained cyclic strength for the specimen with 2 % of fibers was 1.4 times higher than that for the specimen without fiber. In Figures 18and 19, number of cycles to cause liquefaction was defined as the numbers in which double axial strain reached to 5% (DA=5%). Several definition of liquefaction is used in the current design for liquefaction. For example, number of cycles to cause about 1.0 (for example 0.95) of excess pore pressure ratio is defined also. Then difference of the two definitions is compared on Fig.20. As shown in the figure, two definitions become different if the sand contain many fibers. This implies that liquefaction-induced deformation of ground and structures can reduce by mixing the fiber, even though excess pore water pressure ratio increases up to almost 1.O. 0.8
0.6
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Figure 18. Relationships between R and N1 in cyclic triaxial tests for Toyoura sand.
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Figure 19. Relationship between undrained cyclic strength and mixing rate of fiber. 50
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Number of cyclic to cause 92% of excess pore pressure ratio,Nl(u/ 0 =0.95)
Figure 20. Comparison of two definitions of liquefaction.
6 CONCLUSIONS Static and dynamic strength of cement mixed clay and sand reinforced by fibers were studied by conducting unconfined compression tests, split tests, bending tests and undrained cyclic triaxial tests under several rates of fiber and cement. Main conclusions derived through the tests are as follows: 1. Compression, tensile and bending strength increased with containing of fibers for both clay and sand. Especially tensile and bending strength for clay increased about two times. Residual strength was also increased with containing of fibers. The best mixing rate of fiber and cement for the tested clay was 1.5 % and 100kg/m3,respectively. 2. Undrained cyclic strength of cement mixed sand increased with the mixing rate of fibers. The undrained cyclic strength for the specimen with 2 % of fibers was 1.4 times higher than that for the specimen without fiber. However, the undrained cyclic strength did not increased if the sand is not mixed with cement. 175
7 ACKNOWLEDGEMENTS
REFERENCE
The authors would like to express sincere appreciation to Messers K. Takanami and T. Kobayashi, for their great assistance during the tests.
The Japanese Geotechnical Society. 2000. Method for soil testing (in
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Landmarks in Earth Reinforcement,Ochiai et a1 (eds), 0 200 1 Swets & Zeitlinger, ISBN 90 265 1852 8
Practical experience in small scale pullout tests H. Zanzinger LGA, Ceosynthetic Institute, Nuremberg, Germany
E. Gartung LGA, Geotechnical Institute, Nuremberg, Germany
G.L. Sivakumar Babu Civil Engineering Department, Indian Institute of Science, Bangalore, India
ABSTRACT: Pullout response of geogrids in soil is a function of soil type, soil properties, normal stress, type of geogrid, specimen width and length as well as other boundary conditions. Extensive research has been executed in the past into the problem of soil-geogrid interaction, so the fundamentals are well understood. However, the determination of individual project specific design parameters for soil-geogrid interaction analyses cannot be executed with expensive, sophisticated scientific experiments. Simple, inexpensive, small scale pullout tests are recommended for this purpose. The paper points out a number of important details to be observed in routine testing and presents some test results obtained with different material combinations. ried out, modelling the mechanical performance of the members involved under field conditions. Since there is a great variety of GSY reinforcement applications and of different soils, it is necessary to determine the required length of reinforcement anchorage or of the overlap for each project individually unless there is past experience with conditions which fit. The most common way to establish the soil-reinforcement interaction in the limit state is by pullout or by direct shear testing. The present paper refers to practical experience with pullout tests.
1 INTRODUCTION In geotechnical applications reinforcing geosynthetics (GSY) are typically used as strengthening elements for embankment foundations, for supporting structures, for the stabilisation of slopes, as pavement reinforcement or for other assignments. The GSY reinforcement sheets are installed between more or less horizontally placed and compacted soil layers. They retain the soil and prevent it from slipping. For this purpose reinforcements are anchored in stable soil zones. The reliability of the anchorage is crucial for the stability of the structure and/or for its performance. Accordingly, the problem of soil-reinforcement interaction and safety assessment of anchorage has been given due consideration in theoretical and experimental studies and there arc numerous reports on this subject in the literature. A special question of soil-reinforcement interaction arises when extensive areas have to be covered with reinforcement sheets and when it is necessary to transmit forces across the boundaries of individual reinforcement sheets. One way of accomplishing this is by sewing the sheets together. The strength of the seam can then be determined experimentally by tensile tests and the ratio of the seam strength to the strength of the reinforcement can be entered into the design calculation as a reduction factor. Another way of transmitting forces across the boundaries of reinforcement sheets is by friction in overlap joints. To evaluate the interaction of the reinforcement sheets with each other and with the soil surrounding them, pullout tests can be car-
2 PULLOUT-TESTS The question about the required minimum anchoring length for the particular application can be answered by pullout tests with GSY specimens of different lengths. Reducing the length of the reinforcement specimen step by step, the length can be found where the GSY just starts slip. Alternatively, a number of wire extensometers can be fixed to the tested GSY specimen, in order to determine the loaddisplacement performance of the soil-reinforcement system and thus deduce the required anchoring length. Relatively large equipment is needed to carry out these tests. Typical dimensions for such pullout boxes are width B = 0.6 to 1.0 m and length L = 1.0 to 3.0 m (Adanur et al. 1996, Lopes & Ladeira 1996). Tests of this type are quite demanding. That is why they are carried out for research purposes or fundamental studies rather than for day to day construction projects. The costs of such pullout tests are too high for routine applications. 177
ever, it does not apply to geogrids with open spaces between the strands (Jones 1996). In shear tests, the resistance acting at a single GSY/soil or GSY/GSY interface is determined. In contrast, with pullout tests two interfaces between the GSY and the soil are involved. Shear tests on soil-geogrid interfaces do not characterise the pullout resistance correctly for the following reason: when shear tests are carried out on geogrids, the soil above the geogrid is only partially in contact with the surface of the geogrid, but probably to a greater extent with the soil beneath, due to the openings of the geogrid. Since the friction angle between soil and GSYs and the angle of internal friction of the soil may be quite different, the results of shear tests may be misleading when applied to the determination of minimum anchoring lengths of geogrid reinforcements. The test results of pullout tests which are discussed subsequently were executed on a routine basis for road and railroad construction projects.
Figure I . Set-up of a small scale pullout test
To meet practical demands, simpler tests as described in this paper are recommended out with smaller equipment. Shear boxes commonly used for shear tests on GSYs, with dimensions of B = L = 0.3 m, are employed for these routine pullout tests (Figure 1). In pullout tests of GSY specimens from the soil, two different failure mechanisms may occur. Under the applied tensile force the geogrid may either slip out of the soil or it may lose its integrity before slipping and tear. The normal surcharge loads are increased from test to test, as in direct shear testing and the limit tensile force is measured. It is important that the tensile force is introduced into the geogrid by proper clamping device. The free end of the geogrid projecting from the rear of the pullout box facilitates displacement measurements at the rear side. Displacements are also measured at the front side. So the average strain of the geogrid under load can be determined in the soil contact area. The pullout forces and the normal forces are measured. It is very important not only to measure the applied total normal load, but also to evaluate the actual normal stress in the plane of the GSY sheet during the entire pullout procedure. The results of pullout tests are commonly presented as plots of average pullout shear stress ‘I: vs. average normal stress G.
z = F / (2 e B 0 L)
(1)
o=N/(BoL)
(2)
with F = pullout force; B/L = width/length of tested specimen; and N = normal force (corrected for dilatancy).
3 DIFFERENCES BETWEEN DIRECT SHEAR TESTS AND PULLOUT TESTS It is sometimes stated that routine pullout tests are not needed, because the soil-reinforcement interaction can be determined adequately by direct shear tests. This is correct for woven geotextiles. How-
4 FACTORS DETERMINING THE PULLOUT RESISTANCE 4.1 Normal stresses The normal surcharge load is generally applied pneumatically or hydraulically on to the geogrid-soil interface via layers of 5 to 10 cm of soil of well defined compaction and water content conditions above and below the geogrid. The effect of the skin friction between the soil and the walls of the box is compensated by measurements of the load transducers at the bottom of the test set up. Figure 2 shows an example of a pullout test on a GSY embedded in medium sand. The grid spacing of 30 mm corresponds to the wave length of the waves that show up in the force-displacement curves recorded while the geogrid is pulled out of the soil at a constant rate. The first maximum value of the measured pullout force defines the pullout resistance under the particular normal stress. Since the test set up does not permit vertical movements of the upper and lower loading plates, considerable increase of normal stress due to dilatancy effects may be observed (Hayashi et al. 1996). Figure 3 presents an example of a pullout test on a geogrid of 30 mm grid spacing in medium sand. The previously mentioned waves related to the grid geometry can hardly be detected here, but depending on the magnitude of the normal stress an increase of the normal stresses up to about 50% is observed as soon as greater movements do occur. In most geogrid reinforcement applications dilation is not restrained, so the pullout test results have to be corrected accordingly. If the increase in normal stress due to restrained
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4.2 Geogrid tensile strength The pullout resistance can only be measured by the test discussed here, when it is smaller than the tensile strength of the geogrid under the pertinent test conditions. The purpose of the test is to determine the average composite action of the geogrid-soil system under particular conditions (type and state of soil, normal stress etc.). The test has to be prepared accordingly which may require some standard soil mechanics experiments in advance. Experience has shown that the distribution of the normal stress acting on the geogrid-soil interface is not uniform. Because of wall friction there is a stress concentration in the middle part. As a result, the central area of the geogrid strip experiences greater normal stresses than the edges which in turn leads to more efficient retention of the reinforcement in the middle part of the strip. The tensile load is therefore concentrated at the central zone of the test specimen. The edges of a wide strip show a tendency to yield first, while the central zone still holds, so the tensile forces are concentrated in the middle. If the strength of the geogrid is overcome, the central strands tear first. This should be taken into consideration in the preparation of the pullout testing programme. In some cases failure of wide geogrid specimens in tension was observed at only 50 to 60% of the nominal strength. This effect can be influenced in practice by testing specimens of smaller width. 4.3 Geogrid geometry The geometrical features of the GSY e.g. the grid spacing and the granular properties of the soil influence the pullout resistance. Figure 5 compares the results of pullout tests carried out on a geogrid with grid spacing of 20 mm with medium sand, grain size 0 to 2 mm and with crushed basalt rock, particle size 16 to 32 mm.
Figure 4. Pullout stress vs. normal stress (example knitted geogrid 30 mm spacing in medium sand)
dilation of up to 50% shown in Figure 3 would not be corrected for, the pullout resistance would be severely overestimated as can be seen on Figure 4. The stress correction for dilation produces outstanding betand leads to a linear ween pullout resistance and normal load as long as the pullout resistance does not exceed the tensile strength of the geogrid.
Figure 5 . Pullout stress of a knitted geogrid with 20 mm grid spacing in two different soil types
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In both cases the pullout resistance of the geogrid was overcome without the tensile strength of the geogrid being reached. In spite of very different soil particle sizes the frictional component of the pullout resistance appeared to be almost identical, but there is a “cohesion” intercept at the crushed rock line. The pullout resistance of the geogrid in coarse crushed basalt at zero normal stress is caused by interlocking. In the test with the much finer sand, no interlocking effect was observed. Obviously geometrical features, grain shape and the relationship between grain size and grid spacing have an important effect on pullout resistance under low normal stresses.
For any given geogrid the pullout strength depends on the type of soil and the compaction and water content conditions. As an further example, test results are presented for a woven polyester geogrid, tensile strength 80 kN/m in machine direction and 30 kN/m in cross machine direction with grid spacing of 20 mm. The tests are carried out with gravel, sand and clay soil, the graduation curves of which are shown an Figure 6. The pullout resistance for the geogrid embedded in gravel and in sand shows simi lar patterns (Figure 7) as the one for crushed rock
0006001 002
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5 OVERLAP JOINTS Pullout tests can also be used to determine the required overlap width of geogrids when it is necessary to transmit forces across the edges of geogrid sheets. Figure 8 compares the results of two pullout tests. The lower line presents the pullout stress of a single geogrid. The upper line shows the test results of a test with two geogrids of the same type, overlapping in such a way that the lateral strands of the geogrids can interlock. In this configuration the lateral strands of the geogrids that interlock, contribute substantially to the pullout resistance. The geogrid
4.4 Soil types
0001 0002
and sand (Figure 5 ) with a pronounced intercept at zero normal stress for the coarse grained material. Again, the line for sand is parallel to the line for the coarse grained soil like Figure 5. The pullout resistance measured for the stiff clay (water content w = 22%, dry density pd = 1.64 g/cm3) is lower than for the cohesionless soils. Apparently it is controlled by the shear strength of the clay and may depend on the drainage conditions as well as the velocity of pullout.
Figure 8. Pullout stress of a tensional overlapping seam of knitted geogrids
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Figure 6. Graduation curves of three soil types
Figure 7. Pullout stress of a geogrid with 20 mm grid spacing
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Figure 9. Influence of congruent and incongruent overlapseams on Dullout resistance of knitted rreozrids
Din.-
Figure 10. Geogrid overlaps, congruent placement at left, incongruent placement at right hand side
without an overlap joint but otherwise embedded in the same sand shows appreciably less pullout resistance. However, if the geogrid is installed in such a way that the lateral strands are not facing each other, so there is no interlocking, the smoother sides of the geogrid are in contact with each other, then a reduction of the pullout force is experienced. For the overlap joint this means, that the force to be transmitted by friction is smaller than in the case where interlocking occurs (Figure 9). When geogrids overlap, they may be placed in such a way that strands are lying on top of each other in a perfect match, they are congruent as depicted on the left hand side of Figure 10. But they may be placed in such a way, that the strands are not on top of each other, being incongruent, as shown on the right hand side of Figure 10. Test results (LGA 1995) on Figure 9 demonstrate that the pullout resistance differs slightly for these different placement conditions. For an assessment of the force that can be transmitted via overlap joints, the more conservative values of incongruent conditions apply.
6 APPLICABILITY TO OTHER GSY All geogrid pullout tests referred to in the present paper were carried out with knitted and woven geogrids. The manufacturing system for all geogrids was practically of the same type. Tensile strengths
ranged from 30 to 400 kN/m. Accordingly, the results are valid for these geogrids. A transfer of the experiences presented here to reinforcing GSY of other types are not justified.
7 SUMMARY AND CONCLUSIONS Small scale pullout testing of geogrid-soil systems, is a practicable tool for the determination of design parameters of GSY-reinforced soil structures on routine projects. The tests are carried out with the soils from the site in question at the pertinent compaction and moisture conditions. It is inappropriate to estimate the pullout resistance of geogrids on the basis of direct shear tests. Normal stresses acting on the geogrid-soil interfaces have to be corrected for dilation effects. The distribution of the normal stress in the test set up leads to a stress concentration and an associated concentration of tensile forces in the middle part of the test specimen. This should be considered in the preparation of the tests and the size and shape of the test specimens. Pullout tests can also be used to determine the transmission of tensile forces via overlap joints. REFERENCES Adanur, S., Mallik, S. & Zhai, H. 1996. Analysis of geotextile-soil interaction in pull-out tests. Proceedings of the International
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Symposium on Earth Reinforcement, Fukuoka, Kyushu, Japan, 12-14 November, Vol. I , Rotterdam: Balkema. 3-8. Hayashi, S., Alfaro, M.C. & Watanabe, K. 1996. Dilatancy effects of granular soil on the pullout resistance of strip reinforcements. Proceedings of the International Symposium on Earth Reinforcement, Fukuoka, Kyushu, Japan, 12-14 November, Vol. I , Rotterdam: Balkema. 39-44. Jones, C.J.F.P. 1996. Earth reinforcement and soil structures. Thomus Telford Publishing, London, 379 pp.
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LGA 1995. Test reports GE 9408972/A, GE 9408972/B and GE 9408972/C (unpublished) in German. Lopes, M.L. & Ladeira, M. 1996. Pull-out tests for the assessment of soil-geogrid interaction - Influence of some mechanical and physical parameters. Proceedings of the International Symposium on Earth Reinforcement, Fukuoka, Kyushu, Japan, 12-14 November, Vol. I , Rotterdam: Balkema. 89-94.
2 Embankments
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Railroads on piled embankments in Germany: Milestone projects Dimiter Alexiew HUESKER Synthetic GmbH & Co., Gescher, Germany
Wolfgang Vogel Deutsche Bahn AG, Munchen, Germany
ABSTRACT: Embankments on soft subsoil supported by piles and geogrid reinforcement on top of them have important advantages compared to "conventional" embankment foundation: no consolidation time is required (traffic can start immediately after construction), there is no import/export of additional embankment soil to accelerate consolidation or to compensate the settlement, practically no additional settlement occurs under traffic, the interference with the environment is minimized, etc. The application of such solutions is growing recently in Germany. The most important projects of this type being "milestones" for the German Railways (Deutsche Bahn, DB) with high-strength geogrids are presented, demonstrating the development of experience, materials and acceptance. All structures have been approved by the German Supervising Authorities. Long-term measurement results for the "oldest" project are presented also.
I INTRODUCTION Embankments on soft subsoil supported by pointbearing piles or columns have important advantages compared to embankments directly founded on soft soils (today typically embankments with highstrength geotextile in the base to ensure local and overall stability, with or without vertical strip drains to accelerate consolidation): due to the load transfer via the piles/columns into the firm sub-layers the soft subsoil has to carry only marginal additional loads, therefore no consolidation time is required (traffic can start immediately after construction), there is no import/export of additional embankment soil to accelerate consolidation or to compensate the settlement , practically no additional settlement even under traffic occurs if the system is correctly designed, the interference with the environment is minimized, etc. Generally different alternatives are available when setting the embankment on piles, starting with the oldest solution with a concrete slab on top of them. The cost-benefit analysis today results in most cases not in a RC-slab, but in a solution with one or two layers of horizontal high-strength geosynthetic reinforcement on top of the piles/columns (= the base of embankment), which may have a large pile spacing and small pile caps (or no caps at all). The use of last mentioned solution is growing recently in Germany. First concepts were analyzed about eight years ago. About seven years ago such systems were designed and then constructed for the first time for the German Railways (Deutsche Bahn, DB) for heavy loads and fast trains, followed by
others. High-strength biaxial and uniaxial geogrids from 150 kN/m to 800 kN/m ultimate tensile strength (UTS) have been used successfully. Shortand long-term measurement programs were performed and are still ongoing with positive results. The most important projects of reinforced embankments on piles/columns with high-strength geogrids for DB are presented, demonstrating the development of systems, materials, technologies and acceptance. DB as owner has very high requirements concerning safety and serviceability. All structures have been finally approved by the German Supervising Authorities also (Federal Railroad Agency, FRA). The most important long-term measurement results for the ,,oldest" executed project after more than 5 years under traffic are presented also, confirming the long-term stability and serviceability of the system.
2 GENERAL PRINCIPLES The principle of functioning and dimensioning is based on the fact that, after redistribution of stresses (similar to a 3D-arching) in the point-supported embankment body, the major portion of the stresses is transferred directly to the tops of the columns/piles, while the remaining portion (which would otherwise overstress the soft subsoil between the piles leading to local and/or global failure or unacceptable settlement differences) is absorbed by the membrane-type supporting effect of the horizontal reinforcement (Figure 1).
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cal relevance punching is not an issue for stronger reinforced systems. The spread forces (H-forces) perpendicular to the embankment axis below the slopes near their edges have to be overtaken by the reinforcement (no inclined columns for H-forces!). They can be calculated as described e.g. in British Standards Institution (BSI). BS 8006:1995 (1995) using a simplified approach on the safer side. The total active earth pressure between the crest and the base of the embankment is assigned to the geosynthetic layer perpendicular to the embankment axis, resulting in significant tensile forces. Other more precise procedures performed by the authors result in lower spreading forces. In the following chapters five "mile stone"projects for DB of the type mentioned will be shortly described with their most important focal points only. Due to the lack of place some differences in the dimensioning procedures and assumptions in the projects during the years can not be explained in detail herein. Generally, the "German procedure" mentioned above has been used; it is the only one which is being accepted by DB and FRA at the present time. Research and development were and are still ongoing in Germany, based on mathematical analysis (e. g. Alexiew 1996), on the experience today incl. monitoring (e. g. Alexiew & Gartung 1999), and on large-scaled tests (Kempfert et al. 1999). These activities result day by day in a better understanding of the structure's behavior and in optimised design.
Figure 1 . General principles of geosynthetic reinforced embankment on piles/columns.
The German analytical dimensioning procedure being applied with some modifications during the last years is described in Kempfert et al. (1997). It assumes a stress redistribution ("arching") according to Hewlett & Randolph (1988). Thus, it depends on both the embankment and pile geometry and the soil strength. This is one of the most important differences to a popular code, namely to British Standards Institution (BSI) BS 8006:1995 (1995), which (surprisingly) does not take into account the strength (say q) of the embankment soil. Another difference is that a portion of the upwardly directed reaction stress of the subsoil between the piles may be included in the final equation for reinforcementtension as counter-pressure thus reducing the requireed tensile forces. For that purpose a part of the undrained shear strength cu (or su ) of the soft subsoil can be used as described in Kempfert et al. (1997); this is not the case in BS 8006. Note: assuming an upward reaction stress for the entire design life (which is >100 years for permanent structures in Germany) i.e. reducing the calculated force in the geogrids could be risky, because e. g. a decrease of the ground water level even decades after construction could eliminate the upward counterpressure due to settlement of subsoil. Flexible geogrids could follow the settlement to some extend "keeping the contact" (which is by the way an important difference to a RC-slab), but the prognosis is very uneasy. Consequently, this assumption of counter-pressure is being analysed/judged by DB for every single project for the post-construction-stage design. The allowed total strain (short-term & creep) of the geosynthetic reinforcement over the service-life of the structure (>loo years for DB) is being limited by the DB to a maximum of 3% in railway structures, which is definitely a hard restriction on the safer side (e. g. BS 8006 (1995) allows 6%). Some additional special not only railroad-related issues (e. g. "pile-punching" in extremely low embankments) were discussed in Vogel (1995) and motivated a corresponding research in Germany to analyse and take into account the possible problem; meantime it has been done: in most cases of practi-
3 PROJECT WERDER-BRANDENBURG NEAR BERLIN This project was not the first one discussed and developed for and with DB (the analyses for the projects described in Chapter 4 herein started earlier), but it was the first one constructed and put into operation. During the years 1994 to 1995, the one-hundredyear-old railway line between Berlin and Magdeburg was upgraded to withstand a speed of 160 k d h and heavy loads. Soft organic soils were found on parts of the stretch between Werder and Brandenburg. To provide for the foundation of the railway embankment, a conventional total soil replacement was made only in the thinner (about up to 2 m) compressible soil layers. In sections in which the soft soil layers were thicker, a geogrid reinforced embankment on piles was proposed. At the time of deciding to accept the proposal, DB yet had no experience of geogrid reinforced embankment structures on piles as permanent foundation. The subsoil layers comprise organic soils having a depth from 2 m to more than 20 m and typically cu-values scattering from 10 kPa to 25 H a .
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Below, loose, uniformly sand is to be found, which transfers to dense sand or boulder clay as depth increases and can then serve as foundation soil for vertical pile loads. The ground water extends from close to ground surface to a depth of 2 m (Brand1 1994.). The organic soils have a low permeability and a low coefficient of consolidation. Very slender "ductile castiron" driven piles with precast 1.0 m x 1.0 m RCcaps on top were used in a square pattern of 1.9 m axially, with total lengths varying from 10 m to 30 m, while soft subsoil thickness varied from 4 m to 20 m. Both tracks were treated separately to allow to keep the traffic going. The respective working track was secured by a temporary sheet pile wall at the middle of the embankment (Figure 2). Three layers of 5 m wide biaxial geogrids were installed to ensure support and load transfer to the pile caps in both directions. The flexible geogrids (FortracC3 R 150/150-30) are made of high-tenacity coated polyester with low creep (important due to the strict long-term deformation limitations), having an ultimate short-term strength (UTS) of 150 kN/m in both directions and a corresponding ultimate strain of about 13%. Some of the layers were jointed in plant by special high-strength seems. At this time (in 1993-1994), the load bearing performance of the system "embankment soil geogrids - caps - piles - soft subsoil" could not be described definitively in mathematical terms in a "guaranteed way". Additionally for this project only 3 % total (i. e. short-term plus creep) strain of geogrids were allowed for 120 years design life. No difference was made between (short-term) construction stage, start of traffic and (long-term) operation stage, on the one hand, and the stage-corresponding time-dependent geogrid strains, on the other hand. It was a conservative requirement, resulting in a more or less conservative design. Very important issues were the QA of geogrids (control of the prescribed mechanical parameters,
Figure 2. Project Werder-Brandenburg: typical cross-section.
precision of installation etc.) and the QA of earthwork. The piled sections are since about six years in operation, being accompanied by an extensive monitoring program (see last chapter herein), and performing better (less maintenance work required) than the stretches founded on common soil replacement, see above. Both materials and structure have been approved by FRA also. 4 PROJECTS SOUTHERN BYPASS STENDAL The new high-speed rail link Hanover-Berlin for the ICE-trains (Inter City Express), having a speed up to 300 M h , had to bypass the city of Stendal in 1996 from the south, crossing areas with soft subsoil on two sections (designated as PfA 4.3 and PfA 4.6) of some hundred meters each. The soft clayey layers have a thickness of typically about 6 m to 8 m from the terrain to the firm sub-layers consisting of sandy gravel and gravelly sands. The subsoil strength varied in a wide range from cu= 15 to 25 kPa. First prelaminary studies and analyses including a geosynthetic-reinforced system on columns started in the early nineties. In both cases a geogridreinforced embankment on columns was found to be the optimal solution. Final design was performed more or less parallel to the project Werder-Brandenburg by different teams, but the construction started later. This project was a further step in application of such systems, using them for the foundation of a high-speed link now (300 km/h instead of 160 km/h for Werder-Brandenburg). The rails had to be set on a ballast bed. As vertical bearing elements cemented stone columns without pile caps in a triangular pattern and an axial spacing of about 1.8 m were chosen having a diameter of approx. 0.6 m and being founded in the firm sub-layers. The axial spacing was maximised for using the full column bearing capacity and saving costs. The most important difference between the sections PfA 4.3 and PfA 4.6 was the embankment height being about 2.5 m for PfA 4.3 and about 1.5 m (only!) for PfA 4.6, and the subsoil conditions, which were a bit better for PfA 4.3 despite the cu-scatter. The allowed total strain of geogrids was limited to 3% for 100 years for this projects also (see Werder-Brandenburg), but in that case this requirement seems to be sound (highspeed-trains !). For PfA 4.3 high-strength flexible uniaxial geogrids FortracC3 R 200/30-30, having an UTS of 200 kN/m in longitudinal direction (machine direction, MD) and an ultimate strain of about 12% were used. For PfA 4.6 the design calculations resulted in two layers of specially produced semibiaxial geogrids FortracC3 R 400/200-10 with 400 kN/m and 200 kN/m UTS in machine (MD) and cross-machine (CD) direction respectively (Figures 3 & 4). 187
remarkable in these cases. Both structures have been approved by FRA also.
5 PROJECT KORGRABEN (STATION RATHENOW) In this case the high-speed ICE-link Hanover-Berlin crosses in the region of the Rathenow rail station a longer area of soft subsoil with a thickness varying from 0.5 m to about 6 m from the terrain downward (old flat river bed filled by young soft sediments). The ground water level ranges from 2 m to 3 m below surface. The firm soil layer in depth consists of gravelly sands. At this segment of the link the rails had to be installed on an infinite concrete slab ("concrete slab track"), which is the most actual concept of DB for ICE-trains. All versions of this track system are more sensitive to settlements than the common ballast-bed system: thus, embankment and foundation deformations are rigorously restricted. An additional problem was the low railroad level equal to the terrain, dictating together with the high ground water level the thickness of the bearing structure. Consequently, the system chosen comprises a relatively thin geogrid-reinforced soil body (not really embankment, but a block embedded in the ground) set on lime-cement stabilized soil columns as vertical bearing elements founded in the gravelly sands (Figure 5). The columns are positioned in a semi-triangular pattern with about 1.6 m axial spacing and a diameter of approx. 0.6 m. The design analysis asked due to the extreme deformation restrictions for a geogrid providing a total strain of < 1.5% for hundred years design life. A specially produced uniaxial geogrid FortracO R 800/100-20 A with an UTS of 800 kN/m in longitudinal direction (MD) and an ultimate strain of only 3% was installed in two layers parallel and cross to the rail axis respectively (Figure 5). In this case no additional reinforcement for spreading forces was needed because the system is completely embedded in the surrounding soil.
Figure 4. Project Southern Bypass Stendal, segment "PfA 4.6"
The main reason for such a high-strength doublelayered reinforcement at PfA 4.6 was the small embankment height, resulting in an unfavorable stressdistribution i.e. in a low "pile efficiency" (Hewlett & Randolph 1988): a high portion of the stresses is not born directly by the columns, but have to be born first by the geogrids overbridging the columns. Further factors were the unfavorable subsoil parameters, generally the safety philosophy of DB, especially for ICE-trains, doubts about "punching" still available at that time (Vogel 1995) and, last but not least, the lack of really founded experience (design was performed in 1994-1995; the monitoring at WerderBrandenburg had just started, see above). Both sections have been instrumented with flexible horizontal inclinometers at the level of column tops and reinforcement, and geophones (acceleration gauges) for monitoring the static and dynamic behavior of the structure. After completion of the entire ICE-link Hanover-Berlin in Summer 1998, the performance of both systems was tested by trainpassing under increasing loads and velocities up to 330 km/h. Deformations, geogrid strains (derived from the horizontal inclinometers) and oscillation velocities were far below the allowed values. Last measurements after more than two years of ICE-traffic indicate very low deformations and geogrid strains (mostly < 1%). It seems (unfortunately) that the scatter of subsoil parameters mentioned above makes a precise back-analysis for PfA 4.3 and PfA 4.6 not realistic. Generally, the system resources seem to be
build up from top to bottom: superstructure: concrete slab track sandlgravelly sand crushed 0.7 rn geogrid lengthwise same fill 0.3m geogrid crosswise sand 0.1 rn
Figure 5. Project Korgraben (Rathenow station): cross-section
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(decomposed peat). The ground water reaches often the embankment toe. Local and overall stability was insufficient, serviceability doubtful. Therefore, the following work had to be done: first, widening of the crest of embankment from 10.0 m to 11.3 m to produce the required new standard profile for a two-track line, and second, increasing the stability of the embankment and its foundation taking into account a new train speed of 160 km/h. For ecological reasons widening of the embankment base was not allowed. The solution found to be optimal is depicted in Figure 6, which is selfexplaining. It comprises a partial new embankment on geogrids and cemented stone columns ("increasing global stability"), which has at the same time over-steep geogrid-reinforced "green" slopes ("widening the crest without widening the base"). For reinforcement on top of the columns highstrength uniaxial geogrids FortracO R 400/50-30 T (UTS 400 kN/m) and FortracO R 150/30-30 T (UTS 150 kN/rn) cross and parallel to the embankment axis respectively were used (the cross-geogrids are stronger due to the spreading forces, see Chapter 2). Due to the existing old part of embankment beneath reinforcement level and pre-consolidation some upward counter-pressure on geogrids in the space between columns was (carefully) assumed, reducing the calculated tensile forces (see Chapter 2). The slope reinforcement consists of different layers of lighter FortracO. According to EBGEO (1997) the dimensioning was based on Bishop's circle analysis and on block-sliding analyses. Construction was completed in Summer 1999, test drives by heavy trains were performed in Autumn 1999, regular traffic started in December 1999. Vertical and horizontal deformations and deflections are being measured. After three measurement sessions the registered values are negligible.
The structure was instrumented similar to the Southern Bypass Stendal Projects (see above). Immediately after completion in 1997 tests by a special very heavy oscillating equipment were performed, simulating ICE-train drives for many days, and varying frequencies and amplitudes due to the critical character of the structure (soft subsoil of varying depth, small thickness of reinforced earth block, high speeds and sensitive concrete slab track). Neither deformations, nor accelerations succeeded the allowable values. Additionally a temporary embankment was set on top of the system for 3 months simulating a train-equivalent surcharge. The maximum deflection of the geogrids between the columns was < 10 mm, and no deflections on the surface could be registered. In Summer 1998 the structure underwent additionally the same real train-drive procedures like the Southern Bypass Stendal. After that the materials and the system were finally approved by DB and FRA and put into operation.
6 PROJECT HARPER MUHLENBACH In 1998 DB arranged to extend the important westeast rail link between Uelzen and Stendal near the former boarder between West and East Germany. This railway line was first put into one-track-service in 1873 and was subsequently extended to two tracks between 1900 and 1906. In 1945, traffic was suspended in the occupation-zone border region. The Section 51 at "Harper Miihlenbach" of about 0.5 km length is located on an embankment up to 6.5 meters high and 10 m wide at the crest. There had been no traffic or maintenance work there for more than 50 years. It comprised insufficiently compacted sand with silty components on soft foundation soil
Figure 6. Project Harper Miihlenbach: concept, stages, Components
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Figure 7. Werder-Brandenburg: recent deformations of 1" geogrid just on top of pile caps (note the cap contours).
Figure 8. Werder-Brandenburg: recent deformations of 3rd geogrid 0.5 m above caps (note the smoothed contours).
It is the first DB-project combining a partial new "reinforced embankment on columns" and "geogrid reinforced slopes" to an integral structure. More details on design considerations, dimensioning procedures and construction techniques are reported in Alexiew et al. (2000). The combined structure has been approved by DB and FRA.
(see Figure 2 also), and, in fact, the actualshape of the geogrids and some tilting of the cap plates. (Note: the horizontal and vertical scales are very different!). It should be pointed out that the deformations depicted do not reach the ballast bed. REFERENCES Alexiew, D. 1996. Modified Redistribution of Tensile Forces of Biaxial Span Reinforcement Between the Pile Caps. HUESKER Synthetic, internal report (unpublished). Alexiew, D., Gartung, E. 1999. Geogrid Reinforced Railway Embankment on Piles - Monitoring 1994-1998. Proceedings Geosinteticos '99. I" South American Symposium on Geosynthetics: 403-4 1 1. Alexiew, D., Pohlmann, H., Lieberenz, K. 2000. Railroad Embankment with Reinforced Slopes and Base on Stone Columns. Proc. 2Ild European Conference on Geosyiithetics EiiroGeo2000, October 2000: 239-244. Bologna: Patron Edi tore. Brandl, H. 1994. ABS Berlin to Magdeburg (km 37.060 to km 59.4 19) Werder to Brandenburg section. Geotechnical report on soil characteristics (in German, unpublished). British Standards Institution BS 8006: 1995. 1995. Code of Practice for Strengthened/Reinforced Soils and Other Fills. London: BSI. EBGEO Emfehlungen fur Bewehrungen aus Geokunststoffen. 1997. (Recommendations for Reinforcement with Geosynthetics, in German). German Soc. Soil Mech. Geot. Eng. (ed.) Berlin: Ernst & Sohn. Hewlett, W.J., Randolph, M.F. 1988. Analysis of Piled Embankments. Ground Engineering, Vol.21, No.3: 12-18. Kempfert, H.-G., Stadel, M., Zaeske, D. 1997. Berechnung von geokunststoffbewehrten Tragschichten uber Pfahlelementen. Bautechnik, Vol. 74, Heft 12: 818-825. Kempfert, H.-G., Zaeske, D., Alexiew, D. 1999. Interactions in Reinforced Bearing Layers over Partially Supported Underground. Proceedings of the 12"' European Conference on Soil Mechanics arid Geotechnical Engineering: 15271532. Rotterdam: Balkema. Vogel, W. 1995. Application of a New Construction Method for the Substructures of Railroads. Analysis Report. Deutsche Bahiz AG (Gerinan Railroads) (in German, uiipublished).
7 WERDER-BRANDENBURG: MONITORING OF THE "OLDEST" PROJECT IN OPERATION At the time of deciding to accept the geogridreinforced embankment on piles for this project in 1994, DB as yet had no experience of geogridreinforced embankment constructions on piles. DB and FRA called for verification of the concept and certification of stability and serviceability for a monitoring program. It includes two comprehensively instrumented scientific measurement crosssections. The behavior of the structure has been systematically observed since 1994. A large quantity of vertical and horizontal inclinometers, deflectometers and resistive strain gauges on the geogrids have been installed. It is the most detailed, precise and long lasting measurement program worldwide for systems of the type discussed, being performed by the LGAGeotechnical Institute in Nuremberg. Meantime measurements are running since about 6 years under traffic. The static geogrid strains are max G 1.5% tending asymptotically to values < 2%, and their increase tends to zero, the max dynamic strains are < 0.02%. More detailed information including earlier stages, too, can be obtained e. g. from Alexiew & Gartung (1999). The long-term monitoring has confirmed the stability and serviceability of the structure Figures 7 & 8 (courtesy LGA) show typical results for the settlements in two different geogrid levels
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Final design of an overbridging for railways endangered by cavities at Groebers W. Ast Technical University of Applied Sciences, Biberach, Germany
J. Sobolewski HUESKER Synthetic GrnbH & Co. KG, Gescher, Germany
J. Haberland GLOETZL Gesellschaft fur Baumesstechnik mbH, Rheinstetten, Germany ABSTRACT: Sinkholes beneath roads and railways can cause serious difficulties in some regions in Germany. The introduction of high strength geosynthetic reinforcements and new polymers created better possibilities to reduce the foundation risk. The actually being under construction works railway node at Groebers includes 8 tracks and is about 800 m long and 120 m wide. This new railway section is situated in a postmining area. Two of the tracks belong to the high speed line (300 k d h ) . A new developed overbridging system consisting of warning layer, two orthogonal installed geogrid layers (ultimate tensile strength: 1200 kN/m) and a cement stabilized bearing layer (to achieve a stable arch) was built and tested in 1998. The results of this full-scale test were decisive for the approval by German Railways. Some results of the test and of static calculations performed by FEM for the final project will be presented in the paper.
1 INTRODUCTION The new railway node at Groebers, being constructed today, is located on an old post mining area prone to subsidence. The 8 tracks should be founded on an embankment with the width at the base of ca. 120 m and the height of 4-5 m. Two of these tracks will be used for the high speed line (300 k d h ) . The observed and predicted diameter of sink holes in this post mining area was estimated for the design purposes with up to 4,O m. Unfortunately, the location of the all possible existing cavities in the underground (depth of about 30 m under surface) is not known. During the first stage of the project many foundation methods ware analyzed (for example: deep foundation on piles, stiff reinforced concrete plate, dynamic compaction after deep excavation, etc.). All of these methods were too expensive or practical not acceptable. Finally, especially for this project developed concept was chosen. This complex concept is based on geosynthetic reinforced gravel cushion coupled with an arching effect in the upper positioned stabilized bearing layer (CSBL). The overbridging system for the Railway Node Groebers consists of two main protection measurements: -
injection of cement slurry in the cavities with the known position, for example: old pits or mine galleries (this phase of the project is now in the finishing stage)
-
construction of a special developed geosynthetic overbridging system including a warning layer on the base and a cement stabilized bearing layer above the geosynthetic reinforcements.
This system was designed for the duration of 60 years and must be able to overbridge sink holes with a diameter up to 4,O m 1 month long (i.e. during the injection works needed for filling the developed cavity). The integrated warning layer must be able to detect the position and the diameter of sink holes and to control the propagation of the occurred cavity. The required resolution of the designed warning system was defined as: 1 signal from 1 m2. The geosynthetic reinforced gravel cushion (GRGC) (founded on the warning layer) serve as a tie rod supporting the stable arch in the CSBL. The allowable deformation on the track level after the sink hole development was defined according to the German Railway Regulations modified as follows: - (Ask) I 1: 500 with: L - track interval (1500 mm), As I 3,O mm - allowable difference of settlements, (especially for this project modified value). The extension limits of the overbridging system in the cross-section and in the longitudinal axis were estimated basing on the results of the full-scale field test and results of static calculation. The rules of the limits of the system in the cross-section of the railway embankment are presented in the Figure 1.
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layer should be as well a drainage for the rainwater, especially during installation works.
2 DESIGNED OVERBRIDGING SYSTEM 2.1 Cement stabilized base layer In order to achieve a stable and weather resisting base for construction works the overbridging system will be founded on a cement stabilized base layer with the thickness 0,40 m, Figure 1 and Table 1. The excavated existing clayey silt (middle plastic, consistency: weak to stiff) will be reused with a cement amount up to 4,5 %. The cement stabilized base layer will be inclined in order to drain off the rainwater. 2.2 Leveling layer
2.3 Warning layer On the exact profiled surface of the leveling layer a warning layer (a special developed composite) will be installed in panels (max. 5 m x 40 m) and put together in situ to a large matrix, which will cover the whole foundation base. The warning layer consists of two non-woven in-between them an orthogonal grid (0,25 m x 0,25 m) of electrical wires (PTFEcoated) is fixed. The warning layer should be understood as a large matrix, in which the position and the state of each node will be held under control. For that purpose each wire will be separately connected to a computer controlled device. The warning system will be additionally supported by extensometers (glass fiber PP-coated). This special designed by GLOETZL warning layer should monitor the deformation of the embankment base. The change in the electrical resistance of the wires and their rupture will be exactly detected. The location of the defected node points will be positioned and displayed in a clear graphical form, like a topographical map
2.4 Ballasting layer In order to force the deflection of the warning layer simultaneous with the development of sink holes and to achieve the break of the wires as soon as possible, the warning layer will be ballasted by a gravel layer (0/16 mm) with the thickness of 0,30 m. 2.5 The first geogid layer (installed crosswise)
On the cement stabilized base, a leveling layer of gravel (0/16 mm) with the thickness of 0,lO m is provided. Aside from the leveling function, this
On top of the ballasting layer the first geogrid layer will be installed crosswise to the longitudinal axis of
Table 1. Requirements on parameters of soil layers of the embankment structure Designation
Soil art Proctor's density DPR
(5%) Blanket layer
KG 1-
103
Unit Weight % (kN/m3) 123
Angle of Friction
Parameters Cohesion Ch
'PI
(")
242
(kN/mZ)
Elasticity Module Eh (MN/mZ) 2 100
Water Permeability"' kF (m/s) I l,o 10-6 '
KG 2* 100 123 242 2 100 2 l,o 10-> SU/TL'* min. 4,5 % 100 1 22,5 2 35 2 150 2 350 k Pectacrete 100 Gravel 0/32, 100 523 242 2 100 2 1,o. lo-s KG 2") TL 918 062 Ballasting layer Gravel 0/16 100 223 242 2 100 2 l,o lo-s Leveling layer Gravel 0/16 100 123 242 2 100 2 1,o. lo-s Cement stabilized SU/TL** 97 122,5 135 2 150 2200 base layer min. 4,5 % Pectacrete - TL 91 8 062 - Technical terms of delivery of soil materials for railways structures (Requirements of German Railway) ** - SU/TL- clayey silt or silty clay, classification according to DIN 18 196 - kF- according to DIN 18 130 Frost protection layer Cement stabilized bearing layer Bedding layer
'
'
<>A
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outside the edge of the most external sink hole, the embankment with an overlap of 0,25 m. No conwhich should be overbridged by the first geogrid nections or overlaps are allowable in the unroll dilayer, Figure 1. rection, i.e. only one roll for the whole length in the given position of cross-section can be installed, max 2.6 Bedding layer roll length 150 m. To optimize the installation works and to minimize the usage of geosynthetic products, The first geogrid layer will be covered by the bedthe installation plans for each product were preding layer with the thickness of 25 cm consisting of pared. The configuration of the embankment is also round shaped gravel 0/32 mm and compacted to 100 presented in special prepared cross-sections, which % of Proctor's density. are positioned in distance of 25 m. According to static calculation and the performed 2.7 Second geogrid layer (installed longitudinal) field test the following requirements for the geogrid used were defined: The second geogrid layer will be installed in the ultimate tensile strength: Fk 21200 kN/m (md) longitudinal direction with cross overlaps 0,25 m. Fk 2100 kN/m (cmd) The length of the overlaps in the longitudinal direcEmax = 2,5 rt 0,5 % (md) elongation at break: tion (the main tension direction) was ordered to be Emax = 5 rt 1,0 % (cmd) not lower than 1I , 0 m (LA = 7,O m, required an(md) - machine direction chorage length (7,O m) and D =4,0 m, predicted di(cmd)- cross machine direction ameter of sink hole). Due to required high friction design tensile strength: F B2~500 kN/m (md) by between the overlapped geogrids (& = 37,s") an intotal elongation ct 2 1,7 % (for loading time t = 1 terlayer of gravel (0/32 mm) with the thickness of 10 month) cm was designed. increase of the elastic elongation after 105 loads cycles and load level 500 k 100 kN/m: A E ~ 2.8 Upper bedding layer -50,20 % (md) The second geogrid layer will be covered by a gravel mesh size: 2 10 mm. (0/32 mm) layer with the thickness of 30 cm. The These requirements fulfills a geogrid with ultimate upper surface of this layer presents the upper surface tensile strength 1200 kN/m - 100 kN/m made in of the GRGC. Unless mechanical function this layer (md) from ararnid and in (cmd) from PVA (polyviserves as a buffer against alkaline water, which can nyl alcohol). The similar product was already used escape in small amounts from the above lying in 1994 to protect the embankment of the Federal CSBL. Road B 180 near Eisleben in Germany against danger of subsidence (Alexiew 1998). 2.9 Cement stabilized bearing layer (CSBL) The anchorage length of the geogrid incl. the This layer has a very important function, because wrapped around part was estimated to be not less here a stable arch with defined geometry must be than LA= 7,O m. This anchorage length is required
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layer, on the both geogrid layers and in the cement stabilized bearing layer (CSBL): - 60 Displacement tranducers on 1. and 2. geogrid layers, for deflection and elongation of geogrid layers. The first geogrid layer with installed measurement devices is exemplaryly presented in Fig. 4 - 6 Plastic tubings for settlement measurements with horizontal inclinometer probe. Tubings are embedded in the interlayer of gravel on 1. and 2. geogrid layer, in the body and on the upper surface of CSBL - 120 earth pressure cells for biaxial stress measurements in different layers across the sinkhole - 8 triaxial geophone installed in different layers. The following stress, strain and settlements were monitored by GLOETZL (Ast, unpubl.): - settlement on the upper surface of frost protection layer and within the deeper layers of CSBL and the interlayers on the geogrids - compression stress in the cement stabilized bearing layer (CSBL) above the geogrid layers - compression stress at the edge of the cavity - tensile force in the anchorage zones of geogrid layers outside the cavity - velocity of vibration of the cement stabilized bearing layer.
formed itself and to be stable during the design time of 1 month. The requirements on the mechanical parameters of this layer are compiled in the Table l . The second important part of requirements relating to geometry is given below: - thickness in the active stress zones under tracks: min. 2,95 m, Figure 1 - the thickness outside the active stress zones under tracks: min. 1,95 m, for further details, Figure 1 . 2.10 Non-woven- separation layer On the top of CSBL or cement stabilized embankment a non-woven (PP, 350 g/mz) will be installed for separation and protection of the prepared surface of cement stabilized embankment. 2.1 1 Frost protection and blanket layers The requirements on these “classical layers “ of the railway tracks are based on the German Regulations, Table 1. 2.12 Track for high speed line First of all the tracks including the high speed line will be constructed as ballasted tracks for the speed up to 250 k d h . The option with non-ballasted track designed for the speed of 300 km/h was taken into account in the design.
3 FIELDTEST The objectives of the performed full-scale test were as follows (Ast & Watzlaw 1999): - proof of the serviceability the systems after the development of sink hole with the predicted diameter of 4,O m - verification of the bearing behavior and loadbearing capacity of the system, especially the geometry of the arch and the tensile forces in geogrids - verification of hard- and software of the warning layer. The full-scale test was located in the axis of the high speed line and was designed for the two planned tracks, Figure. 2. For the simulation of the cavity a rectangular pit with dimensions: 4,O m x 9,O x 1,50 m was excavated an twenty one water filled rubber cushion were installed on its bottom. The free space between cushions was filled with gravel 0/8 mm. Finally, the temporary sheet pile wall was removed, Figure 3. The construction works of the embankment (layer by layer) were supervised by EBA (Federal Railway Approval Office) and all delivered products were controlled by the supervisor according to DIN 18 200. During construction works many geotechnical measuring devices were installed above the warning 194
At the first stage of the test a cavity with the dimensions 4,O m x 4,O m as the result of the discharge of water from 9 rubber cushions was simulated. After 7 days from the development of the cavity, the static load with q = 55 kN/m2 on the top of embankment was applied and the dynamic loading test started. Finally, the cavity was enlarged to dimensions 4,O m x 9,O m and the dynamic loading test continued. Figure 5 presents the used equipment DYSTAFIT with the diameter of loading plate of 2,5 m and vibrating mass up to 12,000 kg, (Neidhart & Watzlaw 1998). The dynamic loading test was performed with frequencies of 10 - of 27,5 Hz, mainly with 27,5 Hz. The measured inducted velocity of vibration was up to 30 m d s . The corresponding dynamic maximal tensile force inducted by the equipment was not higher than 464 kN/m. After the enlarging of the cavity to dimensions: 4,O m x 9,O m, the protection system remained still stable. The energy transmitted into the embankment during the dynamic loading test was equal to the whole energy caused from passing trains during one months of operation of the both tracks, i.e. was equal to the energy transmitted in the embankment during the design time of the overbridging system. After finish of the dynamic loading
test the embankment was cut through in the crosssection to the bottom in order to verify the measured parameters and to establish the real geometry of the arch, Figure 6. Some results of the measured parameters are given in the Figure 7. They correspond very well with the predicted values of the static calculations. 4 SOME ASPECTS OF FINAL DESIGN The final design bases both on the requirements defined in the EBA Approval (Ast et al. in press) and the performed static calculation for the final geometry of the railways embankment. The mean objective of the FEM-calculations was to establish the geometry of stable arch, to verify the compression stress in the arch, to predict the deformation on the track level and to estimate the required tensile force of reinforcement. In reality the final design calculations present a more detailed repetition of the static calculations conducted in the previous stages of the project, (Ast, unpubl.). The static calculation were performed using PLAXIS 7.0. Some relevant crosssections, i.e. one and two tracks lines, in each case for few selected positions of sink hole, were examined. The reinforcement was modeled by an elastic band (width: 1 ,O m) with the stiffness module: - for immediately developed load, i.e. direct after opening in the base JK= 65,900 kN/m - for the long time loading (in this case I month) JL = 29.41 1 kN/m. The cavity was modeled as longitudinal gap with the width 4,O m, it means that the static calculations were performed for a planar problem. Principal the design based on DIN V 1054 - 100 (Edition 1995) and the German Recommendation, EBGEO 1997. The required design tensile strength of reinforcement was estimated for the predicted geometry of the stable arch using results of FEM-calculations, s. Figure 8 and 9. The principle of the estimation of the 195
After the examination of the static calculation and project documentation by EBA (Federal Railway Approval Office) the EBA - Building Permit for the start of construction works in the first Part (Field 1 and 2) at the end of 2000 was given. In this project about 250,000 m2 of the aramid geogrid with ultimate tensile strength of 1200 kN/m and about 80,000 m2 of warning layer by very strong requirements on quality will be installed. The quality management and the supervision plan of construction works and materials are very clearly defined.
5 CONCLUSION The developed and verified in the full-scale test protection system presents a new foundation method of railway embankments in area prone to subsidence. This combined system: a geogrid reinforced gravel cushion (GRGC) and a stable arch in a cement stabilized bearing layer CSBL) enables to overbridge sink holes by very small track deformations (allowable inclination of tracks due to settlement difference: up to 1500). The computer operated warning layer integrated in this system holds the foundation base under control and locates the position and geometry of developed sink holes. The during the study stage, full-scale test and final design developed procedures enable to construct railway embankments with the presented overbridging system as a safe an economic engineering structure. REFERENCES
tensile force in the reinforcement is presented in Figure 9. The obtained maximal values of the tensile force are following: F B ~ ,=K489 kN/m (short-time) and F B ~ ,=L374 kN/m (design time - 1 months loading time). For the check of the arch stability in CSBL the developed overbridging system (arch and tie member, Figure 9) was analyzed by FEM, using shear parameters of soil layers given in Table l , (Ast & Hubal, in press.) The max. inclination of the tracks for the serviceability state was estimated with 1500, i.e. equal to the above given allowable value.
Alexiew, D. (1997) Bridging a sink-hole by high-strength highmodulus geogrids, Geosynthetics '97, Long Beach, USA Ast, W. (1998a), unpubl. Felderprobung Groebers, Geotechnische Stellungnahme Ast, W. (1998b), unpubl. Sicherung von Bahnanlagen im Altbergbaugebiet bei Groebers gegen Erdfaelle; Sicherheitstheoretische Analyse Ast, W. & W. Watzlaw (1999). Eisenbahnbau in Altbergbaugebieten, Planungszwaenge- neue Bautechnik, 25. Lindauer Bauseniinar. Hrsg. Fachhochschule Biberach, Wissenschaft und Baupraxis, Bi berac h/Riss Ast, W. (2001), H. Hubal & P. Schollmeier, in press. Bewehrter Erdkoerper mit Erdfall Warnanlage fuer den Eisenbahnknoten Groebers, ETR, Darmstadt Ast, W. & H. Hubal (2001), in press. Geogitterbewehrter und zementstabilisierter Eisenbahnunterbau in einem Erdfallgebiet, KGEO 200 I , MuenchenEmpfehlungen f i r Bewehrungen aus Geokunststoffen-EBGEO. Dt. Ges. fur Geotechnik e. V. (DGGT), Berlin: Ernst, 1997 Neidhart, T. & Watzlaw, W. (1998). Ueberpruefung der dynamischen Untergrundstabilitaet und Optimierung von Bodenverbesserungsmassnahmen, Baugrundtagung, Stuttgart
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
A 112-m high geosynthetic-reinforced residual soil slope S.H. Chew, G.P. Karunaratne, S.A. Tan & Y.T. Seah National University of Singapore, Singapore
C.T. Ho, S.K. Lim, W.H. Ho, K.Q. Ho & J.Wei Housing and Development Board, Singapore
ABSTRACT: A 12-m high steep slope was to be cut in residual soil to make way for the construction of a new road in the Bukit Panjang district, Singapore. Numerous technical and environmental requirements imposed construction of a retaining structure for the resulting steep soil slope. Being in the vicinity of a nature park, the proposed structure should blend aesthetically with its environment. It, too, had to be inclined at about 60 degrees to create the needed space. Field instrumentation comprising total pressure cells, piezometers, standpipes, tensiometers, inclinometers and settlement devices were installed and monitored regularly. A large number of strain gauges were also installed on the geotextiles to monitor the strain development with construction and subsequent slope behavior. This paper presents the monitored construction, field pullout tests and preliminary short-term performance of the 12m high geosynthetics reinforced slope in residual soil.
1 INTRODUCTION
2 SITE DESCRIPTION
Geosynthetic-reinforced slope systems have become a popular choice over many traditional earthretaining systems. Recent research (e.g. Rao et al., 2000; Koerner, 1998) has shown realization of substantial cost savings with geosynthetic-reinforced systems when compared with reinforced concrete wall systems especially for tall wall heights more than 3m. In this project, a 12m high slope inclined at about 60 degrees to the horizontal and spanning over a length of 60m was proposed in conjunction with a road widening project associated with a nature park. A study conducted at the National University of Singapore (Chew et al., 1998) on the use of geotextiles in poorly draining residual soil revealed the feasibility of using composite geotextiles with adequate in-plane transmissivity to reinforce soil slopes in cohesive soils. Therefore, composite geotextiles with less-than-high-quality fill has a high potential for success. In the current project the excavated in-situ residual soil was considered as backfill to save cost and endorse waste minimization, while encouraging plant growth for the nature park, which creates an aesthetically pleasing “green” slope, and minimal environmental impact. In the course of construction, the geotextiles reinforcement as well as the soil slope was heavily instrumented for monitoring the behavior of the system during construction and in the long term. Field pullout tests were conducted and preliminary results from these observations are reported in this paper.
This project is situated in the Bukit Panjang district in Singapore (Figure 1) and the underlying soil profile is developed on Bukit Timah Granite formation. Bukit Timah Granite consists of igneous plutonic rocks, mainly granite, adamellite and granodiorite, while its top portion comprises heavily weathered residual soils. A thorough soil investigation was conducted prior to the commencement of the project. The soil borings revealed that the site consists of up to 3m of fill at the surface (RL 140.7m), a yellowish silty sand subsoil of about 6m, followed by 7m of reddish sandy silt with granite as the bedrock. The soil above the bedrock was considered to be residual soils evolved from weathering of granitic rock. In-
Figure 1. Geology of Singapore and the location of Bukit Panjang District.
197
stallation of standpipes showed that water level is found at a depth of about 9m (RLI 32m). Table 1 summarizes the effective shear strength parameter (c7 and +,) obtained from high quality consolidated drained test with undistributed soil samples.
density of modified AASHTO compaction test. After compaction, reinforcements were placed with the principal strength direction perpendicular to the face of the slope. This procedure was repeated for each successive layer of soil (See Figure 3). Around instruments lightweight compaction equipment was used to prevent damage. At the end of construction, instead of a face wrap, geosynthetics cells with turfing would be installed on the slope to minimize sloughing and erosion, and in addition to encourage greenery. Throughout the construction process, precautions have been taken to ensure proper handling of geosynthetics during delivery, storage and construction. Quality control testing has been emphasized. Sensitive instrumentation including stain gauges mounted on the geotextiles would be protected from damage due to construction.
Table 1. Soil and rock properties Soil Type Depth (m) Residual soil 0-3 Silty sand 3-9 Sandy silt 9-1 I 11-16 Granite >I6
(degrees) Cohesion, c’ (kPa). Organic humus soil 33 23 34 18 15 31
__
3 CONSTRUCTION SEQUENCE The designed slope, consisting of three 4m-high terraced reinforced slopes, is 12m in height, and spans over a length of 60m. The slope is l(h): 2(v) with berms of 1.5m width. The reinforcement used was a composite material consisting of a mechanically bonded non-woven geotextile reinforced by a series of uni-directional high strength polyester yarns. Three different grades of geotextile are employed in this project. The geotextile Type C with the highest tensile strength (200kN/m) is used in the lowest terrace, Type B (150kN/m) in the middle terrace, and Type A (75kN/m) on the top. Each terrace has 7 layers of geotextiles installed at a spacing of 600mm. A sketch of the proposed slope is shown in Figure 2. The existing slope was first excavated to level of the proposed road extension (RL 130.7). A filter drain with granite aggregates wrapped in a filter geotextile was then laid to grade with provision connection to an inclined filter at the back of the fill. The backfill was then placed and compacted with 60 Ton vibratory rollers. The backfill soil was compacted in lifts of 300mm to a minimum 90% dry
Figure 2. Schematic diagram of the proposed slope.
4 FIELD INSTRUMENTATION In this project, an extensive array of field instruments for monitoring ground conditions were installed and monitored regularly. Total pressure cells, pneumatic piezometers, standpipes and tensiometers were used to monitor soil pressure and pore pressure in the reinforced soil slope. Soil movements were captured with inclinometers and settlement plates. Furthermore, strain gauges also mounted onto the geotextiles to examine the strain development with construction and subsequent performance of the slope. Not only does the strain measurement in the geotextile allow the evaluation of the geotextile performance, but it also provides an alternative method of assessing the behavior of the reinforced slope. A total of 146 post-yield large-strain electrical resistance gauges, capable of reading strains up to 20%,
Figure 3. Laying of geotextiles at the base while protecting the cut slope in the front of the nature park.
198
were installed in the geotextile. The reliability of these gauges would depend on the effectiveness of the waterproofing barrier, the capacity to withstand construction stresses during backflling and the installation technique.
5 FIELD PULLOUT TESTS A series of field pullout tests were designed right from the outset of this project. Three pieces of geotextiles of 0.3 m wide and 3 m long were extensively instrumented with strain gauges and horizontal telltales, and were placed midway between two reinforcement layers. The main objective of this test was to evaluate the actual pullout capacity of the soilgeosynthetics interface. Many factors such as degree of compaction, soil properties, strain rate, in-situ water content and suction in the soil would affect this interface behavior. Hence, the interaction coefficients obtained from the in-situ tests can be used to evaluate the correlation between field and largescale pullout tests conducted concurrently in the laboratory using the same soils and reinforcements. The three series of pullout tests conducted covered the soil at its relatively dry condition (with back filter drain working properly) as well as the soil subjected to adverse ponding condition simulating clogged up inclined filter drain during the lifetime of the slope. All the field pullout tests were conducted with an overburden of about 2.5 m of backfill soil above the test pieces. A pullout resistance at this small overburden pressure will thus yield a conservative estimate of the ultimate capacity under actual site conditions where the overburden will be much larger. 6 RESULTS The construction of the wall is progressing at the time of this publication with the first terrace of 4m constructed and the three pullout tests completed. Therefore, the results presented here will be based on the progress made up to this stage. In the tropical climate of Singapore, one of the biggest concerns in slope construction is the high rainfall intensity and water table. Especially in this reinforced slope project where poorly draining residual soil is used, rainfall and groundwater are likely to pose a problem in terms of slope stability during and after construction. After the first 4m-high terrace was built, the slope experienced the tropical monsoon for 2 moths amounting to a rainfall of about 835mm. With the objective of studying the effect of excessive rainfall and high groundwater on the reinforced slope throughout this period, all instruments were monitored regularly.
Behind the reinforced slope, a water standpipe installed to a depth RL124.4 registered no water initially but after 2 months, water rose to 7.4m to RL131.8 where the base level of the slope is at RLl30.7. Within the geosynthetics reinforced slope, no water had been registered throughout the entire 2month period. This would probably be due to the effective drainage system that was constructed behind the slope. A total of 9 piezometers, pneumatic and vibrating wire type, installed at several levels all registered 0 kPa. However, tensiometers recorded an initial average reading of about 18 kPa of suction, which gradually increased to about 34 kPa of suction in the first month where weather was still relatively dry. At the end of the second month, the tensiometers registered about 23 kPa of suction. Settlement plates were placed to monitor the vertical displacement of the 4m terrace. An average of 8mm settlement was recorded during construction. The inclinometer behind the slope registered a movement of 20mm over the entire construction period of approximately 3 months. The inclinometers within the reinforced slope had registered about 5mm. These movements were within the allowable limits. Total pressure cells were placed at 2 levels in the reinforced slope. The total pressure cells at RL13 I .6, with an overburden of 3.lm of surcharge upon completion of the 4m terrace, registered 24 kPa, whereas at RL132.1, with a surcharge of 2.6m registered 36 kPa. One possible explanation for this phenomenon where the total pressure cells at a lower level (with a higher surcharge) registered a lower pressure than one at a higher level (with a lower surcharge), could be due to the non-uniform compaction of the soil at these two locations if not for the differences in soil stiffness. Due to the presence tensiometers and pore pressure transducers near the area where the total pressure cells are located, hand held roller compactor was employed which hindered proper compaction as compared with pneumatic rollers. The results obtained from the strain gauges are presented in Figure 4. Only selective layers of geotextile had been instrumented. It can be seen that the lower layers of geotextile experienced a higher strain increment where the maximum force was about 62 kN/m. Maximum strain occurred at about 2-4m from the face of the slope. The potential failure plane would follow the locus of the peak strain across the geotextile layers. Field pullout tests showed that 0.3m wide test specimens registered a displacement up to a distance of 1500mm within the soil slope. The maximum pullout loads exerted at the pullout clamp are found to be 70 kN and 55 kN for dry and wet conditions of soil respectively. A more complete analysis of these tests is out of the scope of this paper.
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last for a substantially long period. The survivability of the strain gauges depends significantly on the installation technique (which will determine the effectiveness of the moisture barrier and adhesive) and the level of construction stresses. An improved technique from the strain gauging technique proposed by Chew et al. (1999) had been adopted in this project. To date, 90% of the gauges had survived the three months of construction period. Preliminary pullout tests at low overburden pressure demonstrated that the pullout capacities not less than 233 kN/m and 183 kN/m can be realized with a 3m long geotextile for dry and 80% saturated residual soil respectively. The high strength and permeable geotextiles should be used in such soils. Figure 4. Strain development across geotextile
8 ACKNOWLEDGEMENTS 7 CONCLUSIONS The co-operation of Wee Poh Construction CO (Pte) Ltd, the main contractor, in carrying out the construction, instrumentation and pullout tests for the project is gratefully acknowledged. Assistance given by the specialist contractor Presscrete Engineering Pte Ltd for the successful pullout test is also acknowledged.
Despite the partial completion of the project, several conclusions could be drawn from the instrumentation and pullout tests. From the two months of intensive tropical rainfall, the usage of non-free-draining fills, which was a major concern since the dissipation of pore pressure is critical in slope stability, did not affect the pore pressure build up unnecessarily. When a geotextile reinforcement with high in-plane permeability is deployed together with a filter drain at the back of the slope, problems of high ground water could be alleviated. This observation is supported by the reasonably low pore pressure readings from the tensiometers throughout the entire twomonth period. The trend in the strain developed in the geotextile yields not only the force experienced in the geotextile (essential in assessing the breakage capacity of the geotextile); it is also a good indicator of the potential failure plane of the slope. A sub-objective of this project is also to study the long-term survivability of the electrical resistance gauges. Generally, strain gauges in the field do not
REFERENCES Chew, S.H., Tan, S.A., Loke, K.H., Delmas, P.H. & Ho, C.T. 1998. Large scale pullout tests of geotextiles in poorly draining soils. Proc. 6"' Intern. Corzf. or1 Geosynthetics, Atlanta, August 1998. 2: 821-824 Chew, S.H., Wong, W.K., Ng, C.C, Tan, S.A. & Karunaratne, G.P. 1999. Strain gauging geotextiles using external gauge attachment method. In Peter E Stevenson (ed.), Grips, Clamps, Clamping Techniques and Strain Measurement for Testing of Geosynthetics. ASTM STP 1379. 97-1 10. Koerner, R. M. 1998. USA study shows geosynthetics reinforced retaining wall costs are lowest. IGS News. 14(3):6-7. Rao, G.V. & Nirwal, R.K. 2000. Life cycle costing of geosynthetics reinforced soil walls. Proc. Geosynthetics Asia 2000, Kuala Lumpur, Malaysia, 2: 3 1-36.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Use of inclined test to assess stress mobilization of liner on slope J.P. Gourc, P. Villard & R. Reyes Ramirez Lirigm, Grenoble, France
N. Feki Faculte' des Sciences, Sfax, Turzisie
L. Brianqon & H. Girard Cemagref; Bordeaux, France
ABSTRACT: For liner systems on slopes, a separation of the functions of the different geosynthetics is generally proposed. The geomembrane acts as the sealing layer while a geotextile reinforces the stability of the soil veneer layer. This paper is dedicated to the key-issue : how to obtain from laboratory accurate friction interface relationships, since tensile force mobilization is very sensitive to the interfaces behaviour.
1 INTRODUCTION Composite liners systems are widely used for river banks, dams, reservoirs, landfill caps (Gourc et a1 1998). The watertightness fonction is provided by a geomembrane with a soil veneer as protection. But in the past, many local failures of this system were observed, due to the sliding of the soil layer on the geomembrane slip surface or tensile failure of the geomembrane due to the friction tangential stresses at the interface with the soil veneer (Girard et al 1990). The updated design suggests to separate the functions of the different geosynthetics : while the geomembrane (GM) acts as the sealing layer, a geotextile (GTX) of reinforcement ensures the stability of the cover soil. Systems with intermediate component (geospacer) for drainage are not considered here. However the distribution of forces within each component is complex and results mainly from the tensile stiffness (J) of the geosynthetics and frictional interaction between components. The authors previously presented a FEM approach adapted to this problem, and compared with experimental results obtained at several monitored sites : sloping side of the bottom barrier of a landfill (Villard et a1 1997, 1999 and 2000 for a large experimentation in progress). In the present time, the role of every component is clearly understood, but the evaluation for design of the tensile force actually mobilized in the different geosynthetics remains very difficult to predict, due to extreme sensitivity to the friction relationships. As in any problem of reinforcement, ultimate limit state is relatively easy to consider but serviceability limit state where elongation and relative displacements at the interface are predominant, a very accurate knowledge of the friction relationships and also laying of geosynthetics conditions are required. In this framework, the interpretation of the field ex-
perience of the figure 1 (Cahors, Brianqon 2002) is in progress on the sloping bank of a reservoir two different liner systems are monitored, bituminous geomembrane (GMb) for the trial P1 and polypropylene geomembrane (GMpp) for P2 associated to a geotextile non-woven reinforced by fibers (GTX). Forces in the geosynthetics (fixed at the top edge) and displacement are collected while loading the granular material layer up the slope (increasing Lc value). On figure 2, tensile forces in the geosynthetics are plotted versus Lc.The efficiency of the geotextile to sustain the soil cover weight is clearly demonstrated, since the reported tensile value in the two geomembranes is very low. On the other hand, the tensile force in the geotextile associated to the GMpp (P2)
Figure 1. Field experimentation: Tensile mobilization for rising cover soil and two different liner systems (this experhentation was realized thanks to the participation of the owner - ASF : Autoroutes du sud de la France- and of 2 producers- Bidim Geosynthetics ans Siplasr)
20 1
0
1
2
3
4
5
6
7
8
9
gref apparatus (Girard et al., 1990) and the Lirigm Grenoble Standard Inclined Plane (IPS), (Gourc et a1 1996, Lalarakotoson et a1 1998). In case of low normal stresses, the characterization of the liner interfaces falls outside the scope of the standard Shear Box (SB), generally used for a pressure range of 25 to 200 Wa. Then, the inclined plane apparatus is more appropriated. As the Lirigm IPS, the Cemagref IPH, provides during the progressive inclination j3 of the box, the monitoring of the tensile forces by clamping the sheets at their head and the monitoring of the displacement or force (if attached) on the upper box (figure 4). In addition the IPH allows to induce different hydraulic conditions on the interfaces. The surface of contact betyeen the upper box and the base geosynthetics is lm-. The test of reference with the upper box empty and attached at the top gives the evolution of the force Fo required to sustain the box alone (figure 5) :
10
Figure 2 . Field experimentation: Tensile mobilization in the geosynthetics for the two liner systems.
is twice as large as the force in the case of GMb (Pl). This is correlated without any doubt to the interface (GTX/GM) properties. The laboratory tests exhibited below are dedicated to these materials. So in our opinion in the present time it is reasonable to admit that, even if by inverse calculation it is possible to justify the tensile mobilization in the different components, geosynthetics actual tensile values are generally impossible to foretell with accuracy. As large scale experimentations are costly and heavy to implement, a joint research program was carried out, based on specific laboratory facilities and associated finite element code (Goliath), to predict the behaviour of liner systems on slopes.
Fo = Wb . sinp - Fr
(1) Fr corresponding to the residual friction of the box guides is relatively low, and Wb is the weight of the empty box (1.28 kN). Figure 10 exhibits the measures collected during a test likely to simulate partially the conditions of the large scale experimentation of the figures 1 and 2. The upper box is filled with sand (thickness 0.23 m, yh = 16.6 kN/m3) and the two geosynthetics are fixed ahead. The main features of the liner system behaviour are as following :
2 EXPERIMENTAL SIMULATION OF THE LINER SYSTEM SLOPE KINEMATICS
A specific inclined plane apparatus (figure 3) was designed at the Cemagref Bordeaux (BrianGon 2002) to characterize geosynthetics interfaces and to simulate sliding. This device, called Inclined Plane for hydraulic applications (IPH) is an evolution of the first CemaFigure 4. Monitoring in the Cemagref Inclined Plane (IPH).
Figure 5. Tangential active force during sliding of the box without soil (IPH).
Figure 3. Cemagref Inclined Plane facility (IPH).
202
-for an inclination less than 12.5", surprisingly there is no tensile mobilization of the geotextile (GTX) for sustaining the upper box. The assumed cause of this phenomenum is the initial laying of the geotextile without prestressing. The modelisation (figures 10 and 11) below will confirm this assumption. For p> 12.5", the increasing value of the tensile force in the geotextile during the plane tilting corresponds to sliding at the interface GTX /GMpp. The tensile force value in the GM is also increasing but remains very low. -Sliding of the upper box U is also increasing but remains under control until p = 30" corresponding to its global sliding. The limit equilibrium formula provides values of the friction angle Cpg for the two interfaces : tan $g = (Ws. sin 0
+ Fo) / Ws . cos p
Figure 8, for the GMWGTX interface, exhibits the level of repeatability of tests. A summary of the friction angles at the different interfaces and for the different devices is presented on table 1. A good compatibility is obtained between Table 1. Interface friction for different interfaces and different facilities.
Interface GMpp/GTX GMb/GTX Sand/GTX o/cosB (kPa)
SB 11.9" 15.6" /
>50
Lirigm Cemagref -~
IPS 16" 23" 39.3" 3.2
IPH 16,Y 23" 38" 3.9
(2)
With Fo (figure 5 ) and Ws soil weight=3.9kN/m
p = 12.5 p = 30"
GTX/GMpp GTX/sand
Cpg = 16.5" Cpg = 38"
In conclusion this experience provides values of the limit friction angle but the progressive mobilization of the shear stresses at the interfaces seems questionable. More informations could be got by a meticulous observations of standard tests as generally implemented for the design of liner systems on slope.
Figure 6. Friction tests between geomembrane and geotextile in the Lirigm Standard Shear Box (SB).
3 EXPERIMENTAL SIMULATION OF THE LINER SYSTEM SLOPE KINEMATICS Standard tests were carried out at the Lirigm on the materials considered above, both using an Incline Plane (IPS) and a shear box (SE) of (0.3 X 0.3 m2) . As previously indicated, Shear tests are performed under normal stresses higher than 25 !@a. So results are only indicative, since normal stresses in the conditions of figure 1 (field) and figure 3 (laboratory simulation) are lower than 5 kPa. However the present authors (Lalarakotoson et a1 1999) report compatibility between friction angle values obtained with SB and IPS ($g decreases with increasing 0). Results for GM/GTX interfaces are reported on the figure 6 and the table 1. For the same interfaces, tilting tests (IPS) were performed : GM is glued on the upper plate ( no upper box and so no residual friction due to the guides : at the slip angle, p = @g. In reference to the surprising behaviour observed on the test of figure 10, GTX which in the standard case is glued on the lower support, was only fixed ahead (figure 7) : before global sliding obtained for the same limit inclination, displacement U of the upper plate is significantly higher if the GTX is not glued (progressive tensile and shear mobilization : elongation and distortion of the GTX).
203
Figure 7. Lirigm Standard Inclined Plane (IPS): influence of the geotextile contact with the rigid support (GMppETX).
Figure 8. Repeatability of tests with the Standard Inclined Plane (E'S) on GMWGTX interface.
the two Inclined Planes. However it's worth noting that in the present time in the IP test, only limit friction angle for continuous sliding is used. The information corresponding to the progressive displacement before sliding is useless. Figure 8. Repeatability of tests with the Standard Inclined Plane (IF'S) on GMb/GTX interface.
4 NUMERICAL SIMULATION OF THE LINER SYSTEM SLOPE KINEMATICS A numerical approach of the test of the figure 10 is performed using the FEM code Goliath of Lirigm (Villard et a1 1999). The main characteristics of the model for geosynthetics and interfaces are recalled on figure 9 and table 2. The problem is arising of the Model 2 is an artefact to simulate wrinkles in the GTX before tilting. The compatibility is actually better for the tensile force in the GTX (and GMpp) and also for the upper box displacement (figure 1 1). Table 2. Main parameters of the model.
GTX GMPP Sand/GTX GTX/GMpp GMpp/support
J, Jz E, kN/m kN/m % 624 6 0.4 15
4%
UP
'
mm
38 16.5 20
Figure 11. Comparison between experimental (PH) and theoretical behaviours (model 2).
5 CONCLUSIONS Composite liners systems with geosynthetics are widely used on slopes. However the distribution of forces within each component is complex. The evaluation of friction interfaces relationships from Inclined Plane tests is efficient only at the sliding limit state. Displacement before sliding remains unknown and the application of high-performance numerical codes is problematic as long as significative progress will not be obtained on interface friction relationships.
10 1 2
Figure 9. Numerical approach (Lirigm): modelisation of geotextile tensile behaviour and friction behaviour.
Figure 10. Comparison between experimental (IPH) and theoretical behaviours (model 1).
REFERENCES Brianqon L. (2002). Stabiliti des dispositifs comportant des giosynthitiques sur pentes - Etude d partir du plan Incline. These Universitaire de Bordeaux (To be printing). CEN. European Standard - Geosynthetics - Determination of friction characteristics -Parts 1 and 2 :Shear BodInclined plane test. Ref. No prEN IS0 12957-2:2000E. Girard H.,Fischer S . and Alonso E. (1990). Problems of Friction Posed by the Use of Geomembranes on Dam Slopes Examples and Measurements. Geofextiles and Geomembranes, 1990, vol. 9, p. 129-143. Gourc J.P., Lalarakotoson S., Muller-Rochholz H. and Bronstein (1996). Friction Measurement by direct Shearing or Tilting Process-Development of an European Standart. Proceeding of Eurogeol, 1I' European Geosynthetics Conference, Masstricht, Netherlands, Oct. 1996, pp. 1039-1046. Gourc J.P., Villard P. and Thomas S . (1998). Geomechanical aspects of the use of geosynthetics in landfill. Third International Congress on Environmental Geotechnics, Lisbao, 7-11 September 1998, Portugal, pp. 1371-1374. Lalarakotoson S.,Villard P. and Gourc J.P. (1998). Geosynthetic Lining Systems Reinforcement by Geosynthetic Inclusion". 6th International Conference on Geosynthetics, Atlanta, March 1998, Vol. 1, Georgie USA, pp. 487490. Lalarakotoson S., Villard P. and Gourc J.P. (1999). Geotechnical Testing Journal. Shear Strength Characterization of Geosynthetic Interfaces On Inclined Planes. GTJODJ, Vol. 22, No 4, December 1999, pp. 284-291. Villard P., Gourc J.P. and Feki N. (1997). Anchorage strength and slope stability of a landfill liner. Geosynthetics'97 Conference Proceeding, Vol. I , Long Beach, California, USA, March 1997, pp. 453-466.
204
Villard P., Gourc J.P. and Feki N.(1999). Geotextiles and Geomembmnes. Analysis of Geosynthetic Lining Systems (GLS) Undergoing Large Deformations. Vol. 17, No 1, February 1999, pp. 17-32.
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Villard P., Gourc J.P., Reyes Ramirez R., Thomas S . and Nesma F. (2000). Stability of different inclined cap liner systems-landfill field trials. EuroGeo2000, Proceedings of the 2"dEuropean Geosynthetics Conference, Bologna, Italy, 15-18 October 2000, Vol. 2, pp. 523-526.
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Performance of geotextile reinforced slopes subjected to seepage flow A. Hiro-oka, M. Kobayashi, H. Nagase & K. Shimizu Department of civil engineering, Kyusyu Institute of Technology,Japan
H. Fujiwara Sewage WorksBureau, Fukuoka Municipal Government (Formerly Kyushu Institute of Technology),Japan
ABSTRACT: A series of centrifuge model tests were carried out to investigate the behavior of slopes reinforced with geotextile subjected to seepage flow, and to evaluate the contribution of reinforcement to the stability of these slopes. It was found that the smaller shear deformation in the toe area of embankment and the smaller settlement and fewer crack at the top of embankment were observed with the narrower reinforcement spacing. The factor of safety obtained by means of limit equilibrium could assess the condition of a reinforced slope to a certain extent; however, it could not accurately evaluate the stability of embankments for the reason that in those stability analyses the deformation of embankment that is required to produce the reinforcement force of geotextile was not taken into account. As part of investigation in this study, the model tests were conducted with the viscous fluid in order to check the similitude in seepage phenomenon between the prototype and the model. 1 INTRODUCTION
A lot of slopes fail due to a long or heavy rain in Japan during rainy or typhoon season. Therefore, many researchers have studied on the failure mechanism of slopes due to rainfall and prediction methods of slope failure. Takemura et al. (1994) investigated the effectiveness of centrifuge model test using 1/50 scale model of embankment which object of large scale model tests conducted by Sugiyama et al. (1993), subjected to seepage flow in 50g centrifugal acceleration field. Kimura et al. (1991) carried out a series of centrifuge model tests to investigate the failure mechanism of fills subjected to various intensity of rainfall. Hiro-oka et al. (1999) studied the influence of soil shear strength and inclination of slopes on mechanism of failure due to seepage flow. Various types of grand flows in centrifuge model were studied by Goodings (1994). In many cases of these slope failures, the decrease of the slope stability was caused by the reduction of the soil strength due to the increase of its degree of saturation or the decrease of the effective stress due to the rise of ground water level in the slope. The reinforcement with geotextiles has been often adopted to increase the stability of slope. The characteristics of friction between soil and geotextile were examined by many researchers (e.g. Ingold, 1982 and Myles, 1982). Zornberg et al. (1998) conducted a series of centrifuge model tests to investigate the performance of slopes reinforced by two-types of geosynthetics and various reinforcement spacing and two-types of back fill conditions, and to evaluate the stability of
slopes by means of limit equilibrium. However, the performance of geotextile reinforced slopes subjected to seepage flow has been not sufficiently understood because the mobilized reinforcement forces in addition to tensile force acted on geotextile itself were too much complicated under such condition. In this study, a series of centrifuge model tests were carried out to investigate the behavior of slopes reinforced with geotextile subjected to seepage flow, and to check the similitude in seepage phenomenon between prototype and model in the case with the viscous fluid. In order to evaluate the stability of reinforced slopes, after making clear the friction between soil and geotextile was examined by means of direct shear test, the limit equilibrium analyses were also carried out. 2 MATERIAL PROPERTIES 2.1 Soil material The artificial mixture, which consists of 50% of silty sand sampled in Kyushu Institute of Technology and 50% of Toyoura sand, and it classified as SF and the properties are shown Table 1. The internal friction angle and cohesion of soil obtained by direct shear test were also shown in Table 1. 2.2 Geotextile The geotextile shown in Figure 1 was installed in the model embankment. It is non-woven fabric that consists with 100% polyester and 0.06 mm thickness. 201
Table 1. Physical properties and strength parameters of artificial mixture. Specific gravity, Gs Effective grain size, Dlo(mm) Uniformity coefficient, U, Sand fraction (%) Silt fraction (%) Clay fraction under 5 p (%) Maximum dry density, pdmax (g/cm3) Optimum moisture content, woPl(%)
2.67 0.13 14.0 67.5
25.0 7.50 1.85 16.7
Dry density, pd(g/cm3) Degree of compaction (%) Coefficient of permeability, k (x 103c d s ) Initial condition Cohesion (Wa) w=IO(%) Angle of friction (degree) Cohesion (kPa) Sr=lOO(%) Angle of friction (degree)
Wide-width strip test under unconfined condition was conducted to examine the tensile strength of geotextile with 200 mm wide and 100 mm long specimen, which was based on the size of IS0 10319 tensile test. The unconfined ultimate tensile strength and the strain breakage were 0.27 kN/m and 13 % respectively. The tensile strength converted to prototype was 13.5 kN/m if it used at 50g centrifugal acceleration field.
3 DIRECT SHEAR TESTS 3.1 Test procedures and conditions The direct shear tests were carried out to investigate the soil shear strength and frictional properties between soil and geotextile. Figure 2 shows the test setup with soil-geotextile shear tests. Both of shear boxes was 6 cm diameter and 1 cm depth. The porous stone, on which the filter paper and geotextile were pasted, was placed in the lower box, for the purpose of corresponding the shear surface with contact surface between soil and geotextile. The mixture with 10 % water contents, was threw into the upper shear box and compacted with dry density pd=1.40
1.40 75 3.88 20.8 33.0 4.50 33.0
g/cm3. The shear tests were carried out under displacement-controlled conditions that shearing at the rate of 0.25 &min. In the case with saturated conditions, the shear boxes were covered with vinyl bag, which filled with water, in order to keep specimen saturated during shear tests. The shear was continued until horizontal displacement reached 7 mm or shear force became constant. 3.2 Test results and discussions Figure 3 shows the relationships between shear stress and effective stress obtained by direct shear tests. According to this figure, in the case with soil the internal friction angle are identical to each other condition, though the cohesion decreases as increasing the degree of saturation. On the other hands, in this figure, it is pointed out that the friction angle increases with installed the geotextile and cohesion decreases, irrespective of degree of saturation. It may be the influence of the roughness of the geotextile. It is considered that the geotextile made decrease the suction of soil at contact surface and it cause to reduce the cohesion, and increasing friction angle may be caused by the geotextile firmed the engagement with soil. However, the detail discussion could not present, because the frictional properties were depend on the combination of geotextile and soil and only one combination was investigated in this study.
Geotextile (Non-woven fabric) 20 X 2 0 (mm)
Figure 1. Geotextile.
Figure 2. Direct shear tests.
Figure 3. Relationships between shear stress and effective stress.
208
Table 2. Test conditions. Test code
Type of geotextile
Tensile strength (lolNrn)
Reinforcement spacing (cm)
Non-woven fabric F Non-woven fabric F Non-woven fabric F
0.93 (46.5) 0.93 (46.5) 0.93 (46.5)
5.8 (290) 5.8 (290) 2.9 (145)
so sov S3F s3Fv S6F
Figure 4(a). Test set up for the cases with water.
Pore fluid Water Viscous fluid Water Viscous fluid Water
Figure 4(b). Test setup for the cases with viscous fluid.
corresponding to the radial variation of the centrifugal acceleration. At the toe area, the coarse sand, which was covered with non-woven fabric, were placed as drainage. In order to reduce the friction between soil-mass and box to minimum, the inside walls and a front window was well-lubricate with silicon grease. The white Kaolin clay are painted onto the models using a template to improve observing deformation of embankments in flight. A pore water pressure transducer was placed on the bottom of water supply tank, and 2 linear variable displacement transducers (LVDTs) were placed over the top of embankment for measuring its settlement. After completion of model embankment, the test setup was mounted on the K.I.T. Centrifuge and centrifugation was conducted at 50g. Supplying water to the water tank behind the model embankment from laboratory floor through hydraulic slip ring at a rate of 5 m d m i n to carried out seepage tests and the tests were continued until slope failed clearly or the water level reached the total height of embankment. In the case with viscous fluid, although rate of supplying fluid in the water tank was identical to another series, water level in the tank has not kept increasing during seepage tests because the fluid has too much viscosity. Hence, the water level was raised 5 cm at once and left this condition alone until the output of pore pressure transducer installed model embankment was assumed to be steady. To determine the ground water level at a lower stream, the drainage was made by holing the backside wall of model container at the height of base part of embankment.
4 CENTRIFUGE MODEL TESTS 4.1 Test procedures and condition Table 2 presents the test condition in this study. Tensile strength and reinforcement spacing in the bracket in this table shows the value converted to proto type scale. The model embankment was built in a steel-made box with 450 mm in length, 350mm in depth and 150mm in width, which has a water tank inside for supplying water to model embankment. The test setup is illustrated in Figure 4(a). To construct a base part of embankment, the mixture was placed in a model container and compacted by a bello-fram cylinder into a 2.0cm thick layer with a predetermined dry density, pd=l.40 g/cm3, and this was continue until the total thickness reached 10 cm. On the other hand, in the case with viscous fluid, a concrete block was placed in the model container as a base part shown in Figure 4(b). 3 pore water pressure transducers were placed on the top of the base part of a model embankment that corresponds to the shoulder, center and toe of a slope. After this work, the mixture was threw into the model container again and compacted by a bello-fram cylinder with 2.9cm thickness. The geotextile was installed on the 2.9 cm or 5.8cm thick layer of model ground in the case with reinforced slopes. These works were continued until the total thickness reached 30.4 cm and then the layer was cut into a model embankment with 176 mm height by using a tempIate so that the surface of model embankment, and base ground which consist with soil in itself, have a curvature 209
4.2 Test Results and Discussions In the cases with non-reinforcement and water, the catastrophic failure, which occurred along a large well-defined slip surface suddenly, was observed as shown in Figure 5, however, except for this case, model embankments were not failed catastrophically when the ground water level at the upper stream of embankment reached the limit height. Hence, the performances of embankment until test stopped were discussed below. Relationships between the settlements of crest of embankments (LVDT1) and piezometoric heads at the water supply tank (PPl) in the case with water were illustrated in Figure 6. The arrows in this figure indicate when the crack was observed at the top of embankments in each case. It is pointed out that both of the behaviors of the settlement until the water level in the tank reached about 10 cm are similar. It indicates that the added tensile resistance by the reinforcement requires its deforming to some extent in order to make the slope stability increase. In the case SO, higher rate of settlement with increasing head at the tank was observed than in other two cases. The stability of this slope decreased due to increasing the soil mass with rising ground water level and decreasing the shear strength with increasing the degree of saturation and these caused the lateral deformation of embankment with developing the shear deformation at toe area. Subsequently the tension crack was observed at the crest of embankment as shown in Figure 7. It is also pointed out that the smaller shear deformation in the toe area of embankment and smaller settlement and fewer cracks at the top of embankment are observed with the narrower reinforcement spacing. The reinforcement force well contributed to the stability of the slopes. Figure 8 shows the relationships between the settlements at the top of embankment and heads at the water supply tank in the cases with viscous fluid. It is found that the settlements with raising ground water level are almost similar in SOV and S3FV. It indicates that there is no effect of reinforcement for decreasing the settlement, however, in Figure 7, which showed the observed deformation of embankment during seepage, smaller shear deformation in the toe area was observed with reinforced embankment. It implies that the tensile resistance was contributes the stability of slope. Figure 9 is comparing the profiles of the ground water surfaces recorded by photographing during seepage tests for each case. It is found that the lower ground water level at the toe area is observed in the case S3FV. It is considered that the transmitivity of non-woven fabric made increase the permeability of embankment at the stratum of installed the geotextile, especially, the lowest geotextile made the effect of toe drainage increase. Furthermore, it is pointed out that the ground water level at the toe area against the up-
210
Figure 5. Final collapse of embankment (SO)
Figure 6. Relationships between the settlement at the top of embankment and head at the water supply tank.
Figure 7. Observed deformation.
per stream area in the cases with viscous fluid was much lower than the cases with water. It may be the influence of the viscosity of pore fluid. It is considered that the ground water surface in the model embankments determined by water level at the upper
Figure 8. Relationships between the settlement at the top of embankment and head at the water supply tank.
stream if the fill consisted of well-infiltrated material or the pore fluid had less viscosity. On the other hands, the fill material had the less infiltration or pore fluid had the more viscosity, the ground water level at the lower stream might have the more effect on the profiles of ground surface in the embankment. Thus, the ground water surface in this test series was not in agreement with each other.
Figure 9. Variation of ground water surface of embankment
5 STABILITY ANALYSIS 5.1 Calculation method On varying conditions of ground water surface shown in Figure 9, stability analyses were attempted by means of Modified Fellenius method. In these analyses, angle of shear resistance and apparent cohesion listed in Figure 3 were adopted and calculations were made for the geometry condition just at the moment when the base layer was saturated shown in Figure 9. The ground water surface just before failure in the case SO was adopted in S3F and S6F in order to evaluate the reinforcement force. On the other hand, in the case with viscous fluid, the ground water surfaces just before the end of test were adopted in the calculations. According to Figure 10, the factor of safety was calculated by following formula,
F , = ( E M R +ZAM,)lXM,
Figure 10. Evaluation of reinforcement.
der to decide whether geotextile pull out or break at failure, the minimum value of frictional resistance and tensile strength of geotextile Tf are compared. The minimum value of those was added to the moment of resisting slope failure by reinforcement MR as followed formula,
(1)
here, MR and MO, MR means the moments of resisting slope failure and the moments of driving slope failure, and the moments of resisting slope failure by reinforcement respectively. The evaluations of reinforcements are explained in detail below. On the assumption that the frictional resistance acted on both of surface of geotextile, to determine the direction of pull-out, the sums of frictional resistance R/j corresponding to the effective overburdenpressure inside the slipping soil-mass and Rzj corresponding to the one outside it were calculated. In or-
here, i means a number of slice and W means a weight of slice. The cracks observed at the top of embankment in the case SO and S3 were taken into account in the analysis, which on the assumption that the one crack reach the uppermost geotextile in the case S3 each other.
21 I
5.2 Results and discussion
6 CONCLUSIONS
The minimum safety factors and critical circles in several cases were illustrated in Figure 11 and 12. It is pointed out that the critical circles in the case SO and S3F are similar each other. It implies that the stability of embankment in S3F decreased to a certain extent though the catastrophic failure was not observed in this case. In the case S6F, it is found that the critical circle is largest and crosses the uppermost geotextile near its edge. It suggests that the internal stability of this embankment is sufficient. In the cases with viscous fluid, it is found that these embankments were still stable according to the values minimum safety factor. On the critical circle of S3FV, its excluding from the reinforcement area of embankment indicates the stability of embankment was enough like a case S6F. It may be caused by the profiles of ground water surface in the embankments, especially, the ground water level at the toe are of the slopes. It implies that to keep the water level at the toe area low make the stability of embankment subjected to seepage flow increase.
A series of centrifuge model tests were conducted to investigate the performance of geotextile reinforced slopes subjected to seepage flow and to try the evaluation of stability of reinforced embankment by means of limit equilibrium. Following conclusions were obtained. A smaller shear deformation in the toe area of embankment and smaller settlement and fewer cracks at the crest of embankment were observed with the narrower reinforcement spacing of reinforced slopes subjected to seepage flow. However, the added tensile resistance by the reinforcement requires its deforming to some extent in order to make the slope stability increase. The factors of safety obtained by means of limit equilibrium could assess the condition of a reinforced slope to a certain extent. However, it could not accurately evaluate the stability of embankments for the reason that in those stability analysis the deformation of embankment that is required to produce the reinforcement force was not taken into account. REFERENCES
Goodings, D.J. 1994. Implication of changes in seepage flow regimes for centrifuge models. Proc. of Centrifuge’94 393398, Rotterdam: Balkema. Hiro-oka, A., M.Kobayashi, H.Nagase & K.Shimizu 1999. Centrifugal modeling of slope failure due to seepage flow. Proc. Symp. Recent Development of Theory and Practice in Geotechnology: 150-157. Ingold, T.S. 1982. Some observation on the laboratory measurement of soil-geotextile bond. Geotechnical Testing Journal, Vo1.5, N o 3 4 57-67. Kimura, T., J.Takemura, N.Suemasa & A.Hiro-oka 1991. Failure of fill due to rain fall. Proc, of Centrifuge’91: 509-516, Rotterdam: Balkema. Myles, B. 1982. Assessment of soil fabric friction by means of Figure I 1. Critical circles and minimum safety factors in water shear. Second International Conference on Geotextiles, Las series. Vegas: 787-79 I . Sugiyama, T., H.Muranishi, K.Kagawa, K.Kusano & K. Mizushima 1993. Estimating the timings collapse of embankment slope based on experiment of large scale model. Proc. Annual Meeting JSSMFE, 2: 2167-2168. Takemura, J., T.Kimura, A.Hiro-oka & H.Muranishi 1994. Failure of embankments due to seepage flows and its countermeasure. Proc. of Centrifuge’94: 575-580, Rotterdam: Balkema. Zomberg, J.G., N.Sitar, J.K.Mitchel1 1998. Performance of geosynthetic reinforced slopes at failure. Journal of Geotechnical and Geoenvironmental engineering, Journal of Geotechnical Engineering and Geoenvironmental Engineering, Vol. 124, No.8: 670-683. Zomberg, J.G., N.Sitar, J.K.Mitchell 1998. Limit equilibrium as basis for design of geosynthetic reinforced slopes. Journal of Geotechnical and Geoenvironmental engineering, Joumal of Geotechnical Engineering and Geoenvironmental Engineering, Vol. 124, No.8: 684-698. Figure 12. Critical circles and minimum safety factors in viscous fluid series.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 902651 863 3
Drainage effect of geosynthetics in high and very wet embankment S. It0 & Y. Yokota Maeda Kousen CO., Ltd, Japan
M. Inagaki, A. Morikage & Y. Kumagai Magara Construction CO., Ltd, Japan
K. Kawamura Kanazawa Institute of Technology, Japan
ABSTRACT: Construction a high embankment using high water content clay, a belt drainage material is used as a drainage system in the embankment. And placing geosynthetics obtains a good result due to the accelerated consolidation effect of the drainage material. Based on the field observation results so far, horizontal displacements both during and after the construction period were found to be significantly small compared to the vertical displacements, and the effect on confining the displacement of the embankment (reinforcing effect) was considered to exist. Then a plate-type strain meter was set on the belt drainage material placed in the embankment, the stress generated in the drainage material was measured, and the reinforcing effect was confirmed. In addition, displacement measurement at the surface of the embankment by the inver wire extensometer , and field displacement measurement in the embankment by the inclinometer, and the settlement plate, were carried out. This paper reports on the discussion on the drainage effect and the reinforcing effect of the belt drainage material placed in a staggered arrangement in the embankment, from the measurement results. Here describes the drainage effect and the reinforcing effect of the belt drainage material placed in a staggered arrangement in the embankment.
1 INTRODUCTION When constructing a high embankment using high water content clay, countermeasure shall be taken in order to stabilize the embankment. One of the countermeasure is to placing geosynthetics having drainage mechanism in the embankment. As this method has an advantage of utilizing low quality materials generated at site for embankment, it can be said as a effective method field observation and environmental points of view. In actual construction, two belt drainage materials having different properties in permeability and stiffness were placed in an about 20 meter high embankment, and field observation of the embankment was carried out.
2 OVERVIEW OF MEASUREMENT 2.1 Configuration of the embankment Configuration of the embankment is 20 m high, 1:2.0 slope inclination, as shown in Figurel. The drainage material was placed at every 5 m height, and at every 2 m horizontal interval, in a staggered arrangement. Drainage material A of high stiffness was mainly placed at the section No.1, and drainage material B of low stiffness was mainly placed at the section No.2.
Figure 1. Configuration of embankment.
213
Table 1. Measurement scheme.
Measurement method Section No. 1
Settlement plate
Displacement at surface Electro optical
0
0
Section No.2
0
0
Settlement
2.2 Measurement
Displacement in ground Inclinometer 0
Strain of drainage material Strain gauge 0
3 MATERIAL PROPERTIES
The measurement carried out at each section is shown in Table 1. Interval of the measurement was once per day, and it started immediately after the setting. After the completion of the construction, the measurement was continued at properly extended intervals. 2.3 Measurement apparatus As the drainage material is a composite material, it
is not feasible to attach a strain gauge directly. Therefore, a polyethylene sheet on which the strain gauge was attached, was sandwiched with aluminum plates, and this device was fixed to the drainage material with bolts, and measurement was carried out (as see in Figure 2, Figure 3).
3.1 Embankment material Embankment materials were soil material aggresively weathered, or become clayly from tuffy breccia and andesite, and soft rock material. The natural water content of the material was high such as from about 50 % to 70 %, and the decrease in strength due to re-mixing was significant. Properties of the embankment material are shown in Table 2. Table 2. Properties of embankment material Soil Name of sample Density pf soil particle 2.630-2.691 p s (dcm’) Natural water content Wn (%) 53.7-68.1 Gravel 35.1-65.6 2-75mm (%) Particle Sand Size 75 nl-2mm (%) 24.5-26.3 Fine particles smaller 9,9-39,2 than 75pm (%) Liquid limit 75.0-99.3
Soft rock 2.677-2.695 8.3-26.4 70.1-18.6 7.5-18.6 2.9-4.0
-
WL@)
Plastic limit Wd%)
Plasticity index In(%)
Maximum dry density pdnlar(dcm3) Optimum water content wopt (or)
Figure 2. Outlook of strain measurement.
Figure 3. Detail of strain measurement.
214
48.3-62.9
-
26.7-46.0
-
0.908-1.068
1.461-1.840
44.0-58.0
15.2-24.5
3.2 Horizontal Belt drainage material 3.2.1 Drainage material A As shown in Figure 4.(a), it is a plate-like drainage material consisting of HDPE rib-type structure wrapped with PP spun bonded nonwoven fabric. Its width is 50 cm, and its thickness is 0.7 cm. As shown in Figure 5, its tensile strength is 16.0 kN/m.
and has its tensile strength of 13 W/m. Strain is larger than that of the drainage material A. It has low strength, and low stiffness material properties. shows properties Of each drainage Inaterial.
3.2.2 Drainage material B As shown in Figure 4(b), it is a plate-like drainage material made of polyester nonwoven fabric in which polyester hollow tubes are embedded at 10 cm interval in a width direction. It is 50 cm wide and 0.4 cm thick,
Figure 4.Detail of drainage material.
Figure 5. Tensile test result.
Table 3. Properties of drainage material. Item Tensile strength (Wm) Strain (%)
Width (cm) Height Section area (cm) Section area (Cm7 Core material section area Inner permeability coefficient (cdsec) Flow rate (cms/sec)
Drainage material A
Drainage material B
16.0
13.0
12.5
27.0
50.0
50.0
0.7
0.4
35.0
20.0
17.45
-
-
0.25
150
0.16
Remarks
0
0 0
Strain rate 5 %/min
@ How rate test Earth cover: 20kN/m2 Hydraulic gradient: 4%
215
4.2 Displacement measurement at surface
4 MEASUREMENT RESULTS 4.1 Measurement results Figure 6 (a) and (b) show results of the settlement measurement at the section No.1 and No.2, respectively. At the section No.1, settlement of 61.4 cm was observed for the embankment height of 20 m, which was a consolidation settlement about 3 % of the embankment height. Consolidation was almost completed in about 40 days at either section, which was within the whole construction period of the embankment. Therefore, it was confirmed that a sufficient consolidation effect was obtained at either section. The magnitude and speed of settlement at both sections were about the same.
216
From the results of the displacement measurement at the surfaces by the sensors set on each shoulder of the embankment, the maximum horizontal displacement was about 2.5 cm at the section No.1, and about 4.0 cm at the section No.2, both to the slope side, which were very small displacement compared to the embankment height. Only the horizontal movement at the 3 rd shoulder took place toward the opposite of the slope side, and the similar behaviors were observed at every section. However, judging from the displacement vector shown in Figure 7, vertical displacement dominates more than the horizontal displacement at the section No.1, whereas at the section No.2, horizontal displacement dominates. Thus, as though the similar
Figure 8. Result of displacement measurement in ground (section No. 1).
4.4 Strain measurement of drainage material
Figure 9. Strain distribution of drainage material.
deformation patterns are shown at those two sections, the section No.2 is considered to exhibits the more dangerous behavior. Though the difference is minimal, the absolute horizontal displacement at the section No.1 is smaller than that at the section No.2. If this can be attributed to the difference in the stiffness and the strength of the drainage material at each section, the drainage material is considered to resist against the tensile stress, and to confine the deformation.
4.3 Measurement of displacement in the ground Figure 8 shows the distribution of the displacement
Figure 9 shows the strain distributions at the embankment height of 10 m, and at 47 days after the completion of the embankment. The circular line shown in the figure is the one gives the minimum safety factor. The strain in the drainage material tends to exhibit its maximum either immediately after, or at several days after the completion of the embankment, and to decrease as time passes after that. For the strain distributions in the first and third layers, the maximum strain takes place at the circular failure line. For that in the second layer, an increase in strain is shown at the fixed part in the outer side of the circular line. Its strain is about 2 %. Estimated from the results of the tensile test, this holds tensile stress of about 6 kN/m. Therefore, from the design method of the reinforced embankment, this contribute little to the stability of the embankment. However, as the horizontal displacement is very small, the drainage material is considered to be effective in confining the displacement. 5 CONCLUSIONS
in the ground measured at the section No.1. In this figure, taking the ground displacement at the completion of the embankment (embankment height of 20 m) as the initial value, the differential displacement at 10 days, and 40 days after the completion are exhibited. As shown in the figure, divided by the potential circular sliding line, the upper part displaced toward the slope front side, and the lower part displaced toward the opposite side. It is thus confirmed to become almost the same result as the circular case.
In this study, drainage materials having different drainage capacity and stiffness were used and compared. As a result, for all drainage materials, the settlement was almost completed within the construction period of the embankment, and sufficient consolidation effect was obtained. At the section where the drainage material of high stiffness was placed, the horizontal displacement was found smaller, and deformation was more confined than the drainage material of lower stiffness.
217
The stress level generated in the drainage material was about 6 kN/m, and the drainage material is considered to work as a reinforcing material, and to be effective in being integrated with the embankment F~~~ this study, it was verified that the drainage material could act as reinforcing material, other than its original drainage purpose.
218
REFERENCES M. Kamon, M. Mimura, T. Kato, & T. Akai; “Finite Element Analysis of GHD-reinforced Soft Clay Embankment”, Proceedings of 13 th Geosynthetics Symposium, :1-12,1998 M. Inagaki, A. Morikage, Y. Kumagai, Y.Yokota, S . It0 & K, Kawamura; “Effeecr of Be(t Drainage Material in High Embankment Using High Water Content Clay ”,Proceedings of 15 th Geosynthetics Symposium, :50-57,2000
Landmarks in Earth Reinforcement, Ochiai et a/. (eds), 02001 Swets & Zeitlnger, ISBN 90 2651 863 3
A full-scale field trial of electrokinetically enhanced cohesive reinforced soil using electrically conductive geosynthetics C.J.F.P. Jones & R.C.Pugh University of Newcastle upon Tyne, UK
ABSTRACT The use of cohesive fill in reinforced soil construction is not permitted by most design codes, including the current British Standard on reinforced soil BS 8006: 1995. This paper reviews a full-scale field trial utilising electrokinetics to improve the strength of reinforced cohesive fill in situ within a reinforced soil wall, allowing construction with a material that is currently classified as unsuitable. In the trial the cohesive fill was dewatered using horizontally placed electrokinetic geosynthetic electrodes and drains at an applied potential difference of 30 volts within a conventional, sandbag faced, wraparound reinforced soil wall. Results show that the fill within the electrokinetically treated zone of the structure was significantly improved, resulting in a stable structure.
1 INTRODUCTION The ability of electrokinetic phenomena to transport water, charged particles and free ions through fine grained, low hydraulic permeability soils has been well established following their discovery by Reuss (1809). Electrokinetic phenomena will occur in any soil. However, in medium to coarse-grained soils they provide a less effective transport mechanism than conventional hydraulic flow, due to the high hydraulic permeability (kh) of these soils, in comparison to the electro-osmotic permeability (k) which is used by electro-osmotic flow (Nettleton et al, 1998). In 1939, Casagrande (1952) demonstrated that applying electrokinetics to fine-grained soils with high water contents resulted in an increase in the effective stress within the soil through the generation of negative pore water pressures, increasing its shear strength to such a degree that steep cuts for railway cuttings were able to remain stable. Since then there have been many applications of electrokinetic phenomena in field projects including: improvement of excavation stability, electrochemical induratiodhardening, fine-grained soil stabilisation, consolidation, densification and electro-remediation (Pamukcu, 1996). This paper describes the first known employment of electrokinetic phenomena to reinforced soil.
2 ELECTRO-OSMOTIC THEORY
tion takes place. Positive ions (cations) are attracted to the cathode and repelled from the anode and negative ions (anions) are forced in the opposite direction. As the ions migrate they drag with them their water of hydration and exert a viscous drag upon the free pore fluid around them. Since there are more cations than anions in a typical soil, composed of negatively charged clay particles, there is a net flow of pore fluid towards the cathode (Pamukcu, 1996). Although the Helmholtz-Smoluchowski theory (Helmholtz, 1879; Smoluchowski, 1914) was one of the earliest suggested for electro-osmosis it is still one of the most widely used (Mitchell, 1993). The electrical condenser analogy adopted by this theory assumes that the soil capillaries have charges of one sign on or near the surface of the wall (-ve) and countercharges (+ve) concentrated in a double layer protruding a small distance from the wall, the remaining void is assumed to be filled with free pore fluid, as shown in Figure 1. Upon the application of an electrical potential difference across the system the mobile shell of counter-ions is assumed to drag water through the capillary by plug flow, resulting in a high velocity gradient between the two plates of the condenser. The rate of water flow is controlled by the balance between the electrical force causing water movement in one direction and friction between the liquid and the wall in the other. The overall flow (qA) generated by the application of a potential difference (A) may be expressed as (Mitchell, 1993):
AV qA = k, -A
AL
When a direct current (D.C.) electrical potential difference is applied across a wet soil mass ion migra219
Figure 1. Helmholtz-Smoluchowski Model for electro-osmotic flow (After Mitchell, 1993).
where k, is the electro-osmotic permeability of the soil; AVIAL is the electrical potential gradient; and A is the cross-sectional area of the soil sample across which the potential difference is applied.
3 COHESIVE REINFORCED SOIL AND ELECTROKINETICS Many codes of practice do not permit the use of cohesive soils in the construction of reinforced soil structures for permanent works. The reasons given are the potential problems of low strength, high moisture content, creep and low bond strength between the reinforcement and the soil (Jones et al, 1997). Laboratory and full-scale field trials of reinforced soil incorporating non-permeable reinforcement have demonstrated that the incorporation of reinforcement into the soil can result in a reduction of the overall strength of the structure. This is due to a rise in the pore water pressure in the vicinity of the
soillreinforcement interface during shearing as observed by Ingold (1981). If permeable reinforcement is used, as opposed to non-permeable, then, although the structure may be approaching drained conditions at the soilkeinforcement interface, the remainder of the cohesive fill may be experiencing undrained conditions and may be failing. The fullscale trail embankment reported by Tatsuoka and Yamauchi (1986) using permeable reinforcement demonstrated this phenomena. Figure 2a shows the reinforcement configurations for both sides of the reinforced trial embankment and Figure 2b shows the resulting displacements of the embankment. It is apparent from the deformed shape of the embankment that the left-hand side reinforcement spacing was too large and hence the pore water pressures were unable to dissipate during loading, thus causing the wall to deform. The right-hand side, however, had a closer spacing of the reinforcement and hence each reinforcing layer took less load. In addition the drainage path length for the dissipation of excess
Figure 2a & b. Full-scale reinforced cohesive embankment (After Tatsuoka and Yamanuchi, 1986).
220
pore water pressure was reduced, thus preventing a build-up of excess pore water pressure; as a result the wall underwent less deformation. From this and other published literature (Murray and Boden, 1979; Ingold, 1979&b; Heshmati, 1993; Boardman, 1999) it is apparent that if the undrained shear strength (c,) and pore water pressures within reinforced cohesive fill can be controlled then the use of cohesive fill in reinforced soil would be a more viable option. Electrokinetic phenomena have been shown to dewater cohesive soils successfully, thus causing a decrease in water content and an increase in c, through an effective overconsolidation of the cohesive soil in the vicinity of the anode. Additionally, due to a decrease in the water content the pore water pressure coefficients A and B are reduced, such that any increase in total stress generates a smaller increase in excess pore water pressure (Craig, 1992). Furthermore, due to electro-chemical changes that take place within the soil during electrokinetic treatment, an increase in shear strength of up to 80% greater than that which can be obtained from an equivalent increase in effective stress alone may be achieved (Bjenum et al, 1967). Hence there is an apparent synergy between these two technologies. The use of electrokinetic phenomena requires the use of electrodes to apply the electrical potential difference across the soil mass to be treated. Conventional metallic electrodes experience oxidation at the anode resulting in degradation of the electrode itself and an increase in the soil electrode resistance. This results in an overall loss of electrical efficiency within the system. The electrodes utilised in the trial were electrokinetic geosynthetic (EKG) materials (Nettleton et al, 1998). The benefits of using EKGs as opposed to metallic electrodes include: flexibility of the electrode allowing easy manhandling and installation during construction, ease of connection to power supply, greater surface area for soil/electrode contact
and better durability. The EKG electrodes used in the trial are described in Rowe and Jones (2000).
4 CONSTRUCTION OF THE WALL The wall was constructed using a wraparound design, utilising sandbags for the front face to temporarily retain the cohesive and granular fills. The ends of the cohesive trial wall were retained using conventional reinforced soil blocks. A small trial section was constructed at one end of the wall contemporaneously with the main trial; electrodes and drains were also incorporated into this zone but no electricity was supplied to this zone such that it acted as a control. This area was retained on one side using geosynthetic gabions. Figure 3 shows a schematic of the overall trial wall. The main cohesive trial section of the wall was subdivided into three zones, with each having an electrode spacing of 1.2m, 0.8m and 0.4m respectively. Geosynthetic drains were placed midway between the electrodes to give a drainage path for the excess pore water pressure. The reason for different electrode spacing was to achieve different electric field intensities, thus a variation in AV in Equation 1 could be achieved using a single power source. the electrical potential applied across the electrodes was 30 Volts D.C. This gave voltage gradients of 0.45, 0.6 and 0.83V/cm based upon the anodekathode spacing. The actual 2-D potential gradients calculated using finite difference analysis were slightly less than this due to the loss of potential difference around the electrode/soil interface. The wall was constructed using a staged construction technique, such that if a single lift of clay was constructed and dewatered vertically by electroosmosis applied via horizontally placed electrodes and drains. Once this lift had been successfully treated then the next lift was constructed. and so on
22 1
Figure 4. Detail of single lift. Table 1. Percentage improvement in shear strength of first lift after 278 hours of treatment.
until the full height of the wall was achieved, a total of 8 lifts. Figure 4 shows a detail of the first lift in the cohesive region of the wall.
Electrode spacing 1.2m zone 0.8m zone 0.4m zone Control zone
5 MONITORING AND RESULTS During the construction of the wall the voltage, current and generator fuel consumption were monitored for each lift. In addition the undrained shear strength of the cohesive section was monitored using a hand shear vane at different treatment times. The results from the hand shear vane for the first lift of construction taken at a depth of 0.25111 into the clay are given in Figure 5. A summary of the percentage improvement in strength at 0.25m and 0.5m depth into the first lift for all cohesive zones and the control zone is given in Table 1.
Figure 5. Shear vane results at 0.25m depth for first lift of wall.
222
0.25m depth 16% 36% 99% 27%
0.5m depth 7% 31% 57%
4%
6 CONCLUSIONS The electrokinetically treated zones of the wall showed increased improvement over the control zone in the wall as demonstrated in Figure 5 and Table 1. With successive lifts further improvement took place in underlying lifts due to surcharge loading and drainage through the redundant EKGs and drains. The results have also demonstrated that the electrode spacing and applied voltage gradient is
critical in order to achieve successful treatment of the cohesive fill in situ. The initial results demonstrate that electrokinetic techniques can indeed be used to improve the performance of reinforced cohesive soil. The forensic study of the trial is continuing.
7 ACKNOWLEDGEMENTS The Authors would like to express their gratitude to EPSRC for their ongoing support in this research. In addition the Authors would like to thank Kvaerner Cementation Foundations Ltd, Tensar International Ltd, CAPITOL, Naue Fasertechnik GmbH, Okasan Livic CO Ltd and NEW Associates for the provision of materials, funding and advices. REFERENCES Boardman, D I. 1999. Investigation of the consolidation and pull out resistance characteristics of the Paradrain geosynthetic. Research Contract Report, November, University of Newcastle. Bjermm, L, Moum, J and Eide, 0. 1967. Application of Electro-osmosis to a Foundation Problem in a Norwegian Quick Clay. Gkotechnique, 17,214-235. Casapnde. 1952. Electro-osmosis stabilisation of soils. Journal of Boston Society of Civil Engineers, January, 39,51-83. Craig, R F. 1992. Soil Mechanics. 5th Edition, Chapman & Hall, London. Helmholtz, H. 1879. Wiedemanns Annalen d. Physik. 7, 137. Heshmati, S. 1993. The action of geotentiles in providing cornbined drainage and reillforcement to cohesive soil. PhD thesis, University of Newcastle upon Tyne, UK.
Ingold, T S. 1979a. Some observations on failure mechanisms in reinforced clay. Proc. VI Panamerican Conference on Soil Mechanics and Foundation Engineering, Lima, Peru. Ingold, T S. 1979b. Reinforced clay - a preliminary study using the triaxial apparatus. Proc. Int. Conf. on Soil Mechanics and Foundation Engineering, Brighton. Jones, C J F P, Fakher, A, Hamir, B and Nettleton, I M. 1997. Geosynthetic material with improved reinforcement capabilities. Proc. Int. Symp. on Earth Reinforcement. Fukuoka, Japan, 2, 865-883. Mitchell, J K. 1993. Fundamentals of Soil Behaviour. 2nd Edn., Pub. John Wiley & Sons Inc, New York, USA. Murray, R T and Boden, J B. 1979. Reinforced earth wall constructed with cohesivefill. Colloque International sur le renforcement des sols: terre armee et autres techniques. Paris Association amicale des ingenieurs anciens eleleves de LEcole nationale des ponts et chaussees. Nettleton, I M, Jones, C J F P, Clarke, B G and Hamir, R. 1998. Electrokinetic Geosynthetics and their Applications. Proceedings of the 6th International Conference on Geosynthetics, Atlanta, Georgia, USA, 2, 871-876. Pamukcu, S. 1996. Electro-chemical technologies for in-situ restoration of contaminated subsurface soils. The Electronic Journal of Geotechnical Engineering, available at http:/l39.78.66.61/ejge/ Reuss, F F. 1809. Sur un nouvel effect de l'klectricitk glavanique. Mkmoires de la SocietC ImpCriale des Naturalistes de Moscu, 2,321-331. Rowe, K and Jones, C J F P. 2001. Geosynthetics: innovative materials and rational design. GeoEng 2000, Melbourne, Australia, 1124-1 156. Smoluchowski, M. 1914. In Graetz, L (Ed.) Handbuch der Elekmrizitot und Magnetisums. V01.2, J A Barth, Leipzig, Germany. Tatsuoka, F and Yamauchi, H. 1986. A reinforcing method for steep clay slopes using a non-woven geotextile. Geotextiles and Geomembranes, 4,241 -268, Elsevier.
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lS5N 90 2651 863 3
High embankment of clay reinforced by GHD and its utilities M. Kamon Disaster Prevention Research Institute, Kyoto University, Kyoto, Japan
T. Akai & A. Matsumoto Technology Research Institute of Osaka Prefecture, Osaka, Japan
S . Suwa, M. Fukuda & T. Simonodan Geo-Research Institute, Osaka, Japan
S . Yanagihara, Y. Nambu, K. Iwata & M. Matsushita Okumura Construction, Obayashi Construction, Toy0 Construction, Toyobo, Osaka, Japan ABSTRACT: Over ten years, the research on the behavior of three prototype test fills of clay soils reinforced by the geosynthetic horizontal drain (GHD) has been lasted. Some results have been already published. In this paper, fundamental data referred to strength of filled clay, frictional coefficient and durability of filter effect of GHD are focused the checking method of permeability of installed GHD at the site is pointed out for investigating clogging condition of filter. Another efficient role in the clay utility such as the counter measure is proposed analytically for dynamic motion of sandy embankment. 1 INTRODUCTION
3) Role of clay materials partially installed at a foot of fill is stressed on seismic stability of embankment.
The volume of surplus clays delivered from construction sites is increasing year by year. However, their properties are so complicated that they have become difficult to use as fill materials. This is because they have properties such as high water content, fine grain size distribution, weak structural characteristics and so on that scatter across a broad range. However, with the exception of weak structural characteristics and low bearing capacity, clay has beneficial properties with regard to engineering characteristics. It has especially been shown to have high durability against erosion, high strength for liquefaction and strict water proofing characteristics. These specific properties could be valuable, if combined in fills with other materials. Even though it has weak structural characteristics, there are many examples of reclaimed island using clay of high water content developed as foundation material for lower building facilities. It proves that the reclaimed clays can be changed to have many functions, if its water content is reduced by vertical drain method based on the consolidation theory. To broaden the utilization field for the surplus clay, it is especially important to show whether high, steep and stable embankment can be constructed even though using clay of various water contents. On this purpose, the prototype embankment tests were performed using soft clay, but reinforced by GHD. Emphasis is placed on the following issues in this research.
2 FILLING OF CLAY REINFORCED BY GHDs Based upon the measured data, the behavior of embankment during and after completion relating to the strength, water content and flowing properties was described. These data obtained by the large scale of embankment test can prove that surplus clay has many functions to make an embankment being stabilized (Kamon et al. 1994, Kamon et al. 1996). Figure 1 shows the scale of fills, types of arranged GHDs, and material properties relating to the tensile strength and elongation. Every test used the clay of water contents ranging from 40% to 100%. The heights of fills are 3m to 10m with the slants of 1:0.6 to 1:1.5. Figure 2 describes the representative GHD materials, plastic core with nonwoven fabric type, reinforced nonwoven fabric type and tufted pile fabric type. Their tensile strengths range from 17 kN/m to 82.8 Mv/m and their elongations from 11% to 90%. In fact, it is evident that they have lower strength and higher extensibility compared to the characteristics of geogrids. Among three cases, the case 2 was forced to collapse by the surcharge acting on the head surface of the fill after reaching the height of 3 m. The two other embankments remained stable after the passage of 5 or 7 years, even though the height of 10 m test embankment has the steep slope of 1 :0.6.According to the site observation, there was no sign of the development of a failore or erosion in the future.
1) Increased strength of filled soft clay taken through prototype fill test is shown. 2) How to research durability of filter effect of installed GHD at a site is proposed.
225
3 INCREASED STRENGTH OF CLAY
4 ENGINEERING PROPERTIES OF GHD
Figures 3 and 4 reveal the profile of water content and cohesive strength (c,,) that obtained at the site of case 2 during filling and after the compulsory failure. The solid lines are the data during filling and the dotted lines correspond the relationship after the failure. Both figures make it clear that the clay soil has high water content ranging from 80% to 90% and low strength less than about 60kPa. It was impossible to fill over 3m if not be reinforced by GHD. Each zone was successfully filled to the height of 3m with a steep slope of 1 : O S and collapsed by the intensity of the surcharge estimated. Data of water content proved that their values have a tendency to decrease with depth, and the strength of clay increases with depth. These phenomena were caused by the consolidation so that the facility of GHD was clearly confirmed.
To improve the design method of fill reinforced by GHD, the knowledge of the coefficient of friction between a filled soil and GHD, its creep properties and its permeability are required. Figure 5 shows the frictional characteristics of GHD against the clay and fine sand. The brackish marks show the maximum frictional relationship between GHD and clay. On the other hand, the white marks reveal the maximum frictional pressure between GHD and sand. And the dotted lines are the inner frictional characteristics of the find sand itself. The coefficient of friction ranges from 0.3 to 0.4, and are grouped together in spite of the difference of GHD. On the contrary, the characteristic of friction is predicted to depend on the soil classification from Figure 5. Because the trends of increasing friction are similar with each type of GHD, the difference among characteristics of soils appear more influential than the difference in materials. Figure 6 shows the creep characteristic of GHD. The creep strength will take a more important role than the tensile strength, because the influence of the
Figure 3. Profile of water content at the site of case2.
Figure 5. Frictional properties between GHD and surrounding soil.
Figure 4. Comparison of strength between during filling and after failure.
Figure 6. Creep properties of GHD.
227
ductivity of GHD in plane, a = the horizontal area of GHD for water flowing, and A = the sectional area of a pipe. Figures 9 and 10 give the test results performed at the case 1 site and case 3 sites. Figure 9 contains the test results obtained at the non-reinforced zone that is shown by the mark (A).The other thin solid lines with the mark (U,+) are the data at the site reinforced by the GHD. The set of thick lines are the relationship relating to the hydraulic conductivity of various orders of GHD predicted by Equation (1). According to the trend appearing in Figure 9, the coefficient of permeability is estimated at 0.01 c d s , on the other hand, the properties of GHD are predicted to have the value around 0.5 c d s . Figure 10 shows the test results obtained at the case 3. The solid lines with the brackish marks are the relationship obtained at the site in the case 3 reinforced by GHD of plastic core type shown in Figure 2, and the dotted lines with the white marks are the data at the site in the case 3 reinforced by GHD of the reinforced nonwoven fabric type in Figure 2. Two thick lines without mark are the predicted rela-
maximum strength characteristic will emerge well beyond large strain, on the contrary, the creep effect shows up even under small load intensity.
5 PERMEABILITY EFFECT OF GHD Water pouring tests were carried out through poly vinyl chloride pipe at the sites. This test was performed to investigate the permeability characteristic of GHD installed horizontally. As shown in Figures 7 and 8, the corresponding hole was dug out to insert a pipe for pouring water. The pipe must be in contact with the surface of GHD at the end. Poured water flows along the plane of the GHD as shown in Figure 8. This relationship between height of water level in a pipe and elapsed time is derived as the Equation (1) on the assumption of the simple flow like Figure 8. h h,
h 4kn’t = 1+h, h,A2
-- ln(-)
where h = water head at elapsed time t, ho = an initial value of the water head h, k = a hydraulic con-
GHD : Distance of front : Sectional area of pipe GHD in plane : Sectional area of
Figure 8. Direction of water flowing
228
Figure 10. Decreasing of height of water level in the pipe at case 3 test site.
Figure 1 I . Shape of slope attached by clay block
tionship by Equation (1) on the assumption of the hydraulic coefficient of 0.5 and 1 c d s . On the basis of this trend, the coefficients of permeability of two GHDs are estimated to range from 0.5 c d s to I c d s . These values clearly maintain the initial permeability, even though these tests were conducted a few years after completion of filling.
6 COMPOUND SLOPE WITH CLAY BLOCK ATTACHED AGAINST EARTHQUAKE To overcome the problem of increasing surplus clay, it is necessary to study the structure of slope that will be resistant against liquefaction if facilitating the strength of clay. Figure 11 shows the corresponding slope for analysis. Figure 12 is the result of the analysis. Figure 12 includes the results given by Taniguchi et al. (1985) applied for the 1984 Nagano west side earthquake. Their corresponding slope has the length of about 200m and the thickness of sliding block of about 20m. Comparing to the simple slope shown in Figure 1 1 , it is much larger. However, the tendency of their research shown in Figure 12 is similar to the results from the simple slope. Based upon the results of the analysis, the clay block can enhance the level of the slope stability, even though the small block is only placed at the foot of a slope.
Figure 12. Role of clay block placed at the foot of slope against earthquake.
2) A way was proposed to research permeability characteristics of GHD installation and it was confirmed that the properties of GHD change little from the initial conditions after installation. 3) A clay block installed at the toe of a slope can be effective in preserving a slope of road against earthquake. REFERENCES Kamon M., Akai T., Fukuda M. & Yaita 0. 1994. Reinforced Embankment Using Geosynthetic Horizontal Drains. Fifth International Conference on Geotextiles, Geomembranes arid Related Products: 791-794. Kamon M., Akai T., Fukuda M. & Nanbu Y. 1996. In situ failure test of higher water content soft clay embankments reinforced by GHDs. International Syniposium Earth Reinforcement (Is-Kyushu): 114-119. Taniguchi E., Kubota T. & Kuwahara T. 1985. Slope failure at Matsukoshi by the Nagonoken-seibu Earthquake. Journal of Soil and Foundation Vo1.33, No. 1 I: 59-65.
7 CONCLUSIONS In this paper, the engineering properties of the full scale clay fill reinforced by GHD were examined analytically. Conclusions can be summarized as follows. 1) Through the results of three prototype filling model tests, it was found that the slopes filled by clay remain stable and have durability. 229
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Tension in geosynthetic liner based on hyperbolic interface response K.V.S. Krishna Prasad & M.R. Madhav I.I.T., Kanpur, India
J. Kodikara Victoria Univ.of Tech., Melbourne, Australia
A. Bouazza Monash Univ., Clayton, Australia
ABSTRACT: The paper presents an analysis for the estimation of tension in a geomembrane liner considering lower interface response of the liner with the underlying soil to be hyperbolic. The governing equation is non-dimensionalised and solve numerically with the finite difference approach. A parametric study and few design charts have been presented for ready use.
1 INTRODUCTION There has been significant growth in the general awareness towards the environmental aspects of the habitat in the last two decades primarily because of the enormous increase in the quantities of waste being produced. Proper management of these wastes is absolutely essential or else their harmful effects can b e quite devastating. Amongst various types of wastes, disposal of municipal solid wastes (MSW) constitute a major challenge to geoenvironmental engineers. Apart from its potential to damage the environment, MSW remains highly visible. Remedial alternatives for minimizing their impact on the environment are reduction at source and/or containment with proper treatment. Engineered landfills are considered as an answer to the above problem. One of the chief components of an engineered landfill is the liner system that prevents the migration of the leachate or harmful gases into the surrounding soil. A liner system may comprise of a combination of barrier materials such as natural clays, amended soils, flexible geomemebranes or combinations there of. Geosynthetic clay liners (GCL) composed of a geomembrane and a clay liner are frequently utilized for the containment of the leachates. These GCL placed on a slope is anchored at the crest level. Construction involves placement of soil and waste layers up to a design height because of which the GCL is subjected to down-slope shear stresses which induce tension in the geomembrane that needs to be estimated and suitably provided for in the design.
2 REVIEW OF LITERATURE Construction of landfills involves placement of geomembrane liners on slopes and anchored at the
crest. Though the primary function is containment of the leachate, Sharma & Lewis (1994) report tension failures of the liner. Rupture leads to leakage of leachate thus nullifying the purpose of providing the liner. Koerner & Hwu (1991) analyze the tension induced in the geomembrane in terms of shear stresses developed on its upper and lower faces. The stability of cover soil on liner system with geomembrane is analyzed by the FEM by Wilson-Fahmy & Koerner (1993). Kodikara (1996, 2000) presents two simple approaches for the estimation of tension in the geomembrane for the initial linear deformation and the final plastic condition of the interface stresses. This paper presents a unifying approach that incorporates a continuous hyperbolic interface response. 3 PROBLEM STATEMENT Landfill slopes are formed in one or several benches with berms providing anchorage for individual geomembrane lining segments. Subsequent construction involves filling in stages, successively covering each bench. In this process, the weight of the overburden materials above a particular liner is transferred to the underlying clay base through the various layers, viz., waste, soil, geomembrane, etc. resulting in substantial down slope shear being applied to the upper interface of the liner that induces tension therein. For the analysis, a single material representing all the overburden materials above the liner and an idealized model featuring a single bench is depicted in Figure 1. In the figure, L is the total length of the liner, H - the height of overburden, 0 and fi - the angles of inclinations of the liner and the top of the landfill respectively, y - the unit weight of the 23 1
where t and E are the thickness and modulus of elasticity of the liner and K, - the coefficient of lateral stress and Fx= xsin0-xcos0tanp. Eq. 3 is nondimensionalised as
r
overburden and U - the displacement of liner at a distance, x, from the anchorage. The shear stress, q, displacement, U, response of the lower interface between the liner and the clay, is assumed to be represented by a hyperbola (Figure 2) and the relation expressed as 2,
=k,u/(l+k,u/zf}
(1)
where k, and TIP are the initial slope and the maximum stress respectively. TIP is given by
1
where F1= (F3-Pl) tan 61 + CI, F2 = (1-K,) sin 20 (H* jos2 0 + sin' 0) [H* + FX]. y , = y L2/tE, H* = WL, C1= call$.,, PI = pl/yL, X = x/L and U = u L . The tension, T, per unit width, in the reinforcement is du
T=tE-
dx
(5)
The governing equation is solved for the boundary conditions: at x = 0 (i.e. at the point of anchorage) u=O (no displacement) and at x=L (at the bottom of the liner), T = 0 (no tension).
4 RESULTS where onand PI are the normal stress and pore pressure respectively at the lower interface, 61and cal are intrinsic friction angle and adhesion between the liner and the clay at the interface. Following Kodikara (1996), considering the force equilibrium of an infinitesimal element (Figure 3) and simplifying, one gets
232
Eq. 4 has been solved by the finite difference method to obtain the displacements and tension at various points along the liner. The length of the liner was discretised in to a number of elements varying from 10 to 100. No increase in accuracy of results was achieved for n values greater than 100. Hence n equal to 100 was adopted for further analysis.
A parametric study was carried out for the following ranges of parameters: 'i: 12-18 kN/m3; cal: 0-20 kPa; p1: 0-20 kPa; 61: 10'-25'; 8: 5'-30'; p: -20' to +30°; H: 0-10 m; L: 1-100 m; t: 0.5-5.0 mm; E: 100-500 MPa; K,: 0.3-0.6 and kr: 103-105kN/m3. The ranges of non-dimensional parameters were calculated from the above ranges of parameters. The solutions based on the proposed approach for were validated with those from Kodikara (1996) and verified to be in close agreement. The variations of normalized tension, T* (=T/tE) and normalized displacement, U (=u/L), with distance from the anchor point, for the parameters listed therein, are presented in Figures 4a and b. The parameter, h, represents the effects of unit weight, y, of the overburden material and of the length, L, of the
233
liner. As expected, the tensions are maximum (Figure 4a) at the anchor point and decay very rapidly with distance. This effect can be correlated with the full mobilization of shear stress on the lower interface as a consequence of large mobilized displacements. However, the maximum value of tension in the liner increases with h, the value increasing from 0.003 for h equal 80 to 0.065 for h equal to 1500 reflecting the influences of either longer liner or that of heavier overburden. In contrast, the liner displacements appear to follow a linear increase beyond the rapid initial increases. Results for different angles of inclination of the liner, 8, (Figure 5), depict an interesting trend. The values of tension increase with 8 for 8 values up to 35' and decrease with 8 for higher values. The variations of T* with distance extends to farther points away from the anchorage.
5 DESIGN CHARTS A designer is often interested in the maximum value of tension so that he may choose an appropriate liner from the large variety available in the market. The variations of maximum tensions with h, K,, H*, and 8 are presented in Figures 6 to 10 respectively.
x,
6 CONCLUSIONS A simple approach is proposed for the estimation of tension in a geomembrane used as a liner in a MSW landfill project based on a hyperbolic response of the lower interface with the soil underneath. The governing equation is normalized and solved using the finite difference method. The solution obtained agrees closely with the results of Kodikara (1996) for the initial linear and the final fully plastic interface responses. A parametric study and design charts are presented quantifying the effects of stiffness of the interface, height of fill above and the length of liner, coefficient of lateral stress coefficient, height of fill above the anchorage point, inclinations of the liner and of the top of the landfill, interface strength parameters, etc.
REFERENCES Kodikara, J. 1996. Prediction of tension in Geomembranes placed on landfill slopes. In M. Kamon (ed.) Environmental Geotechnics: 557-562. Rotterdam, Balkema. Kodikara, J. 2000. Analysis of tension development in Geomembranes placed on landfill slopes. Geotextrles arid Geomembranes. 18:47-61. Koerner, R.M. & B.L. Hwu 1991. Stability and tension considerations regarding cover soil on Geomembrane lined slopes. Georextiles and Geomembranes. 10:335-355. Sharma, H.D. & S.P. Lewis 1994. Waste containment systems, waste stabilization and landtills: design and evaluation. New York, John Wiley & Sons, Inc. Wilson-Fahmy, R.F. & R.M. Koerner 1993. Finite element analysis of stability of cover soil on geomembrdne lined slopes. Proc. Geosynthetics '93, Vancouver: 1425-1437.
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Landmarks in Earth Reinforcement, Ochiai et al (eds), 02001 Swets & Zeitlinger, ISBN 902651 863 3
Trial construction of arching structure by using large-sized soilbags Tetsuya Kubo, Yoshihiro Yokota & Syuuji Itou Maeda kousen Co., Ltd, Japan
Liu Sihong & Hajime Matsuoka Nagoya Institute of Technology, Japan
ABSTRACT: Soil bag has been confirmed to perform excellent due to the restriction effect of the geosynthetics (of the bag material). From this fact, it comes to an idea if any arch structure could be built with soil bags, instead of the rigid material such as stone or concrete that have been conventionally used. However, it is difficult to build an arch structure of actual scale with the soil bag of an ordinary size. Hence, in order to confirm the feasibility on construction of the actual arch structure and its behavior, an experiment that large size soil bag instead of the ordinary size was used to build an actual size arch structure (R=5m), was conducted. At the final experiment, stress and displacement took place in the arch structure, and the strain worked in the geo-synthetics, were measured, and its feasibility whether it could be constructed as a structure. was confirmed. 1 INTRODUCTION
2.2 Measurement In the experiment, earth pressure activated on the arch structure, strain of the geo-synthetics forming the “soil bags”, and displacement of the arch structure were measured. The locations of the measurement apparatus are shown in Figure 1.
Arch structures have been used for the structures such as bridges and tunnels for very long time. In order to form the arch, mainly rigid material such as stone and concrete have been used. Whereas “soil bags”, which can be easily made at site, have been used for various purposes in the civil construction. Recently, it is confirmed that improving the quality of the “soil bags” by packing the soil into the geo-synthetics, turned out to be greatly effective by research done by Matsuoka et al. Therefore, the authors considered that the “soil bags” could be used as material for the arch structure, and carried out an experiment of actual size. Here it is described on the evaluation of its constructability and the results of its observation.
2.2.1 Earth pressure cell Earth pressure cells were set in the basement of the arch structure (earth pressure cells No.] and NO.^), and at the inside of the 9 th large soil bags from the bottom (earth Pressure cell No.3 and No.4). The earth Pressure cell consisted of four load cells, and it can measure the earth Pressure at the upper and lower faces, and at the inner and outer faces.
2 CONTENTS OF EXPERIMENT A centre was used in order to construct the arch structure. The centre was set at 50 cm raised position by four hydraulic jacks, which would be able to be removed at last. Time dependent behavior was observed by measurement from the setting of the “soil bags” until the removal of the centre. 2.1 Figures of the arch and the soil bags The arch structure built in the experiment is shown in Figure 1. As for the “soil bags”, box-shaped soil bags and wedge-shaped soil bags were used. Filler material used was crushed stone (C-40).
Figure 1. Arch figure and location of measurement apparatus.
235
2.2.2 Strain gauge Strain gauges were set inside and outside of the arch at the 6 th and 9 th large soil bags from the bottom. 2.2.3 Displacement seismograph and target Displacement seismograph used was of wire-type displacement seismograph. It was set at the centre (displacement seismograph No.7-1) when the centre was settling, and it was set at the 14 th large soil bag from the bottom (displacement seismograph No.7-2) after the centre was removed. In addition, targets were set at the front face of the arch structure, and the displacement of the arch structure was measured by using transits. 3 EXECUTION PROCEDURES OF THE EXPERIMENTS The execution procedure of the experiment is shown in Figure 3. Pictures during the construction are shown in Figure 4 to Figure 7.
236
Tablel. Characteristicsof geo-synthetics. Item
embedded in the base ground treated with stabilization, and its set condition became of buried type, and little effect was observed at the end of the arch.
Characteristics
Material
Polvester
Mesh size
(mm)
7.0 X 7.0
Mass Tensile strength
(s/mZ)
Elongation ratio
(%)
700 98.0 X 98.0 22.0 x 22.0
(kN/rn)
4 MEASUREMENT RESULTS AND DISCUSSIONS 4.1 Earth pressure 4.1.1 Behavior at construction f o r large soil bags Figure 8 show relationships between each earth pressure and time passed while setting large soil bags. The earth pressure inside of the soil bags (earth pressure cells No.3, No.4) increased significantly at 10 th from the bottom. Instead, the earth pressure at 7the end of the arch (earth pressure cells, No.1, No.2) did not shown as much increasing tendency as that shown by the earth pressure inside. This is considered because that the earth pressure cell was
237
4.1.2 Behavior when centre is lowered Figure 9 shows the relationship between the change in earth pressure accompanied with the lowering of the centre, and the vertical displacement at the centre. The earth pressure inside the soil bags (earth pressure cell No.3, No.4) tended to increase significantly with an increase in the vertical displacement. This is considered because that the large soil bags set in the upper part leaned against the centre, and that the earth pressure increased as the centre lowered. Whereas the earth pressure at the end of the arch (earth pressure cells No. I , NO.^), shows a tendency of gradual increase. This is also considered to be the same effect of the setting as mentioned above.
4.1.3 Earth pressure distribution in the section Figure 10 shows the relationship between the earth pressures measured by the load cells inside and outside of the panel type earth pressure cells No.2 and No.4, and the time passed. Each earth pressure cell tends to show that the earth pressure generated at the inside is the lager.
bag from the bottom (T-4), whereas a tendency of some settlement was observed in the upper soil bags. The settlement took place after the removal of the centre, was considered to be caused by a phenomenon that interlock (or cohesion) between the neighboring soil bags became strengthened. There was a tendency of settlement observed at any target until about 30 days, but that was observed to cease after that. The magnitude of the settlement in accordance with the lowering of the centre was about 0.4 m as the maximum settlement.
4.1.4 Strain of geo-synthetics Figure 11 shows the relationship between the strains at the inside and outside of the arch in the geosynthetics during the lowering and after the removal of the centre, and the vertical displacement. The strain of the geo-synthetics at the inside of the soil bags is very small and tension (+ symbol: tension) occurs. This is because soil bags are compressed and bended laterally. As seen in Figure 12. From this, stress is confirmed to be concentrated at the inside of the arch, and it is considered that the filler material which would otherwise deform, is constrained by the geo-synthetics.
5 FINALREMARKS 4.2 Displacement of the arch structure
As stated in Figures 10 and 11 compression stress is consentrated at the inside of arch structure. This means that arch action is constructed at the inside of the structure. The displacement behavior shown in Figure 12 also supports this aspect. If this arch structure was constraimed laterally, more complete arch action would be observed. From this experiment, the proposed method was found to have some room for improvement in its constructability, but it was confirmed to be able to build it. In addition, the arch structure consisting of the large soil bags using the geo-synthetics, was found necessary to allow some deformation. However in reality, as an arch effect due to the lateral earth pressure and the soil itself is expected, it is considered to be able to build a structure exhibiting little deformation, and high stability.
Figure 13 shows the behavior of the arch structure expressed by the measurement of the targets set in the large soil bags. From this Figure, little settlement was observed for the soil bags lower than the 6th
REFERENCES S. Okuda, H. Matsuoka, & S. lwai 1993,“A Discussion on Strengthening of Bearing Force by Confining A Part of Ground‘‘, 48 th Annual Convention, JSCE, III-554, :11421143. T. Ono, H. Matsuoka, & T. Takizawa 1994, “A Discussion on Bearing Force Test for Foundations of Various Shapes and Bearing Force Strengthening Test with ‘Soil Bags’ “, 49 th Annual Convention, JSCE, 111-392, :774-775. S. Yamamoto, & G. Matsuoka 1995,“Compression Test of Soil Bags Using DEM and Simulation of Bearing Force StrengtheningTest with Soil Bags”, 30 th Geotechnical Engineering Research Meeting, :1345-1348. G. Matsuoka, Liu, H. Kodama, & K. Yamaguchi AU Annual Convention H. Matsuoka, Chen., H. Kodama, Yamaji & R. Tanaka “Mechanical Properties of Soil Bags and Its Pressure Test”, 35 th Geotechnical Engineering Research Meeting.
238
Landmarks in Earth Reinforcement, Ochiai et a/. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Study on the critical height of fiber-reinforced slope by centrifuge test G.X.Li,Y.X.Jie & G.Z. Jie Department of Hydraulic Engineering, Tsinghua University, P.R.China
ABSTRACT: Centrifuge model tests were carried out to study the behavior of fiber-reinforced cohesive steep slope. It has been found that unlike unreinforced one, reinforced soil still retains a certain tensile strength in the crack zones, which increases greatly the critical height of the steep slope. A simple method to estimate the critical height of the steep slope is also presented in this paper.
1 INTRODUCTION
2 TEST MATERIALS
Fiber-reinforced soil (Texsol) is a promising geotechnical material. It can be considered to be a composite made of soil and continuous threads of synthetic fibers. To obtain such a material, a number of threads are penumatically or hydraulically projected on soil in a movement at the extremity of a conveyor belt or at the vent of a pipe used to build a hydraulic filI(Leflaive 1998). It has excellent mechanical, hydraulic and environmental performance in civil engineering. The soil used to produce texsol is mostly natural sand. However, cohesive soil should also be paid enough attention to, especially in places where there is short of sand. Tsinghua University has conducted a series of tests on the static, dynamic and hydraulic behavior of fiber-reinforced cohesive soil (Zheng 1992, Li et al. 1993, Li et al. 1995, Zhang 1995). It has shown by the tests that fiber-reinforcing increases greatly the shear resistance of the cohesive soil, and improves its plasticity under tensile stress. The fiberreinforced cohesive soil has excellent performance in resistance against hydraulic fracture, too. In this paper, centrifuge model tests are carried out to study the fiber-reinforced and the unreinforced cohesive soil slope. A cohesive soil with two dry densities pd=1.65g/cm3 and pd=l .55g/cm3 was used in the tests. Failure process was observed on CCTV camera. It was found that the reinforced slopes have different failure pattern in comparison with unreinforced ones, and they still retain a certain tensile strength in the crack zones, which increases greatly their critical heights. A simple method is also proposed in the paper to estimate the critical height of fiber-reinforced steep slope.
The physic properties of the soil used in the tests are listed in Table 1. The thread used is polypropylene fiber. The average tensile strength is O.O6N/d, and the average Yang’s modulus is 0.8Nld. Triaxial compression tests were performed to determine the cohesion and friction angle of the soil and fibers. The samples, loomm in diameter and 200mm in height, were prepared with two density, p d = l .65$jcm3 and pd=l .55g/cm3, respectively. The content of polypropylene fibers used in the soil is 0.2% by weight. The test results are shown in Table 2. Table 1. Physic properties of the soil. Phvsic DroDertv Specific gravity (G,) Plasticity index (I,,) Optimum water content (a&,,) Maximum dry density (pdmUx)
Value 2.71 14.9 17% 1.7g/cm3
Table 2. Index properties of shear strength.
Dry density Adhesion strength c g/cm3 kPa 1.55 (Unreinforced slope) 51 (Reinforced slope) 78 1.65 (Unreinforced slope) 68 (Reinforced slope) 100
Friction angle Q 27.1 27.1 27.6 27.6
3 CENTRIFUGE MODEL TEST The geotechnical centrifuge at the University of Tsinghua used in this study has a capacity of 50 g-
239
tons. All model slopes were 303x200 mm in area, and 350mm in height. They were constructed in a rigid aluminum container with inside dimensions of 600mm by 200mm in area, by 300mm in depth. The inside vertical side boundaries of the container were sprayed with silicon and overlaid with a thin plastic film to reduce boundary friction effect. One side of the container is fixed with plexiglass, so that the behavior of the model slope can be observed through CCTV camera. Each test involved loading the model by gradually increasing its self-weight till failure occurred. The increment of each step of loading the acceleration was 5g or log (g is the acceleration of gravity). The next loading step was conducted after the slope displacement almost ceased. The test results are shown in Table 3 and in Figure 1. Table 3. Results of centrifuge model tests. Reinforced slope Unreinforced slope 1.55 1.65 1.55 1.65 Drvdensitv(dcm? Acceleration when 1OOg* 12Og* 45g* 60g* failure occurred Critical height H,,(m) 35.0 42.0 15.7 21.0 Depth of tension (m) 14.4 17.5 9.4 11.9 Zone a** * g is the acceleration of gravity, which equals to appromately 9.8 1 meters (32 feet) per second per second. ** How to obtain the depth of tension zone & is described below.
It showed from the model tests that unlike the unreinforced ones, fiber-reinforced cohesive steep slopes fail gradually. There are a family of failure surfaces in the reinforced cohesive slope. The failure surfaces developed progressively. Cracks appeared initially at places near the front side of the slope, then they developed gradually further into the slope. Most of the cracks are vertical. The two parts separated by cracks were still connected with each other by fibers. However, unreinforced slopes collapsed abruptly at certain acceleration (see Figure 1). 4 A SIMPLE METHOD FOR THE CALCULATION OF REINFORCED SLOPE The apparent cohesion and friction angle of reinforced soil can be determined through triaxial compression tests, as shown in Table 2. The depth of tension zone & can then be obtained:
z,=-
2c
YJK, The results are presented in Table 3. For unreinheight forced -% is 6o percent Of the Hc,tested, while for reinforced ones, it is only 40 percent.
Figure 1. Failure patterns of the reinforced and unreinforced cohesive steep slopes (--------indicates the slide surface used in the calculation).
240
If cracks develop in the unreinforced cohesive soil slope, it usually means that tensile resistance reduces to zero. However, fiber-reinforced slope is not the case. The separated parts at each side of the
The force polygon is shown in Figure 3. Following Coulumb’s theory, the critical height of reinforced soil slope can be calculated, the results of which are presented in Table 4. Table 4. Calculation results of the critical height of reinforced steep slope.
(dcm’) c (Wa) ZO (m) 0, (Wa) T (W
w (W c (W
H,, calculated (m) H , , tested (m)
Slooe 1 1.55 78 14.4 52.7 759 4092 1518 31 35
Slooe 2 1.65 100 17.5 61.2 1071 5250 2046 35 42
5 CONCLUSION Fiber-reinforced is a promising material. Fiberreinforced slope hardly collapses when failure occurs, except that cracks develop progressively in the soil. However, unreinforced slope collapses abruptly when fails. In crack zones, fiber-reinforced soil has still a certain tensile strength resulting from the tensile strength of fibers. It is the reason that the reinforced slope has a much higher critical height in comparison with the unreinforced one. cracks are connected with each other by fibers. Fibers can sustain tensile stress, so that the reinforced soil can still retain a certain tensile strength when cracks develop in the soil. The tensile strength increases when tensile strain increases. The uniaxial tension test results of fiber-reinforced and unreinforced cohesive soil are shown in Figure 2. The potential slide wedge of the Texsol slope is shown in Figure 3. The depth of the tension zone (or crack zone) & can be achieved by Equation 1. The vertical crack zone constitutes one of the slide surfaces. The other slide surface intersects with the crack zone and passes through the toe of the slope, and it makes an angle of (45’+@2) with the horizontal plane, see Figure 3. The tension force T acts on the wedge at the crack zone, it can be obtained by:
T = Zoor
in which c, is the tensile strength of fiberreinforced soil, (3) where Ac is the cohesion difference between the reinforced and the unreinforced soil, and v is the friction angle of the Texsol.
6 ACKNOWLEDGEMENT The support of China National Science Foundation under award No.50099620 is greatly acknowledged. REFERENCES Li, G.X., Chen, L. & Zheng, J.Q. 1993. Experiment study on fiber-reinforced cohesive soil and its elastic-plastic analysis. Proceedings of I“’Asia-Oceania International Symposium on Plasticity, Beijing, 1993: 361-368. Li, G.X., Chen, L., Zheng, J.Q. & Jie, Y.X. 1995. Tests on fiber-reinforced soil. Journal of Hydraulic Engineering (Chinese) 6: 3 1-36. Zheng, J.Q. 1992. Experimental study on properties of resisting fracture and hydraulic fracture of fiber-reinforced cohesive soil. Paper presented to Tsinghua University at Beijing, P. R. China, in partial fulj5llnient of the requiremerits.for the master’s degree. Leflaive, E. 1998. Texsol: Already more than 50 successful applications. Proceedings of the International Geotechnical Symposium on Theory and Practice of Earth reinforcement: 541-545. Zhang, X.J. 1995. Experimental study on the static, dynamic characteristics and fracture properties of fiber-reinforced soil. Paper presented to Tsinghua University at Beijing, P. R. China, in partial fulfillment of the requirements for the master’s degree.
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Landmarks in Earth Reinforcement, Ochiai et al (eds), 0 200 1 Swets & Zeitlinger, ISBN 90 265 I 863 3
Response of geosynthetic reinforcement to transverse force M.R. Madhav Department of Civil Engrg., I.I.T., Kanpur, India
H.B. Poorooshasb Department of Civil Engrg., Concordia Univ., Montreal, Que., Canada
N. Miura Institute of Lowland Technology, Saga University, Saga, Japan
ABSTRACT: Reinforced earth fills, ground and slopes counteract the destabilizing forces by mobilising tensile forces in the reinforcement. In most studies, only the pull out resistance due to axial pull is considered. In this paper, a new approach is presented for the analysis of sheet reinforcement subjected to transverse force. Assuming a simple Winkler type model for the response of the ground and the reinforcement to be inextensible the resistance to transverse force is estimated. The response to the applied force depends not only on the interface shear characteristics of the reinforcement but also on the deformational response of the ground. A relation is established between pull-out resistance and transverse free end displacement. A parametric study quantifies the contributions of depth of embedment, length and interface characteristics of the reinforcement, stiffness of the ground, etc. on the over all response. tance of the reinforcement to pullout and not the transverse one even though in some of the methods of stability analysis, the inclination of the reinforcement force (Figure 2) is considered (Bergado and Long 1997) to vary between the direction of the reinforcement and the tangent to the slip surface. In this paper, a method is presented for the estimation of the pull out capacity of sheet reinforcement to transverse force.
1 INTRODUCTION Reinforcement of soil and of ground have become extensive and very commonly preferred alternative to enhance the performance of the earth structures in the former case and of the in situ ground conditions in the latter. Thus reinforced earth retaining walls, embankments, slopes, foundation beds are commonly adopted while nailing is chosen to stabilise slopes and excavations. Reinforced earth structures have been observed to perform better under seismic conditions. The reinforcement in all the above instances is in the form of strips, bars, grids or sheets and fabricated or manufactured from metals or geosynthetics. The reinforcement is presumed to restrain tensile deformations of the soil and thus increases the over all resistance of the composite soil through interfacial bond resistance but limited by its own tensile strength. The bond resistance that operates in reinforced soil is determined either by direct shear or by pull out tests (Jewel1 1996). Considerable literature is available (Juran et al. 1988, Hayashi et al. 1994, Alfaro et al. 1995, etc.) on the test procedures, analysis and interpretation of pull out tests. However, the kinematics of failure are usually (Figure 1) such that the failure surface intersects the reinforcement at an oblique angle. The reinforcement is subjected to both axial and transverse components of the force by the sliding mass of soil. Most available theories for the analysis and design of reinforced soil structures consider only the axial resis-
Figure 1. Kinematics of reinforced embankments
243
2 PROBLEM FORMULATION AND ANALYSIS Fig. 3a depicts a sheet reinforcement of length, L, embedded at depth, H, from the surface, in a soil with a unit weight, y, subjected to a transverse force, P,, at one of its extreme end. The interface angle of shearing resistance between the reinforcement and the soil is @r. The response of the reinforcement to the transverse force is to be obtained in terms of a relation between the force P, and the normal displacement, WO. The model proposed for the analysis is shown in Fig. 3b. The reinforcement and the underlying soil response are represented respectively by a rough membrane and a set of Winkler springs. Figure 3c represents the deformed profile of the reinforcement. qt and qb and Tt and z b are the normal and shear stresses acting on the top and bottom surfaces respectively of the reinforcement. The displacement - normal stress relation of the soil is characterised by the relation
Q = ksw
(1) where k, is the modulus of subgrade reaction (the interaction parameter of the Winkler springs) and w the transverse displacement. Considering an infinitesimal element (Figure 3d) of length, Ax, unit width, the inclinations of tensions acting in the reinforcement at distances x and x+Ax, are T and (T+AT) and 0 and (0+A0) respectively. The horizontal and vertical force equilibrium relations for the element are (T + A
0 s ( @ + Ae)- T C O se(q,+q,)tan@,.A x = O
(2)
and (T+A T)sin(@+A@)-TsinB-
(3) (qb-qi).A xro Equations (2) and (3) on simplification reduce to
cos@--
dT
dx
- T sin 8 - (ql + q,,)tan@r= 0
(4)
and sin@-++cos@-(q, dT
-ql) = O
dx Multiplying equation (4) with cos0 and equation (5) with sine and adding the two, one gets
similarly, multiplying equation (4) by sine and equation (5) by cos8 and subtracting the latter from the former, one gets
Figure 3. Definition sketch (a) reinforcement subjected to oblique force, (b) model, (c) deformed profile and (d) forces on an element
-T-
244
d@ - (4, + q,,)tan& sin8 (q/,- q,)cos8 = 0 (7) dx
+
dw But tan8 =-
dx
d2w -T-+k,.w=O dx2 and
The boundary conditions are: at x=O, the tension in the reinforcement, T, is zero, and at x=L, the displacement W=WO and the applied transverse load, P,, obtained from the vertical equilibrium of forces as
0
Non-dimensionalising Eq.s (9) and (10) with X=x/L, W=W/WO,and T*=T/yHL, one gets d2W - T "-+,UW dX
3 RESULTS
d8 and -=cos28-
d2w and the dx2 Winkler spring response to the increment in normal stress, (qb-qt) is equal to k,.w. Substituting for these in equations (6) and (7) and simplifying for small values of 8, the coupled governing equations for the reinforcement under transverse force are derived as
dx
=0
dT' --@WOW +2}tanq?
The solution to the problem described above is obtained by solving the finite difference equations for displacements, transverse force and the tension in the geosynthetic. To check the accuracy of the solution the number, n, of elements in to which the length of the geosynthetic is discretised is varied. The results did not show any further improvement for n > 100. Hence, n=100 has been adopted for further analysis. The transverse force is calculate for a specified value of w&. Parametric studies have been carried out for w&=O.OOl-O.l; H=1-10 m; d-2-8 m; &=2Oo-4O0 and ~ 1 . 5 - 2 kN/m3. 0 The variation of normalised transverse force, P* with normalised front end displacement, WO,is depicted in Fig.4 for $,=25O. For low values of p (=k,LlyH)<1000, implying short reinforcement or large depths of embedment, the transverse force increases linearly with the displacement. The curves tend to become concave upward for p > 1000 indicating that larger forces are required to mobilise larger displacements. Longer reinforcement or reinforcement placed at shallow depth tends to deform significantly at larger displacements requiring greater forces to be mobilised. The variations of displacement profiles with distance for WO= 0.01 and +30° are shown in Figure 5 for different values of p. For very large values of !-t
(12)
dx
where p=k&/yH and Wo=wo/L. The boundary conditions become: at X=O, T*=O and at X=l, W=Wo and I
P*=pWo 1W.dX where P*=P,/yHL. 0
As the coupled equations can not be solved analytically, a finite difference approach is adopted. Eq.s (1 1) and ( I 2) in finite difference form become respectively
and
q:, = -1~ W , W , + 2}ta114~+ T,* n
where AX=l/n and n - the number of sub-elements in to which the reinforcement strip is divided into, Wr and T h e respectively the normalised displacement and normalised tension at node 5'. The normalised transverse force, P* is obtained from
Figure 5. Displacement profiles for qr= 30'
245
4 CONCLUSIONS Reinforcement in reinforced earth constructions and nails in nailed soil structures are rarely subjected to pure axial pull-out force. The kinematics of the problem often dictates a non-axial movement of the sliding soil and an imposition of an oblique force on the reinforcement. In this paper, an analysis of a reinforcement sheet embedded in soil at depth to a transverse force is proposed modelling the soil response by a set of Winkler springs. The governing differential equation is normalised and solved numerically to obtain normalised force versus normalised tip displacement relationships, normal displacement profiles, for a range of parameters considered. The former relation has been shown to be non-linear in view of the relative compressibility of the soil, andor due to relative length or embedment of the reinforcement. REFERENCES
Figure 7 . Normailised transverse force versus
Alfaro, M.C., Miura, N. and Begado, D.T. 1995. Soil-geogrid reinforcement interaction by pull out and direct shear tests. Geotechnicul Testing J., ASTM: 18(2):157-167. Bergado, D.T. and Long, P.V. 1997. Discussion leader's report: embankments. In H.Ochiai, N.Yasufuku and K,Omine (eds) Earth Reinforceinentt2:1015.1022. Hayashi, S., Makiuchi, K. and Ochia, H. 1994. Testing methods for soil-geosynthetic frictional behaviour - Japanese Standard. Proc. 5"' Int. Con$ on Geotextiles, Geomembrunes und Related Products:1:411-414. Jewell, R.A. 1996. Soil reinforcement with geotextiles. Publ. No. 123, CIRIA, 332p. Juran, I., Knochennmus, G., Acar, Y.B. and Arman, A. 1988. Pullout response of geotextiles and geogrids. In R.D.Holtz (ed.) Geosynthetics for Soil Iinprovement, Geotech. Spl. Publ. no.18:92-I 11.
Q,
the displacements are localised near the free end, the rest of the reinforcement remaining undisturbed. However, for p < 1000, the displacements progress towards the farthest end. The effect of the interface angle of shear resistance on displacement profiles is not very significant (Figure 6). The normalised transverse force varies almost linearly with the interface angle of shear resistance (Figure 7).
246
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Model experiment and analysis of sandwich earth fill reinforced with geosynthetics H. Nagashima Graduate School of Marirze Science and Engineering, Nagasaki University, Japan Nihonchiken Co. Ltd, Japan
Y. Tanabashi & H. I Civil Engineering Dept., Faculty of Eng. Nagasaki University, Japan
N. Fujise & H. Nakahara Nihonchiken Co. Ltd, Japan ABSTRACT: The one of the issues in Japan is to reuse marginal soils such as high water content volcanic cohesive soils and/or construction by-products which have never been used for fill materials. This paper proposes the conventional sandwich method conbined with geosynthetics for obtaining the long term stability due to both drinage and reinforcing functions of gravel sandwich layers and geosynthetics respectively. This paper describes the results from two series of steep earth fill model test. One is for sandwich earth fills without any geosynthetics to obtain the effect of the gravel layer itself. Another is for sandwich earth fills reinforced with geosynthetics varying coarce-grained soil layers to obtain the both fffect of the coarce-grained soil layers and geosynthetics. This paper also conducts Finite Element analysis to develop the simulation model for the proposed sandwich earth filling method conbined with geosynthetics. The results of the model tests have shown the sufficient effects of the proposed filling method. 1 INTRODUCTION
Table 1. Model experiment sheFification.
Recently, the increase of the surplus soil from a construction site, especially soft soil and clay, has become the one of the serious issue in Japan. It is difficult to keep the disposal ground for the such surplus soil and the disposal cost of these soils becomes expensive, too. Therefore it is a social demand to utilize such soils effectively. Recently, it has become necessary to use even high water content volcanic soil as earth fill materials. In addition to the shearing resistant force and drainage function in the sandwich method, instant increase of tensile strength by reinforcement materials is expected to provide both short and long term reinforcing functions. In this study, the authors observed the behavior of the model earth fill made by Sandwich method, in case of different thickness of the coarse-grained soil layer, and several reinforced cases. Also, the authors observed the effect of the thickness of coarsegrained soil layer on to the tensile resistance. 2 MODELTEST
Name Scale Experiment box width and shape of earth fill Condition of volcanic COhesive soil Condition of sandy soil Reinforcement material
I Contents I 1/10
I experiment box width crest 0.6m width of the base 0.9m height 0.5m w=47%, p=1.7g/cm3 , c= 1-2kPa w=3%, p=l .5g/cm3, (p=35" Geogrid Kz-2-20' T=2kN/m, E=4.5x 1O'MPa
by the air control was applied to simulate the load by the fills on top of that. the authors observed the behavior against the surcharge up to 47.5 kPa (equivalent to a 25m high earth fill). Table 1 shows the experimental specification. 2.2 Test materials The material of earth fill model was volcanic cohesive soil with high water content sampled from Mt.Aso, and the material of coarse-grained layer was sand. The material of reinforcement was geogrid.
2.1 Experimental method
2.3 Measurement installments
The authors assumed a case of an earth fill for road. Since usual compaction method can not be applied, the model was made by the handiwork at the medium density. A earth fill model was constructed to height H =0Sm as a base fill, and uniform surcharge
Figure 1 shows the arrangement of measurement installments. The authors installed the earth pressure meters and the pore water pressure meters in three places, and set up the displacement gauges in five places to measure vertial and lateral displacement at 247
Figure 3. Relationship between surcharge and fill settlement. Table 3. The comparison of the fill settlement and maximum lateral displacement. Maximum ment(rnm)
I NT I
S2N
I 0.49 97.68 I 0.87 54.86
0 0 0
119.04 1.00
0
58.22 0.49
@
54.78 0.46
@
28.94 0.24
@
-~~
The settlement of case S I N is 50-61% of that in case NN. The settlement in case S2T is 27-35% of that in case NN. The authors found the sandwich method with reinforcement materials is effective to restrain settlement. Whereas the case of clay with reinforcement materials inserted, the effect is smaller in restraining settlement.
Figure 2. The experimental cases (6 cases).
the top of the earth fill. In case of reinforcement earth fill, the authors installed the strain gauge in reinforcement materials. 2.4 Experimental cases The purpose of this experiment is inspection of the behavior of earth fill in different thickness of sand layer and combination effect of inserted reinforcement materials. Table 1 shows the six experimental cases, and Figure 2 shows the cross-section views of those models. In this model experiment, the following cases were examined and the results were compared.
3.2 The maximum lateral displacement in the slope (the upper step) Figure 4 shows the comparison of each case of maximum lateral displacement (the top). The tendency of maximum lateral displacement is same as the amount of settlement, and those in the case NN and NT are especially large. In case NN, the lateral displacement is 119mm at the load 47.5kPa, showing the upper part the laying earth was transformed greatly.
Table 2. The experimental cases. Case division
T ( reinforced soil )
3 EXPERIMENT RESULTS 3.1 Amount of settlements Figure 3 shows the comparison of the settlement of each case and Table 3 shows maximum settlements. The settlements in case NN and NT are especially large. The amount of settlement in case NN is 111 mm at the final surcharge 47.5 E a . 248
Figure 4. Relationship between suecharge and maximum lateral displacement of slope.
Figure 6 . Relationship between fill settlement and maximum slope lateral displacement /fill settlement. Figure 5 . Slope deformation of final state (lateral displacement).
The lateral displacement in case S1N is 49% of that in case NN. The lateral displacement of case S2N is 83% of that in case NN, and the lateral displacement of case S I T is 23% of that in case NN, the lateral displacement of S2T is 24% of that in case NN. So the lower clay layer of S2N, SIT, and S2T, did not move much in lateral.
3.3 Lateral displacement distribution in the slope (final states) Figure 5 shows final states of the fill slope deformation in each case. In no reinforcement cases, the deformation of the slope in case NN is significantly. On the other hand, the slope in case S1N (Sandwich) becomes flat, and the maximum lateral displacement is about half of that. The maximum lateral displacement in case S2N, where sand layer is thick, is about same as that in case SIN, but it is characteristic that displacement of lower clay is extremely small. The maximum lateral displacements in case S I N and S2N are about 50% of that in case NN. These results show the sandwich method of construction is effective for restraining slope deformation. On the other hand, in reinforcement cases, the lateral displacement in case NT is smaller than that in case NN (about 85%). The displacements in case S I T and S2T are extremely small. The authors found the transformation restraint effect was small by laying reinforcement materials in a single clay layer, but transformation was remarkably restrained by laying reinforcement materials in the sand layer of sandwich method. And, in case of using the reinforcement, the authors found the difference of displacement by thickness of the sand layer was small. Accordingly, the sandwich method of construction with reinforcement materials in thin sand layers was found effective for transformation restraint and stability of high earth fill.
3.4 Evaluation of the slope stability Figure 6 shows the relationship between the fill settlement and maximum slope lateral displacement per fill settlement at each 3.5 kPa increment up to 47.5 kPa. In case NN and NT, the curve incline changes steep at low surcharge of less than 20 kPa, there is a sign of the destruction in early stage. In case SIN, the curve incline change steep at 25 kPa, that is the sign of destruction. However, in case S2N, increase of incline is slow even over 30 Wa, and the settlement increases slow (in this case, there is the destruction tendency only upper part sand layer). On the other hand, in case S I T and S2T with reinforced sandwich method, the curve incline does not change and there is no sign of destruction. In Figure 8, the reinforced Sandwich method (case S I T and S2T) shows the large effect on the increase of stability. 3.5 Mobilized tensile force in reinforcement materials Figure 7 shows tensile force in reinforcement materials. The authors measured tensile force in reinforcement materials by strain gauge in case NT, SIT, and S2T, which are the case that reinforcement materials are inserted. Tensile force almost increases in a constant ratio in all cases at the surcharge up to 47.5 !@a. In case NT, the largest tensile force 0.9kN/m is measured at the distance of 50cm from the edge of the slope, which is significant at the center of loading plate. This means the tensile force is about 50% of reinforcement strength. The largest tensile force in case S I T is 0.45 kNlm at 45cm distance from the edge of the slope. The authors found that stresses are more spread and peak plateau is wider compared with case NT. In case S2T, the largest tensile force is 0.45 kN/m at 37.5cm distance from the edge of the slope. The authors found that the peak plateau is more spread than case SIT, and the peak is lower toward edge. Thicker the sand layer is, the nearer the maximum tensile force is to the slope edge, shallower the deformation zone becomes. 249
terials inserted in thin sand layer have enough reinforcing effect.
3.6 Earth pressure Figure 8 shows the relationsip between surcharge and earth pressure curve. In case of no reinforcement in the lower layer No.1, earth pressure in case S1N increases. The authors think vertical earth pressure by sand layer restraints. The authors think, the reason that earth pressure does not change in case NN is due to the large horizontal deformation by earth fill-material made from clay only. On the other hand, in cases of reinforcement, NT, SIT, and S2T, earth pressure increase linearly, in spite of different gradients. The authors think earth pressure declines in case S2N, because of circular slip of internal sand layer. Comparing with and without reinforcement, earth pressure is larger with reinforcement. The authors think vertical earth pressure decrease due to lateral flow in case of no reinforcement. In case of lower layer No.2, earth pressures increase linearly in all cases. Comparing with and without reinforcement, earth pressure in reinforced case is larger. The authors think this is due to the confining effect of the reinforcement materials. In case of upper layer No.3, earth pressure decrease after the peak in all cases, except case SIT. The authors think earth pressure decreases due to circular slip in the case of NN and NT with only clay.
3.7 Pore waterpressure Figure 9 shows surcharge-pore water pressure curve. In case of lower layer No. 1, the pore water pressure decreases after a peak in all cases. The authors think this is due to the drain effect generally. In case of lower layer No.2, it is difficult to drain because measurement position is lower right in the structure. In cases except S2N and S2T, the authors think the pore pressure increases linearly. In case
Therefore, the thickness of the sandwiched sand layer has no effect the tensile force and the tensile force is relatively small in case of the thin (6cm) sand layer. So, the authors found reinforcement ma-
Figure 8. Surcharge -Earth pressure curve.
250
Figure 9. Surcharge - Pore water pressure curve.
sandwich method (case S 1 T, deformation is 35% of that in case NN) than thick sandwich method (case S2T, deformation is 50% of case that in NN). S 1N>S2N>S 1T @ The maximum tensile force in reinforced materials is largest in case NT with 0.9kN/m, and is same in case S1T and S2T with 0.45kN/m. The thicker the sand layer becomes, the nearer to the slope edge the maximum position is.
S2N and S2T, there is a tendency that pore pressure decreases after a peak due to effect of the thick sand layer. In case upper layer No. 3, the pore pressure decreases after a peak in all cases. It is considered that this cause is the drainage effect.
3.8 Summary of experimental result
0Amounts of deformation are in the following order. N N > m > S 1N>S2N>SlT>S2T The deformation in case NN made from clay only is extremely large. However, it is possible to decrease the deformation to some extent if the sandwich method adopted (the S series) (the deformation is 50-60% of that in case NN). Furthermore, it is possible to decrease the (S2T deformation extremely by inserting reinforcement materials (the T series) in the sand layer (the deformation is 27-35% of that in case NN). However, the effect in case NT, which is not sandwich method, is small and the deformation in case NT is 85% of that in case NN. @ The effect of reinforcement by inserting reinforcement materials in clay is small (the deformation is 85% of case NN). In case SIN, sandwich method in thin layer is rather effective (the deformation is 50% of that in case NN). NT>S 1N @ Scales of circular slip are in the following order. NN>NT>S lN>S2N There is no evidence of sliding in case S 1T and S2T. @ In case of no reinforcement, the restraining effect of sliding deformation is smaller in S1N (the deformation is 60% of that in case NN). When they are reinforced by the sandwich method, the restraining effect of sliding deformation is larger in thin
4 PULLING TEST OF REINFORCEMENT MATERIALS The authors conduced the tensile strength test (pull out speed is l m d m i n ) , in order to determine the friction between reinforcement materials and soil, and calculated the apparent cohesion c*, the coefficient of reduction a, and the gradient of straight line p. The maximum tensile shear strength can be calculated by the followi?g equations. T~~~ =oN-tanp+c =a*oN*tan@c* By these equations, the maximum tensile shear strength in volcanic ash soil is T max =3.6 (Wa), and, that of sand soil is z max = 0.366 ON tan35.05(") +0.82 (kPa).
5 VANETEST Natural water content of volcanic ash soil used in this model test was about 45%, and this shear strength was about c=IO kPa. It is desirable to reduce strength around 1/10. Therefore, in order to make consistent samples, the authors set up the target water content (47%), and strength around z=1-2 Wa. As a result the authors found this cohesion is c =OS-2.5 Wa.
25 1
Table 5. The comparison of the experimental values and FEM analytical results. Fill settlement
I
1
(mm)
Experimental values
FEM
1
A maximum lateral displacement(mm) Experimental values
1
SIN S2N SIT S2T
Figure 10. Deformational analysis.
25 21
31
6 ANALYSIS 6.1 Stability analysis The authors conducted stability analyses assuming a circular slip surface for three cases NN, SlN, and S2N in accordance to C =5 kPa. The safety factor of each case is roughly equal to 1, so It is considered that Analysis result is consistent with the deformation behavior by experiment (figure 10). 6.2 FEM analysis parameters
Figure 1 I . The example of deformation analysis results.
The authors assumed the constitutive model of soil is linear elastic body. Table 4 shows analysis parameters. Modulus of elasticity of volcanic cohesive soil is determined based on the results from parametric study.
values of lateral displacement are all smaller than experimental values, except sandwich method by reinforcement (case S I T and S2T). However, the settlement clearly shows the effect of reinforcement. On the other hand, lateral displacement proves the effectiveness of sandwich method, but it does not prove the effectiveness of reinforcement.
, Table 4. Analysis parameters. The earth fill materials
Youn 's modulus E( kPa) Densit of soil (kN/m.) The oisson's ratio v
Volcanic cohesive soil V
17.0 0.45
Sandy soil S 5,000 15.0 0.35
7 CONCLUSION
The authors found following results in this test 0 The construction of high earth fill with steep slope is possible by using reinforcement materials to the traditional sandwich method. 0 The sandwich method with reinforcement materials in thin sand layers is very effective in decreasing the deformation. 0 Tensile strength in reinforced materials of sand layer is not very mach influenced by the sand thick.
6.3 Comparison of the experimental values and FEM analytical results. Table 5 shows the comparison of the experimental values and FEM analytical results (Linear elasticity analysis). Also Figure 11 shows the example of deformation analysis results. The settlement is close to the experimental value, except the cohesive soil with a single layer (case NN and NT). But, the analytical
252
REFERENCES The Geogrid workshop 1990. A guideline Geogrid cobstraction method. Japanese geotechnical society Kyushu branch office. 1995. A design and execution of the special ground of Kyushu and Okinawa.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Reinforcing function of a liner system by geotextile and geogrid S. Nakamura Rivive Division, Okasan Livic Co., Japan
S. Imaizumi Graduate School of Engineering, Utsunomiya University, Japan
K. Kuzumaki Graduate School of Engineering, Utsunomiya University, Japan
ABSTRACT: When the base of landfill subsides, deformation occurs in the geomembrane which is used as a component of liner. Usually, non-woven textile is being used to protect a geomembrane from mechanical damage. This non-woven geotextile seems to function as reinforcing the geomembrane when it deform. A geogrid if it was used together with the non-woven textile may emphasize this reinforcing. That influence was examined here. For example a geogrid is used together in HDPE. Assuming the amount of maximum distortion with only geomembrane is 100, it is amount with not only but geogrid becomes 53. A geogrid is thought to show effect by not only the frictional decrease of the geomembrane but also rigidity increasing.
1 INTRODUCTION A geomembrane is being used as a main component of a liner system in waste landfill. This functions to prevent the polluted water (leachate) from flowing into underground. When local subsidence may happen at the bottom of waste landfill, deformation of the liner occurs. Then strain creates within the geomembrane. As for the deformation and/or the strain of geomernbrane caused by subsidence of the base, some evaluation-methods have been proposed. Usually, a geomembrane is placed on the base ground together with the non-woven textiles to prevent it from mechanical damage. The authors thought that use of the geogrid as well as geotextile may emphasize the function of protecting and reinforceing the geomembrane. In this paper, they conducted the trapped door tests of modeled liner which was composed of geomembrane made from HDPE (High Density PolyEthylene) or FPA (Flexible Polypropylene Alloy) underand overlying the geotextile, and of geomembrane underlying geotextile and overlying geogrid.
Figure 1. Outline of device.
overlying the liner system including geomembrane. The sand layers with a thickness of 20cm were underlaid and overlaid the liner system. The dimension of the liner system is a length of 180cm and a width of 80cm. An air-bag made of rubber was installed on the upper layer of the sand. This air-bag can apply the sand layer the pressure up to 255kPa by the air compression machine. The steel-cap was fixed to the container by the bolts. Two kinds of geomembrane were used. One is HDPE (High Density PolyEthlene) with a thickness of 1.5 mm, and the other is FPA (Flexible Polypropylene Alloy) with a thickness of 1.5mm. Forty-two strain gauges in total were pasted on the both surface of geomembrane. These points are center of geomembrane, each 5cm far (10 points), and each lOcm far (10 points) toward the outside. Therefore,
2 THE OUTLINE OF THE EXPERIMENT 2.1 Device and used material The outline of the device is shown in Figure 1. This device is composed of steel container having a length of 250cm, a width of 90cm and a depth of 70cm. The bottom board of the container is divided into three parts. The center part of the board with a length of 30cm can settle by the jack and the motor. This simulate the local subside of the base ground 253
axial strain and bending strain can be estimated at 21 points of geomembrane. The sand used for the experiment is made from crushed stone. This sand was sieved for the grain size to distribute between 8 4 0 and ~ 74pm. It's internal friction angle that was determined by tri-axial compression test is about 48". The used geotextile as protection layer is a stapled non-woven geotextile with a thickness of 10 mm reinforced by short fiber. Geogrid is also used in the experiments. This geogrid made from polypropylene was enlarged biaxially to form big net. The physical and mechanical properties of these materials are listed in Table I. Table 1. Properties of geosynthetics. Geornernbrane
HDPE
FPA
Density (g/cm3)
0.95
0.9
Tension strength (20°C)(MN/m2)
33.5
Breaking elongation (20°C)(%)
860
Liner expansion coefficient ( X 10-4"C)
1.65
22.7 720 1.2
Elastic modulus (20°C)(MN/m2)
484
81.9
Geogrid
longitudinal transverse
20.0
Reference strength (kN/m)
10.0
Creep strength
3.0 6.0 Polypropylene
(kN/m)
Quality of the material
where geogrid was placed on the geomembrane and geotextile was placed under the geomembrane were Case.4 and Case.8 for HDPE and FPA. First, the sand was poured into the container for its thickness to be 20cm. Then its surface was smoothed. When the geotextile was used as bottom protective layer, it was placed on the sand. Then geomembrane was spread on the geotextile. In case the geotextile was not placed, the geomembrane was spread directly on the lower sand. Then, after geotextile or geogrid was placed on the geomembrane, upper sand layer with a thickness of 20cm was compacted. In case no geotextile or no geogrid was placed on the geomembrane, upper sand layer was directly compacted on the geomembrane. The wire that was used to measure vertical displacement of geomembrane was glued to the center on its bottom surface. This wire was connected to displacement transducer outside of the container through the settling board. Forty-two strain gauges were lead to the data logger outside of the container through its side-wall. The pressure of 49kPa was applied on the upper sand layer through the air bag. Center part of base board, then, was lowered by electrical motor screw at a rate of l m d m i n . The amounts of strains and vertical deformation were measured and recorded each 0.5mm of base settlement. It lasted until the amount of lowing of board reached to 65mm.
Non-woven rreotextile Thickness (mm)
10
Density (g/cm3)
0.1 2
Tension strength (2OOC)(MN/m2)
1.27
Breaking elongation (20°C)(%)
154
Permeability coefficient (cm/sec)
3 RESULT AND DISCUSSIONS
3.1 Vertical displacement and elongation of the geomembrane
2.65x10-'
2.2 Testing cases and procedure Table 2 showed the testing cases. The tests where no protective layer was used were case1 and case5 for HDPE and FPA, respectively. The tests where geo textiles were placed both on and under the geomembrane were case2 and case6 for HDPE and FPA. The tests where geogrid was placed on the geomembrane were case3 and case7 for HDPE and FPA. The tests Table 2. List of the experiment. Case Geornembrane Geotextile Geogrid Pressur Temperature (kPa)
("C)
15.4
HDPE
Without
Without
49
HDPE
With
Without
49
19.7
HDPE
Without
With
49
-8.4
HDPE
27.1
With
With
49
FPA
Without
Without
49
2
FPA
With
Without
49
27
FPA
Without
With
49
-7.1
FPA
With
With
49
5.2
254
Firstly, axial strains at 2lpoints were estimated as an average value of strains measured on top and bottom surfaces of geomembrane at given settlement of central base board. Then the axial strains were integrated along the axial of geomembrane to result in its elongation. Figure 2 shows changes of the amount of vertical displacement (4, elongation (2AL) of geomembrane, and the ratio of (2AL)l (4 versus settlement of base board. In case of HDPE, the vertical displacements of geomembrane are somewhat smaller than the settlement of base and the trends are almost similar whether there are protective layer or not. In case of FPA, which is more flexible than HDPE, it can be recognized that vertical displacement of geomembrane is almost equal to the settlement of base when no protective layer was placed. However, when some protective layer as geotextile or geogrid was placed on or/and under the geomembrane, vertical displacement beyond about 30mm of base settlement is somewhat smaller than that in case of without any protective. In both cases, among three types of placement of protective layer, geotextiles on and un-
Figure 2. The amount sinking of Base and Geomembrane.
der the geomembrane seems to give a bit smaller vertical displacement. As for the elongation, it was largest for both case of HDPE and FPA when no protective layer was placed. It can be found that the elongation is the smallest when geotextile was placed under geomembrane and geogrid was on the geomembrane. An amount of the latter is about 50% of the former in case of HDPE and about 40% in case of FPA. Therefore, placing the geotextile and/or geogrid on and under the geomembrane functions to reduce the vertical displacement of geomembrane when the base subsides partially and then results in a reduction of elongation. This is due to the increase of rigidity of liner system including geomembrane.
3.2 Distribution of the axial strain of geoiizenzbrane Figure 3 shows the distribution of axial strain of geomembrane when the elongation is 5mm, where the axial strain was estimated as a mean of an amount of strain gauges pasted on top and bottom surfaces. The cases where no protective layer was used show the largest value of strain at center point both for HDPA and FPA. The cases where geotextile was placed under the geomembrane and geogrid placed on show the smallest value. However, the axial strain creating at the point far from the center, for example 50cm far, the former cases seem to show the smallest. AS the reason for this difference, it is thought theoretically that the case without any pro255
Figure 3. Distribution of axial strain at elongation of 5mm.
tective layer has lower rigidity and high froctional coefficient between geomembrane and adjacent material. It is also seen that the range of strain distribution of HDPE is somewhat larger than that of FPA. This is due to the fact the rigidity of HDPE is much higher than FPA. To see the effect of placing the protective material on axial strain distribution, an amount of strain at given point with protective layer was divided by an amount corresponding point incase of without any protective layer. Figure 4 shows the distribution of the ratio (%). The ratio is less than 100 within the central 40cm and more than 100 beyond it. This trend is somewhat considerable in case of FPA, having lower rigidity. As for the type of materials of protective layer, placing the geotextiles both on and under the HDPE geomembrane gives the most marked redistribution of the axial strain, for example, 53% at the central point. Placing geogrid on and geotextile under the FPA geomembrane give also remarkable effect of strain redistribution, as 59% at central point. Placing the geotextile only under the geomembrane seems to do a little effect of redistribution, as 83% and 70% for HDPE and FPA respectively. Using the geogrid on the geomembrane, it is though that rigidity of the liner system, of course, may increase. But the frictional coefficient on the surface may not change because the sand can contact
Figure 5. Axial- and bending strain versus base settlement. (HDPE)
256
with the surface of geomembrane through open area of geogrid. As the result, the effect of redistribution of strain is as similar as the case with placing the geotextiles on both sides.
~i~~~~4. Distribution of strain ratio.
3.3 Distribution of the bending strain of the geomembrane v
Figure 5 shows the change of axial- and bendingstrain at the points 0, 15, 25 and 35cm far from the central point for HDPE geomembrane with settlement of the base board. At the central point, axial strain in case of without any protection layer gives the highest values. The axial strain decreases in order of the case of with geotextile under HDPE, the case of with geotextiles on both sides, and the case of with geogrid on and geotextile under HDPE. At the point of 25cm far, the case of with geotextiles on both side gives the smallest strain. As to the bending strain, positive value means that the geomembrane deforms concavely. When the geomembrane was spread without any protective layer, the bending strain beyond 30mm of base settlement at points of 15cm indicates negative one, this means it deforms convexly, though other cases indicates positive one. These behavior come from the difference of their rigidity. Figure 6 shows the distortion of FPA for each position of 0,15,25 and 35cm from the center. Bending strain is stable in the position of 15cm at the time of a sinking 30mm of case7 which a geogrid was used for. The value of the bending strain is reduced at the stage. They fall down slowly as for which case as well after the value of the bending strain increased
Figure 6. Axial - and bending strain versus base settlement. (FPA)
257
from the negative value and became maximum distortion value. The bending strain of case6 in which a nonwoven textile was used for is big, and it is moving to the negative side in the position of 25cm. It shows a tendency of being the same even in the position of 35cm. As friction is small, it is because the influence of sinking appears in the wide range. And, these mean it deforms convexly, though other cases indicates positive one. Axial strain shows the increase which is comparatively simple. The value of the axial strain of case5 is big and the influence of the non-woven textile and the geogrid is big in the central point around the cenLnr
LGL*
4 CONCLUSION The authors conducted trapped door tests of geomembrane sandwiched by sand-layer. Material of protective layer that was installed between geomembrane and sand was changed such as geotextiles Onand under the geomembrane, geogrid under the geomembrane, and geotextile under and geogrid on the geomembrane. Two types of geomembrane, HDPE and FPA, were used. Main results obtained -from the experiments are follows.
( I ) When a non-woven textile and a geogrid are used, the amount of sinking of the geomembrane and stretch can be restrained. As for the early stages of sinking, the effect by the non-woven textile is bigger than the effect of the geogrid. However, sinking in the one which a geogrid was used for when sinking grew more than 40mm. (2) Bending strain is big, and proceeds at the negative side with the progress of sinking when the distortion of HDPE is seen in the position of 25cm from the center. It shows a tendency of being the same even in the position of 35cm. On the other hand the absolute value of the bending strain which a non-woven textile and a geogrid were used for is bigger than the case in which they are not used. If a non-woven textile and a geogrid are used, the influenced range of sinking grows big. The size of the influence range is geogrid + non-woven textile, geogrid, nonwoven textile and use-less in the order. (3) The distortion of FPA falls down in the position of 15cm after the value of the bending strain increased from the negative value as for which case as well and became maximum distortion value. The bending strain of case6 in which a non-woven textile was used for is big, and it is moving to the negative side in the position of 25cm. It shows a tendency of being the same even in the position of 35cm. As friction is small, it is considered that these phenomena are because the influence of sinking appears in the wide range. (4) Axial strain shows the comparatively simple increase. The value of the shaft distortion of case5 is big, and the influence of the non-woven textile and the geogrid is greatly shown in the position of 15cm around the center.
25 8
( 5 ) When the distribution of the distortion is researched, distortion in the center decreases, and both geomembranes are on the increase in the circumference part. Assuming the amount of distortion of case1 is 100% in HDPE, the amounts of each distortion of case2, case3 and case4 become 83%, 53% and 59%. Supposing the distortion of case5 regarding FPA is 100, the amounts of each distortion become 68% (case6), 66% (case7) and 59% (case8). On the other hand the size of the distortion far from the center grew conversely large. FPA of effect on a decrease of the maximum distortion by the non-woven textile and the geogrid was bigger. Particularly, the effect of the geogrid was shown with FPA remarkably. As for this, rigidity is enhanced by the geogrid regarding FPA more than HDPE. It can reduce the friction to install non-woven textile or geogrid at landfill sites, which can also increase their rigidity. Therefore the result stated above is important regarding the safety of the liners at the landfill sites. REFERENCES Giroud,J,P. 1995. Quantification of Geosynthetic Behavir, Proc. of the 5'" Int. Con$ on Geotextiles, Geomembmnes and Related Products, Special Lecture & Keynote Lectures, 23-24, Singapore. Imaizumi,S. Yokoyama,Y. Takahasi,S & Tuboi,M. 1996. Elastic Formula for Pull-out Behavior of Embedded Geomembrane, Proc. of the 12'" Southeast Asian Geotechnical Corzf., Vol. 1, 57-62, Kuala Lumpur. Imaizumi,S. Futami,T. & Nomoto,T. 1998. Tensile behavire of embedded geomembrane subjected to differential settlement of base ground, Proc. oftlie 13'" Southeast Asian Geotechnical Con$, Vol. 1, 325-330, Taipei.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
The design of steep slopes constructed from cohesive fills and a geogrid P.J. Naughton University College Dublin, Ireland, formerly Terram Ltd, UK
R.A. Jewell Consultant, Belgium
G.T. Kempton Terram Ltd, Mamhilad, Pontypool, UK
ABSTRACT: The shear strength and consolidation properties of compacted cohesive fills are reviewed and are found to vary greatly between different soil types. Recent research on dissipating excess pore water pressure in steep slopes constructed from cohesive fill is also reviewed. A simple and practical design method for constructing steep slopes from cohesive fill is proposed. The design method estimates both the time to dissipate excess pore water pressures and the magnitude of the resultant settlements. It also estimated the required transmissivity required by the drainage elements. Guidance is also given on controlling seepage into the reinforced soil block. interaction coefficients between cohesive fills and geogrid that may be used for design. A design example is presented to illustrate the design method, and the range of strategies for construction. These range from relatively fast construction that results in higher construction pore water pressure, to relatively slow construction but with relatively little pore water pressure generated in the slope at any stage. Faster construction requires additional reinforcement compared with slower construction. A related problem for a steep slope constructed from cohesive fill is the possible seepage of water into the reinforced block over the life of the structure from the unreinforced fill that is retained. Such seepage could cause an increase in pore water pressure in the reinforced fill, reducing stability and potentially causing deformation. Guidance is provided on how such seepage from around the reinforced block may be controlled.
1 INTRODUCTION Free draining granular fill is conventionally used for the construction of steep reinforced slopes. Such fills have relatively high shear strength characteristics and their free draining properties minimise pore water pressures in the slope. Research (Zornberg & Mitchell, 1994, Kempton et al., 2000) and long term case histories (Inada et al., 1978 and Fukuoka, 1998) have shown, however, that finer grained and cohesive fills can be used to construct satisfactory steep reinforced slopes, as long as adequate drainage is provided within the slope. The use of a cohesive fill in a reinforced slope would typically result in the generation of excess pore water pressures during construction. Such pore water pressure reduces the shearing resistance in the fill, and reduces the bond between the geogrid and the fill. Continuing deformation and settlement of the slope can result. This paper reviews the strength and consolidation properties of cohesive fills as they relate to their potential use in reinforced slopes. Results from a laboratory research program (Kempton et al., 2000) into the ability of a geogrid combining reinforcement and drainage functions to dissipate pore water pressures in clay soil are presented and analysed. Based on the research program, a simple and practical design method is presented which deals with the excess pore water pressures and settlements that would develop at each stage of fill construction. The design method is considered to be applicable to both saturated and partially saturated cohesive fills. Recommendations are made on the
2 THE PROPERTIES OF COHESIVE FILL The shear strength and consolidation properties of compacted cohesive fill can vary greatly. The shear strength properties are a function of the plasticity index, mean effective stress and soil density, while the consolidation properties greatly depend on the grading of the fill. Jewell (1996) provides guidance on the selection of appropriate strength parameters for cohesive fill. However, a simple and practical method to relate the state of a compacted cohesive fill with a peak angle of friction and dilation has not yet been developed. Typical values for the angle of 259
6 No clogging or wash through of fines into the
friction of compacted cohesive fills from the literature are presented in Table l . It should be noted that strength parameters can vary considerably between cohesive fill types. The coefficient of consolidation, CV can also vary greatly between fill types. Table 2 presents values of CV for various cohesive fills determined in the laboratory (CV Lab) and from site measurements (CV In situ). It should be noted that the CV values of marginal fills could be significantly higher than those of cohesive fills.
drain over the test period was noted.
4 OVERVIEW OF THE DESIGN OF STEEP SLOPES WITH COHESIVE FILLS Codes of practice differ in their treatment of cohesive fills. Table 3 summarizes four codes widely used in design. While both UK coded allow for the use of cohesive fill the two US codes specify maximum values of fines. It should be noted that ever these requirements could be met by a wide range of slow draining or marginal fills.
Table 1. Strength properties of compacted cohesive fills. Soil Name Keupar Mar1 Fort Creek Dam Lmcate-leBarcares
Plasticity Index %\
$' (")
19
27
-
31
23.5
20
21
25
15
c' (Wa)
Reference Table 3. Approach to the use of cohesive fill adopted by codes of practice.
Cox (1978) Lieszkowszky (1978) Thompson et al. (1978)
Code of Practice BS 8006 (1 996) UK
HA 68/94 (1994) UK
Table 2. Consolidation properties of compacted cohesive fill, after Vaughan et al. (1978). Soil Name Usk Fill Selset Llyn Brianne Derwent
Plasticity Index (%) 8 14 10 19 - 22
Cv Lab (m2/ year) 8.9 0.9 - 3.1 15-20 1.3
FHWA (1998) US
Cv In Situ (m2/year) 12.3 I .4 - 3.6
NCMA (1997) US
-
Requirements Cohesive fill maybe used in conjunction with an appropriate reinforcement for reinstatement or new slopes. No comment on cohesive fills but does not prohibit their use. Allows soils with a gradation of up to 15 % passing No 200 sieve (0.075 mm). Allows soils with a gradation of up to 35 % passing No 200 sieve (0.075 mm).
1 - 1.7
5 DESIGN METHOD 3 RESEARCH ON THE USE OF A COMBINED REINFORCEMENT DRAINAGE GEOGRID
Based on the research presented by Kempton et al. (2000) a design method has been developed for constructing steep slopes from cohesive fills. The aim of the design method is to dissipate any excess pore water pressures present in the slope during the construction stage. This results in an increase in the shear strength of the fill and enhanced bond between the reinforcement and the fill. It also allows vertical and horizontal deflections to be controlled as construction proceeds. The design method is shown diagrammatically in Figure 1.
Kempton et al. (2000) presented the results of a laboratory investigation into the ability of a new combined reinforcement drainage geogrid to dissipate excess pore water pressures in cohesive fills. The laboratory study used English China Clay (coefficient of consolidation 1.3 to 2.3 m2/year). The main conclusions of that study are: The new combined reinforcement drainage geogrid was seen to dissipate the excess pore water in the fill to 20 % of its initial value in 32 hours, It was noted that the initial excess pore water pressure in the immediately vicinity of the new combined reinforcement drainage geogrid only reached 40 % of the applied stress, Dissipation of excess pore water pressures occurred on both sides of the new combined reinforcement drainage geogrid even though the drainage channel was only on one side, Increased pullout resistance was recorded both after full and partial dissipation of the excess pore water pressure, Adequate transmissivity is provided to remove water from soil even at low hydraulic gradients,
5.1 Dissipation of excess pore water pressure generated during construction Analyses of the dissipation test results (Kempton et al., 2000) reveal that the dissipation of excess pore water pressures using the new combined reinforcement drainage geogrid can be reasonably modeled using the following equation:
where, FDiss = a factor of safety applied to the calcu-lated dissipation time, C = is a constant de-
260
test data to field conditions. At present a value of FDISS= 2.0 is recommended for use in design. It is also proposed to limit the lift height to a maximum of 0.5 m. An average degree of consolidation of 80 % is desirable at the end of constructing an individual layer. When a new layer is constructed an excess pore water pressure will be generated in the previously constructed layers. The excess pore water pressure in any layer immediately after placing a new lift is:
termined from Figure 2, CV = the coefficient for consolidation (m2/year) of the fill. The factor of safety F D I ~isSincluded to account for uncertainties in the determination of CV, variability in the cohesive fill and extrapolation of Estimeie li@:ghtof eacn lift. (maaimuin recc r n m m W i height 0 5 rnl
Epwp in layer i = h y + ( n - l)[$~r,,~] where, h = the height of each layer, y = unit weight of the fill, n = number of layers constructed, rui= is the pore water pressure parameter in layer i.
5.2 Calculation of vertical deflection and volume of water leaving each layer. Settlements in each layer of the slope will arise as dissipation of the excess pore water pressure proceeds. In practice the magnitude of the settlement will be small and will generally be corrected automatically as construction proceeds. The magnitude of the settlement in each layer can be calculated from:
at each lit;
Gesign slope using eftecti\e stress analysis, and c r, tG be selected based ui7 degee cif pot-2 watsr pressure dissrptiorr ~&ieveci,rriiniiium iu = 0 2 \r
Error! Objects cannot be created from editing (3)
field codes.
where, 6h = the settlement of the layer, h = the initial height of the layer, mv = the coefficient of volume compressibility, Aa,' = the change in the vertical effective stress and is equal to
4 Check transrnismty of combined reiiiioicemmt Jrainage geogrid
\
I
/
AD,,'= $Z+q
(4)
where, y = is the bulk unit weight of the fill, h = the lift height, q = the surcharge due to construction traffic. Where the degree of saturation of the fill is greater than 90 %, the volume of water leaving any soil layer can be determined from the magnitude of the vertical settlement per unit plan area:
Figure 1 . Flow chart of design method.
v = &*loo0
(5)
where, V = the volume of water, in litres, leaving a soil layer, 6h = the settlement of the layer in metres.
5.3 Designing the slope for the ultimate limit state Dissipation of the excess pore water pressure during the construction phase will increase the shear strength of the cohesive fill. An effective stress analysis, using the constant volume angle of friction, &', can therefore be used in designing for the ultimate limit state. A minimum ru value of 0.2, corresponding to a degree of consolidation of 80 %, is recommended for all designs.
Figure 2. Constant C for various lift heights.
26 1
The coefficient of bond between reinforcement and the soil is directly proportional to the angle of friction of the soil. Values can be determined for a particular fill by carrying out shear box testing or alternatively for the new combined reinforcement and drainage geogrid the values in Table 4 can be used. The coefficient of direct sliding for the new combined reinforcement drainage geogrid is 0.7. Table 4. Values for coefficient of bond for different ranges of 4’. Angle of friction (4’) < 20 20 - 25 2.5 - 30 > 30
Figure 3. Slope used in design example.
Coefficient of bond 0.4 0.5 0.6 0.7
Table 5. Consolidation properties of the fill. Confining stress (kPa) Cv (m’/year) m, (m’/MN)
5.4 Checking that the transrnissivity of the geogrid is adequate
0 - 40 10 0.5
40 - 75 8 0.4
75 - 100 6 0.3
proposed to construct the slope using vertical lifts of 0.5 m. Using the proposed design method the dissipation time for each layer can be calculated, the resultant settlements estimated and the transmissivity checked. The above values have been determined for each layer after the construction of layer 10, the final layer in the slope and are presented in Table 6. The longest dissipation time is estimated at 103 hr (4.29 days) in the bottom two layers and the maximum vertical settlement of 2.25 mm occurs in layers 7 to 10 at the top of the slope. The total vertical settlement after constructing the final layer is 18.9 mm. The maximum transmissivity required after constructing this layer is 0.036 l/m.hr. The hydraulic gradient has been calculated in the body of the slope 4 m from the face. In the design of steep slopes using this design method it is necessary to establish a balance between dissipating excess pore water pressures and arriving at a satisfactory construction time. Figure 4 shows the excess pore water estimated at the base of the slope after construction of the first 3 lifts, 1.5 m high slope. While dissipation of excess pore water pressure to 70 % considerably reduces the construction time it results in a residual pore water pressure in the slope that may cause
The transmissivity of the new combined reinforcement drainage geogrid at different confining stresses and hydraulic gradients are presented in Kempton et al. (2000). When the volume of water leaving the soil and the time for consolidation to occur are both known, the required transmissivity can be calculated and compared with the transmissivity available. The transmissivity needs to be checked twice, firstly immediately after construction of a layer, when the hydraulic gradient is high but the confining stress is low and secondly at the end of construction when the hydraulic gradient is low but the confining pressure is high. If the transmissivity provided is inadequate then the lift height should be reduced.
6 A DESIGN EXAMPLE USING THE NEW DESIGN METHOD A design example is presented to demonstrate the proposed design method. Figure 3 shows a 5 m high steep slope inclined at 70’ to the horizontal. The bulk unit weight of the fill is 18 kN/m3 and the consolidation properties are given in Table 5. It is
Table 6. Dissipation times, vertical settlement and transmissivity after placing the final lift. Layer 1 Height of slope (m) 0.5 Confining stress (kW 90 Dissipation time (hr) 103 Vertical settlement in each layer (mm) 1.35 Transmissivity (Vm.hr) 0.013 Hydraulic gradient after 1 hr Hydraulic gradient after 80 % dissipation
2
1 .o
3 1.5
4 2.0
5 2.5
6 3.0
7 3.5
8 4.0
9 4.5
10 5.0
81 103
72 77.6
63 77.6
54 77.6
45 77.6
36 62.1
27 62.1
18 62.1
9 62.1
1.35
1.8
1.8
1.8
1.8
2.25
2.25
2.2.5
2.25
0.013
0.023
0.023
0.023
0.023
0.036
0.036
0.036
0.036
0.225 0.018
262
Figure S. Development of seepage pressures inside the reinforced block.
Figure 4. Excess pore water pressure at base of slope after constructing 3 layers (1 .S in).
short term instability. Dissipation to 90 % increases the construction time but reduces the residual pore water pressure in the slope. The effects of dissipating the excess pore water pressure to 70 and 90 % of the initial value are summarized in Table 7. In the situation where a construction time of 135 hr and 90 % dissipation is required the height of each lift should be reduced to 0.24 ni.
Figure 6. Control of seepage into reinforced block.
Constructing base and back drainage outside the reinforced block, Figure 6, can best control un wanted seepage. This method of construction has a number of advantages: The base and back drainage can be independently designed to control any seepage conditions present. The drainage component in the reinforcement block can be designed for a short design life, until the excess pore water pressure in the slope is dissipated. This will reduce the cost of the combined reinforcement drainage geogrid as one drainage configuration can be used to dissipate excess pore water pressures rather than having a range of drainage elements to meet particular site conditions. Removal of the seepage water before it enters the reinforced block will remove any danger of piping channels developing which could have an adverse effect on the stability of the reinforced zone irrespective of the quantity of reinforcing or drainage elements present in this region. It is recommended that suitable granular fill with adequate permeability wrapped in geotextile is used for both the back and base drain.
Table 7. Summary of dissipating excess pore water pressures to 70 and 90 C/o of initial value.
Construction time (hr) Average residual excess pore water pressure (kPa) Average ru value in layer 3 at end of construction
70 % dissipation
90 % dissipation
132.4
278.6
1.87
0.50
0.42
0.1 1
7 ASSOCIATED PROBLEM OF CONSTRUCTING STEEP SLOPES FROM COHESIVE FILLS
An additional problem associated with the construction of steep slopes from cohesive fill is the development of seepage pressures in the reinforced block, Figure 5 . Seepage pressures can lead to a reduction of both the shear strength of the fill and the bond between the reinforcement and the fill.
263
Zornberg, J. G. & Mitchell, J.K. 1994. Reinforced soil structures with poorly draining backfills. Part I: Reinforcement interactions and functions. Geosynthetics International, l(2): 103 - 148. Jewell, R. A. 1996. Soil Reinforcement with Geotextiles. CIRIA Special Publication 123, London. Kempton, G.T. Jones, C.J.F.P. Jewell, R.A. & Naughton, P.J. 2000. Construction of slopes using cohesive fills and a new innovative geosynthetic material. EuroGeo 2, Bologna: 825 - 828. BS 8006 1995. Code of practice for strengthenedreinforced soils and otherfills. British Standard Institution, London. HA 68/94 1994. Design methods for the reinforcement of highway slopes by reinforced soil and soil nailing techniques, Design manual for roads and bridges. HMSO, London. FHWA 1998. Mechanical stabilized earth walls and reinforced soil slopes design and construction guidelines. Elias, V. & Christopher, B.R. Authors. FHWA-SA-9607 1, Washington. NCMA 1997. Design manual for segmental retaining walls. Collin J.G. Ed. Herndon, Virginia. Cox, D.W. 1978. Volume change of compacted clay fill. Clay Fills, ICE London. Lieszkowszky, I.P. 1978. Fort Creek Dam - impervious clay core. Clay Fills, ICE London. Thompson, G.H. & Herbert, A. 1978. Compaction of clay fills in situ by dynamic compaction. Clay Fills, ICE London. Vaughab, P.R. Hight, D.W. Sodha, V.G. & Walbancke, H.J. 1978. Factors controlling the stability of clay fills in Bntain. Clay Fills, ICE London.
8 CONCLUSIONS The properties of cohesive fills are presented and shown to have a wide range of shear strengths and generally low permeability leading to long pore pressure dissipation times. Recent research has shown that a new combined reinforcement and drainage geogrid can dissipate excess pore water pressures during the construction period, resulting in an increase in the shear strength and enhanced reinforcement soil interaction. A new simple and practical design method is presented which can be used to estimate the time to dissipate excess pore water pressures generated during construction. Methods of dealing with advancement of wetting fronts and seepage into the reinforced soil block are also presented.
REFERENCES Fukuoka, M. 1998. Long-term deformation of reinforced cohesive soil fills and walls. 6"' International Conference on Geosynthetics, Atlanta: 8 11 - 814. Inada, M. Nishinakamura, K. Kondo, T. Shima, H. & Ogawa, N.1978. On the long-term stability of an embankment of soft cohesive volcanic soil. Clay Fills, ICE London.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Numerical analysis of reinforced embankments on soft soils C.T. Sa, E.M. Palmeira, L.M.A. Dellabianca & A.R.S. Fahel Department of Civil and Environmental Engineering, University of Brasilia, Brasilia, Brazil
ABSTRACT: This paper presents a numerical study on the performance of reinforced piled embankments and reinforced retaining walls on compressible foundation soils. Parameters such as pile spacing, foundation stiffness, wall height and number and stiffness of reinforcement layers were varied. The results obtained allowed the identification of relevant parameters and the importance of a proper modelling of the reinforcement. The behaviour of case-histories in the literature support the findings of the numerical analyses 1 INTRODUCTION Geosynthetic products can be effectively used to reinforce embankments on soft soils. The presence of the reinforcement increases the factor of safety of the embankment, allows a faster rate of construction, steeper embankment slopes and minimises fill consumption and construction time. In most of the cases it is required that the geosynthetic reinforcement presents large tensile strength and tensile stiffness to fulfil its role as an effective reinforcing element. Geosynthetic layers can also be combined to piles with caps at the base of the embankment. In this case the piles transfer the fill loads to a deeper and stronger soil layer beneath the soft deposit, reducing embankment settlements. This type of solution is of interest not only for problems of embankments on soft clay deposits, but also for embankments on collapsible soils. This is the case in some regions in the city of Brasilia, Brazil, where vertical deformations due to soil collapse can reach as much as 15%. The presence of the reinforcement layers can provide a better distribution of vertical load to the piles and to reduce the vertical stresses transmitted to the soft soil, if an effective arching mechanism can be mobilised in the fill material. The design of piled embankments with reinforcement is still a complex task, where many variables involved are still difficult to quantify. This is the case of the required reinforcement strength and tensile stiffness, number of reinforcement layers and spacing between reinforcement layers. Some limit equilibrium methods are available in the literature for the design of piled embankments with and without reinforcement (Terzaghi 1943, John 1987, Hewlett and Randolph 1988, BSI 1995 and Russel and Pierpoint 1997). However, the complex nature of the interaction between materials in this type of problem
265
limits their application. This favours the utilisation of more powerful numerical tools, such as finite difference or finite element methods to predict embankment behaviour. Nevertheless, even the latter methods have their own limitations to capture the actual mechanism of soil reinforcement in this type of situation. This paper presents some numerical studies of geosynthetic reinforced embankments built on compressible foundations. The presence of piles at the base of the embankment and how numerical simulation relates to the observation of the performance of real structures is also approached in the present paper.
2 ANALYSIS OF PILED EMBANKMENTS WITH GEOSYNTHETIC LAYERS 2.1 Characteristics of the cases analysed Figure 1 shows a typical cross section of a piled embankment with reinforcement layers. The aim of the study at this stage is to assess the behaviour of the region around a pile in the central part of the embankment. The computer code FLAC (Itasca 1995) was used in the numerical analyses. The characteristics of the materials used in the simulations are summarised in Table 1. A linear elastic analysis was used at this stage under plane strain conditions. The piles (0.3 m diameter) were modelled as equivalent walls to satisfy plane strain conditions. Interface elements were employed at the pile-soft soil and fillreinforcement interfaces. The adhesion between pile and soft soil was assumed as equal to 10 kPa (zero interface friction angle). The construction of the embankment was staged, similar to the conditions found in real works. It was assumed a value of
1 -Without reinforcement
Figure 1. Geometrical characteristics of the problem.
3 - Pile and cap only 5a - 3 reinforcement layers
Table 1. Material properties used.
(J = 4000kN/m) - beam element 5b - 3 reinforcement layers (J = 4000kN/rn) - cable element
Property Young modulus (MPa) Poisson coefficient Unit weight (kN/m3) Thickness (m) Note: (*) Pile diameter.
Fill 40.0 0.3 19.6 6.0
Soft Soil 2.0 0.3 14.7 6.0
Pile 1.4 x 10'
2 - Pile only 4a - 1 reinforcement layer (J = 300kN/rn) - beam element 4 b 1 reinforcelnent layer (J = 300kN/rn) - cable element
Figure 2. Average vertical displacement on the foundation surface between pile caps.
24 0.3"'
piles and caps reduced the vertical displacement even further (up to 20 to 50%) for lower values of d. As d increases the benefit of the presence of piles and one layer of reinforcement is reduced. For the case of 3 layers of a stiffer reinforcement (J = 4000 kN/m) a significant reduction of the vertical displacement of the foundation surface can be noted, even for larger values of d. Figure 2 also shows that the differences coming from modelling the reinforcement by different types of elements were smaller than 20%. The influence of the number and stiffness of the reinforcement layers on the vertical displacement of the foundation soil surface can be visualised in Figure 3, as a function of the distance between pile caps. It can be observed that the greater the stiffness or the number of reinforcement layers the smaller the vertical displacement for values of d below 2 m (H/3). For larger values of d the stiffness of the reinforcement is less important to the reduction of vertical displacements than the number of reinforcement layers. The reinforcement tensile stiffness and number of reinforcement layers have a marked effect on the average vertical displacement at the embankment surface between pile caps, as shown in Figure 4, for d = 2m. It should be noted that the number of reinforcement layers influences more the vertical displacements than the reinforcement stiffness for values of J above 1000 kN/m. Figure 5 shows the influence of reinforcement stiffness and number of reinforcement layers on the average vertical stress on the surface of the foundation soil. Reductions of vertical stresses from 15 up to 83%, with respect to the situation of the foundation soil alone, can be observed for d < 2 m (d/H<0.33), depending on the number of reinforce-
Young modulus for the layer of soil below the soft soil equal to 60 MPa and Poisson ratio of 0.3. The caps, when present, were assumed rigid with dimensions 1 x 1 x 0.5 m (width x length x height). The spacing between piles was varied to investigate the influence of this factor on the general behaviour of the system. The following values of pile spacing were analysed: 1.8, 2.4, 3.0 and 5.2 m. The spacing values of 1.8 and 2.4 m would yield pile efficiencies of 90 and 75% by the design methodology proposed by Hewlett and Randolph (1988). The number and tensile stiffness of the reinforcement layers were varied as part of the investigation. Problems with up to 3 reinforcement layers above the piles were analysed. The spacing between reinforcement layers was kept constant at 0.25m. The values of reinforcement tensile stiffness (J) used were 300, 1000,4000 and 40000 kN/m. The reinforcement layers can be simulated by FLAC as cable elements or beam elements. Both types of elements were used to assess the influence of the type of element on the results obtained. Additional information of the analyses carried out can be found in Sa (2000). 2.2 Results obtained Figure 2 shows the variation of average vertical displacement of the foundation surface between pile caps as a function of the distance between cap faces (d, in Figure 1). A 15 to 21% reduction of settlements was can be observed for the case of the embankment with piles and caps in comparison to the embankment on the soft soil alone up to distances between caps of 1.5m. The addition of one layer of reinforcement (J = 300 kN/m) to the system with
266
Figure 5. Average vertical stress on the foundation soil surface.
Figure 3 . Influence of reinforcement stiffness and number of reinforcements on foundation surface average settlements.
Figure 6. Influence of the relative stiffness of the piles.
Reinforcement stiffness, J (kN/m)
figure. It can be observed that the increase of the relative stiffness of the pile above a value of 100 had a marginal effect (less than 5 % ) on the reduction ofvertical stresses transferred to the foundation surface. The effect of pile stiffness increase was slightly more relevant when the reinforcement layer was present. Figure 7 presents the variation of the maximum mobilised reinforcement force versus reinforcement stiffness for the case of an embankment with piles, caps and three layers of reinforcement, for d = 2m. The results show that the distribution of tensile force between layers is not uniform, with the reinforcement layer at the bottom being 3 times more loaded than the top reinforcement layer.
Figure 4. Average vertical displacement at the fill surface for different numbers of reinforcement layers (d = 2m).
ment layers and reinforcement stiffness. For d = 4.2 m (d/H = 0.7) and piles with caps with three layers of a 4000 kN/m stiff reinforcement the reduction of vertical stresses is of the order of 33% in comparison to the case of the foundation soil alone. The influence of the ratio between pile modulus and foundation soil modulus on the average vertical stress on the foundation soil surface is presented in Figure 6 for d = 2.0m. The presence of piles and caps only and piles, caps and a layer of reinforcement, with J = 1000 kN/m, are considered in that 267
Figure 8. Reinforced wall on compressible foundation. Table 2. Characteristics of the walls analysed
Figure 7. Forces mobilised in the reinforcement layers.
Wall 1 2 3
3 REINFORCED WALLS ON COMPRESSIBLE FOUNDATIONS The behaviour of reinforced soil structures such as the one presented in Figure 8 was also investigated in the present work. The computer code FLAC was also employed in the numerical simulations. It is important to point out that the reinforced walls were designed using the program Reslope (Leshchinsky 1995). The aim was to have a preliminary deformation analysis of a reinforced wall designed using current limit equilibrium methods and the influence of the foundation soil stiffness on the general behaviour of the structure. The general characteristics of the walls studied are presented in Table 2. Three wall heights were investigated and the reinforcement layout chosen was the one given by the program Reslope for the condition of uniform reinforcement spacing and length (stronger and stiffer reinforced mass). The backfill material was assumed as having a Young modulus of 20 MPa, Poisson coefficient of 0.3 and unit weight of 18 kN/m3. The properties of the foundation soil in each case are presented in Table 3. Three foundation soils with Young modulus values ranging from 10 to 120 MPa were analysed, yielding to different levels of foundation stiffness. Both elastic and elastic-plastic analyses were carried out. Table 4 shows the parameters adopted in the elastic-plastic analyses conducted. The construction of the wall was staged, simulating real construction conditions. The reinforcement tensile stiffness (J) values used in the analyses were 1200, 1800 and 3600 kN/m. Interface elements were used between soil and reinforcement layers. The friction angle at these interfaces was adopted as 31". Different values of the interface stiffness (normal, Kn, and shear, K,) were used. For a rigid interface both the normal and the shear stiffness of the interface were equal to 99000 MPdm. Based on direct shear tests (Tupa 1994), more realistic values of interface stiffness were used 268
H(m) 3,O 6,O 12,o
s,(m) 0,3 0,3 03
n 10 20 40
B (m) 1,7 3,4 63
Notes: (1) H = wall height, s, = reinforcement spacing, n = number of reinforcement layers, B = reinforcement length; (2) See also Figure 8. Table 3. Properties of the foundation soil. Foundation type A B C
Young Modulus (MPa) 120 60 10
Poisson Unit weight (kN/m3) ratio 0.3 22 0.3 19 0.3 17
Table 4. Parameters used for the elastic-plastic analysis. Soil Fill Foundation A Foundation B
C
(KPa) 5 -
S" (KPa) 1200 600
@ (deg.) 35" 0 0
E (MPa) 20 120 60
Notes: c = soil cohesion, S, = soil undrainded strength, @ = soil friction angle and E = soil Young modulus.
in some of the cases. In these cases the values of K, and K, adopted were 9000 and 30 MPdm, respectively. Additional information on these numerical analyses can be found in Dellabianca (1999). 3.1 Results obtained Figure 9 shows the results of horizontal wall face displacements (normalised by the wall height) for different values of foundation stiffness. It can be observed that the softer the foundation soil the larger the wall horizontal displacement, particularly close to the wall toe. This behaviour has been also identified in model tests (Palmeira and Monte 1997) and in real structures, as will be discussed later in this work.
sion, close to the wall face. This can be attributed tothe pattern of horizontal displacements in the foundation soil. A similar behaviour was also observed when the number of reinforcement layers was varied in wall 2 (with varying safety factors andor conditions assumed in the design of this wall with program Reslope). For reinforcement layers at midheight and at the top of the wall the values of foundation stiffness used in this work had little effect on the distribution of tensile forces and maximum tensile force in the reinforcement. For the conditions of the walls there was little difference between predictions using a linear elastic and an elastic-plastic model, as can be seen by the results of normalised wall face horizontal displacements in Figure 12, except for the regions close to the ground surface and wall base.
The distribution of vertical stresses along the base of the reinforced wall 2 on foundation soil B can be seen in Figure 10. The distribution of vertical stresses is rather uniform along most of the length of the wall base. For regions close to the wall face it can be observed the influence of the presence and value of the interface elements stiffness on the results obtained. Figure 11 presents the distribution of tensile loads along the reinforcement length at the base of wall 2 (K, = 9000 M P d m and K, = 30 MPdm) for different values of foundation stiffness. It can be noted that the foundation stiffness can have a major influence on the tensile loads at the base of the structure. For the softer foundation condition the reinforcement was subjected to compression, rather than ten
Figure 11. Distribution of tensile forces in the reinforcement at the wall 2 base. Figure 9. Horizontal displacements of the wall face - Wall 2.
Figure 12. Elastic versus elastic-plastic analysis - Wall 1 and foundation soil B.
269
4 CASE-HISTORIES BEHAVIOUR
5 CONCLUSIONS
To illustrate the link between numerical simulations and real structure performance, some case-histories are presented and their general behaviour evaluated. The reinforced structures to be discussed below are piled geotextile reinforced abutments built on soft soils as part of the crossings of the rivers Itariri, Subauma and Sauipe in the BR- 101 highway, along the Brazilian north-east coastline. Soft organic clay deposits are common in the region, with undrained strengths varying typically between 10 to 60 kPa. Detailed information on these case-histories can be found in Fahel (1998) and in Palmeira and Fahel (2000 and 2001). The Itariri, Subauma and Sauipe abutments heights are 3.8, 1.75 and 2.0m, respectively. The former two abutments were built on piles with caps. The pile diameter was 0.25m and the piles crossed the entire soft soil thickness. The pile caps measured 1 x 1 x 0.3 m and the pile spacing was 1.5m (d = 0.5 m). The Sauipe reinforced abutment was built directly on a 4.5m thick layer of sand overlying the a 5.7m thick soft clay foundation. Rather extensible woven (Subauma wall) and non woven geotextiles (Itariri and Sauipe walls) were used. For the pile spacing and cap geometry used, the settlements of the piled abutments were negligible, which is consistent with the results of the numerical analyses presented earlier. Relative displacements between wall panels and rotation of the wall face were observed for the Sauipe abutment (built without piles) as shown in Figure 13 (a), as well as a significant vertical settlement (29 cm) close to the bridge (Fig. 13 b). The flexibility of the reinforced soil mass was a major component for the reduction of damages to the abutment due to settlements. After rather minor repairs, these embankments have been performing well since.
This paper presented a numerical analysis study of reinforced embankments and walls on compressible foundation soils. The results obtained showed the benefits of the combination of reinforcement layers and piles beneath the embankment. The analysis also compared different ways to model the reinforcement layer and some differences of results can be observed, although of limited relevance. The use of interface elements is very important for an accurate modelling of the reinforcement. Horizontal wall face displacements and reinforcement forces are influenced by the foundation stiffness. For the conditions investigated in this work the use of a conventional limit equilibrium design approach yielded rather stiff walls when uniform reinforcement lengths and spacing were employed, even for the softer foundation soil.
Figure 13. Damages to a reinforced abutment caused by foundation settlements (Fahel 1998).
270
REFERENCES BSI 1995. Code of practice for strengthenedheinforced soils and other fills. British Standards Institution, UK. Dellabianca, L.M.A. 1999. Numerical analysis of geosynthetic reinforced structures. MSc. Thesis, University of Brasilia, DF, Brazil (in Portuguese). Fahel, A.R.S. 1998. Instability and construction problems in geosynthetic reinforced abutments. MSc. Thesis, University of Brasilia, DF, Brazil (in Portuguese). Hewlett, W.J. and Randolph, M.F. 1988. Analysis of piled embankments. Grouiid Eizgiizeering (21)3: 12-18. ITASCA 1995. FLAC 3.30 user’s manuul. Itasca Consulting Group Inc., Mineapolis, Minesota, USA. John, N.W.M. 1987. Geotextiles. Blackie and Sons, NY, USA. Leshchinsky, D. 1995. ReSlope: supplemental notes. Dept. of Civil Eng., University of Delaware, Newark, USA. Palmeira, E.M. and Fahel, A.R.S. Effects of large differential settlements on embankments on soft soils. EuroGeo 2000, Bolonha, Italy, Vol. 1 : 261 -266. Palmeira E.M. and Fahel, A.R.S. 2001. Lessons learned from failures of wall facing units in two geotextile reinforced walls. Book Lessons Learned from Fuilures, Editor: Jean Pierre Giroud, IFAI, USA (in press). Palmeira, E.M. and Monte, L.M. 1997. The behaviour of model reinforced walls on soft soils. Geosynthetics ’97, Long Beach, CA, USA, Vol. 1: 73-84. Russel, D. and Pierpoint, N. 1997. An assessment of design methods for piled embankments. Gi-ourzd Engineering (30)lO: 39-44. Sa, C.T. 2000. Numerical analysis of piled geosynthetic reinforced embankments on soft soils. MSc Thesis, University of Brasilia, DF, Brazil (in Portuguese) Terzaghi, K. 1943. Theoretical soil mechunics. Wiley and Sons, New York, USA. Tupa, N. 1994. A study on bond strength and soilreinforcement interaction. MSc. Thesis, University of Brasilia, DF, Brazil (in Portuguese).}
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swefs & Zeitlinger, ISBN 90 2651 863 3
Trial construction of the reinforced river dike and its performance K. Sawada, A. Yashima, Y. Sat0 & Y. Fujita Gifu University, Gifu, Japan H. Maeda Maeda Kosen Co., Ltd., Japan
N. Matsumoto Mitsui Chemicals Industrial Products, Ltd., Japan
A. Hazama Mitsubishi Chemical Functional Products Inc., Japan ABSTRACT: As a new application of the reinforced soil, a method called reinforced river dike is proposed. In order to establish a design and construction method for the proposed reinforced river dike, a real scale field construction and field monitoring were conducted at Shin-Sakai River in Gifu Prefecture, Japan. Many in situ measurements, which include the pore water pressure within the dike, the water level of the river, the deformation of the dike, and the flow speed of the river, were carried out to clarify the deformation mechanism of reinforced river dike. In the test embankment, some of the filling materials were drawn out in a large scale. Based on the observed data, the reason for the soil draw-out and the corresponding countermeasure are discussed in detail, During a heavy rain from Sept. I I to 12, 2000, the test river dike with the reinforced soil was immersed with flood. Although the soil draw-out of the filling materials and the considerable deformation of the dike took place, it did not fail in whole. From this experience, the stability of the proposed reinforced river dike is confirmed. river, the deformation of the dike, and the flow speed of the river, were also carried out to clarify the deformation mechanism of the reinforced river dike. The observed data are explained in detail along with the introduction of the structures of the reinforced river dike. In the test site, various kinds of filling materials and protect net against the soil draw-out were used. Some of the filling materials were drawn out in a large scale. Based on the observed data, the reason for the soil draw-out and the corresponding countermeasure are discussed in detail. A heavy rain that broke the local record of history occurred from September 11 to 12, 2000. During the rain, the test river dike with the reinforced soil was immersed with flood. Many measuring facilities flowed away and the top surface of the dike was severely eroded. Although the soil draw-out of the filling materials and the considerable deformation of the dike took place, it did not fail in whole. Based on the neighboring observed data including precipitation data and measurement of river water level changes, the damage process of the dike and its mechanism are explained in detail.
1 INTRODUCTION As a new application of the reinforced soil, a river dike embankment method with geo-grid reinforcement is proposed. There are some uncertainties in the design concept of the reinforced soils used in the river dike. Such factors as the limitation of the length of geo-grid reinforcement in rive dikes, the quality of filling material, the design the river-water flow speed, the resistance of the front unit against floating logs and boulders, and the plantation on the dike surface for preventing soil draw-out, should be investigated. The design of river dike is actually based on the sophisticated design concept, which provides insight into the failure mechanisms of the dike. The main possible failure mechanisms include: 1) overflowing, 2) piping, 3) sliding of dike slopes, 4) erosion of dike slopes, 5 ) soil draw-out from the dike, 6) squeezing, 7) seepage, etc. The main purpose of this research is to establish a design and construction method for the newly proposed reinforced river dike by conducting a real scale field construction and field observation at Shin-Sakai River in Gifu Prefecture, Japan. The test dike has a length of 19.2 m and a height of 2 m with a steep gradient of 2: 1. Within the 19.2 m length of the dike, the length of geo-grid, the back-filling materials, the surface-processing materials, and the material of dike wall were changed to investigate their characteristics against the aforementioned problems. Many in situ measurements, which include the pore water pressure within the dike, the water level of the
2 RESEARCH METHOD A research group organized by Maeda Kosen Co. Ltd., Mitsui Chemicals Industrial Products. Ltd., Mitsubishi Chemical Functional Products Inc. Co. Ltd. and Gifu University is working on reinforced
27 3
Figure I . Location of the test river dike.
river dike. (Maeda(2000), Sawada et a1.(2001)) In the research project, a real-scale test on the river dike with geo-grid reinforcement was conducted in Shin-Sakai River in Gifu Prefecture, Japan, in order to establish a reliable design and construction method for reinforced river dike, considering thereservation of natural environment. The trial river dike locates at the right bank of Shin-Sakai River with a gradient of 2:l. The location of the test dike is shown in Figure 1. The length of the test bank is 19.2 m and the height of the bank, from the normal water level to the top of the bank, is 2.0 m. It is divided into three sections. Sections A, B and C, are divided according to the different materials of geogrid, the length and the pitch of the grid. Figure 2 shows the overall view of the tested river. The river dike is embanked with 1.1 m in depth in Sections A and B and 1.5 m in Section C. The lengths of geo-grid are arranged in the same as the embanked depths in Sections A, B and C. The back-filled materials of the dike are sand above water lever and gravel beneath the water level. The tensile strength of the geo-grid is about 35 kPa. There is a metallic cover grid on the dike slope, taking a Lshaped structure whose gradient is the same as the 272
dike, that is, 2:1, as shown in Figure 3. Figure 3 shows the metallic cover, the geo-grid, the soil draw-out-proof material and the back-filling materials. In order to investigate the effectiveness of the coating used for the metallic cover, two kinds ofcoating, that is, electronic coating and PE (Polyethylene) coating are used. Because the main attention is paid to the deformation of the dike, various measuring instruments are installed to measure the deformation of the surface of the dike and the failure mechanism of the reinforced soil dike. Figure 4 shows the outlet of the measuring instruments and the dike composition.
After May, no remarkable deformation was observed. The deformation is considered to be the compression of the dike itself. As will be discussed later, the self-compression of the dike was finished around in September when the dike was fully immersed in water because of the flood caused by the heavy rain. The dike itself, however, did not fail anymore.
4 RESULT OF ELECTRO OPTICAL DISTANCE MESUREMENT AND ITS ANALYSIS The deformation of the reinforced river dike was also measured with electro optical distance technique on April 6, May 18, July 4 and November 6, 2000. The deformations of reinforced soils and concrete-retaining sheets were measured by using barshaped targets inserted in the dike. From the measured horizontal displacement of Section A, it is recognized that the dike tends to move towards the river. The reason is that the newly constructed dike is not stable and easy to be affected by the water flow of the river and the underground water within the dike. Furthermore, settlement was also observed in Section A. The horizontal displacement and the settlement in Sections B and C are not so large as compared to those in Section A. Similar results are also obtained from the measurements with usual measuring instruments.
3 MEASURING RESULTS AND DISCUSSION The field measurements started on February 26, 2000. The measuring items are listed in Figure 4, which include the pore-water pressures in Sections A and B, water levels of the river at Sections A and C. The measurement stopped on September 12,2000 due to a heavy rain during which the measuring instruments flowed away. From the observed data, which include the horizontal displacement of the dike, the pore-water pressures within the dike and water levels of the river and ground, it is found that those data are co-related to the change of the temperature to some extent. Therefore it is necessary to remove the influence of the temperature in processing the data. By adopting a filter which can remove the one-day-cycle change in temperature, the highfrequency variation existing in the observing data can be clearly removed, as shown in Figure 5. From the observation, it is found that the deformation mainly occurred during the first three months, that is, from February to May.
5 SOIL DRAW-OUT FROM THE DIKE Since the beginning of the measurement in April, the water level experienced several big changes that resulted in a maximum value of several 10 cm of
Figure 4. Position of measuring machine and outline of facility.
273
Figure 5 . Results of various measurements of the river dike.
draw-out of filling materials from the slope surface of the dike. In order to investigate the situation of the draw-out, a frame with 10x10 cm grid was laid on the dike surface and the depth of the draw-out at every cross was measured by inserting a bar perpendicularly to the frame. It is found that serious drawout occurred at the downstream side of concrete retaining sheets. The maximum values of the draw-out at Sections A, B and C are 55 cm, 45 cm and 65 cm, respectively. Figure 6 shows the distribution of the draw-out in each section. The reason why such a phenomenon occurred is that the resistances of the concrete retaining sheets and the dike surface are 274
quite different and consequently a turbulent flow happened at the downstream side of the sheet, resulting in the draw-out. Especially in Section A where a larger draw-out occurred as compared to the other two sections, the pipe was installed for setting up the measuring instrument at the concrete retaining sheet, which results in a large turbulent flow. In other wards, if any extruded part of a structure exists, a turbulent flow will happen behind the extruded part, resulting in the draw-out of the materials. Meanwhile, at the boundaries between different cover materials of the surface, a relatively larger draw-out was also observed. The reason is thought
Figure 6. Soil draw-out areas at each section and depth of cavity.
to be that the boundaries are easy to be eroded by the flowing water and become the origin of the drawout. During the observation, it is found that the drawout at the second and third layers was most serious. The explanation of the phenomenon can be given clearly, that is, when the water level of the river increases due to flood, water flowed deeply into the reinforced soils through these two layers boundary. After the flood went away, the water within the soils ff owed out and took away the soils.
From the erosion of the top of the dike due to the flood, it is understood that in designing a reinforced soil dike, the erosion caused by flood should be considered. In present study, concrete retaining sheets are installed in the dike with each 6 m interval. Therefore, even if the retaining ability and endurance of the dike are confirmed during the flood caused by the heavy rain, it is still needed to confirm whether the retaining ability and endurance of a dike is still kept after a serious flood if the concrete retaining sheets are not installed.
6 DAMAGE DUE TO HEAVY RAIN
7 PLAN OF REBUILDING NEW DIKE
A serious flood due to a heavy rain happened from September 9 to 12, 2000. Because of the flood, the test site was also overflowed, the data logger flowed away and the top of the reinforced soil dike was eroded. Figure 7 shows the eroded top surface of the river dike at Sections A, B and C. The draw-out on the slope surface of the dike wall was also observed. The dike itself, however, was not destroyed and the retaining ability and endurance of the dike are remained in work. Figure 8 shows the situation of the dike one day after the flood. It is known from the figure that the water level was still very high, only 1 m below the top of the dike although the highest flood had passed one day before.
Based on the damage of the first trial river dike, the construction of a geo-grid reinforced river dike is newly planned. In the planning of rebuilding, different checking items are proposed for each sections, for instance, the soil draw-out will be checked in Section A; the total safety of the embankment will be checked in Section B; the retaining effect of the surface coating will be checked in Section C. Of course, in the present case, suitable instruments will be installed on different sections according to their measuring items, based on which careful discussion will be done to find the weak points of the reinforced river dike which acts as a retaining structure of a river. Furthermore, as aforementioned, concrete
275
failure test, a dead weight will be loaded on the top of the dike to simulate a possible external force that acts on the reinforced river dike. It is expected that by these field tests, it will be easy for us to design and construct a reinforced soil dike in the future.
8 CONCLUSIONS In this study, a geo-grid reinforced river dike was proposed. In order to investigate the real behavior of the proposed river dike, a real scale field test was conducted. From this study, the following conclusions are derived. From the observation and various kind of measurement, it is found that the deformation mainly occurred during the first three months that is from February to May 2000. Because of the change in the water level of the river during site investigation term, it is found that serious soil draw-out occurred at the downstream side of concrete retaining sheets. A serious flood due to a heavy rain happened from September 9 to 12, 2000. Because of the flood, the test site was overflowed and the top of the reinforced soil dike was eroded. The dike itself, however, was not destroyed and the retaining ability and endurance of the dike are remained in work. In constructing a new river dike, a failure test will be conducted on the dike to investigate the mechanical behavior of the dike. From the results of failure tests, it will be easy for us to design and construct a reinforced soil dike in the future. ACKNOWLEDGEMENT Gifu Prefectural government is deeply appreciated for providing the research group with the trial site at Sin-Sakai River.
Figure 8. Situation of the dike one day after the flood
retaining sheets will not installed at the boundaries between different sections this time because it will enhance the strength of a dike. It will be installed only at the beginning and ending sides of the test area. In constructing a new river dike, the old one will be destroyed and removed. Therefore, it is a good chance to investigate the intact state of the old dike. A failure test will be conducted on the dike to investigate the mechanical behavior of the dike. In the
REFERENCES Maeda, H. (2000), The river revetment where it is built by the reinforced soil structure, JCZGS.Vol.16,No. 1. Sawada, K., Yashima, A., Sato, Y., Fujita, Y., Maeda,. H., Matsumoto, N.and Hazama, A.(2001), Soil draw-out from the reinforced river dike and its behavior during floods, Proc. of the 36"' Jripun Natiorzul Cotlference on Geotechnicul Eng iiieeritig , in printing .
276
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Analysis of reinforced slopes and walls using Horizontal Slice Method M. Shahgholi & A. Fakher Department of Engineering, University of Tehran, Iran
C.J.F.P. Jones Department of Civil Engineering, University of Newcastle, UK
ABSTRACT: In this paper a new limit equilibrium method of analysis is presented. The method is identified as the Horizontal Slice Method (H.S.M.). In this method horizontal slices are used in place of vertical slices to analyze the stability of reinforced and unreinforced slopes and walls against gravity and earthquake loads. Comparative analyses using the Horizontal Slice Method and an established computer program show good agreement, and the method can be shown to produce advantages over existing limit equilibrium methods. been adapted to any type of failure surface and soil. Unfortunately the number of unknown parameters with the vertical slice method is greater than the number of equations, and accordingly it is necessary to make simplifying assumptions to reduce the number of unknowns. A number of authors have presented vertical slice methods of analysis. The procedures differ principally in the equilibrium requirements which they satisfy and the manner in which they handle interslice forces which are normally dealt with in terms of vertical and horizontal components (Sharma, 199l), Table 1. In addition to conventional analysis, limit equilibrium methods can be used for the pseudo-static analysis of slopes against seismic loads and for the and a dynamic earth pressure component is added to the static earth pressure forces to determine the required reinforcement force. In the analysis of the stability of reinforced soil slopes the tension forces in the reinforcing elements need to be considered. Due to the method of construction and the usual orientation of the reinforcement, these forces are usu-
1 BACKGROUND A number of methods are available to analyse the stability of slopes. The majority of these may be categorized as limit equilibrium methods (Fang 1991). In practice, limit equilibrium methods have advantages over other methods of analysis, including: i. the methods are simple; ii. the results derived are reliable; iii. the material properties required for the analyses are limited and can be easily obtained. Limit equilibrium methods can be divided into two main groups. The first group considers the equilibrium of the whole failing mass, assuming a failure surface. These methods are suitable for the analysis of homogeneous soils and specific failure surfaces. In the second group, a sliding wedge or "active" mass is divided into a number of vertical slices and the equilibrium of each individual slice considered. This procedure, known as the method of slices, has
Table 1. Characteristics and assumption of some vertical slice methods of analysis. Equilibrium conditions CM=O CM=O (overall) (individual) No Fellenius (1936) Yes Method
CFx=O (individual) No
CFy=O (individual) No
Shape of failure surface
Assumptions
Circular
Resultant of side forces is parallel to base of each slice Vertical side forces neglected Location of side force resultant on sides of slices can be varied A pattern of variation of side force inclination from slice to slice is assumed Side forces of all slices are
Bishop (1955)
Yes
No
Yes
No
Circular
Janbu (1954)
Yes
Yes
Yes
Yes
Any
Morgenstern & Price
Yes
Yes
Yes
Yes
Any
Spencer
Yes
Yes
Yes
Yes
Any
277
Figure 1. Forces acting on a single horizontal slice cotaining reinforcement
ally assumed to act horizontally. The limiting force developed in any reinforcing element, (t,), is the lesser of the rupture strength of the reinforcement or the pull out resistance, Figure 1. It can be seen from Figure 1 that the orientation of the reinforcement has a direct influence on the interslice forces and that the reinforcement tensions are added unknowns in the vertical slices method analysis. As a result the vertical slice method is not particularly suited to the analysis of reinforced soil slopes.
that no interslice forces are generated by the reinforcements. The following assumptions are made:
2 HORIZONTAL SLICE METHOD OF
Thus, if the failure wedge is divided into N horizontal slices, there are 4N unknowns which can be determined by 4N equations, and a complete formulation is possible, as detailed in Table 2. The complete formulation can be simplified if only vertical equilibrium is considered for individual slices together with overall horizontal equilibrium for the whole wedge, no account being taken of moment equilibrium. In this case the number of equations and unknowns are reduced to 2N+l. Table 3.
i.
The vertical stress on an element in the soil mass is equal to the overburden pressure. ii. The factor of safety (F.S.) is equal to the ratio of the available shear resistance to the required shear resistance along the failure surface. iii. The factor of safety for all slices is equal. iv. The failure surface can have any arbitrary shape but it does not pass below the toe of the slope or wall.
ANALYSIS The limitations of the vertical slice method for the analysis of reinforced soil can be resolved by the use of horizontal slices, identified as the Horizontal Slice Method (HSM). In this method, a failure surface is assumed and the failure wedge divided into a number of horizontal slices. The forces that act on each slice are shown in Figure 1. From Figure 1 it can be seen
Table 2. Equations and unknowns of complete forinulation of Horizontal Slice Method of analysis. Equations CF,=O (for each slice) CF,O (for each slice) CM=O (for each slice) =z /(for each slice)
F.S. Sum
Number N N N
Unknowns Horizontal intcrslice force Normal forces upon base of each slice Shear forces upon base of each slice Location of normal forces
Number N- 1 N N N
Factor of Safety Sum
1 4N
N 4N
Table 3. Equations and unknowns of simplified formulation of Horizontal Slice Method of analysis. Equations CF,=O (for whole wedge) CFsO (for each slice) =z/(for
each slice)
F.S. Sum
Number 1 N
Unknowns Normal forccs upon base of each slice Shear forces upon base of each slice Factor of Safety
Number N N
Sum
2N+ 1
1
N 2N+ 1
278
3 EVALUATION OF RESULTS
Therefore, from Figure 1:
C Fy = 0 (for each slice) 3
In order to evaluate the Horizontal Slice Method the analysis of a typical reinforced soil wall was undertaken and compared with the results produced by an established analytical computer program ReSlope, (Leshchinsky 1997, Ling et al. 1997). Details of the wall are given in Table 4. The value of Ctj max determined using the ReSlope program can be compared with the values of Ctj max determined using the Horizontal Slice Method for different values of Kh and cp, Table 5.
Vi+l -Vi - ( l + K v ) W i + S i s i n a i + N i c o s a i = O
z
Z,
= f (for each slice) 3
F.S. 1 S. = -(cb,+ N , tan @) ' F.S.
Fx = 0 (for whole wedge) 3
Table 4. Details of reinforced soil wall.
where Vi = vertical interslice force; Hi = horizontal interslice force; K, = vertical seismic coefficient; Kh = horizontal seismic coefficient; Wi = weight of slice; Ni = normal force upon base of slice; Si = shear force upon base of slice; tj = tensile force of reinforcement; zr = required shear stress; zf = failure shear stress; c = cohesion of soil; cpi = angle of friction of fill; F.S. = factor of safety; m = number of reinforcement layers; N = number of slices; bi = length of base of slice; cli = angle of base of slice.As a result Si can be derived as a function of the F.S. using Equation (2). Si is derived from Equation (2) and substituted into Equation (1). Ni is derived as a function of the F.S. as follows:
y -yil +(l+K,,)W,--sinaI Cbl F.S. tan @ sina, + c o s q F.S.
NI =
Height Unit weight of fill (y) Cohesion of fill (c) Friction angle of fill (cp) Horizontal seismic coefficient (Kh)
i=l
i=l
(4)
Having determined Si and N, the value of F.S. can be determined using Equation (3) when Ct, is known and vice versa. It should be noted that vertical interslice forces ( V , and V1+l) could be calculated by integration of overburden pressures on horizontal borders. As an example, for a wall with horizontal soil surface, V, is equal to (1 +K,)yh,l, where y is unit weight of soil, h, is depth of slice and 1, is length of the slice.
= 5m = 18kN/m3
=O = varies 20"-45" = varies 0-0.3
4 CONCLUSION The Horizontal Slice Method overcomes the inherent difficulties in adopting the vertical slice method of analysis for the design of reinforced soil structures, in particular: i. There are no interslice forces developed by the action of the reinforcement. ii. Different seismic accelerations at different heights of the soil structures can be modelled. The results of a trial analysis of a reinforced soil structure subjected to seismic forces agree closely with the results produced using a log spiral assumption of a failure plane. REFERENCES Fang, H-Y & Mikroudis, G.K. 1991. Stability of earth slopes. Foundation engineering handbook (2nd Edition), H-Y Fang (ed.), New York: Van Nostrand Reinhold: 379-409. Leshchinsky, D. 1997. Reslope. Ceotechnical ,fabric report 15(1 ): 40-46. Ling, H.I., Leshchinsky, D. & Perry, E.B. 1997. Seismic design and performance of geosynthetic-reinforced soil structures, Geotechizique 47(5): 933-952. Sharma, H.D. 1991. Embankment dams. New Dehli: Oxford and IBH Publishing COPvt Ltd :359.
Table 5. Comparision between Ct, ,nilx (kN) calculated using ReSlope and Horizontal Slice Method of analysis. cp
Kh
0 0.05 0.10 0.15 0.20 0.25 0.30
20 ReSlope HSM 110 110 119 119 128 128 137 139 151 151 167 167 187 187
25 ReSlope HSM 95 91 99 99 110 107 119 116 126 127 137 139 153 153
30 ReSlope HSM 74 75 81 82 90 89 99 97 106 106 117 117 128 128
279
35 ReSlope HSM 63 61 68 67 74 74 81 81 90 89 99 98 106 108
40 ReSlope HSM 50 49 54 54 59 60 65 67 74 74 81 82 88 90
45 ReSlope HSM 38 39 43 43 47 49 54 55 63 61 70 68 74 75
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Geogrid-reinforced road embankment over an old dump G. Stolarski Project supervising engineer, LGA Geotechnical Institute, Nuremberg, Germany
E. Gartung Chief geotechnical engineer, LGA Geotechnical Institute, Nuremberg, Germany
ABSTRACT: A new road designed in the German State of Thuringia crosses an old unconsolidated waste dump area which had been in operation as an uncontrolled landfill before 1950. Analyses of the waste material showed a low content of organic components in the fill body. However, due to some chemical contamination of the fill, there was concern about the expenses which would have arisen for proper handling and deposition of the waste material and for the decontamination of the site, if a partial replacement of the waste with inert soil would have been executed. The favoured solution was, to keep the waste in place and design the road embankment accordingly. For this purpose, the compressibility of the fill was determined by two large scale trial loading tests with geogrid-reinforced test embankments. The test results supported the design of the embankment which is under construction at the present time. The geogrid reinforcement was essential in reducing the total settlements and in particular in smoothening the differential settlements.
In Germany difficult soil conditions are frequently encountered in the construction of new roads. Near urban areas space is limited, and often it is not possible to select alternative road alignments to avoid foundation problems, which arise when unfavourable ground conditions are encountered locally. In such situations more elaborate engineering solutions are called for, which may benefit from the application of geosynthetics.
When the excavations for construction clay were undertaken back in 1850 to 1880, the explored open pits had been protected against floods by a dam along the river bank which partly still exists, but which is now hidden by backfill. Furthermore, historical records tell that a small creek existed in the area, but nobody knows exactly where it was. So the ground below the foundation of the embankment is very heterogeneous and its mechanical properties can hardly be predicted with the desirable degree of confidence.
2 LOCATION OF THE SITE
3 SITE INVESTIGATIONS
At Saalfeld, a small town in the State of Thuringia, Germany, a two lane road had to be constructed. Parallel to an existing railway embankment which is a little higher than the new road embankment, the alignment crosses the flat valley of the river Saale. In the past, the valley floor had been partially excavated for the recovery of clay, used in ancient construction. The excavated pits had then been refilled with all kinds of materials. The road alignment connects two bridges and must fit with the junctions on both sides of the valley. So it was not possible to realign it and optimize the location of the new road embankment with regard to the ground conditions. The new road had to traverse the fills. So a technical solution had to be found to cope with the unfavorable foundation situation over a length of 170 meters.
The waste analysis comprised historical research, sampling and chemical testing. It lead to the conclusion that the waste-backfll is very heterogeneous. The oldest parts of the fill consist of domestic waste from the nineteenth century. Later all kinds of debris and residues were dumped, especially ashes and rubble from buildings which were destroyed during the second world war (1939 to 1945). The waste dumping operations ceased in 1950. Then the entire area was covered with a layer of sand and gravel of variable thickness up to 3 meters. It was subsequently used as a training range for trucks. Due to these activities the waste fill and the predominantly cohesionless cover soils were compacted rather erratically.
1 GENERAL
3.1 Waste analysis
28 1
3.2 Geological conditions The schematic soil profile below the road embankment can be described as follows: From the ground surface to 1,0 to 1,5 meters nonuniforly compacted cohesionless sand and gravel fill - Heterogeneous, predominantly inorganic waste material until 4 to 6 meters below ground surface - Soft, grey alluvial clay, thickness 0.5 to 1.5 meters - Gravel layer of 3 meters - Sandstone bedrock below the gravel layer
-
The water content of the deposited waste material varies with its composition and the access conditions of precipitation and surface runoff. Some leachate was encountered at the base of the fill above the impervious clay layer. A continuous ground water table was located in the lower gravel layer. There is a second ground water table at greater depth inside the sandstone. The waste deposit appears to be well sealed at the base by the underlying clay layer. The investigations for the new road focussed on the local situation and did not include an assessment of the environmental hazard of the entire dump area which extends far beyond. As for the construction of the road, it was decided by the authorities in charge, that the waste should remain in place, because the natural clay at the base acts as an efficient seal. 3.3 Solutions.for the road foundation According to the design, the surface of the new road will be 3.5 m above the current ground level. The existing fill is considered an incompetent ground for the foundation of the embankment. Without special treatment the anticipated settlements would be on the order of 20 cm to 40 cm with probably quite substantial settlement differences over short distances. Several alternatives of foundation concepts were evaluated under the aspects of tolerable settlements and differential settlements, environmental issues and costs. One option with respect to small, well controllable vertical displacements seemed to be a geogrid reinforced embankment on piles. However, besides high expenses, the disadvantage of the deep foundation would be, that the piles would perforate the natural clay barrier below the waste body in order to reach the bearing stratum below. This was considered unacceptable, because it might lead to a contamination of the ground water. As indicated already, the option to excavate the waste, move and deposit it at an engineered landfill site and replace it with inert soil material, would have caused high costs, additional environmental and technical problems and thus was considered inappropriate. The most promising concept was to
place the embankment with additional surcharge well before installation of the road surface in order to preload the ground and thereby reach a high degree of consolidation of the waste material and the underlying soft clay. A reduction of total settlements and elimination of intolerable settlement differences was to be achieved by an adequate soil reinforcement. Since the mechanical behaviour of the waste material could not be predicted on the basis of past experience nor could it be determined by laboratory experiments, it was decided to execute large scale loading tests at the site. 4 TEST ARRANGEMENT
4.1 Trial embankments
To study the behaviour of the ground under the designed road embankment of 3.5 meters height and to evaluate the effect of an additional surcharge, two test embankments were erected. Embankment No. 1 was approximately square in plan, 6 meters high, 5 by 5 meters at the crown and about 25 by 25 meters at the base. Embankment No. 2 was 3 meters high, 4 by 4 meters at the crown and 18 by 18 meters at the base. The average unit weight of the compacted well graded sand fill was 1.72 Urn3. The distance between the test embankments was sufficient to avoid interaction. According to the ground exploration it was anticipated that test embankment No. 1 was placed where the thickness of the compressible layers was largest. At the base, test embankment No. 1 was reinforced by a single sheet of high tenacity polyester geogrid with a laboratory-tested short term uniaxial tensile strength of 150 kN/m in both directions. The reinforcement of test embankment No. 2 consisted of the same type of geogrid, but with a tensile strength of 150 kN/m in the main direction and 30 kN/m in the perpendicular direction. At the base of test embankment No. 2 two layers of geogrid were installed perpendicularly to each other. 4.2 Instrumentation The area for the trial embankments was carefully selected and surveyed. A system of survey points for the measurement of elevations was installed around the trial embankments and on the adjacent railway embankment (Fig. 1). Horizontal inclinometers were selected to serve as the main measurement instruments for the determination of deformations. They were placed at the base of the embankments below the geogrid layers (Fig. 2). The inclinometer pipes were installed parallel to the main axis of the road and perpendicular to it. One additional inclinometer measures the settlements along an axis parallel to the road axis under the slope of the embankment No. 1.
282
Figure 2. Instrumentation tools: I - Horizontal inclinometers, E - Earth pressure cells, D - Strain gauges.
- the varying depth of the former clay pit which determines the thickness of the fill - the varying thickness of the soft clay below - positioning of the road embankment at the periphery of the landfill near the buried slope of the ancient clay pit - ancient facilities such as a trench with a sever pipe placed decades ago in the waste body with inadequate backfill.
Earth-pressure cells were installed for the measurement of the vertical contact pressures between the embankment and the ground. The geogrid sheets were equipped with strain gauges to determine the strains and deduce reinforcement forces from these data. Measurement of the ambient temperature and temperature below the embankment completes the monitoring system. The layout of the instruments is presented in Fig. 2. 4.3 Boundary conditions and eflects on settlement
5 RESULTS
The heterogeneity of the waste body is probably the governing factor with regard to the expected differential settlements. However, the following boundary conditions may also have a significant influence on the development of the deformations under load:
Readings were taken at all instruments for 600 days at varying time intervals. During the erection of the embankments monitoring was done currently, then at daily, weekly and finally monthly intervals. 283
5.1 Settlements The plot of inclinometer data obtained below embankment No. 1 in the direction perpendicular to the road axis shows a very smooth development of the settlement profile (Fig. 3). There are no indications of erratic settlement differences due to irregularities in the compression of the waste layer under load. Evidently, the geogrid reinforcement in combination with the compacted well graded sand fill serves its purpose successfully. All monitored settlement profiles support this observation, although some differences were noticed in the shape of the settlement trough. The maximum settlement below the higher embankment No.1 amounts to 27.5 cm. Below embankment No. 2 the maximum settlement is only 8 cm. The difference between both test embankments results form the different heights and the different loads associated with them. It is also influenced by the difference in the thickness of the waste deposit below the trial embankments (5 m thick under embankment No. 1, only 3.5 m thick under embankment No. 2). Furthermore, there may have been different local moisture contents inside the waste. Below embankment No. 1 the former creek is suspected which may partially still be an active flume, this may have played a role in the compression of the fill. During construction of the embankment about 30% to 40% of the total settlement took place. 10 weeks after completion approximately 80% of the total settlement had occurred. The development of settlements with time is plotted on Fig.4. It can be seen, that the settlement due to the placement of the road embankment presently under way, follows the pattern of the test embankment and that the magnitude of the observed deformations falls in between the lines for both test embankments. It was interesting to notice that the rate of settlement of the smaller trial embankment was initially higher than the rate of settlement of the larger embankment .
The settlements of the test embankments are nearly equal in both measured main axis directions with some minor deviations only due to the reasons mentioned in chapter 4.3. The additional inclinometer under the trial embankment No. 1 determined a differential settlement perpendicular to the road axis within a range of 6 cm at 6 m distance (l%), which would have to be considered significant for the asphalt pavement, if it would occur after completion of the road. But knowing the rate of settlements and the time when time-dependent settlements cease, this can be taken into account in the construction sequence. The calculated elongation of the geogrid amounts to 0.05 mm, which is a small acceptable value with respect to the performance of the reinforcement . The influence of the embankments on the adjacent area remained only marginal. At the toe of the large embankment No. 1 maximum settlements of 17 mm to 54 mm were measured, while at a distance of 5 m from the toe of the embankments no settlements nor any heave could be detected. The placement of the trial embankment had no influence on the nearby railway embankment. This observation was important for the preparation of the construction procedures. No significant creep deformations were observed.
5.2 Vertical soil stresses at the base of the test embankments The earth-pressure cells placed horizontally at the base of the embankments indicated the following maximum contact stresses at the level of the former ground surface: embankment No. 1 o = 57 kN/m2; embankment No.2 o = 47 kN/m2. It is interesting to note, that the maximum values measured do not correspond with the calculated peak values below the centre points of the embankments (expected were 105 kN/m2 and 52 kN/m2 respectively) Since the stresses are distributed quite uniformly in the area of measurement, it is proposed in the interpretation of the measured data, that there is an arching effect within the embankments, specially pronounced in the larger one and that the geogrid reinforcement facilitates this arching with the associated stress distribution at the base of the embankment.
5.3 Strain of geogrids The measured strains are small. For example, below embankment No. 2 for the uniaxial geogrid (150/30 kN/m tensile strength) the measured strains are varying locally between 0.2%0to 2%0.The average strain of the 12.5 m long geogrid deduced from the measured settlements was approximately 0.2%0. At the perimeter of the geogrid sheet values of 0.4%0were measured. The ultimate strain of the geogrid at tensile failure in a laboratory test was 13%.
Figure 3. Settlement below test embankment axis I 1.1
284
5.4 Temperature
has been under construction. The monitoring program using 6 horizontal inclinometers reveals that at the structure behaves as predicted. (Fig. 4).
The variations of the temperature registered during the test period remained moderate and had no influence on the measured data.
6 CONCLUSION AND CLOSING REMARK By large scale testing with two trial embankments on an old waste deposit it was possible to provide reliable information about the compressibility of a heterogeneous fill. The test results were used as the basis for the design of the geogrid-reinforced road embankment. It was thus possible to avoid removal of the waste. The geogrid reinforcement played a key role in the structural performance of the embankment by eliminating intolerable settlement differences and reducing total settlements. It also facilitated a favourable rather uniform stress distribution at the base of the embankment. Currently the road
Figure 4. Settkment Versus time.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Loading test of earth flow prevention embankment reinforced with geosynthetics N. Tatta, Y. Yokota, S. Ito & T. Kubo Maeda Kosen Co., Ltd., Japan
K. Arai Fukui University, Japan
ABSTRACT: Traditionally stiff structures like concrete retaining wall have been employed for preventing earth flow due to slope failure, avalanche of earth and rocks, and so on. Flexible structures like embankment reinforced with geosynthetics have the following advantages. I ) Soils excavated on the site are effectively used for the structural material. 2) The structures can be applied at a narrow space and a steep slope ground. 3) The structures are well matched to natural environment. 4) It is easy to repair and maintain the structures. At the present stage, however, the displacements and failure strength of the reinforced embankment structures have not been clarified yet when shock loads or dynamic loads are applied. This paper reports the result of field model test of reinforced embankment against quasi-static lateral loads.
1 INTRODUCTION
2 OUTLINE OF EXPERIMENT
The experiment reported herein was such that a semi-static load was made to act on a geosyntheticsreinforced embankment structure to take a field observation and to verify the effectiveness of the reinforced embankment utilized as a protection retaining wall. Two types of model test are performed. One is a small type, the height of which is 2m. The other is a full-scale model test, the height of which is 6m. Some of model tests investigate the effect of prestress applied to embankment. The pre-stress is employed for preventing vertical separation of embankment. In the model test, displacement and earth pressure of embankment, stresses on geosynthetics, and so on are monitored. An outline of the experiments and test results involved are reported hereunder.
The geosynthetics which had the highest tensile strength of 35kN/m (strain 5 % ) was used for the experiment. For wall surface material, an L expanded metal (h = 0.25m, 0.5m) was applied as slope unit. (See Figure 1.) The embankment material applied was clayey sand. Table1 shows the material properties. With the natural ground taken for reaction force, a linear semi-static load was applied to an embankment on one side by means of a hydraulic jack. The horizontal force applied to the embankment was measured with a load cell while the horizontal displacement was determined with a displacement gauge located at the loading point of embankment.
2.1 Experimental methods
Figure 1. Slope unit.
287
structure is the same as the structure in model 3 except the size (see Figure 4). The geosyntheticsreinforced embankment was prestressed with a compressive prestress to set up a reinforced embankment.
Table 1 . Properties, embankment materials. Item Cohesion c' Angle of shear resistance Unit weight Water content Density of soil particle Maximum dry density Optimum water content
4'
Property Ow/mL 37.27" 15.87kN/m' 20.0% 2.662gl m3 1.654glcm' 19.1%
3 TESTESULTS 3.1 Model experiment
An earth pressure gauge was installed to grasp a vertical earth pressure and a horizontal earth pressure. Measurements, moreover, were taken, with a strain gauge attached to grasp the tensile force that would act upon geosynthetics and on a prestressed reinforcement. 2.2 Experiment for three models Model 1: Extrusion test without geosynthetics. Model 2: Extrusion test with geosynthetics (see Figure 2). Model 3: Bending test with geosynthetics and prestress (see Figure 3). By using the expanded metal, these model structures were constructed and thoroughly compacted at each layer. In model 3, geosynthetics was prestressed with a compressive force to check the effect of pre-stressing. The monitoring points are shown in Figure 2 and 3.
1) Extrusion test (model 1 and 2) Figure 5 shows the results of an extrusion experiment on extrusion horizontal stress and horizontal displacement. Though slightly, a difference could be observed. As shown in Figure 6, the strain gauge attached to geosynthetics showed a tendency to shrink until the extrusion load had reached the maximum displacement. The larger the displacement, however, the more the material tended to elongate. According to Figure 7, the earth pressure gauge read a rise of vertical earth pressure presumably because geosynthetics restrained a deformation of the embankment. Under the influence of such restraint, the vertical earth pressure could be deemed to rise. Figures 8 and 9 show the situations upon completion of the test. Model 1, which did not have geosynthetics, lay out in the interior, caused a significant crack to appear at the upper part of the embankment, with the slope unit falling down in front. 2) Bending test (model 3) After the embankment had been constructed, it was prestressed at 30kN. Figure 10 and 11 show the changes in earth pressure and in strain of geosynthetics.
2.3 Experiment for actual size structure The experiment for actual size structure was performed for proves the effect of retaining structure reinforced with geosynthetics and pre-stress. The basic
Figure 2. Shape of models 1 and 2 .
288
289
Figure 1 1. Prestressing load and strain of geosynthetics.
From Figure 12, it may be gathered that the extrusion stress began to slowly drop from a peak of 15kN/m2. When the displacement reached 430 millimeters, the embankment as a whole toppled over in the form of one solid. (Refer to Figure 13.) As shown in Figure 14, the prestressed reinforcement strained by 0.87% in front and by 0.67% in rear, thereby generating a bending stress. The geosynthetics in the interior of the embankment had a shrinkage take place as shown in Figure 15, especially shrinking more conspicuously at the center than elsewhere.
290
3.2 Actual size experiment An actual size embankment was constructed and prestressed over a range of 35kN to 50kN as divided in four cycles. (Refer to Figures 16.) As shown in Figure 18, the pressurizing time decreased stage by stage while the load increased stage by stage. Figure 19 shows the strains that appeared on geosynthetics in the first cycle of prestressing Reinforcement No.”. From the Figure, it may be gathered that prestressing caused geosynthetics to generate a strain. Nevertheless, strains were found concentrated on geosynthetics in the upper stages (2-4 and 2-5). This tells us that prestressing does effectively act on the embankment at the upper part only. The results of extrusion test are as follows. As gathered from Figure 20, the horizontal extrusion stress was maximized when the horizontal displacement reached approximately 500 millimeters. When the horizontal displacement reached 1,600 millimeters, the embankment toppled over, with approximately a quarter of the embankment height acting as fulcrum. (Refer to Figure 17). Figure 21 shows a curve of extrusion displacement vs. strain of No.3 Rod. The prestressed reinforcement had a strain take place at the upper part and shrinkage at the lower part. The earth pressure gauge did not read any rise of either horizontal or vertical earth pressure. For rupture, the reinforcement did not break but was found to have bent at a height of 1.5 meters.
Figure 15. Horizontal displacement and strain of geosynthetics.
Figure 17. Topple-over situations.
29 1
Figure 2 1. Horizontal displacement vs. strain of pre-stressed reinforcement.
4 CONCLUSION
A series of the model experiments reported hereinabove have resulted in the findings that the embankment reinforced with geosynthetics raises the shear resistance as compared with an unreinforced embankment. It could be verified, moreover, that an addition of prestressing would improve the effective integration. In the actual size experiment, it was impossible to spread the effectiveness of prestressing all over the embankment because of an insufficient load andor of a small reaction plate area. As a solution to the problem, therefore, prestressing was carried out step by step while constructing the embankment so that the prestressed stress could be made to remain. From the results referred to above, it may be gathered that there is a possibility of utilizing the geosynthetics-applied protection retaining wall satisfactorily in the future. Figure 20. Horizontal displacement vs. horizontal extrusion stress.
292
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Deformation analysis of PLPS GRS bridge pier during construction and in service T. Uchimura, M. Shinoda, M.S.A. Siddiquee & F. Tatsuoka University of Tokyo, Tokyo, Japan
ABSTRACT: Preloaded and prestressed (PLPS) reinforced soil method aims at substantially increasing the stiffness and decreasing the residual settlement of reinforced soil structures. The first prototype PLPS reinforced soil pier was constructed for a railway bridge in 1996. As a part of a research program to develop a methodology for predicting the time-dependent behaviour of such reinforced soil structures, one of threecomponent rheology models, named “New Isotach Model”, is used to analyze the observed time-dependent (viscous) load-deformation behaviour of the PLPS pier during its preloading procedures. and 1,280 kN, respectively. Before constructing the pier, an about 9 m-thick very soft clay layer was improved by constructing a set of 0.8 m in-diameter cement-mixed soil columns in-situ. The whole crosssection of a 1 m-thick surface clay layer immediately below the pier was improved by cementmixing method to form a reaction layer for applying vertical compressive load to the backfill. The lower ends of four steel tie rods with a nominal yield tensile strength of 1,034 kN were anchored into the cement-mixed soil columns for a length of 4 m. A
1 INTRODUCTION Preloaded and prestressed (PLPS) reinforced soil method aims at substantially increasing its stiffness and decreasing the residual settlement of reinforced soil structures by applying preload and prestress in the vertical direction[ I]. Soil structures constructed by this method can be used as important permanent structures to support heavy load without harmful transient and residual deformation. Because soil structures are more flexible than R C structures, they may not be damaged by foundation deformation if it is not excessive. Therefore, a pile foundation, which is required to support RC structures, could become unnecessary. Besides, the construction of reinforced soil structure could be much more cost effective if inexpensive backfill soil is available on site. The first prototype PLPS geogrid-reinforced soil (GRS) bridge pier was constructed to support a pair of temporary railway girders, each 16.5 m in length, in the summer of 1996 in Fukuoka City, Japan (Figure 1) [2]. This pier has been open to service for more than three years since the summer of 1997. One of three-component rheology models, named “New Isotach Model”, is used to analyze the observed load-deformation behaviour of the PLPS pier during its preloading procedures.
2 CONSTRUCTION OF THE PLPS PIER 2.1 Construction of the pier backfill The PLPS geogrid-reinforced soil bridge pier is 6.4 m x 4.4 m in cross-section, and 2.7 m in height. The design dead load by the girder weight and the design live load by train including impact load are 196
Figure 1 . PLPS reinforced soil pier for a railway bridge.
293
well-graded gravel of crushed sandstone (D,,, = 30 mm; D ~ =o 0.9 mm; U, = 16.5; and @ =60° at 0'3 = 50 kPa by triaxial compression tests) was used for the backfill. The backfill was constructed with a help of gravel-filled bags, stacked along the periphery of each gravel layer and wrapped-around with reinforcement. The geogrid reinforcement used was polyvinyl alcohol coated with polyvinyl chloride (PVC). The nominal rupture strength is 73.5 kN/m and the nominal stiffness is 1,050 kN/m at tensile strains less than 1 %. In each of the horizontal two orthogonal axes of the pier, the reinforcement layers were arranged in the same way as usual geosynthetic-reinforced soil retaining walls with a fullheight rigid facing having the same height as the pier, constructed under plane strain conditions. The vertical spacing of reinforcement layers was 30 cm. As each cross-section, having one pair of wall faces, of the pier was designed independently, by overlapping the two cross-sections, the actual average vertical spacing of reinforcement layers became 15 cm. 2.2 Preloading procedures
k vertica'i pre'ioaa 01Z,4WuUdq,equiviient to an average vertical pressure of 200 kPa, was applied to the backfill through the top reaction block by using four hydraulic jacks installed at the top of the tie rods. On the first day, the preload was applied stepwise. During the process of increasing by 200 kN, the load was kept constant for 30 or 60 minutes at each step to allow creep deformation to occur (A in Figure 2). Then constant preload of 2400 kN was applied only during daytime for 2 weeks (B in Figure 2). The backfill was compressed by 8 mm during preloading. Then the load was released to 970 kN (C in Figure 2), and the top ends of the tie rods were fixed to the top RC block to maintain the compressive stress (i.e. prestress) in the backfill. Finally, full-height rigid facings were cast-in-placearound the backfill. Since then, nearly constant prestress has worked on the backfill during service (D in Figure 2).
3 ANALYSIS OF BEHAVIOUR DURING PRELOADING
3.1 New isotach model The long-term deformation of PLPS reinforced soil structures is affected by time-dependent properties of the backfill material and the reinforcement under static and dynamic loading conditions. If the prestress decreases due to creep deformation of the backfill and/or the reinforcement, the stiffness decreases and the transient deformation by traffic load may increase, resulting in large residual deformation. Then, the prestress decreases further, resulting into a phenomenon of vicious spiral. Therefore, it is essential to keep high prestress in the backfill during the life time of structure. As the first step of the study on this topic, it was attempted to analyze the time-dependent deformation of the PLPS reinforced soil pier during preloading (A in Figure 2) based on the new isotach model proposed [ 3 ] . The new isotach model is a kind of three component rheology model (Figure 3). According to this model, the total strain increment dE consists of elastic strain increment dE" in the component (A) and irreversible strain increment dE" in the components (B) and (C); that is,
+ ddr (1) For the component (A), which is non-linear elastic, the following stress dependency is assumed: d& = dEe
where d o is the stress increment; d$ is the elastic strain increment; 00 is the reference stress; E(o) is the Young's modulus as a function of o;E0 is equal to E(o0);and m is a constant. This equation is based on experimental studies (Tatsuoka et. a1.[4]). The total stress o consists of the time-independent stress d in component (B) and time-dependent stress U'' in component (C); that is, 0-=
Figure 2. Compression of the pier by preloading.
d+ d'
Figure 3. New Isotach model.
294
(3)
where 6 is a function of irreversible strain E" independent of its rate if'if the sign of i" does not change. If the sign of &" changes; i.e. in case of unloading or cyclic loading, the function 6= 6(E" ) changes dependin on the loading history. The details of function ( E " ) is unknown at this moment and should be determined based on the results from relevant material tests. c",which controls the timedependent aspect of the stress-strain behaviour, is a function of E" and &" . Here, the following function is assumed after Tatsuoka et. a1.[3].
J
ov( E" ,€" ) =of(E" ) *gv( ) €Ir
where a ,m and i;rf are the material constants. That is, the compoents (B) and (C) are not independent from each other, as Eq. (4) contains d(E" ). The function gl,(i" ) is a Inonotonically increasing function, which becomes 0 when &" = 0 and a when it' = +W. Therefore, we obtain nearly = d(E" ) under extremely slow loading (2"is nearly zero), CT increases as &" increases, and nearly (T- (1 + a )d(E" ) at infinitively high 2'. As the stress-strain relationships are always located between the curves of L T = ~ ( E "1 and o= ( I + a ) 6(&I),'they are called the lower and upper bounds, respectively. According to the new isotach model, the creep behaviour of the materials are explained as follows (Figure 4a). Suppose that the current irreversible strain is E" under a given constant stress c o e e p . Then the timeindependent stress component is given as d = 6(E" ), and the time-dependent stress component is obtained
, = IS,,.^, - J(E"). For the known value as ( T " ( E ' ~i".) of gcreep and the known initial value of E " , the initial can be obtained. During the creep stage, value of iir as E" increases, &"- decreases according to the decreases in d'.After an infinite time period, the stressstrain state reaches the lower bound and stays there.
3.2 Estimating the parameters for elastic component The parameters in Eq. (2), 00, E0 and m, were estimated as follows, based on the results from a triaxial test on a specimen of the backfill gravel used in the PLPS pier (Figure 5a). The specimen had a rectangular prismatic shape with dimensions of 30cm x 30cm x 60cmH. Both the axial and the lateral strains were measured accurately by using LDTs. The material was very densely compacted (yt = 2.2 W/m3 with a water content of 3.0 %). This density is very likely similar to that of the pier backfill. In order to evaluate the elastic Young's modulus, cyclic axial stresses with a small amplitude (10 kPa) were applied at several stress states during monotonic loading at a constant confining pressure of 49 kPa. Figure. 5b shows the relationships between the obtained Young's modulus and the axial stress in a log-log plot. By linear fitting, the following values were obtained: E0 = 610 MPa for = 100 kPa, and m = 0.63. Using these values, the elastic strain increment dce was estimated by Eq. (2), and then the irreversible strain dc"' was estimated as the difference between total and elastic strains increments (Figure 5a).
Figure 5. Triaxial test result on the backfill soil of the PUPS pier:
Figure 4. a) Creep deformation according to the new isotach model. b) Assumption with respect to the lower bound for creep analysis.
ti ~ 295
,
~
~
~
~
~
~
~
~
s
3.3 Estimating the parameters for time-dependent stress component
&Lf
The parameters of gv( kir) in Eq. (4), a , m and were estimated as follows, also using the results from the triaxial test shown in Figure 5. During monotonic loading, the deviator stress, q = 200 kPa, was kept constant for 1 hour. Figure 6 shows the detailed behaviours. The irreversible strain rate gfr was obtained by numerically differentiating E ' ~with time (Figure 6b and c). Figure 6d show the plot of gir against E". In the model, the relationship between iirand E" is expressed as follows, which is equivalent to the Eq. (4).
ocreep where x = g ,( k L r=) -1
of( E i r )
By fitting Eq. (5) to the plot in Figure 6d, the parameters can be determined. However, the function of(&")in Eq. ( 5 ) is unknown. Therefore, it was assumed that ' ( & " ) is approximately linear for a limited range of strain for which E" changes at each creep loading stage. Then x in Eq. (5) can be expressed as follows (Figure 4b):
is the irreversible strain when the stresswhere ,E;/ strain state is on the lower bound, and A is the slope of the lower bound that is linearly fitted for a stress rate around the current stress level. This assumption is acceptable because, for usual geomaterials, the stress-strain relationships before failure is smooth without sudden changes in its tangent modulus. By fitting Eqs. (5) and (6) to the data plotted in Figure 6d, the values of the five parameters, a , m , , /z and E;,>, , were obtained. The values of a , m and E,,; are considered to be specific to the mate&,I;,
rial, while /1 and E;[, change at each creep stage. It was shown that several different combinations of parameters gives similarly good fitting to the same plot (Table 1). The relations using the parameter combi-
Figure 6. Details of triaxial creep test on the backfill soil. a) Deviator stress vs. time relation. b) Total and irreversible strain vs. time relation. c) Rate of irreversible strain vs. time relation. d) Irreversible strain vs. its rate relation.
nations with, for example, a =0.09004 and a =I0 are also shown in Figure 6d. It is possible that there exist a set of parameter values which give similarly good fitting for each values of a . One of the reasons is that the range of E" encountered in the data shown in Figure 6d is much more narrow than the full range dealt with by Eqs. ( 5 ) and (6) The other reason may be that there is a large redundancy among the five parameters. It is also possible that other models are more relevant to be present case. Dan et al. (2001)[5] used the so-called general TESRA model to simulate triaxial compression test reults of a dense well-graded gravel. This point should be studied more. Herein, the combination with a =I5 was used because for a reason mentioned later.
Table I . Obtained parameters of time-dependent irreversible component
a
rn
.ir Eref
0.09004 0.5 1 5 I0 15
0.0137 0.00589 0.00414 0.00194 0.00 15 0.00 13
0.000224188 0.000090750 0.000072450 0.000029659 0.000018 I06 0.000015034
296
226.934 5 18.0 726.5 1623.6 235 1.8 2898.8
0.097 16 0.09861 0.09889 0.10022 0.101 18 0.10168
3.4 Estimating the parametersfor time-independent stress component (lower bound) The lower bound function for the PLPS pier during preloading was estimated as follows. The elastic and irreversible deformations of the pier during the first day of preloading were estimated by using the parameters obtained above as shown in Figure 7. The deformation is expressed in the strain averaged for a backfill height of 2.4 m, and the load is expressed in the average stress estimated at the mid-height of the backfill (cross-sectional area is 22.04 m2). In the lower part of Figure 7, the relationships between the irreversible straind' and its rate 2" , obtained in the same way as shown in Figure 6d, are shown. Eqs. (5) and (6) are fitted reasonably well to the relationship between E" and &" for the respective creep stage. Here, fixed parameters a =15, m =0.0013 and g;;, =0.000015034, from Table 1, were used, while /z and were changed to optimize the fitting. As mentioned before, E;! I presents the ultimate value of E" when the stress-strain state reaches the lower bound after an infinitively long period of creep. These ultimate points are denoted by the symbol "x" in Figure 7. The following lower bound function was obtained by fitting to these ultimate points.
of( I " ) = (34.84 + 644.1 E" )/( 1+2.241E" )
(7)
Figure 8 shows the ultimate points at the respective creep stage obtained by using the same method for different values of parameter a , 0.09004, 10 and 15. The arrows represent the instantaneous tangent slope of the lower bound, 1,obtained at each point. With a = 15, the arrows are directed to the ultimate point at the immediately preceding creep stage, validating the value of a = 15. With a = 10, nearly the same results are obtained. On the other hand, wiht a = 0.09004, the directions of the arrows are nearly horizontal, showing that the value of a = 0.09004 is not appropriate. The method described above may be one of the methods to choose the most appropriate parameters from possible values as listed in Table 1. Following the preloading in the first day, the tie rods were fixed to the top reaction block by using nuts for a period of one night with some compressive stress remaining in the backfill (A in Figure 8). For this period, the backfill expanded in the vertical direction due to creep recovery. According to the new isotach model, creep recovery can take place only when the stress-strain state is below the lower bound. Thus the lower bound should be located above A in Figure 8. This may be another information to determine the parameters.
Figure 8. Lower bounds obtained by using different parameters.
3.5 Simulation of the behaivour during preloading Using the parameters obtained above, the behaviour of the pier backfill during preloading was simulated (Figure 9). The recorded tme history of the average stress was given to simulate the irreversible strain. The amount of the simulated creep deformation during each creep stage agrees very well with the respective measured value. However, the global irreversible strain is over-estimated. The stress-strain relationships when the load was increasing rapidly are noticeably different between the measurement and the simulation. It is likely that this difference is due at least partly to that the full detailed measured relationships could not be obtained due to a too long sampling interval (2 minutes). Especially in estimating the parameters for the time-independent stress component (the lower bound), the data used for curve fitting for each creep loading stage are too few in number and too much fluctuating (Figure 7). This may have caused a large error in the estimated lower bound. Figure 10 shows the simulation obtained by using a modified lower bound (the value of sf( E " ) has been multiplied by 1.2) with the same values for
297
pier backfill was proposed. However, this simulation method is not sufficient in the following important factors, which should be studied in the next step. The new isotach model should be fully validated based on results from a comprehensive series of accurate tests of the backfill and reinforcement materials. The model is one-dimensional in the sense that only the relationship between one stress component (vertical stress) and the one strain component (vertical strain) is simulated, ignoring the horizontal stress and strain relationship. The latter is essential to evaluate the time-dependent behaviour of the reinforcement placed in the horizontal direction and the interaction between the backfill and the reinforcement. The simulation by the new isotach model of the observed time-dependent behaviour during unloading, which is essential for the proposed PLPS construction method, should be attempted. A model that can simulate and predict the transient and residual deformation of the backfill by traffic load, which applies a great number of small cyclic load to the backfill for a very long duration, should be developed.
REFERENCES
Figure 10. Lower bounds obtained by using different parameters.
the other model parameters as the case of Figure 9. The simulation and measured values agree with each other to a better than Figure 9, but such a procedure of determining parameters as above is logically inconsistent. Further study on this method based on more accurate data from element tests on the materials are necessary.
4 CONCLUSIONS A method of numerical analysis based on the new isotach model of the time-dependent behaviour of the
298
Tatsuoka,F., Uchimura,T., and Tateyama, M. 1997. Preloaded and prestressed reinforced soil, Soils und Fouizdutions, 37(3): 79-94 Uchimura,T., TatsuokqF., Tateyama,M., Koga,T. 1998. Preloaded-Prestressed Geogrid-reinforced Soil Bridge Pier, Proc. 6th lnt. Con$ on Geosynthetics, Atlanta, V01.2: 565572. Tatsuoka,F., Uchimura,T., Hayano,K., Di Benedetto,H., Koseki,J. and Siddiquee,M.S.A. 1999. Time-dependent deformation characteristics of stiff geomaterials in engineering practice, the Theme Lecture, Proc. of the Second Internationcil Conference on Pre-failure Defonnution Cliaracteristrcs of Geoinaterials, Torino, Bulkenza (Janziolkowski et cd., eds.), Vol. 2 (to appear). Tatsuoka,F., Jardine,R.J., Lo Presti,D., Di Benedetto,H. and Kodaka,T. 1999. Characterising the Pre-Failure Deformation Properties of Geomaterials, Theme Lecture for the Plenary Session No.1, XIVIC on SMFE, Hunzburg, Vol. 4: 2 1 29-2 164. AnhDan, LQ., Koseki,J. and Tatsuoka,F. 2001. Viscous deformation in triaxtial compression of a dense well graded gravel and its model simulation, submitted to Summary Book of TC29, ISSMGE.
Landmarks in Earfh Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
A diagrammatic evaluation of geo-composites for reinforcing cohesive soils Kazuya Yasuhara & Satoshi Murakami Department of Urban & Civil Engineering, Iburaki University, Hituchi, Ibaraki, 316-8511, Japan
Chandan Ghosh Graduate Student, Graduate School of Science & Engineering, Ibaraki University, Hitachi, Iburaki, 3168511, Japun
Juan Recio Molina Graduate Student, Graduate School of Science & Engineering, Ibaraki University, Hituchi, Ibaraki, 3168511, Japan ABSTRACT: Two diagrams are proposed to ensure the advantageous features of geo-composite on improvement of stability, stiffness and permeability of cohesive soils. The one is to plot the increase ratio in bearing capacity against the one in ground reaction coefficient, which was obtained from the results of model footing tests in the laboratory. The other is to correlate the change in permeability with the one in transmissibity in geo-composite before and after model footing tests. It is emphasized from two diagrams that placement of thin sand layers above and under geocomposites enables to maintain permeability and transmissibity of geo-composite as well as gives rise to marked improvement of stability of cohesive soil embankment.
1 INTRODUCTION Among the recently developed geo-composites, a geo-composite with a non-woven fabric combined with a vvoven fabric in-between has been used for reinforcing marginal or cohesive soils (Hirao et al., 1992 ; Hirao et al., 1996 : Tanabashi et al., 1998 ; 2000 : Yasuhara et al., 1999). This kind of geocomposite is characterized by having the potentials not only of high tensile and frictional resistance but also of vertical and horizontal drainage potential. Those are very beneficial in reinforcing the marginal soils which are not suitable for constituting ground or embankment with no improvement. In order to confirm the serviceability of the geocomposite for reinforcement of embankment made of Kanto loam which is a typical volcanic-origin silty soil in Japan, a family of small-scaled model tests was carried out. The current paper describes the results from these model tests with interpretation for possible application to fields. The two diagrams are proposed from the test results to ensure the advantageous features of geo-composite on improvement of stability, stiffness and permeability of cohesive soils. The one is to plot the increase ratio in bearing capacity against the one in ground reaction coefficient, which can be obtained from the results of model footing tests in the laboratory. The other is to correlate the change in permeability with the one in transmissibity in geocomposite before and after model footing tests. It is emphasized from two diagrams that placement of thin sand layers above and under geo-composites enables to maintain drainage potential defined by 299
permeability and transmissibity of geocomposite as well as gives rise to marked improvement of stability of cohesive soil embankment. 2 OUTLINE AND SCHEME OF EXPERIMENTS Key sketches for laboratory small-scaled model tests on the Kanto loam embankment reinforced with two or three planar geosynthetics (unwoven geosynthetic (UG) and geo-composite (GC)) layers are illustrated in Fig. 1. Kanto loam used for all the model tests is a volcanic-origin silty soil with 2.69 kN/m3 for specific density of solid particle, 94% for liquid limit (LL) and 29 for plasticity index (Ip). In test series for Case A, embankment was prepared by tamping compaction of this Kanto loam with 40 to 50% as initial water content. On the other hand, embankment and level ground for Case B were made by consolidating slurry of Kanto loam with twice of LL as initial water content in order to make sure of the drainage effect for geosynthetic more clearly. Water content of Kanto loam in these tests was about 72% on the average. In one case among the test series for both Cases A and B, sand mat was placed above and beneath geosynthetics to increase the effects of reinforcement and to avoid clogging due to intrusion of finer into the texture of geo-composite. Model embankments (Case A : 21.8 cm base, 11 cm top width, 18 cm high, 10 cm thick and 1V : 0.6H slope, Case B : 45.8 crn base, 23.1 cm top width, 37.8 cm high, 10 cm thick and 1V : 0.6H slope) encased in acrylic box was prepared with help of a spacer block. For comparison with those effects in embankment, as shown in Fig. 1,
model tests on level ground (Series B-2) were also included in the test series for Case B as B-2. The sample for embankment for Case A series was left for saturation for about one week, followed by pre-compression up to 9.8 E a . As is seen in Fig. 2, most of pre-compression occurred within a day. Since the spacer block was present, less settlement took place. However, presence of a nonwoven geosynthetic and a geo-composite combined without sand mat on embankment made lesser pre-compression settlement than that for unreinforced embankment. This may be mainly attributed to the reinforcement effect. On the contrary, presence of sand mat along with geo-composite has led to comparatively larger settlements. This may be due to the drainage effect and/or faster dissipation of excess pore pressures resulting in consolidation. Exact nature of such phenomena cannot be ascertained from the present test results, because the embankment was
left for free drainage after one week for saturation. After the pre-compression was completed, the spacer disk was removed and the top face of the embankment was loaded with a 5cm strip footing under the constant rate of displacement. The top layer of reinforcement was placed at 9.45cm depth for Case B series of tests. As there was not so much precompression settlement, which was mainly due to the presence of the spacer block as shown in Fig. 2.
Figure Ic. Model tests (Series B-2).
Figure la. Model tests (Series A).
300
ratio for bearing capacity and stiffness can be defined from Fig. 4: Bearing capacity index : R, = qur/qun
Figure 2. Time vs pre-Compression settlement of embankment with spacer block.
Stiffness improvement index : Rk = Kir/K,n (2) where qur and qun are bearing capacities with and without reinforcement which correspond to S/B equal to 0.3. Although this value of 0.3 is arbitrarily selected for the sake of convenience, this procedure is used consistently through the present study for determining bearing capacity in any case where the determining bearing capacity in any case where the peak value of q is not observed in load versus settlement curves as is shown in Fig. 5 through Fig. 7. Ki, and Kin in Eq. (2) are ground reaction force coefficients with and without reinforcement. The results determined from S/B versus 2qlyB curves (Fig. 5 to Fig. 7) in model tests for two cases on embankment with and without reinforcement are plotted in the form of the relation between Kir/Kin and qur/qun in Fig. 8 for Cases A and B-1, and Fig. 9 for Case B-2. It is indicated from both figures that: Geocomposite contributes to larger improvement for both bearing capacity and stiffness of embankment than unwoven geosynthetic does. Placement of sand mats above and beneath geocomposite increases both bearing capacity and stiffness. However, less increase is observed in the case of adopting the unwoven geosynthetic. This advantage in geo-composite may be attributed to the fact that the pull-out resistance force of geo-composite against sand is much higher than that against Kanto loam as is shown in Fig. 10. In addition, regarding this benefit in geo-compoiste, it can be also envisaged that the bending stiffness of the geo-composite is larger than that of unwoven geosynthetic. This effect should work out more markedly due to combination with sand.
Figure 3. Simple laboratory tests for permeability.
This tendency must be helpful for understanding the results of loaded tests after precompression which will be later described. After each model test is terminated, then a piece with 10 cm width and 25 cm length, was cut out from the geo-composite which was placed among modeled embankment and then was used for relatively simplified testing equipment for investigating the changes in permeability and transmissibity of the material. The testing apparatus is shown in Fig. 3. 3 EVALUATION OF ADVANTAGEOUS FEATURES IN GEOCOMPOSITE FROM MODEL TESTS 3.1 Evaluation of hearing capacity and stiflness Fig. 4 schematically illustrates the vertical load versus settlement relation which is obtained from model footing tests from simulated level ground and/or embankment at laboratory with and without reinforcement using geosynthetics. Note that the vertical axis for settlement and the horizontal axis for load are given by normalizing S and q by B and (1/2)yB, respectively, where B is width of loading plate and y is unit volume weight of soil used. The following two parameters which designate the improvement
(1)
Figure 4. Key sketch for load versus settlement curve.
30 1
Figure 5. Effect of sand mat on reinforcement.
Figure 7. Effect of geo-composite on reinforcement to embankment and level ground.
Figure 8. Bearing capacity ratio versus stiffness ratio relation (A,B-I).
Figure 6. Effect of geosynthetic on load-settlement curve.
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consolidation in slurry Kanto loam. This tendency should be interpreted together with the results in pre-compression characteristics as is shown in Fig. 2. In other words, this implies that acceleration of consolidation must be a key for increasing reinforcement effects in the case of marginal soils with high water content. 4) In comparison with both cases of level ground and embankment included in Case B-2, reinforcement effect in stiffness using geo-composite is more marked than that of embankment using unwoven geosynthetic. This effect is more clear in level ground than that in embankment. This advantageous feature of geo-composite is probably caused by the higher bending stiffness than that of unwoven geosynthetic. 3.2 Evaluation of drainage potential It is expected that geo-composite is endowed not only by reinforcement but also by drainage effects designated by permeability and transmissibity. Permeability and transmissibity are indices for drainage potentials through vertical and horizontal directions against the geo-composite, respectively. Potentials in those permeability and transmissibity of geocomposite are required to be retained during the life of embankment or ground. However, since geocomposite is normally sandwitched in the horizontal direction by soils in embankment, these potentials are deteriorated from time to time, mainly due to clogging which is caused by intrusion of finer into textures of geocomposite. As an attempt to avoid this inconvenience, sand mats are placed at both sides above and beneath the geocomposite as was shown in Fig. 1. For characteristic evaluation of clogging effect, the horizontal permeability coefficient as well as the vertical permeability coefficient before and after model tests were investigated using the simple devise shown in Fig. 3. The results from both investigation are demonstrated in Fig. 11 which is plotted in the coordinate by the ratios of both coefficients designated by k&n and f3,/en,respectively. Fig. 11 points out that:
Figure 9. Bearing capacity ratio versus stiffness ratio relation (B-2).
1 ) clogging of the geo-composite without thin sand mat (indicated by 1, 2 and 3 in the figure) is significant while the same with it (indicated by 4, 5 and 6) is insignificant. 2) the in-plane permeability of geocomposite is reduced more markedly than the cross-plane permeability, which means that clogging should affect transmissibity of geo-composite.
Figure 10. Results from pull-out tests.
3) On the other hand, as well as for compacted embankment, geo-composite plays a role in marked improvement, particularly in stiffness of Kanto loam embankment prepared by pre-consolidating from slurry condition. However, placement of sand mat produces an increase in bearing capacity than in stiffness. The difference of the sand mat effect on improvement observed between Fig. 2 and Fig. 9 may be ascribed to insufficient pre-
The difference of the above-stated clogging with and without sand mats is recognized in Photo 1 which evidences that a part of geo-composite gets stained after model tests are finished, in comparison with that before model tests. In other words, this is no more than the existence of finer of Kanto loam being intruded into geo-composite. Such a diagram as in 303
understanding the advantage of geo-composite in permeability potential and for exploring the way how to maintain the drainage potential of geocomposite. 4 CONCLUSION
1) Two diagrams are proposed to ensure the advantageous features of geocomposite on improvement of stability, stiffness and permeability of cohesive soils based on the results from laboratory model tests for embankment and level ground. 2) It is emphasized from two diagrams that placement of thin sand layers above and under geocomposites enables to maintain permeability and transmissibity of geo-composite as well as gives rise to marked improvement of stability of cohesive soil embankment. 3) In order that the use of geo-composite is successful in increasing stability and saving costs, the couple effect of geo-composite reinforced finegrained soils needs elaborate investigation and numerical analysis. Besides, more specific experimental techniques are required to obtain the interaction parameters between soils and geocomposite.
Figure 11. Clogging effect of exhumed geocomposite.
(a) Staining condition of geo-composite with and without sand mat
REFERENCES Hirao, K., Yasuhara, K., Takaoka, K., Nishimura, J. and Tanabashi, Y. 1992. Laboratory model tests on the application of composite fabrics to soft ground, Proc. IS-KYUSHU92, Vol. 1, 601-606. Hirao, K., Yasuhara, K. and Tanabashi, Y. 1996. Effect of bending stiffness of geotextiles on bearing capacity iniprovement of soft clay, Proc. ISKYUSHU96, Vol. 1, 59 1596. Hirao, K., Yasuhara, K., Tanabashi, Y., Ochiai, H. and Yasufuku, N. 1997. Bearing capacity improvement of soft clay reinforced with geogrids, J. JSCE, No. 582/111-41,35-45 (in Japanese). Tanabshi, Y., Yasuhara, K., Hirao, K.,' Kiyokawa, N. and Itoh, H. 1998. Improvement of bearing capacity of soft clay using geogrids, Proc. 6th Interizatioizul Cot$ on Geosynthetics, Vol. 1, 895 - 890. Yasuhara, K., Hirai, T., Tanabashi, Y. and Hirao, K. 1999. Geosynthetics requirements for reinforcing soft cohesive soils, Geosyntherics Engineering Jourml, Japan Chapter of IGS, Vol. 14,277-283 (in Japanese).
(b) Staining condition of geo-composite in each layer of embankment without sand mat Figure 12. Difference of stain in geo-composite.
Fig* l1 that plots both the ratios Of cross-plane permeability and in-plane transmissibity of geecomposite before and after model tests is helpful for
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Design of geosynthetic-reinforced veneer slopes J.G. Zornberg University of Colorado at Boulder, USA
S. Somasundaram Advanced Earth Sciences, USA
L. LaFountain New Cure, Inc., USA
ABSTRACT: This paper provides a framework for the design of steep reinforced veneer slopes such as soil covers in landfill facilities. Instead of using geosynthetic reinforcements along the veneer slope, an approach that becomes unsuitable for high slopes, the proposed framework analyzes the use of horizontally placed inclusions within the veneer slope. Analytical expressions are derived, which are useful for preliminary parametric evaluations and final design. A reinforced veneer was designed using the approach proposed herein for the cover of steep slopes at the Operating Industries, Inc. (011) Landfill. Design criteria include requirements that the final cover should control percolation, resist erosion, and it should remain stable for both static and seismic conditions. The cover selected and constructed includes a 1.8-m thick layer of selected soil that, in order to satisfy stability criteria, was reinforced using horizontally placed geogrids. The geogrid reinforcements were embedded into the underlying solid waste mass in order to provide adequate pullout resistance. Construction of the reinforced veneer at this site was completed recently, and involved stripping the in place cover soil, screening the soil for reuse, and placing an engineered evapotranspirative cover reinforced with geogrid layers. California. The final cover system of this hazardous waste landfill is an evapotranspirative cover. The design criteria for the North Slope were that the cover should control percolation, resist erosion, have a static factor of safety of at least 1.5, and lead to calculated deformations in the event of the design earthquake of less than 150 mm. Compliance with the stability criteria was achieved by including geogrids at 1.5-m vertical spacing within the cover and by embedding the geogrids into exposed refuse. The reinforced veneer was constructed on slopes as steep as 1.5 horizontal to 1.0 vertical to reinforce landfill slope sections as high as 65 m.
1 INTRODUCTION The design of veneer slopes (e.g. steep cover systems for waste containment facilities) poses significant challenges to designers. The use of uniaxial reinforcements placed along the slope (under the veneer and above a typically strong mass of soil or solid waste) and anchored on the top of the slope has been a common design approach. However, this alternative may not be feasible for steep, long veneer slopes. If the veneer slope rests on top of a comparatively stronger mass of soil, rock, or solid waste, an innovative, alternative approach consists of using uniaxial reinforcements placed horizontally (rather than along the slope) and anchored into the underlying mass. A framework for analysis and design of veneers reinforced using horizontally placed inclusions is presented in this paper. The framework is useful not only for final design, but also for parametric evaluations that may be performed to define the reinforcement requirements for this system. This approach is particularly suitable for reinforcement of landfill soil cover systems, constructed on top of comparatively strong solid waste, and for stabilization of protective soil layers (colluvium) on top of more resistant bedrock or residual soils. A reinforced veneer approach was used to stabilize the final cover system for the North Slope of the Operating Industries, Inc. (011) Landfill in southern
2 ANALYSIS OF GEOSYNTHETICREINFORCED VENEER SLOPES 2.1 General considerations This section presents an analytical framework for quantification of the reinforcement requirements for reinforced veneers where reinforcements are placed horizontally and embedded into a comparatively strong underlying mass. An infinite slope configuration is considered for evaluation of stability. Although different definitions for the factor of safety have been reported for the design of reinforced soil slopes, the definition used in this study is relative to the shear strength of the soil:
305
FS =
Available soil shear strength (1) Soil shear stress required f o r equilibrium
This definition is consistent with conventional limit equilibrium analysis, for which extensive experience has evolved for the analysis of unreinforced slopes. Current design practices for reinforced soil slopes often consider approaches that decouple the soil reinforcement interaction and do not strictly consider the factor of safety defined by Equation (1). Such analyses neglect the influence of reinforcement forces on the soil stresses along the potential failure surface and may result in factors of safety significantly different than those calculated using more rigorous approaches. Considering the normal and shear forces acting in a control volume along the veneer slope (or infinite slope), and assuming a Mohr-Coulomb shear strength envelope, Equation (1) can be expressed as:
FS =
c + ( N I L ) tan 4 SIL
where N = normal force acting on the control volume; S = shear force acting on the control volume; L = length of the control volume; c = soil cohesion; and @ =soil friction angle.
where W = weight of the control volume; p = slope inclination; T = veneer thickness; and y = soil total unit weight. From Equations (2), (3), (4), and ( 5 ) , the classic expression for the factor of safety FS,, of an unreinforced veneer can be obtained: FS,, =
C +-tan @ y L T sinp tanp
2.3 Reinforced veneer In the case of a reinforced veneer (Figure 2), the shear and normal forces acting on the control volume are defined not only as a function of the weight of the control volume, but also as a function of the tensile forces that develop within the reinforcements. For the purpose of the analyses presented herein, the reinforcement tensile forces are assumed horizontal and represented by a distributed reinforcement tensile stress t, which corresponds to a uniformly distributed tensile force per unit height. In this case, the shear and normal forces needed for equilibrium of a control volume are defined by: S=Wsinp-tHcosp
(71
N = W c o s p + t H sin p
(8)
H = L sin ,8
(9)
2.2 Unreinforced veneer In the case of an unreinforced veneer (Figure l), the shear and normal forces required for equilibrium of a control volume can be defined as a function of the weight of this control volume. That is:
where H = vertical component of the length of the control volume. From Equations (2), ( 5 ) , (7), (8), and (9), the following expression can be obtained for the factor of safety FS, of a reinforced veneer:
S = W sinp
(3)
N = W cosp
(4)
t s i n p tan@ y L T sinp tanp y T C
tang +-+--
W=yLT
FS, =
Figure 1. Unreinforced veneer.
Figure 2. Reinforced veneer.
t
1 - -1 cos p YT
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(10)
The equation above can be simplified by defining the normalized distributed reinforcement tensile stress t'@ (dimensionless), as follows: i t t = -c o s p YT Using Equations (6) and (1 1) into Equation (10) leads to:
FS, =
FS,, +t* tan p tan 4 1 - t'
(12) Figure 3. Reinforcement requirements to achieve a factor of safety of 1.5 (soil cohesion = 5 kPa; soil friction angle @ = 30").
Equation (12) provides a convenient expression for stability evaluation of reinforced veneer slopes. It should be noted that if the distributed reinforcement tensile stress t equals zero (i.e. in the case of unreinforced veneers), Equation (12) leads to FS,. = FS,, . When the soil cohesion c equals zero, Equation (10) can be simplified as follows:
FS,.=-[tan4 I + t * tan'p tanp l-t I
underlying mass, evaluation of reinforcement vertical spacing, and analysis of seismic stability of the reinforced veneer. These important considerations are beyond the scope of this paper but should be accounted for in the design of reinforced veneer slopes.
j
3 CASE HISTORY: GEOSYNTHETICREINFORCED VENEER SLOPE FOR A HAZARDOUS WASTE LANDFILL
2.4 Determination of Reinforcement Requirements Reinforcement requirements needed to achieve a target factor of safety FS,. of the reinforced veneer, expressed in terms of the normalized required distributed tensile stress t',.erl , which can be derived from Equation (12): i
trcq
= FS,
A reinforced veneer was constructed as part of the final closure of the Operating Industries, Inc. (011) landfill. This case history highlights the final closure of a hazardous waste landfill where the severe site constraints were overcome by designing and constructing an alternative final cover incorporating horizontal geosynthetic veneer reinforcement.
FS, - FS,, + tan p tan 4
Similarly, reinforcement requirements that are needed to achieve a target factor of safety FS, , expressed in terms of the required tensile stress t,, can be obtained from Equations (1 1 ) and (14) as follows:
3.1 Site description
The 60-hectare south parcel of the 011 landfill was operated from 1948 to 1984, receiving approxitan 4 mately 30-million cubic meters of municipal, indusl?q -L--trial, liquid and hazardous wastes. In 1986, the land' y L T s i n p tanp y T frcq = fill was placed on the National Priorities List of cosp FS, + tan p tan 4 Superfund sites. Beginning in 1996, the design of a Equation (15) can be used to assess the reinfinal cover system consisting of an alternative forcement requirements for a given soil shear evapotranspirative soil cover was initiated, and substrength and veneer configuration (slope inclination sequent construction was carried out from 1997 to and veneer thickness). For example, Figure 3 shows 2000. The refuse prism, which occupies an area of the reinforcement tensile stress required to achieve a about 50 hectares, rises approximately 35 m to 65 m factor of safety FS, =1.5 in a veneer slope where the above the surrounding terrain. Slopes of varying soil shear strength is characterized by a cohesion c = steepness surround a relatively flat top deck of about 5 kPa and a friction angle $ = 30". The figure shows 15 hectares. Slopes on the north and east are generthe various combinations of veneer thickness and ally the steepest with considerable portions of the veneer slope inclination that satisfy the design criteNorth Slope as steep as 1.5:1 (horizonta1:vertical). rion. Additional aspects that should be accounted for in 3.2 Design criteria the design of reinforced veneer slopes include the evaluation of the pullout resistance (i.e. embedment The final cover design criteria mandated by the U.S. length into the underlying mass), assessment of the Environmental Protection Agency (EPA) primarily factor of safety for surfaces that get partially into the deal with the percolation performance of the cover, 1 "
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static and seismic stability of the steep sideslopes of the landfill, and erosion control. The percolation design criteria required that the performance of the final cover system be hydraulically equivalent to or better than a layered regulatory cover (prescriptive cover) that includes a 300-mm thick barrier layer with a saturated hydraulic conductivity of lxlO-* m/sec or less. The stability criteria were a static factor of safety of 1.5, and acceptable permanent seismically induced deformations less than 150 mm under the maximum credible earthquake. The basis of the seismic stability criteria is that some limited deformation or damage may result from the design earthquake, and that interim and permanent repairs would be implemented within a defined period. 3.3 Final design One of the most challenging design and construction features of the project was related to the north slope of the landfill. The north slope is located immediately adjacent to the heavily traveled Pomona freeway (over a distance of about 1400 meters), rises up to 65 meters above the freeway, and consists of slope segments as steep as 1.5:1 and up to 30 m high separated by narrow benches. The toe of the North Slope and the edge of refuse extends all the way up to the freeway. The preexisting cover on the North Slope consisted of varying thickness (a few centimeters to several meters) of non-engineered fill. The cover included several areas of sloughing instability, chronic cracking and high levels of gas emissions. The slope was too steep to accommodate any kind of a layered final cover system, particularly a cover incorporating geosynthetic components (geomembranes or GCL). Because of the height of the slope and lack of space at the toe, it was not feasible to flatten the slope by pushing out the toe, removing refuse at the top, or constructing a retaining / buttress structure at the toe of slope. After evaluating various alternatives, an evapotranspirative cover constructed in a monolithic fashion (monocover), and incorporating geogrid reinforcement for veneer stability was selected as the appropriate cover for the North Slope. The evapotranspirative cover had additional advantages over traditional layered cover systems, including superior long-term percolation performance in arid climates, ability to accommodate long-term settlements, construct ability, and ease of long-term operations and maintenance. The selected cover system included the following components, from the top down: 1) vegetation to promote evapotranspiration and provide erosion protection; 2) a 1.2 m - thick evapotranspirative soil layer to provide moisture retention, minimize downward migration of moisture, and provide a viable zone for root growth; and 3) a foundation layer consisting of soil and refuse of variable thickness to provide a firm foundation for the soil cover system. 308
The detailed design of the cover system was Preceded by an extensive laboratory test program to characterize the shear strength, hydraulic characteristics (moisture retention properties and unsaturated hydraulic conductivity), and shrinkage (desiccation cracking) potential of on-site and imported cover soils. The hydraulic equivalence of the evapotranspirative cover to the prescriptive cover was demonstrated by modeling the percolation through both covers. Modeling was performed under simulated rainfall conditions for 30- and 100- year periods, included parametric studies to evaluate the effects of cover thickness and degradation, vegetation, and irrigation. The potential critical modes of slope movement for the North Slope were relatively shallow failures through the cover soils and/or soil/refuse interface. Due to the relatively high strength of the refuse mass, the 1.5:l slopes had a sufficiently high factor of safety against deep-seated movement. Static stability analyses of the cover soil veneer were based on effective stress parameters from backpressure saturated consolidated undrained (C-U) triaxial tests conducted at relatively low confining pressures (24, 48, and 96 kPa) to simulate cover conditions. Pseudostatic stability analyses to support seismic deformation analyses were based on total stress parameters from C-U triaxial tests performed on soaked samples. Unsaturated flow modeling of the evapotranspirative cover on the North Slope indicated that saturation of the cover and seepage parallel to the slope face, which is a usually assumed design condition for surficial stability analyses of slopes, was unlikely to occur. Therefore, stability analyses for the North Slope veneer were performed assuming saturated conditions (back-pressure saturated shear strength parameters and saturated unit weight) for the cover soils, but no perched water table or seepage. Stability analyses showed that for most available monocover materials, compacted to practically achievable levels of relative compaction on a 1.5:l slope (90% of modified Proctor or 95% of Standard Proctor), the minimum static and seismic stability criteria were not met. Veneer geogrid reinforcement with horizontally placed geogrids was then selected as the most appropriate and cost-effective method for stabilizing the North Slope cover. The analytical framework discussed in Section 2 was used in the design. For the given cover veneer configuration (veneer thickness, slope inclination and shear strength of cover soils) the minimum reinforcement required to achieve a static factor of safety of 1.5 was evaluated from charts such as that shown in Figure 3, developed from Equation (15) for a soil shear strength characterized by a cohesion of 5 kPa and a friction angle of 30". The type of geogrid, vertical spacing and minimum embedment that are required to provide the minimum reinforcement stress were then adopted. Figure 4 shows the typical ve-
lifts utilizing scrapers, dozers and sheepsfoot compactors. Since the entire existing cover was stripped off the North Slope surface, the fill included the 1.2 -meter thick evapotranspirative cover and the 0.6meter thick foundation layer, placed as a monolithic 1.8- meter thick veneer. The fill was generally placed along narrow working benches, approximately 4.5 meters in width. When the specified geogrid placement elevation (every 1.5 m vertical intervals on slopes steeper than 1.8:1, and 3m vertical intervals on flatter slopes) was reached, the embedment bench was created as described above and the geogrid was placed over the compacted fill. Geogrid panels were pre-cut to the required length and placed adjacent to each other with a 150 mm overlap. The overlap was in the direction of scraper traffic and the geogrid panels were not attached to the subgrade. This placement pattern was found to be the least disruptive to the geogrid panels when scraper traffic and fill placement on top of the geogrid was undertaken. The geogrid panels were free to slide over one another rather than create ‘waves’ and/or ‘buckle’ when the scraper train was driven over the geogrid layer. The surface on which the geogrid was placed was intentionally kept rough, typically by scarifying prior to placement. This was done to encourage bonding between geogrid and soil and to avoid the formation of horizontal laminations caused by placing geogrids on a smooth surface. To verify the adequacy of the placement method, test trenches were excavated into completed portions of the reinforced cover. Observations showed an intimate contact between the soil and geogrid, and an absence of horizontal laminations or voids adjacent to the geogrid. Construction of the North Slope was accomplished in 12 months. Approximately 500,000 cubic meters of soil and 170,000 square meters of geogrid were placed. Total area of geogrid placement exceeded 9.3 hectares. The maximum height of reinforced portion of the landfill slopes was 55 m (the maximum height of the total landfill slope was 65 m).
Figure 4. Typical veneer reinforcement detail.
neer reinforcement detail selected based on the shear strength of the soils used in construction. The veneer reinforcement consisted of polypropylene uniaxial geogrids, installed at 1.5-meter vertical intervals for slopes steeper than 1.8:1, and at 3meter vertical intervals for slopes between 2:l and 1.8:1. The geogrid panels are embedded a minimum of 0.75 meters into the exposed refuse slope face from which the pre-existing cover had been stripped. The geogrid panels were curtailed approximately 0.3 to 0.6 meters away from the finished surface of the slope cover. This was done to permit surface construction, operation and maintenance activities on the slope face without the risk of exposing or snagging the geogrid. 3.4 Construction The pre-existing non-engineered cover on the North Slope was generally unsuitable to be left in place and was completely removed prior to construction of the new reinforced veneer. The original cover was stripped by a fleet of scrapers and dozers, starting at the top of the slope. Stripping was generally extended to the refuse. The stripped slope face was generally excavated to a 1.5:1 slope with local areas as steep as 1:l The stripped soil was screened through a rotary (Trommel) screen to remove refuse, vegetation and oversize particles, and reused as engineered evapotranspirative cover. Refuse exposed on the slope face was covered with a sprayed-on temporary cover (Posi-shell) to control erosion, refuse migration and odors. Once the entire North Slope was stripped, construction of the geogrid - reinforced veneer commenced from the base of the landfill. To create the geogrid embedment bench into the exposed stripped slope face, further excavation of the refuse and soil had to be undertaken. The excavated refuse and soil from the embedment bench was generally incorporated and compacted into the foundation zone of the cover. The cover fill was constructed in horizontal
4 CONCLUSIONS A framework is provided for the design of steep reinforced veneer slopes such as soil covers in landfill facilities. Instead of using geosynthetic reinforcements along the veneer slope, the proposed framework analyzes the use of horizontally placed inclusions within the veneer slope. Analytical expressions provide the reinforcement requirements for veneer slopes as a function of the soil shear strength and the veneer configuration (thickness and slope inclination). A reinforced veneer was designed using the approach proposed herein for the cover of steep slopes 309
at the Operating Industries, Inc. (011) Landfill. Design criteria include requirements that the final cover should control percolation, resist erosion, and it should remain stable for both static and seismic conditions. The cover selected and constructed includes a 1.8-m thick layer of selected soil that, in order to satisfy stability criteria, was reinforced using horizontally placed geogrids. The geogrid reinforcements were embedded into the underlying solid waste mass in order to provide adequate pullout resistance. The use of horizontally placed geosynthetic reinforcements led to a technically sound and economically feasible approach for stabilization of an
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up to 55 m high soil cover placed over typically 1.5: 1 (H:V) landfill slopes.
5 ACKNOWLEDGEMENTS The authors are indebt to New Cure, Inc., responsible for the site activities described herein, Foster Wheeler Environmental Corporation, general contractor for design and construction, and GeoSyntec Consultants and Advanced Earth Sciences, for contributions to the design and quality control programs of the case history described as part of this paper.
3 Wall structures
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
The performance of buried galvanized steel earth reinforcements after 20 years in service Peter L. Anderson The Reinforced Earth Company North Reading, Massachusetts, USA
John Sankey The Reinforced Earth Company Vienna, Virginia, USA
ABSTRACT: Reinforced Earth@was invented more than 30 years ago by French Engineer and Architect, Henri Vidal, Today there are thousands of Reinforced Earth Structures, reinforced with galvanized steel earth reinforcements, that have been in service for more than twenty years. Recently two of these structures, one in California and one in Virginia, were investigated. Samples of the galvanized steel earth reinforcements and samples of the backfill surrounding the reinforcements were retrieved from each of the structures. Measurements have been taken of the remaining zinc thickness, and the electrochemical properties of the backfill have been tested and confirmed to be within industry standards. The thickness of zinc remaining on the reinforcements is subtracted from the original zinc thickness, and the resulting loss in zinc is compared with the linear loss model used to estimate the service life of the zinc coating. After twenty years in service, the zinc coating is performing better than the loss model and no corrosion of the base metal has occurred. Similar findings are being discovered throughout the United States by Department of Transportation owners, as they unearth galvanized steel earth reinforcements in actual structures that have been in service for many years. These studies suggest that the linear loss model currently in use for the design of Mechanically Stabilized Earth structures may be overly conservative.
1 INTRODUCTION
2 THE CORROSION MODEL
Reinforced Earth@was invented more than 30 years ago by French Engineer and Architect, Henri Vidal. Today there are tens of thousands of Reinforced Earth structures in service worldwide. One very important consideration in the design of Reinforced Earth structures is the service life of the buried galvanized steel earth reinforcement. The corrosion resistance of buried galvanized steel has been studied for nearly a century, beginning with studies launched by The National Bureau of Standards (NBS) in 1910 and reported by Romanoff in 1957 [l]. Terre Armee Internationale carried out additional studies to extend the understanding of underground corrosion as it pertains to Reinforced Earth walls. Results of these studies were published by Darbin in 1986 PI. Today, with thousands of structures having been in service for twenty years or more, there are many opportunities to retrieve galvanized steel reinforcing strip samples that have been in service for many years. The corrosion resistance of the retrieved samples can then be compared with the corrosion model currently used for design.
The following corrosion model is currently used for the design of Reinforced Earth structures. Loss of Zinc (first 2 years): 15 u d y r Loss of Zinc (to depletion): 4 urrdyr Loss of steel (after zinc depletion): 12 urrdyr The above metal loss rates were recommended in 1990 by Elias [3] for design of Reinforced Earth structures reinforced with galvanized steel earth reinforcements in backfill meeting the following electrochemical requirements: Resistivity > 5000 o h d c m @ saturation pH > 4.5 <9.5 If the resistivity at saturation is less than 5000 o h d c m , but greater than 2000 o h d c m , the soluable salt content of the soil should be within the following limits: Chlorides I 100 mg/kg (PPM) Sulfates I 2 0 0 mgkg (PPM) Considering that a minimum zinc thickness of 86 um is required by specification, the zinc coating should be depleted in just 16 years based on the corrosion model.
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range for use within a Reinforced Earth structure. The test results are presented in Tables 1 and 2 for the 1-66 and San Luis Obispo structures respectively. Of course the lineal loss model is a mathematical model for design and not intended to precisely model the actual behavior of the zinc coating. However, if the model greatly under estimates the life of the zinc, the model will likely over estimate the loss of steel during the service life of the structure.
3 RETRIEVAL OF REINFORCING STRIP SAMPLES Galvanized steel reinforcing strip samples and select granular backfill were retrieved from two Reinforced Earth structures after nearly twenty years in service. One of the structures supports Route 101 in San Luis Obispo, California, and the other is located adjacent to Interstate Route 66 in Arlington, Virginia. The San Luis Obispo structure was constructed in 1980, and the samples were retrieved twenty years later in 2000. The 1-66 structure was constructed in 1979 and the samples were retrieved nineteen years later in 1998.
TABLE I : 1-66 ARLINGTON, VA. Description Resistivity PH
4 SAMPLING AND TESTING PROCEDURES One coupon was cut from each of three reinforcing strip samples taken from each structure and labeled for ease of identification. The samples were carefully cleaned of soil with a wire brush, making sure not to remove the zinc oxide on the surface. A thickness measurement was taken at three locations on each coupon using an LVDT extensometer. The extensometer is accurate to within three (3) microns (um). The coupon samples were then put into a saturated solution of ammonium acetate for fifteen minutes. After removing the samples from the solution, they were wire brushed to remove the zinc oxide. Some of the samples were re-immersed and brushed again until all of the zinc oxide was removed. The coupons were rinsed with hot tap water, and then with distilled water, and then dried. A thickness measurement was taken at each of the three (3) locations on each coupon using the LVDT extensometer. On-half of the difference in thickness before and after removal of the zinc oxide represents the average thickness of zinc oxide per side. The remaining zinc thickness was also meassured at the three locations on each coupon using a coating thickness gauge (electrometer). The samples were then carefully measured and weighed prior to removing the remaining zinc by ASTM A90. The samples were then weighed to determine the mass of zinc removed. Based on the density of zinc, the average zinc thickness remaining on each sample is recorded and compared with the coating thickness measured by the thickness gauge. In all cases, the thickness gauge under estimated the thickness of zinc remaining. Samples of the backfill material were retrieved from each structure in the same area as the galvanized steel reinforcing strip samples were retrieved. The backfill samples were tested for gradation, resistivity and pH. The backfill materials were found to be within the acceptable
Well graded sand and gravel with 12.2 percent fines 19,090 OHM-CM @ SATURATION 4.9
AVERAGE THICKNESS OF SAMPLE BY LVDT (pm) After Thickness of Sample As Received Removal of Zinc Oxide Oxide Per Side Coupon 1 5207 5126 41 Coupon2 5215 5131 42 Coupon3 5136 5050 43 Average 5186 5102 42 THICKNESS OF ZINC REMAINING (pm/SIDE) Sample Coupon 1 Coupon2 CouDon3 Average
BY Electrometer 31 36 36 34
By Weight 40 42 46 43
PERFORMANCE OF ZINC COATING Estimated original zinc thickness Average thickness of zinc oxide: Average thickness of zinc remaining: Loss of zinc thickness: Duration in service: Loss of Zinc (first 2 years) Loss of Zinc (subsequent) “‘Specifiedminimum thickness of zinc
Figure 1. - 1-66, Arlington, VA. Samples.
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86 um/side’* 42 u d s i d e 43 um/side 43 um/side 19 years 30 u d s i d e 0.76 um/yr/side
TABLE 2: SAN LUIS OBISPO, CA. Description: Uniform medium sand with 8.5 percent fines Resistivity: 54,000 ohm-cm @ saturation PH 7. I AVERAGE THICKNESS OF SAMPLE BY LVDT (pm) After Thickness of Sample As Received Removal of Zinc Oxide Oxide Per Side 5382 5286 48 Coupon I 5359 5304 28 Coupon2 Coupon3 5385 5298 44 5375 5296 40 Average THICKNESS OF ZINC REMAINING (pni/SIDE) Sample Coupon 1 Coupon 2 Coupon 3 Average
BY Electrometer 66 66 61 64
By Weight 81 71 66 73
PERFORMANCE OF ZINC COATING 1 13 um/side* Estimated original zinc thickness Average thickness of zinc oxide: 40 umlside Average thickness of zinc remaining: 73 um/side Loss of zinc thickness: 40 u d s i d e Duration in service: 20 years Loss of Zinc (first 2 years) 30 umlside Loss of Zinc (subsequent) 0.56 um/yr/side ”Assume = thickness of zinc remaining + oxide thickness
steel was evident on any of the reinforcing strips in either structure. 6 INTERPRETATION OF RESULTS Since the original thickness of zinc put on the reinforcing strip samples is not known, it can only be estimated by adding the thickness of zinc remaining to the thickness of zinc oxide built up on the surface. In the case of the 1-66 samples, the original zinc thickness is estimated to have been about 86 um/ side. This is the specified minimum coating thickness required for earth reinforcing strips. In the case of the San Luis Obispo samples, the original zinc thickness is estimated to have been 113 udside. It is interesting to note that the thickness of zinc oxide built up on the surface of the zinc was about the same (40 um/side) in both the 1-66 and San Luis Obispo structures. As generally agreed by experts of underground corrosion, the loss rate is generally greatest in the first few years and continues to decrease with time. Applying the loss rate of 15 um/year/side for the first two (2) years, one can determine the subsequent loss rate of zinc for the remaining years such that the result is equal to the apparent loss of zinc over the 19 or 20 year period. In both cases, the loss rate is less than one (1) um/yr/side. This loss rate is considerably less than the loss rate of 4 um/side recommended by Elias [3]. 7 OTHER RECENT STUDIES
Figure 2. - San Luis Obispo, CA. Samples.
5 PERFORMANCE OF INVESTIGATED STRUCTURES Reinforcing strips retrieved from San Luis Obispo and 1-66 have a significant thickness of zinc remaining and a significant thickness of zinc oxide builtup on the surfaces of the strips. The zinc oxide essentially protects the zinc, and the zinc protects the base metal (steel) from corrosion. It appears that the zinc coating had not reached its half life in either structure, after nearly twenty years in service. No loss of
Saqiiks et a1 has studied the corrosion performance of galvanized steel strips in nine (9) Florida DOT structures 141. The age of the structures varied from two to seventeen years. For the most part, the backfill environments were within industry standards. The following conclusion with regard to the apparent corrosion rate was reached by Saqiiks: “The results of the field investigation showed that apparent corrosion rates (ACR) of the galvanized reinforcement were very small in most of the elements examined. The average ACR was 1.04 pmly; and 95% of the elements tested had ACR <2.54 pm/y.” William Medford has monitored the corrosion resistance of galvanized steel earth reinforcements in five (5) North Carolina DOT structures [5]. The age of the structures varied from eight to nineteen years. The electrochemical properties of the backfill conformed with the required standards in four out of the five structures. The zinc coating remained on the reinforcing strips in all five structures. The maximum loss of zinc was less than 2 pm/y with the exception of the structure with backfill outside of the required limits.
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8 ADJUSTMENT IN THE LOSS RATES As discussed in FHWA publication No. RD-89- 186 [ 3 ] ,the long term zinc loss rate proposed by Stuttgart University for both non-saturated and saturated soils with resistivities greater than 1000 ohm-cm is 2 u d y r . This loss rate is taken after an accelerated rate for the first few years. Using the Stuttgart model, Elias conservatively recommended that the zinc loss rate be taken as 4 u d y r . In light of the two structures investigated in this paper, the five investigated by Medford and the nine investigated by Saqiiks, it appears that the zinc loss rate should have remained 2 u d y r as recommended by Stuttgart University. Therefore, the loss rates for design should become: Loss zinc (first 2 yrs): 15 u d y r Loss of zinc (to depletion): 2 udyr Loss of steel (after zinc depletion): 12 u d y r Comparing the above model to the results of this study, the loss of zinc after 19 and 20 years in service would be 64 um and 66 um respectively. Therefore at least 20 urn of zinc would remain on the surface of the strip after 20 years in service. The above loss rates are still conservative when compared with
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the findings of this study. However, using the above model in the interim until we can confirm just how conservative it is, is a step in the right direction. Based on the above loss rates the effective life of the 86 um zinc coating will be 30 years instead of only 16. REFERENCES Romanoff, M., Underground corrosion, NBS circular 579, U.S. Department of Commerce, 1957. Darbin M., Jailloux JM, Montuelle J., La perennite des ouvrages en terre armee, Bulletin de Liason Laboratory Central des ponts et chaussees, Paris, France, Jan. - Fev. 1986. Elias V., Durabiilty/corrosion of soil reinforced structures, report No. FHWA-RD-89-186, NTIS, Springfield, VA. December, 1990. SaqiiCs A.A., Scott R., Rossi J., Peiia J.A., Powers R. Corrosion performance of galvainzed strips in Florida mechanically stabilized earth walls. Transportation research board 78"' annual meeting, January 10-14, 1999, Washington, D.C. Medford W.M., Monitoring the corrosion of galvanized earth wall reinforcement. North Carolina Department of Transportation Field Investigation, not publish Presented at 78'h annual meeting of TRB, January 13, 1999, Washington, D.C.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Performance of Mechanically Stabilized Earth walls over compressible soils R.A. Bloomfield, A.F. Soliman & A. Abraham The Reinforced Earth Company, Vienna, Virgina, USA
ABSTRACT: Two projects have recently been completed in New Jersey implementing the use of Reinforced Earth@walls over soft, compressible foundation soils. With differential settlements predicted to be in excess of the standard 1% allowable maximum, portions of a 3,000 square meter project near New York City and a 7,000 square meter project near Atlantic City were designed using an innovative two-stage construction technique. This involved construction of a flexible wire-faced MSE wall that accommodates differential settlement during consolidation of the foundation soils. Following the foundation settlement, standard precast facing panels were connected to the wire facing using a newly developed adjustable connection to complete the finished wall.
1 INTRODUCTION
ground improvement techniques (e.g., vertical wick drains, dynamic deep compaction, high strength geosynthetics, etc.) were utilized to improve the foundation soils below the walls. Successful performance of these structures documents the economic advantage of incorporating specialized mechanically stabilized earth structures in place of traditional cast-in-place walls supported on pile foundations for applications involving high differential settlements. Also, the use of the two-stage wall system in particular, eliminated the need for constructing, and completely removing, conventional temporary earth surcharge embankments with side slopes.
Two large projects have recently been completed in New Jersey implementing the use of Reinforced Earth walls over soft, highly compressible foundation soils. Two different wall designs and construction techniques were used, depending on the amount of differential settlement expected. Conventional Reinforced Earth walls, with precast concrete facing panels and steel soil reinforcements directly attached to the panels, were used in areas where the anticipated differential settlements were within the tolerable limit for the precast panel system (i.e., less than I%, or I meter of differential settlement along a 100 meter wall length). In areas with greater differential settlements, a more flexible, two-stage, Reinforced Earth wall system was developed, which consisted of an initial wire-faced reinforced earth wall and a final precast facing panel attached to it. A 600 millimeter separation is maintained between the two facings during construction and is later filled with gravel. Temporary earth surcharge was placed on top of the wirefaced wall to consolidate the foundation soils and to achieve most of the anticipatded settlement prior to constructing the final concrete facing panels. The advantage of the wire-faced wall is its ability to tolerate more differential settlement than the conventional precast panel system. Due to the extremely low initial shear strength of the foundation soils, some of the walls had to be constructed in several vertical stages to allow the foundation soils to consolidate and gain strength as the fill was being placed. Also, some conventional
2 TWO-STAGE WALL SYSTEM
The two-stage wall system involved initial construction of a Reinforced Earth wire-faced wall, with steel soil reinforcements, to the final height of the proposed embankment. Both the wire facings and the soil reinforcements were designed for the full 75-year design life of the structure. The flexible wire facings were able to tolerate significant amount of differential settlements. The wire facings actually suffered significant deformation and distortion without any adverse impact on the structural integrity and performance of the final wall. Following the completion of the consolidation of the foundation soils, precast concrete facing panels were connected to the wire facings/soil reinforcements with an adjustable connection system, shown in Figures 1 and 2. 317
Figure 1 . Two-Stage MSE wall connection detail.
Figure 2. Connection of precast panels to wirewall.
The adjustable connection system utilized a standard steel coil rod with opposite threads on the two ends of the rod. Coil loops were attached on each end of the coil rod with the opposite threading, resulted in a turnbuckle-type connection. During the initial construction phase, one set of coil loops were slipped onto the connector bar attaching the soil reinforcements to the wire facings, in order to transfer the connection load directly to the soil reinforcements. The coil loop locations coincided with the locations of the soil reinforcements along the connector bar. During the final construction phase, the coil rods were threaded into these coil loops and into a second set of coil loops on the opposite end of the rod. That second set of coil loops were then bolted to
318
the standard tie strips attached to the back face of the precast concrete facing panels. Since the coil loops at both ends of the coil rod were free to rotate in the vertical direction, this connection system allowed for adjustment of the connector rod orientation in the vertical direction (i.e., in the direction of the foundation settlement). Limited adjustment was also allowed in the horizontal direction by this connection system. After the completion of the connection for each course of precast concrete facing panels, the 600 millimeter space between the wire facings and concrete panels was then filled with free flowing gravel to complete the structure.
3 KAPKAWSKI ROAD PROJECT, ELIZABETH, NEW JERSEY In this project, four Reinforced Earth walls (Walls 1 through 4) were constructed along the Elizabeth Waterfront Boulevard and Jersey Gardens Boulevard to provide access from the New Jersey Turnpike to a proposed Mall Development site in Elizabeth, New Jersey, just west of New York City, NY. Walls 1 and 2 were constructed as single-stage walls while Walls 3 & 4 were constructed as two-stage walls. Temporary earth surcharge was not required for any of the walls at this project. Walls 1 and 2 included a total of 170 lineal meters of wall with heights in the range of 3 meters to 10 meters. Geotechnical investigations indicated that the ground surface below Walls 1 and 2 was underlain by a 3 to 4 meter thick layer of waste fill materials. The waste fill materials consisted of loose to dense municipal solid waste intermixed in a matrix of silty sand soil. Underlying the waste fill material, intermittent layers of soft to medium stiff peat and medium stiff silty clay, ranging from 0.3 meters to 1.5 meters in thickness, were encountered. The natural moisture content of the peat layer was found to be near its liquid limit, indicating normally consolidated material with very low shear strength. Total settlements ranging between 0.9 meters and 1.5 meters were estimated to occur due to the construction of the 10 meters high embankment. Due to excessive settlements, a Dynamic Deep Compaction (DDC) technique was recommended to densify the soils/fill, followed by construction of conventional single-stage Reinforced Earth walls. Subsequent to the dynamic compaction, total settlements under the weight of the proposed embankment were estimated to be reduced to a maximum of 460 millimeters (from 1.5 meters originally). A 0.9 meter thick granular mat, reinforced with high strength polyester geogrids, was also recommended to be placed below the embankment to prohibit a possible shear failure through the peat layer and to reduce differential settlements. The objective of the ground improvement techniques was not only to reduce the potential differential settlement but also to enhance long-term global (rotational) stability of the embankment. Actual construction settlements were up to 280 millimeters, at the maximum height of the waWembankment of 10 meters. Within the remaining portions of the wall, actual total settlements were less than 170 millimeters. The calculated differential settlement along the wall face was less than 1%. The cost evaluations performed by the contractor for Walls 1 and 2 indicated that the use of singlestage Reinforced Earth walls in conjunction with
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ground improvements was significantly more cost effective than a cast-in-place wall supported on deep foundations. Walls 3 and 4 were planned adjacent to Spartina Marsh, an existing wetlands area. The walls reached a maximum height of 10.7 meters at the highest end of the embankment. Geotechnical explorations at the wall location indicated highly variable subsurface conditions in the vicinity of Walls 3 and 4. A 2 to 3 meter thick layer of loose to dense silty sand fill was present at the ground surface, overlying marshy soils. Below the fill materials, the borings generally encountered intermittent layers of soft to medium stiff silty clay and peat underlain by medium dense to dense silty sand soils. Bedrock was encountered below these granular layers, about 15 meters below ground surface. Settlement analyses indicated up to 300 millimeters of total settlement but significant differential settlements were anticipated due to the high variability in the soil conditions across the site. Settlement was anticipated to take place over a period of 2 to 5 months after completion of embankment construction. Site restraints as well as the expedited project schedule dictated that the embankment be constructed on the existing ground without any ground improvements. Therefore, a two-stage Reinforced Earth wall system was deemed feasible and was selected, since it has the ability to tolerate significant amount of differential settlements. The two-stage walls were constructed using the technique described earlier in this paper. Actual total construction settlements at the location of Walls 3 & 4 were on the order of 100 to 200 millimeters. Maximum differential settlements along the wall face were computed to be generally less than 1%, with the exception of some localized areas of soft pockets in the near surface soils, where differential settlement slightly exceeded 1%. The reinforced fill used in the construction of the four walls for this project consisted of an open graded crushed stone (locally referred to as No. 57 stone). The particle size of this stone is 19-millime-ters. This project was completed in the summer of 1998. To date, the four walls at this project have performed as expected, with no precast panel misalignment or distress observed due to post constructiodsecondary settlements. During the initial phase of the two-stage walls, only minor bulging (less than 75 millimeters) was observed on the wire facings, with the exception of one local area where bulging on the order of 300 millimeters was observed. This was attributed to poor construction practice (high rate of fill placement and compaction, and use of compaction equipment in close proximity to the wall face).
Figure 3. Settlement plate readings, Wall 3, Kapkowski Road.
4 ATLANTIC CITY-BRIGANTINE CONNECTOR, NEW JERSEY The Atlantic City-Brigantine Connector is the lxgest publidprivate design-build highway project in the recent history of the state of New Jersey. The
Figure 4. Temporary surcharge on wirewall, Atlantic City.
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connector provides a link between the City of Brigantine and the marina-area casinos in Atlantic City, NJ. This project included construction of over 7,000 square meters of Reinforced Earth walls UP to 10 meters in height, at 9 Proposed bridge locations.
Figure 5. Complete two-stage structure, Atlantic City.
Again, both conventional single-stage walls and two-stage walls were constructed. The bridges were supported on conventional stub abutments supported by concrete filled steel pipe piles. The geotechnical investigation indicated most of the site was underlain by highly compressibleAow shear strength clayey soils with up to I meter of total settlement anticipated under the weight of the proposed embankments. Based on embankment heights (up to 10 meters at the bridge approaches) and the low initial undrained shear strength of the foundation soils, the roadway embankments needed to be constructed in several vertical stages to allow the foundation soils to gain strength as the fill was being placed. At several locations on the project the walls were used in conjunction with ground improvement techniques such as vertical wick drains and high strength geotextile base reinforcement. Temporary earth surcharge fills were placed on top of the majority of the walls, as required by the geotechnical design. At four bridge abutment locations the availability of space and time allowed preloading the site with conventional earth surcharge embankments with side slopes. Single-stage wall construction was utilized at these four locations following the removal of the earth surcharge. At an adjacent bridge abutment location, where enough space for a surcharge with side slopes was not available, a single-stage wall was constructed, incorporating a 4.5 meters temporary surcharge above the top of the wall. The use of a single-stage wall was feasible at this location based on an estimated maximum differential settlement of less than 1 % along the wall face. During construction of the wall, there was no evidence of panel mis-
alignment or distress due to the settlement of the soils, confirming the ability of the conventional, single-stage, Reinforced Earth walls to tolerate up to 1% differential settlement, without showing any signs of distress. Over 4,000 square meters of the walls (8 bridge abutment locations and one approach ramp location) at this project were constructed using the two-stage wall system due to anticipated differential settlements being greater than 1% and due to unavailability of space for placement of conventional temporary earth surcharge embankments. Actual construction total settlements along the wire-wall face ranged generally from 600 millimeters to 900 millimeters, as anticipated by the design. Maximum differential settlement exceeded 1 % along the wall face, which was also anticipated by the design. The reinforced fill used in the construction of all walls at this site consisted of a silty fine sand material, containing no more than 15% fines less than 75pm in size. One of the interesting observations on this project was the significant bulging of the wire facings (on the order of 200 millimeters and 400 millimeters) observed in several areas during the surcharge stage of the two-stage walls. The bulging observed exceeded that observed at the Kapkawski Road project described above. The difference in the performance of the wire facings was attributed to the type of the reinforced fill used (crushed stone in Kapkawski versus fine sand in Atlantic City) and the practice used by the contractor to place and compact the fill in the two projects. The Atlantic City project is still under construction, at the time this paper was submitted, with the majority of the precast panels already installed.
32 1
5 CONCLUSIONS Both projects realized significant cost savings by using two-stage Reinforced Earth walls in areas of high differential settlement in place of traditional pile supported cast in place walls, or extensive foundation improvement methods. All walls are performing very well with no precast panel misalignment or distress from secondary settlement. The flexibility of the conventional precast panel system to tolerate up to 1% differential settlement was demonstrated on both projects.
322
The significant deformation and distortion suffered by the wire facings in the initial stage do not have any adverse impact on the structural integrity and the performance of the final wall. Visually, there is no difference in outward appearance of the singlestage or two-stage walls as the precast facing panels are identical. REFERENCES Abraham, A and Sankey, J, 1999. Design and Construction of Reinforced Earth Walls on Marginal Lands. Geotechrzics qf High Water Content Materials, ASTM STP 1374.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets 5: Zeitlinger, ISBN 90 2651 863 3
Flexible facing systems for reinforced soil wall structures characteristics and performance Malcolm Boyd Jabiru Technique Pty Ltd, Sydney, Australia
Pierre Segrestin Freyssinet Intemationnle, Paris, France
ABSTRACT: Facing systems provide a cladding, or skin, to a reinforced soil structure and as such need to be flexible. They need to deform in three directions - transversally, vertically and longitudinally. The ability to deform without loss of function (flexibility) is important to the overall behaviour of the system. It effects the mobilisation of stress in the earth structure and its deformation characteristics. A preliminary quantitative comparison of flexibility highlights the differences between current panel and block systems and quantitative flexibility characteristics are proposed.
1 INTRODUCTION Facing systems for reinforced soil structures provide a cladding, or skin, to the earth structure itself. They are required to retain the earth between the reinforcing layers, and to respond to the behaviour of the earth to which they are attached. Facing may be either hard (eg concrete) or deformable (eg metallic sheet or mesh) or soft (eg geosynthetic or woven wire sheets). Hard and deformable facing systems also serve as a formwork against which the earth backfill is placed and a template for layout of the reinforcement. Flexibility is a critical characteristic of the facing system as it influences the behaviour of reinforced soil structures both during construction (mobilisation of stress) and after construction (modification of stress). Increased connection loads can result from differential settlement of the earth and facing with vertically stiff facing systems. Stress rcdistribution, facing degradation and loss of connection integrity can result from differential settlements along the wall due to construction or foundation variations with longitudinally stiff facing systems influenced by facing unit shape, configuration and joint characteristics.
Transversally, where the facing deforms perpendicularly to its surface (e) Vertically, where the facing is compressed in unison with the backfill Longitudinally, where the facing rotates, or translates, to withstand differential settlements (ta).
(a)
The measurement of deformation is usually defined as in Figures 1, 2 and 3. For transversal deformation it may be defined in absolute terms as a local variation from a designed surface profile, D (Figl), or as a lateral deformation ratio, D/H. Vertical deformation may be expressed as a compression ratio, AH/H (Fig.2).-LongitUdinal deformation may be expressed as a differential settlement ratio, AS/AL (Fig.3).
2 FLEXIBILITY CHARACTERISTICS Flexibility is the ability to deform. The facing system of a reinforced soil structure may have to deform in three directions:
323
Figure 1. Transversal deformation (section).
telescopic arrangement of panels (e.g. 'TerraTrel' wire mesh), compressibility of the joints or pads placed at regular horizontal intervals (discrete concrete panels), placement of each course of units directly on the compacted backfill ("slot storage" panels or from some "green wall" sloping concrete units). Vertical flexibility is an important concern in regard to construction, in conjunction with the quality and compactability of the backfill, and with the compaction procedure. Lack of vertical flexibility may indeed result, during or after construction, in excessive stresses (hence cracking or spalling), or excessive deformation (bulging) of the facing itself, and/or differential settlement between the facing and the reinforcements, leading to local over-stresses at their connections.
Figure 2. Vertical deformation (section).
2.3 Longitudinalflexibility Longitudinal flexibility, parallel to the facing, may be obtained from - variability of the vertical flexibility along the facing, as long as it is available and not absorbed by the compression of the backfill - inclusion of vertical joints at regular intervals (discrete panels, full height panels). Close vertical joints allow the relatively narrow vertical rows of facing units to move with regard to each other as piano keys. More spaced out joints and wider rows may result in a combination of a piano-keys and a fan-shaped effects, with excessive closure or opening of the vertical joints. - (contingent) longitudinal sliding of the facing units on top of each other (blocks), or the distortion of the joint filler or pads in the horizontal joints (discrete panels). Longitudinal flexibility is obviously critical with regard to the anticipated settlements and differential settlements of the foundation soil under the weight of the embankment and reinforced fill structure.
Figure 3. Longitudinal deformation (elevation).
2.1 Transversal.fiexibility Transversal flexibility is obtained from the following - deformability of the material itself (geosynthetic sheets or grids, steel wire mesh), or - shape of the facing units (semi-elliptical steel units), or - joints placed at regular horizontal intervals, provided that they actually work as hinges (discrete concrete panels). Transversal flexibility plays an important role during construction, for the correction of the vertical alignment as well as for the development of predictable horizontal stresses. It may have to play a role after construction, in case of significant post-construction settlements or movements of the foundation soil, inducing deformations of the structure.
3 PERFORMANCE OF FACING SYSTEMS A simple measure of movement capacity can be defined for a general panel configuration as shown in Figure 4 where H = panel height, B = panel width, W = panel thickness and joint widths are h (horizontal) and b (vertical). Flexibility may be represented by - h/W for transverse flexibility, - h/H for vertical flexibility, and - the minimum of h/H or b/B for longitudinal flexibility. Typical values applied to five types of facing panels/blocks are presented in Table 1. These facing
2.2 Verticalflexibility Vertical flexibility is also obtained from the following - deformability of the material itself (geosynthetic sheets or grids), or - shape of the facing units (semi-elliptical steel units), or
324
Table 1. Deformation characteristics, facing systems
H (m)
B (m)
W (m)
h (m)
b (m)
h/W
h/H
b/B
1.50
1.50
0.14
0.020
0.015
14.3%
1.3%
1.0%
Panel, rectangular horizontal
1.50
3.00
0.14
0.020
0.015
14.3%
1.3%
0.5%
Panel, rectangular vertical 1
3.00
1.50
0.14
0.020
0.015
14.3%
0.7%
1.0%
Panel, rectangular vertical 2
6.00
1.50
0.20
0.020
0.015
10.0%
0.3%
1.0%
Block
0.20
0.40
0.30
0.001
0.005
0.3%
0.5%
1.3%
Panel, square
Figure 5. Minimum deformation ratios. Figure 4. General panel dimensions.
types, and dimensions assumed are - square panel, 1.5m (wide) x 1 S m (high) x 140mm (thick), typically the 'TerraClass' system (TAI) - rectangular panel, horizontally oriented, 3.0m (wide) x 1.5m (high) x 140mm (thick) - rectangular panel, vertically oriented, 1.5m (wide) x 3.0m (high) x 140mm (thick) - rectangular panel, vertically oriented, 1.5m (wide) x 6.0m (high) x 140mm (thick) - segmental block, dry stacked, 400mm (wide) x 200mm (high) x 300mm (thick) Joint width are taken as - horizontal joints, 20mm (panels) and lmm (block) vertical joints, 15mm (panels) and 5mm (block). From a transverse flexibility perspective, the panel rotational capacity expressed as the ratio, h/W, is less than 0.5% for the block, but greater than 10% for the panels. This effects the ability of the facing system to accommodate lateral movement as the earth structure is constructed and for the reinforcement to mobilise its resistance. For transverse, vertical and longitudinal flexibility, the joint deformation ratios, expressed as h/W, h/H, or b/B, vary from 0.3% to 14.3%. Only the square panel system achieves a minimum value at, or greater than 1% in all three conditions (Fig.5). With over 10 million square metres in place over 30 325
years, the 'TerraClass' facing panel system may be considered the benchmark for comparing the flexibility of panel wall systems. For longitudinal (differential settlement) capacity, the above comparison does not take into account the interaction between rows, or columns, of panels. The panel layout, in elevation, is critical to overall vertical and longitudinal flexibility. Ideally continuous vertical or horizontal joints are required, however, in practice, only continuous vertical joints can provide relativeIy free movement as horizontal joints are restrained by friction. Vertical oriented panelhlock 'stack' configurations are preferred to horizontally oriented panelhlock 'stretcher bond' configurations for flexibility. The overall wall deformation is also the result of the movement of many panels and the redistribution of strains between panels may result in apparently higher deformations without distress. Examples of gross deformation of Reinforced Earth structures using the 'TerraClass' facing system are shown below. The structure built in Japan (Fig.6), was built over some very soft material under the central section of the wall which was subsequently found to have SPT values between 0 and 3. The central section settled 1.1m over a length of 58m (between vertical separation joints in the wall). The differential
5% 4%
I 3% \
0 2%
1% 0 Yo
0
1
0.5
1.5
L/H
Figure 8. Post-construction lateral deformation.
5 STRATEGIES FOR IMPROVING FLEXIBILITY
Figure 6. Wall deformation, Japan
Improvement in flexibility may be achieved by - articulation of hard panels and blocks through flexible joint design, - promotion of bending with thin sheets or mesh, or - provision for sliding of panel/reinforcement connections in continuous, or rigidly connected facing units Flexible joint design in panel systems is usually achieved with cork strips or rubber pads in the horizontal joints. The use of cork has been generally superseded in order to obtain better durability and performance characteristics. Rubber pads with a two phase load/deflection relationship (Fig.9) allows for improved initial flexibility (during construction) while providing longer term stiffness to avoid concrete to concrete contact under possible severe long term imposed deformations. For block systems, an improvement in vertical flexibility may be achieved by inserting geotextile in the horizontal joints between each block. For a 200mm high block, a residual joint width of 2mm is needed to provide a vertical compression ratio of 1%. For thin flexible sheet, or mesh, facing systems, their vertical flexibility may be promoted by optimising their (section) shape to follow the vertical and horizontal deformations of the earth fill behind. Vidal (1 966) recognised this when he introduced Reinforced Earth and proposed that the semi-elliptical shape of metallic facing panels be proportioned according to the expected ratio of lateral and vertical movement in the earth mass. A simple three-hinge panel model in Figure 10 illustrates the lateral (buckling) deflection behaviour for a plane sheet, under in-plane compression. The initial lateral deflection is shown to be very much greater than the axial deformation. The consequences of this are that - the lateral deflection can rapidly exceed normal wall deformation tolerances, and - the panel may laterally move away from the fill.
Figure 7. Wall deformation, Malaysia
settlement (ASIAL) was over 3.5%, without any significant facial distress or any loss of stability. The structure built in Malaysia (Fig.7), forms part of a major interchange, and settled significantly due to some soft foundations. The differential settlement (AS/AL) observed here is more than 2%, but without distress.
4 PERFORMANCE OF EARTH STRUCTURES Experience has shown that between 1% and 2% allowance for short term lateral movement due to earth placement and compaction is required. Leaning panels back towards the fill typically accommodates this. Reinforcement extensibility also has an impact on potential lateral movement. For typical structureconfigurations, long term lateral movement of between 0.5% (for inextensible reinforcement) and 1.5% (for extensible reinforcement) is possible (Mitchell & Christopher, 1990). Such deformation potential is related to structure configuration as shown in Figure 8 (where L is the length of the reinforcement).
326
Load
Deflection Figure 9. Load/deflection relationship for rubber bearing pads.
Figure 10. Vertical/horizontal deformation compatibility.
For an initial panel height of H,, small initial vertical deformation (h,) will result in large lateral deflection (d,). For example, only 2% vertical compression (hp/Hp) results in 10% lateral deflection (dp/Hp),which is equivalent to lOOmm lateral deflection over a Im panel under 20mm compression. Also, high initial rate of change of lateral deformation to axial deflection (Ad,/Ah,), does not reduce to a value appropriate to a granular fill (typically 0.5) until the ratio of lateral deflection to initial panel height (d,/H, ) is approximately 0.35. This suggests that without other means of vertical flexibility (telescoping panel sheets, or sliding connections), the initial sectional proportions of such a panel must be sufficient to avoid panel separation from the backfill and creation of voids.
ty of systems to deform. Strategies for improving flexibility have been outlined. Further studies are proposed in order to better under stand the interaction of panel and block systemsunder different movement influences, together with the impact on mobilisation of forces in the structure and reinforcement and overall structure behaviour. On the basis of the experience to date, the characteristics of the most common types of facings based on their larger or smaller ability to deform in the three directions are qualitatively described in Table 2. Four degrees of flexibility are considered: good, medium, limited, non existent. This provides a basis for comparison and evaluation of structural impact.
REFERENCES
6 CONCLUSIONS AND RECOMMENDATIONS Flexibility has been defined in terms of the ability to deform transversally, vertically and longitudinally. A preliminary quantitative assessment of the flexibility characteristics of flexible facing systems has been presented which highlights the relative abili-
Mitchell, J.K. & Christopher, B.R. 1990. North American Practice in Reinforced Soil Systems. Design and Pe?-formnnce qf Earth Retaining Structures. Geotechnical Special Publication No.25, Lambe, P.C. & Hansen, L.A., Eds., ASCE, 1990, Cornell University, NY. Vidal, H. 1966. La Terre Armee. Annales de l’lnstitut Technique ciu Batiment et cles Travnux Publics, July-August.
Table 2. Qualitative flexibility characteristics of common facing systems. Type of facing
Cornrnents
Square discrete panels
With mnzpressihle horizontul joints or pads With comnpressible horizontul joints or Good pads Without .sliding, 01’inovahle connections Limited Without s k i i n g , or inovahle connections Limited
Wide discrete panels Narrow full height panels Wide full height panels Wire mesh facing Wire mesh facing + CIP concrete Semi-elliptical steel face Blocks Wrapped around Wrapped around + shotcrete
Transversal flexibility Good
Vertical Flexibility Medium
Longitudinal flexibility Good
Medium
Medium
Good Limited
Non existent Non existent Good Non existent
Medium Limited Good Non existent
Withoutjoint inateriul
Good Limited
Good Non existent
Good Limited
Once covered with slzotcrete
Good Limited
Good Non-existent
Good Non-existent
Once covered with cast in place coilCrete
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Shaking table and numerical modelling of reinforced soil walls M.M. El-Emam, R.J. Bathurst & K. Hatami Geotechnical Research Group, Civil Engineering Department, Royal Military College of Canada, Kingston, Ontario, Canada
M.M. Mashhour Professor and Head, Structural Engineering Department, Faculty of Engineering, Zagazig University, Zagazig, Egypt
ABSTRACT: The behaviour of reinforced soil walls under seismic loading is investigated using physical shaking table tests and numerical simulation, with particular emphasis on wall footing toe restraint conditions. A specially instrumented toe was designed to study the influence of toe restraint condition on model wall behaviour under static and dynamic loading. The model walls were instrumented to measure lateral facing displacement, reinforcement loads, load transmitted to the facing panel toe, and acceleration response along the height of the wall face and within the backfill. After construction, the models were subjected to a stepped amplitude sinusoidal base acceleration. Results from the experimental program were used to verify the accuracy of numerical models simulating seismic response of the shaking table tests using a finite difference-based computer program. The predicted responses from numerical models were found to be in general agreement with measured results. Both numerical and physical test results indicated that the toe restraint condition has a significant effect on the seismic response of the model walls. Some implications of the experimental and numerical modelling results to analysis and performance of propped-panel reinforced soil retaining walls are summarised. reduced-scale wall models to the shaking table base excitations.
1 INTRODUCTION
Seismic design methodologies that are currently used in North America for geosynthetic reinforced soil walls have been based largely on the results of numerical modelling of reinforced soil structures constructed with relatively in-extensible steel reinforcement (Bathurst and Alfaro 1996). Methods based on experience with in-extensible reinforcement are not necessarily applicable to walls using relatively extensible geosynthetic reinforcement products with respect to the required number/strength, location and length of reinforcement layers (Bathurst and Hatami 1998). A research strategy to update current approaches for seismic design of geosynthetic reinforced soil walls is to use numerical models that are validated against the results of carefully conducted shaking table tests on reduced-scale models of reinforced soil walls. This paper presents selected results from a series of reduced-scale reinforced soil retaining wall models that were tested on a shaking table. The model walls were constructed with different toe boundary conditions and loaded to failure using a steppedamplitude harmonic excitation record. Following the experimental tests, a calibrated numerical model was developed using a finite difference-based computer program to simulate the measured response of
2 MODEL SHAKING TABLE TESTS 2.1 RMC shaking table
The Royal Military College shaking table is comprised of a 3 m by 3 m steel platform driven by a 100 kN hydraulic actuator with a rt75 mm stroke range. The maximum payload capacity of the table is 4500 kgf. The table is capable of shaking a full payload at frequencies up to 13 Hz and peak base acceleration amplitudes up to 2g. A data acquisition/control system is used to drive the table using a prescribed displacement function. Test models are confined within a rectangular box 1.4 m wide by 3 m long that is bolted to the steel shaking table. 2.2 Model configurations
A series of fourteen 1 m-high models at 1/6 scale comprise the experimental portion of the research program. In order to predict the behaviour of equivalent (prototype) models at full scale, the model walls were designed in accordance with simulation rules proposed by Iai (1989). The reinforcement vertical spacing for the two models discussed in this paper was S, = 0.23 m. The reinforcement length, L, was 329
chosen to give U H = 0.6, where H is the height of the model (Figure 1). In order to examine the influence of the toe restraint condition on model response, two different toe arrangements were used. The facing toe in different wall models was either hinged (restrained in the vertical and horizontal directions, while free to rotate) or sliding (restrained in the vertical direction only and free to rotate and slide horizontally). 2.3 Materials The backfill soil used in all tests was a commercially available, manufactured clean uniformly graded sand. The primary reason for using this sand was to ensure a repeatable sand placement condition for all tests. The sand has a unit weight of 15.7 kN/m3, peak direct shear friction angle of 52" and dilation angle of 14.5" when placed in a dense condition. At large strain and in a loose condition, the sand has a unit weight of 12.9 kN/m3 and residual direct shear friction angle of 46". A knitted geogrid polyester product with a polyvinyl chloride coating and an aperture size of 45 mm by 25 mm was used as the reinforcement. The tensile strength of the geogrid was 1.8 kN/m at 2% strain and 3.25 kN/m at 5% strain in the loading direction. This particular geogrid was chosen because its stiffness in reduced-scale models falls in the range of typical reinforcement stiffness values at prototype scale. The facing panel was constructed using rectangular hollow steel sections with cross-section dimensions 75 mm by 38 mm. A total of 26 sections were bolted together to form a 1 m-high rigid facing with a thickness of 75 inm and length of 1.4 m. 2.4 Instrumentation Up to 64 instruments were installed in each test wall to monitor the following performance features of the models during construction and dynamic loading: (a) facing horizontal movements; (b) reinforcement displacements and strains; (c) horizontal and vertical
I
Extensometer cable,
-
1.Back ,
reinforcemenl
I
-
\
Slide Rails Load cell o
Extensometer node
ShakingTable
Accelerometer
-
2.5 Base excitation A stepped-amplitude, sinusoidal function with a frequency of 5 Hz was used to shake the models. The actuator stroke was increased at 5 second intervals to generate an equivalent incremental base acceleration of 0.05g until excessive model deformation occurred. The same base excitation history was applied to each model in order to allow quantitative comparisons to be made between different test configurations. 2.6 Test results
back
Due to space constraints, only selected test results are presented here. Typical displacements at the top of the wall facing versus input base acceleration for nominally identical model walls with L/H = 0.6 but with hinged and sliding toe boundaries are shown in
T Reinforcement strain gauge N
toe loads; and, (d) facing and backfill acceleration response at different elevations. A typical instrumentation layout and test wall configuration is shown in Figure 1. Cable-extension position transducers (extensometers) were used to measure facing displacements during base shaking. The body of each transducer was attached to a rigid vertical post mounted on the shaking table. This allowed the devices to record the relative movement of the model facing with respect to the table. Local strain in the reinforcement was measured using strain gauges bonded to the polyester bundles of the longitudinal members. Extensometers were used to measure global strains, which are defined as strains measured over several geogrid apertures. Global strain measurements were used to confirm strain gauge readings and to convert strain values to load. Acceleration response during shaking was measured using seven accelerometers, one of which was attached to the table to record the input base acceleration. Two accelerometers were attached to the facing and four were buried at different locations in the backfill as shown in Figure 1. The forces transmitted from the facing panel to the footing (facing toe) were measured using vertical and horizontal load cells. The toe force was decoupled into horizontal and vertical components by using three linear roller bearings at the base. These roller bearings were used to reduce the friction in the horizontal direction at the toe and to ensure that the entire horizontal component of the toe force was transferred only to the horizontal load cells. Two data acquisition systems were used to take data from 64 instruments at a sampling rate of 150 readings per second per channel. A high sampling rate was required to avoid aliasing effects and to capture the peak values of dynamic wall response induced by base shaking.
Position transducer
Figure 1. Test configuration and instrumentation for model walls.
330
Figure 4 shows the magnitude and distribution of the peak load (static condition and during base excitation) in the reinforcement layers at the wall connections for nominally identical sliding and hinged toe models. The plots for the hinged toe model also include horizontal toe loads. The load distributions are plotted for different input base acceleration magnitudes. The data show that reinforcement connection loads and toe loads increase with magnitude of input acceleration amplitude. The top reinforcement layer did not develop any additional load during shaking. This may be attributed to the small overburden pressure on the top layer (0.15 m from the top of the model) which resulted in slippage of the reinforcement layer in the soil. The reinforcement connection load in the bottom layer increased significantly under severe shaking in the final stages of both test configurations. For base input acceleration values greater than 0.1g, the reinforcement connection loads at the second and third layer below the top of the wall are greater for the hinged toe model than for the sliding toe model. This result demonstrates that the toe boundary condition will influence not only the magnitude of reinforcement loads but also the distribution of loads under large base accelerations.
Figure 2. For the hinged toe model, the facing displacement was primarily a result of facing rotation. However, for the model wall with a sliding toe, the horizontal facing displacement was a combination of facing panel rotation and translation at the facing toe. For these reasons, the model wall with a sliding toe showed greater facing horizontal displacement at the top than the hinged toe model (Figure 2). It is clear from the figure that, the toe restraint condition affects the horizontal displacement significantly at input base acceleration magnitudes greater than 0.3g. At this acceleration level, the base of the sliding toe model started to slide outward dramatically. The variation of the total vertical toe load with input base acceleration magnitude is shown in Figure 3. At the static condition (end of construction), the vertical toe load is about 75% greater than the weight of the facing column. This is due to downdrag forces acting on the back of the facing. Downdrag forces are the result of frictional shear forces developed between the soil and the facing column as well as the vertical component of forces at the connections due to settlement of the soil behind the facing. Down-drag forces increase with acceleration amplitude of base shaking. However, the toe condition does not have significant effect on the magnitude of vertical load developed at the toe.
Figure 4. Measured connection and toe loads for models with sliding and hinged toe boundary vs. elevation at different input base accelerations.
Figure 2. Measured and predicted displacement at the top of the model facing vs. base acceleration amplitude.
Figure 3. Vertical toe load for models with sliding and hinged toe boundary conditions vs. input base acceleration.
Figure 5. Toe load and sum of connection loads normalised to the total earth force for model with hinged toe VS. input base acceleration amplitude.
33 1
Figure 5 illustrates that at the end of construction, the horizontal toe load (in these experiments) is equal to the sum of connection loads. This observation is consistent with full-scale instrumented wall tests that have shown that a restrained footing may contribute significant earth pressure capacity to geosynthetic reinforced walls constructed with a structural facing under static loading (Bathurst and Walters 2000). As the acceleration amplitude increased, a slightly larger portion of the total earth force behind the facing panel was taken by the toe. When the input base acceleration exceeded 0.3g, the reinforcement started to attract more load than the toe. An explanation for this behaviour is that at a base acceleration amplitude of 0.3g, soil-reinforcement interaction is fully mobilised and the reinforcement layers begin to take a larger portion of the load. Nevertheless, over the entire course of the test the toe and reinforcement layers each attracted roughly 50% of the total static and dynamic earth pressure. Outward (away from the soil) acceleration amplification factors plotted against input base acceleration amplitude are shown in Figure 6. It is clear from the figure that prior to an acceleration amplitude of 0.3g, all outward acceleration amplification factors are small but increase significantly thereafter. A possible explanation is that soil dilation during shaking leads to a reduction in stiffness of the system and a corresponding larger amplification factor. Increasing magnitude of amplification factor with increasing base excitation for extensible reinforcement shaking table models has also been reported by Matsuo et al. (1998). These results are in contrast to results of studies using relatively in-extensible metallic reinforced soil wall models that showed that acceleration amplification magnitude decreases with base acceleration amplitude (e.g. Richardson and Lee 1975, Fairless 1989). In general, the outward acceleration amplification factors are larger for the model with a hinged toe than for the model with a sliding toe. This observation is in accordance with
numerical results for prototype-scale propped panel walls subjected to base excitation reported by Bathurst and Hatami (1998). Figure 6 also shows that there is a phase difference between input base acceleration and response accelerations for both hinged and sliding toe model walls. Out-of-phase motions in the soil can be expected to have a marked effect on the distribution of dynamic reinforcement load with time during base excitation (e.g. Steedman and Zeng 1990). 3 NUMERICAL SIMULATION
3.1 Numerical model and material properties The results obtained in the experimental part of this research were used to develop a calibrated numerical model using the finite difference-based, twodimensional stress analysis program FLAC (Itasca 1998). Figure 7 shows the details of a typical numerical configuration that was used in the analyses. The soil was modelled using an elastic-plastic strain softening model with Mohr-Coulomb failure criteria. Material parameters used to model soil were: peak friction angle, &, ; cohesion, c ; residual friction angle, qCv; dilation angle, y~ ; shear modulus, G; and bulk modulus, K. The numerical values for soil parameters were established by calibrating a numerical model to give the same load-deformation response as physical direct shear tests. Direct shear tests were carried out on the same sand prepared to the same density as the shaking table models. The geosynthetic reinforcement was modelled using linear elastic perfectly plastic, two-noded cable elements with axial stiffness, J = 78 kN/m. The cable elements had negligible bending stiffness and compressive strength. The stiffness of the geogrid was determined from load-elongation curves produced from wide-width strip tensile tests. The base input motion was introduced to the wall models by applying a prescribed horizontal velocity and zero vertical velocity to the nodes at the base
Figure 6. Amplification factors for hinged and sliding base models vs. input base acceleration.
Figure 7. Numerical grid for the reinforced soil model wall with fixed toe condition.
332
of the numerical grid. The applied velocity was increased in a stepwise manner to match the experimental input acceleration history until excessive deformation occurred. 3.2 Comparison of numerical and physical results Calculated and measured maximum displacement values versus input acceleration for the hinged and sliding model walls described earlier are shown in Figure 2. It should be noted that the hinged-toe and sliding-toe models became unstable at base acceleration amplitudes equal to 0.4g and 0.35g, respectively. In addition, it can be seen that the predicted and measured facing displacements show good agreement. The figure shows that the sliding toe model wall recorded larger top displacements than the hinged toe model wall in both physical tests and numerical simulations. A comparison of the vertical and horizontal toe loads calculated using numerical analysis and measured in the physical tests is shown in Figure 8. The numerical model results are in reasonable agreement with the experimental data. It can be seen that the numerical simulation captured the trend of slightly lower magnitude of vertical toe load measured for the sliding toe model compared to the hinged toe
Figure 8. Measured and predicted toe loads for model walls vs. input base acceleration.
Figure 9. Measured and predicted connection and toe loads at different input base acceleration for hinged toe model.
333
model (see also Figure 3). This observation may be attributed to a larger amount of soil mass that passed outward beyond the heel of the wall face as a result of the slightly larger rotation of the facing panel for the hinged toe model. The accuracy of predicted connection loads in Figure 9 is reasonably good over the middle of the wall height (up to 0.3g) and at the toe. The predicted top layer response is poor. This is likely the result of reinforcement slip which was not modelled in numerical simulations (i.e. the reinforcement layers were taken as fully bonded to the sand soil). This observation highlights the requirement to include interface slip in numerical models for reinforcement layers under low overburden pressures. However, the numerical results do demonstrate that the facing toe attracted a significant portion of the total horizontal earth pressure at all stages of the simulation and therefore reduced the demand on the reinforcemen t layers. 4 CONCLUSIONS AND IMPLICATIONS TO DESIGN Selected results of shaking table tests on reducedscale, reinforced soil retaining walls with L/H = 0.6 and a vertical spacing S, = 0.23 m are reported together with numerical modelling. The investigation is focused on the influence of the toe boundary condition on lateral displacements, reinforcement loads, toe loads and acceleration amplification in the backfill. The following conclusions can be made: 1. The displacement response for the model wall with a sliding toe was greater than the response of the model with a hinged toe when base acceleration amplitude exceeded 0.3g. 2. The vertical toe load was found to be greater than the facing self-weight throughout each experiment, due to downdrag forces, and increased as the input base acceleration increased. 3. The type of toe boundary condition did not influence the magnitude of vertical toe loads generated at end of construction (static) and under dynamic loading. 4. The footing for models with a restrained toe attracted approximately 50% of the static and dynamic horizontal earth force acting against the back of the facing panel. The magnitude and distribution of reinforcement loads was observed to be influenced by the boundary toe condition. In particular, reinforcement load in the bottom reinforcement layer was greater for the unrestrained toe case compared to the hinged toe case. 5. Input base acceleration was amplified towards the backfill surface. The amplification factor was found to be about 1.2 for input accelerations lower
the walls in this study is opposite to that reported in the literature for metallic reinforced soil wall experiments. Hence, empirical design rules based on reinforced soil walls with metallic reinforcement materials may not apply to walls constructed with relatively extensible reinforcement products.
than 0.3g, and increased sharply to about 2.2 for greater input base acceleration amplitudes. The trend of increasing amplification factor with increasing base acceleration is reversed from the trend reported in the literature for reduced-scale reinforced soil walls constructed with relatively in-extensible reinforcement layers. 6. Overall, the numerical response results of model walls subjected to base excitation were found to be in close agreement with the measured data with the exception of predicted reinforcement loads for reinforcement layers under low overburden depths. The results of this investigation have important implications to the design of geosynthetic reinforced soil retaining walls under static loading and in seismic environments. The contribution of a restrained toe to wall load capacity and the distribution of reinforcement loads is not considered in current pseudostatic design methods. Neglecting the load-carrying capacity of a laterally restrained facing toe may result in over estimation of reinforcement design loads and unsafe design of the footing. Downdrag forces acting against the back of a rigid facing column can be expected to place additional loads on the connections and the footing. These additional loads are not considered in current North American practice and hence current practice may be unsafe in this regard. The acceleration amplification and phase lag along the wall height and in the backfill observed in the physical tests are not considered in current pseudostatic design methods for reinforced soil retaining walls. The trend of increasing acceleration amplification with increasing base acceleration observed for
REFERENCES Bathurst R.J. & M.C. Alfaro 1996. Review of seismic design analysis and performance of geosynthetic reinforced walls, slopes and embankments. Earth Reinforcement, Ochiai H., N. Yasufuku & K. Omine (eds), Balkema: 2, Proc. of Zntern. Sytnp. on Earth Reitif: Fukuoka, Kyusliu, Japan, November 1996 Keynote Lecture: 887-918. Bathurst, R.J. & D.L. Walters 2000. Lessons learned from full scale testing of geosynthetic reinforced soil retaining walls, CeoEng2000, Melbourne, Australia, November 2000, 14 p. Bathurst R.J. & K. Hatami 1998. Seismic response analysis of a geosynthetic-reinforced soil retaining Wall. Ceosyn. Zntern. 5(1-2): 127-166. Fairless, G.J. 1989. Seismic performance of reinforced earth walls. Research report 89-8, Dept. Civ. Eng., Univ. of Canterlmry, Christchurch, New Zealand. Iai S. 1989. Similitude for shaking table tests on soil-structurefluid models in 1g gravitational field. Soils Found. 29(1): 105-1 18. Itasca 1998. Fast Lagrangiun analysis of continua vet: 3.4. Itasca Consult. Group Inc. Minneapolis, MN, USA. Matsuo O., T. Tsutsumi, K. Yokoyama & Y. Saito 1998. Shaking table tests and analyses of geosynthetic-reinforced soil retaining walls. Ceosyn. hitem. 5(1-2): 97-126. Richardson G.N. & K. Lee 1975. Seismic design of reinforced earth walls. ASCE, Proc. .I. Geotech. Eng. Div. lOl(CT2): 167-187. Steedman R.S. & X. Zeng 1990. The influence of phase on the calculation of pseudo static earth pressure on retaining walls. Geotecliizique. 40(1): 101-112.
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Reinforced earth ramps over flexible inclusions in Beirut J.B. ESTA Civil Engineer from the “Ecole Nationale des Ponts et Chausskes ”, Paris. Professor at the “Ecole Superieure d ’Ingenieurs de Beyrouth ”, Lebanon.
ABSTRACT: This paper describes the execution of relatively high reinforced earth retaining walls founded on very compressible soil, the safety factor at the punching being very close to 1. The installation of sand drain columns beneath the walls has considerably ameliorated the foundation soil and has secured in few months the total settlement of 0.5m.
1 INTRODUCTION
2 GEOTECHNICAL DESCRIPTION OF THE SITE
At the beginning of the years ~ O S ’ ,we11 before the war, the Executive Council of the Great Projects of the city of Beirut has foreseen a network of highways cutting out the capital and leaving for road users easy approaches on the South and the North exits. One of them, subject of our study, is the one allowing to connect the two zones surrounding the Beirut River being essentially BOURJ HAMMOUD and its outskirts on one part and ACHRAFIEH on the other. The two branches of the Y are the extensions of the two principal axes connecting the East and the West of Beirut notably the Ring leading out to Charles MALEK Avenue and the SODECO passage leading to President Elias SARKIS Avenue. As Figure I shows, this way is composed of roads of variable elevations including a series of bridges. The very compressible nature of the foundation soil of the plain part has led to the replacement of the sections of the bridges whose foundations were proving very onerous by Reinforced Earth access ramps. In fact to have an idea of the cost of the foundations of the bridges’ supports used on the Y branch thriving from the President SARKIS Avenue, it is appropriate to specify that for each pier 18 piles of 80cm diameter and of a depth varying between 25 and 30m were foreseen; the cost, per support, is about 150,000$. Needless to say that in these conditions a retaining wall solution where applicable was much less expensive. In view of the difficulties presented by the foundation soil, only the reinforced earth by order of its flexibility could afford an economic and elegant solution.
Figure 2 points to the position of the first boreholes realized on the site in 1994 (Bl to B9). These soundings gave a qualitative description of the soil and
Figure 1. General view of project.
335
Figure 3. Typical borehole log. Figure 2. : Boreholes set-up of 1994 : Boreholes set-up of 1996
In these conditions and in view of recognizing the mechanical characteristics of this slightly resistant clay, SOIL MECHANICS has conducted a broader survey including 10 boreholes concentrated in the compressible zone and noted S. 1 to S . l l as shown on Figure 2. The level of the top of these boreholes varies between +6 and +7.7. This survey was done by means of the static dynamic penetrometer ANDINA (No similar engine exists in the world after the economic crisis at the beginning of the years 90s’ in Europe). We hereafter, give one typical section of registered end resistance and lateral friction. Figure 3. The exploitation of these results together with the layers classification based on a normal consolidated clay and applying the relation given in “The Application of Pressuremeter Test Results to Foundation Desing in Europe edited by the International Society for Soil Mechanics and Foundation Engineering, leads to distinguish well differenciated types of soil section where,
( 0 B1-B9 ) (m s1-SI 1)
showed that, roughly, over the lower platform, 1.5m to 3.0m of fill material existed, overtopping an argillaceous creamish deposit, possibly including sand, of a thickness at least equal to 17m. Below, a clayey mar1 was found. The tests on the found soil were rare: some pressuremeters, more or less successfully done due to the enlargement of the hole during drilling, about ten identification tests showing a liquid limit varying between 23 and 52, a plastic limit ranging respectively between 1 I and 28, and a natural water content very close to the plastic limit. Two consolidation tests give the following results: - at 5m depth,
- at 16m depth,
Cc e, dc yd
= 0.078 Cv ~ 1 0cm2 - ~/s = 0.5
Cc e, (s’c yd
= = = =
”
= 70KPa = 18.2KN/m3
0.056 Cv ~ 0.685 I5OKPa 16.2KN/m3
1 cm’/s 0 ~
~
This leads to overconsolidation ratios OCR very close to 1, hence indicating that the clay in presence is norrnaly consolidated. 336
- The first one, between the Corniche du Fleuve and the BADAOUI road, is characterized by following layers :
Om 6m 1 lm 22m-
6m llm 22m 26m
96KPa
5OOKPa
- The second one, between the BADAOUI road and VARTAN road is characterized by the following layers :
Om - l l m 35KPa 350KPa
forced earth serving solely as retainement to the fill. Figure 5. The maximum height of the reinforced earth wall is equal to 11Sm, of which at least l m is for embedding, and the corresponding length of the high adherence galvanized steel strips is equal to 8m. 4 GEOTECHNICAL STUDY
- The third, between the VARTAN road and the railroad is characterized by the presence of a relatively stiff clay, Cu being comprised between 141 and 243KPa. Finally, the water table existed at level +3.5.
4.1 Punching (Ultimate bearing capacity)
3 THEPROJECT
c u = P1/10.
The road project includes four tracks, two of which (the extrems of each side) are being very close to natural ground, and the remaining inner two, cross over a fill and are staggered one with respect to the other. Figure 4 shows an average section where it appeared that the total width of the fill is equal to 17.6m. This fill is held by two extremity walls and by an inner wall, all of them being reinforced earth walls. Elsewhere, at the abutment of the Corniche du Fleuve, the bridge lays over piers leaning against slabs transmitting their load through piles, the rein-
The calculations show that the safety factor at punching is comprised between :
The ultimate bearing capacity quit has been studied by means of the MENARD Pressuremeter method, the Menard limit pressures being calculated from :
1.08 and 1.57 for section I. 0.60 and 0.92 for section 11. 3.35 and 4.68 for section 111. This showed in a clear manner that it was necessary to improve the foundation soil below the reinforced earth walls, - on one hand as to increase the previous factors - on the other hand as to avoid the service settlements which are more important and as to accelerate the speed of these settlements.
Figure 5. Abutment P1 foundation drawing.
Figure 4. Reinforced earth project section.
337
(1)
As we will see hereafter the most suitable technique will consist of realizing sand drains and of substituting the superficial soil, till the water table level, by a well compacted, 2m thick, draining gravelly fill. 4.2 Total settlement The total settlement has been computed by the method indicated by Menard. This method has been preferred to that unidimensional consolidation of Terzaghi for the thickness of compressible layer is in the magnitude of the width of the width of the fill, which is very far from Terzaghi’s hypothesis. The carried out calculations have shown that : - on one hand the settlement’s preponderant values were those of spherical term; i.e, those induced by the exerted loads between the surface and 12m depth. - on the other hand the total settlement in the Corniche du Fleuve-VARTAN road zone varied between 15 and 40cm according to where we stand respectively near to Vartan road or close to the Corniche du Fleuve.
4.5 Stability of great sliding We have verified the stability of sliding along a circle crossing the superficial layers and the reinforced earth fill. The calculations have been conducted by TALREN with the characteristics and safety factors indiis 1.06 cated on Figure 6. The minimum found rmax for short term characteristics of section 11. 4.6 Conclusions The analysis of the previous results shows that it was necessary : - to foresee sand drains as to insure a radial drainage and to accelerate the consolidation. For a four months period of consolidation, these 35cm diameter drains will be in the center of a 2m x 2m square lattice and will be at least 12m deep ; the plan of these drains is given on Figure 7. - to substitute the surficial existing soil down to the water table level by another with better mechanical characteristics and which will be additionally draining. In principal, this excavation was of 3m depth, the substitution gravelly fill being of 2m thickness. See Figure 5 .
4.3 Consolidation period The settlement rate is defined from the time factor Tv given by the relation : TV= CVtfH’
(2)
Where H is the drainage distance. H is being taken equal to 10m. For a consolidation degree Uv of 70%, the time factor Tv given by TERZAGHI’S unidimensional consolidation chart is equal to 0.4.
5 SETTLEMENTS OBSERVATION The settlements have been followed on eight benchmarks of which four between the VARTAN road and the BADAOUI passage.
From which 0.4 = 10‘3 x t ( 1000)’ We derive t = 12 years. 4.4 Acceleration of consolidation To accelerate consolidation it was thus necessary to provide a radial drainage by means of sand drains. Supposing that the radial consolidation factor Cr is equal to Cv and choosing a 4 months consolidation period, the radial consolidation chart of BARRON shows that the radial consolidation degree Ur is equal to 70%, which can be obtained with 35cm diameter drains whose diameter of influence is of 2.5m corresponding to a 2m x 2m square lattice. In these conditions the time factor Tv is equal to 0.0 1 which corresponds to a vertical consolidation degree Uv equal to 15%. The global consolidation degree U is given by the relation (1-U) = (1-Uv) (1-Ur) from which U = 75%.
338
Figure 6. Stability verification by TALREN.
Figure 7. Drawing of the 281 piles of sand.
In this last zone the total settlements having been limited to about ten centimeters at the end of construction, we have stopped their readings and we were interested to follow those obtained in the zone between the BADAOUI passage and the Corniche
du Fleuve where the amplitude was by far greater. On Figure 8 we give an elevation of the wall in this zone and we indicate the position of benchmarks C and E over walls 2 and 4.
Figure 8. Elevation of the wall.
Figure 9. Settlements observation on wall 4.
Figure 10. Settlement observation on wall 2.
339
Table 1. Total settlement and consolidation degree four months after load application. Designation
Bench-mark
Settlement
Settlement
End of loading
four months after
Settlement
U
1,5 year after
%
Wall 4 Wall 4
C E
21.5 33.5
27.5 41.5
31.0 46.0
87 88
Wall 2
C
19.0
23.0
25.5
88
The settlements observation started on September 30 th, 1996 that is as early as the application of the first load and continued as indicated on Figures 9 and 10 during the construction of the structure, and over a one and a half year period after its completion. These curves have shown the following: - During the loading phase the settlement speed is comprised between 4.5 and 5.5mdday at benchmarks presenting an important total settlement (E2 and E4) and is close to 3 m d d a y elsewhere. However, once the load applied the settlement speed at benck-marks E2 and E4 is in the order of 1.5 to 2mm/day during two months and it drops by at least O S d d a y thenafter. - The total loading of the soil is completed on May 20th, 1997, and the total settlement noted on that date was equal to 38.5cm (Bench-mark E, Wall 2); four months later it equaled 47cm and on September 1998 a final reading gave the value of 5 1Scm. The scrutiny of the settlement curve aspect on this mark E shows that we are practically at proximity of an asymptote that would be close to 52 cm for which the consolidation would be in practice terminated. In these conditions, four months after the end of works, the degree of consolidation could be calculated as being equal to 47/52 = 90%, much better than the value foreseen by computation. It is worth to note that at the level of the four bench-marks the total settlement four months after
340
load application as well as the consolidation degree as calculated previously were as shown on Table 1. It becomes evident that the calculated settlements in the most sensitive zone of the structure have values in the same order of that maximum obtained by computation; i.e. 40 cm, but on average the measured settlements are appreciably weaker than the calculated settlements. This could not be otherwise for the calculated value has been obtained by means of correlations which, even standardized, cannot reproduce reality. - The differential settlement obtained is in the order of 2 to 3%, a value ten times greater than accepted for reinforced concrete walls. - The settlement having not led to any tilting of the wall nor to any spectacular fracture of the panels it has been held at wall top by a reinforced concrete beam.
6 CONCLUSION The use of the reinforced earth has allowed in this project to solve a complex technical problem and to realize substantial savings by substituting foreseen deep foundations. Once more if it were still needed to be proven the flexibility of this material and of the concrete panels of the dressed facing wall has permitted to bear important global and differential settlements and this without distinction between retaining walls and bridges’ abutments.
Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Construction of a geogrid-reinforced earth-wall inside a ware-house Franqois Vie1 REHAUAG
+ CO,Erlangen, Germany
ABSTRACT: The construction of a new high-bay warehouse at REHAU's Viechtach factory in the Bavarian forest near the Czech border involved a major height difference ( for incoming and outgoing goods) between the first and the second level ( over 20 meters). Due to the undulating topography and limited space, retaining walls were essential to overcome the difference in levels. A 7.20-meter high geogrid-reinforced wall sloping at 85" was constructed 50 cm behind a facing concrete wall inside the warehouse. This way, no earth was expected to exert pressure against the concrete facing wall. As a second bonus, the reinforced concrete plate base for truck and fork-lift traffic at the second level on the geosynthetic wall structure was reduced in thickness soil was removed during the first summer and stored outside during the winter under rain and snow conditions. The construction inside the warehouse started on April of the second year.
1 INTRODUCTION The new construction of a warehouse in the existing REHAU factory of Viechtach in the Bavarian forest north east from Munich was needed to help the growing demand of goods transport and for logistic. Due to the limited space within the factory area and the level difference due to the topography, a highbay warehouse was designed with 20 meters difference between the first level and the third level.. Owing to the nature of this project, a large volume of earth had to be removed, producing over 4,000 m3 of soil. Initially, detailed plans were drawn up to use reinforced concrete retaining walls. It seemed a good idea to use the soil as fill in a geogrid-reinforced earth-wall instead of dumping it. In total a second solution appeared more economical than the first one. REHAU decided to promote its own technology and use its polyester earth-reinforcing geogrids. A 7.20 meter high geosynthetic retaining structure sloping at 85" including a 90" edge corner was constructed 50 cm behind a facing concrete wall inside the warehouse to prevent earth pressure against the face saving the construction of a consequent concrete retaining structure. The concrete plate on top of the geosynthetic structure at the second level, where truck and fork-lift are circulated, was reduced in thickness. The very cramped conditions inside the warehouse and the 90" corner construction were really a big problem particularly for soil transport, installation and compaction during the construction. The
Figure 1. Situation between the piles.
Figure 2. General situation within the factory with the new high-bay warehouse.
34 1
2 SOIL CONDITIONS The first design included the use of the soil, the mechanical characteristics and the soil performance were tested in laboratory during the first year of the construction. To conserve these characteristics during the winter, it was decided to protect the stored soil against the weather conditions to allow a reutilization. The water content was measured between 8% and 10% (October 1998), the optimum water content was 11.5 %. After a hard winter and a lot of snow fall, the insufficient protection of the stored soil against rain and snow particularly during the snow melt in March, increased the water content till 15 % so that a reutilization was made uncertain. To keep the project still attractive, it was necessary to find a soil alternative with the right capacity to be compacted and economic. The second soil alternative was found on the rubble of the Viechtach's quarry ( silty sand) free of charge at about 4 km from the construction site. In that condition, the project was still economic and interesting, the soil characteristics were the followings: 0
e
* e
Density Optimum water content Friction angle Cohesion
p = 1,955 g/cm3 Wept. = 11,5% cp = 31,8" c = 38 kN/m2
Figure 4. Soil compaction in front.
3 STATICAL CALCULATION AND DESIGN The geosynthetic-wall is considered statically to be a solid-wall as a retaining structure. The calculation is done with the two-wedge analysis and allows to determine the forces required for equilibrium, taking into account the geometry of the slope, the geotechnical properties of the soil, the pore water pressure and the surcharge loading. The calculation determines: 0
0
The settlement over the reinforced geosynthetic-wall was calculated with 7 to 8 cm ( about 1% of the total height of the wall). By starting the design, a supplementary provisory load during two weeks was planed on top of the geosynthetic wall to reduce the danger of settlement. Due to very short construction time, this was deleted. To prevent an overburden of stress and cracks due to settlements in the above installed concrete base plate, it was decided to improve the compaction till 103 % Proctor density. The optimum water content of the rubble was measured with 11,5% in laboratory to achieve this compaction grade. In the quarry, a quality-control was set up, only the rubble with the adequate water content was accepted. At the end of April, the water content on site was measured at 8 to 10 %, also sufficient to allow the utilization of the rubble. The whole construction is founded on the existing gneiss rocks with a compression resistance of 1.O to 1.6 MN/m2.
e e
the principal anchor length of the geogrid the spacing between the geogrid the geogrid strength the tie back length of the geogrid.
The geometrical dimensions within the warehouse are the following: e e
e e
e
Length of the geosynthetic-wall 48 meters Height of the cushion foundation 1.80 meters Cushion foundation with three 60 cm geogrids layers, spacing Height of the wall 7.20 meters Slope angle 85 O Reinforced wall with 14 geogrid layers, spacing 50 cm
Figure 5. Cross section of the geogrid-reinforced wall.
Figure 3. Soil transport between the piles.
342
The geogrids were installed horizontally with a tie back anchor on the front of the wall. The cushion foundation was necessary to achieve a good stability of the three pile-foundation during the construction before any supplementary loads (sideward support) could be applied. The statical calculation has taken into account the following geogrid and factors: 0
e
0
PET geogrid with 40 kN/m strength for the cushion foundation PET geogrid with 80 kN/m strength for the retaining structure Factors of safety: 0 Creep (120 years) 1.75 0 Mechanical damage 1.06 0 Biol. and chemical resistance 1.O 0 Calculation safety factor 1.75
Figure 7. Edge corner of the geosynthetic wall.
vertically directly on the steel plate near the edge of the slope, thirdly wooden planks were installed against the formwork, next the geogrid was laid down with a sufficient length over the formwork for the layer thickness and the tie back length. After laying down and compacting the fill soil, the geogrid was wrapped around the face and fixed in the soil layer with special steel pins. The vertical part of the formwork was removed to allow a reutilization by the next layer. Due to the very narrow condition behind the concrete facing panel, it was not possible to extract the horizontal part of the formwork, this part was not removed from the construction. Step by step, the formwork system was raised to achieve the wall. To prevent a loss of fine sand particles, the wrap around was covered with a geotextile. By the 90" edge corner, the geogrid was installed overlapped one upon another. Due to the cramp conditions of installation and the situation between existing concrete piles, it was very important to choose the right machines for soil installation and compaction. Two wheel loaders and 3 BOMAG compaction machines were used. The installation of approximately 200 to 300 m3 of soil per day was possible. By the last three layers, the soil installation in front of the construction was only possible by hand while no machine could drive under the concrete beam of the second level of the warehouse.
The anchor length was calculated with 4.50 m by the first six geogrid layers (layer 1 to 6) and with 5.40 m by the last 8 geogrid layers (layer 7 to 14). The tie-back length was chosen with 1.50 m for all the geogrid layer. A drainage with a geocomposite on the rear of the wall was installed to collect possible ground water from snow melt behind the warehouse, the water is collected in a base drainage collector and transported outside the construction.
4 CONSTRUCTION OF THE WALL The rubbles were transported direct from the quarry to the construction site and deposited on the foot of the construction outside the warehouse at the start of construction and unloaded inside the warehouse on the top of the construction at the end. For this purpose, the concrete facing panels of the warehouse were installed later and provisory ramps were set up. The wall was built 50 cm behind a concrete facing wall, the wrap around was built with the help of a special movable formwork made of steel profiles. A steel plate was first installed horizontally every 2.0 meters on the previously constructed layer, secondly the movable part of the formwork was fixed
Figure 6. Geogrid installation between the piles.
Figure 8. Installation by hand.
343
Figure 11. Concrete plate on the second level above the geogrid reinforced wall.
Figure 9. Situation at the upper layers.
Figure 12. Gap between the concrete facing and the geogrid reinforced wall.
Figure 10. General view of the construction site.
Between the piles, the geogrid was cutted and fixed all around the perimeter of the piles with steel pins. A gravel layer was well set direct by the last geogrid layer to allow a 100% contact of the concrete plate base with the geosynthetic wall.
5 CONCLUSIONS In a total, 3506 m3of soil and 7000 m2 geogrids were installed. The geosynthetic wall was achieved in 3 and !h months between April and July 1999. The weather conditions were optimal to allow the use of the rubbles without problems. On the other side 742 m3concrete and 80 t steel were economized by the construction of a retaining wall and a concrete base plate.
344
The geosynthetic wall is now always visible, a door allows the easy access to the wall between the concrete facing panel and the geosynthetic face. An air circulation enables the visit and control of the construction. Outside the warehouse, a second similar construction helps the construction of access ramps to the second level of the warehouse. In a total, approximately 5 1.OOO,OO € were saved by using a geosynthetic reinforced wall.
REFERENCES Landesgewerbeanstalt Bayern, Gmndbauinstitut. Nachweis fiir die aussere und innere Standsicherheit von Polstenviinden, Benutzer handbuch.
Landmarks in Earfh Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Bench-type wall with flat slabs and steel bars M. Fukuoka Past Pressident ISSMGE, Tokyo, Japan
K. Kondo & R. Ito Kyowa Concrete Industry CO., ltd, Tokyo, Japan
H. Kawahara & K. Misawa Ukasan Livic CO., ltd, Tokyo, Japan
ABSTRACT: This paper describes test results of a bench-type reinforced soil wall with the combination of concrete slabs and steel bars. A steel bar with a piece of steel plate was inserted from outside through a hole in the slab. Backfill soil expanded horizontally, and the bars resisted expansion of the soil. The backside slope of the backfill was 45" in inclination. Panel-type earth pressure gages were placed there, to measure normal and tangential components of earth pressure. Both horizontal and vertical movements of several points in the backfill were measured. Therefore, relative displacement between bars and soil, and stress distribution in the backfill, could be clarified to some extent. Deformation was observed in early spring due to percolation of melting snow, but no sign of rupture was seen. The state of unequal settlement of soil layers in the backfill was recorded by the deformation of bars. 1 INTRODUCTION A n idea of reinforcing cohesive soils with steel bars used for reinforced concrete was applied to a fill, when a 40m-high fill was constructed on an artificially made soft foundation, in 1968. Inclination of the front face was 34". Concrete bearing plates were fixed to steel bars, to resist horizontal expansion of soils and to transmit the earth pressure to the bars. The bars with plates were embedded horizontally in the embankment. The front slope was covered with grasses. The slope is covered with large trees now. A new type of wall with a steeper face was tested in 1980. Benches were made using concrete panels, steel bars, and anchor plates. In 1983, this type of wall was applied to restore a collapsed slope along a road. The wall was 20m in height and 45" in inclination. Steel plates, for the front face, and steel anchor plates were used. Benches were covered with crashed stones to prevent grasses and trees from growing. No sign of corrosion of steel plates has been observed yet. In 1995, a 10-meter-high wall with a steep slope was constructed. This wall was called ASP, which means "anchored slope protection". Inclination of the front slope was 32". A great number of instruments were used, to monitor the behavior of the fill and reinforcement. Load cells were adopted to both ends of bars, to measure tensile force of the bars. The load cells fixed to the side of anchor plates gave smaller values of tension than
345
one fixed to the side of front plates. A pull-out test was performed with bars. As a result of the tests, the smallest value of resistance was 15kN/m. The bars were 22mm in diameter. From this test, the idea of using steel bars without anchor plates for reinforced soil embankments and walls came to mind. What is described above is written in full detail in my book listed as "Earth Pressure").
2 DESIGN OF THE WALL The test wall was 5m in height and 1:0.3 in inclination. The concrete panels protecting the bench walls were 1 x 2 ~ 0I .m in size, and steel bars were 22mm in diameter. Precast L-type reinforced concrete retaining walls were placed at both sides of the embankment. There was no concrete slab at the bottom of the embankment. The back- side of the embankment or backfill was a slope inclined 45", which is a model of a mountain slope. Earth pressure against the concrete panels and bars was estimated using results of past tests, and a coefficient of earth pressure and unit weight of the soil were applied to this test wall. The largest earth pressure was computed as 17x5~0. 1=8.5kN/m2. Pull-out resistance of 6 kN/m was adopted referring to another test conducted under similar conditions. Then, the safety factor for pull-out resistance was 6x5x2/8.5=7 .A plate-bearing test was performed using concrete plate as a dead
Table 1. Soil properties.
load. Settlement was 4mm against a load of 62.7 kN/m2. From this result, settlement by the test wall was estimated to be no problem. Stability of this wall without reinforcement was examined using the test result of the triaxial compression test. As the result, this test wall was safe even without reinforcement. Horizontal displacement of the wall was calculated using past case records and simple assumption. A key point is at the front face 2.5m above the base. The horizontal displacement at this point is usually about half of the settlement of the crown of the fill. Assuming the settlement of the crown 3% of the fill height, the horizontal displacement is 1.5%. Then, the horizontal displacement is 75mm. The second way to estimate is based on the coefficient of linear deformation by the triaxial compression test, and the effect of the back slope of the backfill. This estimate was 95mm. In conclusion, the estimated horizontal displacement was 75-95
(a) Laboratorv test Grain size: 37-20 mm; 6% ,20-2mm;2%, 2-0.01mm;48%, 0.Olmm- 25%, Sandy loam Liquid limit w~=80% Plastic limit wp=56% Maximum dry unit weight: ymdx=12.14kN/m3 Cohesion: c=23.7kNlm2 Optimum moisture content: wopt=38.7% Angle of internal friction: (p=9 ( tan(p=0.158) Modulus of linear deformation: E=2-5MPa (b) Field test Wet unit weight: 14.94-17.63, mean 16.47 kN/m7 Dry unit weight: 1 1.04-12.66, mean 1 1.59 kN/m3 Moisture content: 36.9-46.4. mean 42.1 % Table 2. List of instruments.
mm.
Item
Instruments
Position
Nm-
Earthpressure
Panel-typed
Badcslope
5
&PIE& gage in soil
Baseoffill
3
Earthpressure gage for base Loadcell Dialgage Levelling Plateandrod Plateandrod
Baseofwall
1
Headofbar Faceofml Topofpmel Steps23,4 Steps 1,23,4
10 5
bers
r e ~ ~ u r e e
3 SOIL PROPERTIES The soil properties of fill material are shown in Table 1. The fill material looked uniform in appearance, but it is not uniform in a strict sense.
Bar tension Wall Horizontal Displacement V d d Displacement Horizontal in backfill Vdcal
4 INSTRUMENTATION Many instruments were used to monitor the behavior of the test wall. Figure 1 shows the types of instruments used for measuring and their location. A twometer section of the center of the wall was used as a test section, and both sides served as dummy. Table 2 gives a list of instruments.
2x5
5 4
5 CONSTRUCTION Construction work was conducted as follows. A simple foundation was constructed under the concrete panel of the first step. The panel was erected
Figure 1. A sketch of the wall.
346
vertically on the foundation. The earth pressure gages and the settlement plate were laid on the ground. The fill was placed layer by layer, and the thickness of the layer was 25mm. As the time of construction was early winter, the soil moisture content was high. Heavy compaction machines could not be used. The used compaction machines were,weight=1-5kN. The lower part of the layer was loose compared with the upper part, and bars were embedded in the loose layer. Cone resistance of a cone-penetrometer (top angle 30°, diameter 12.5mm) was 0.3kN/m2 in the upper part and 1.0kN/m2 in the lower part. For the pull-out test, a smooth skin steel bar (M-bar) and a screw skin steel bar (N-bar) of 22mm in diameter were embedded at the dummy parts of the wall. After finishing the second step of the wall, the pull out test was performed. The pullout resistance was 3 kN with the M-bar, and 10 kN with the N-bar, which were far below the expected value (30kN). The M-bars were used for the lower 2 steps, and the N-bars were used for the upper 3 steps. The pull-out resistance test was conducted at the end of the measurement. The pull-out resistance of the M-bar was 7kN,and that of the N-bar was 58 kN.
inclination. This value is very close to the measured value of 1.7 '. 6.2 Horizontal displacement in the backfill Five plates were buried in the backfill. Two plates were placed near the back slope, to measure displacement of the backfill there. It was found that the back slope did not move. Three plates were buried at points about 2.5m from the face panel. These plates moved only 2-3mm. This shows that the backfill between the plates and the back slope elongated very little. The backfill between the front wall and the 3 horizontal displacement plates elongated largely, and the front panels moved forward as shown in Figure 2. Figure 3 shows a change of length between the front panels and the horizontal displacement plates with the passage of time. An increase in length was found only during winter. As for friction between soil and bars, bars near the wall face work to prevent forward movement of soil and the rest part of the bars work as an anchor. 6.3 Settlement of front panels and settlement plates Results of settlement measurement are given in Figure 4. The settlement of the front panels were much larger than those of the settlement plates, which were installed in the backfill. The front panels settled largely at the time of snow water percolation, in December and April. The amount of ground settlement was about 46mm,which was much larger than the reduction of fillheight 13mm.
6 RESULTS OF MEASUREMENT 6.1 Displacement of the front panel The front panel moved both horizontally and vertically. Figure 2 shows the displacement of panels during and after the construction. Large displacement occurred after the construction. The three months from December to February are winter, and the temperature drops to below zero (coldest: - 18" C), and snow falls. This is the cause of large displacement. Displacements of steps 2 and 3 were larger than those of other steps. The base of the front wall moved both horizontally and vertically. Therefore, inclination of the front face did not change very much. The original inclination was 16.7", and the final inclination was 16.7+1.7=18.4" .The predicted value of displacement was 75mm, which is 1.4" in
6.4 Displacement of the fill top Figure 5 shows the displacement of the fill top. Large horizontal displacement was recorded between 1 and 2m, and percentage of elongation between 1 and 2m is 1.5%. There is no elongation between 4 and 7m. The amount of settlement decreased linearly, influenced by settlement of the base.
Figure 3. A change of length between front panels and the horizontal displacement plates with the passage of time. Figure 2. Horizontal displacement of the front panels.
347
6.5 Tension of bars Load cells fixed to the tops of the bars to measure tension of the bars. Figure 6 shows the records of measurement at the end of construction (Nov. 8,1999) and at the time of observation (May 1, 2000). Tensile forces went down suddenly and recovered soon in December, but they went down remarkably and never recovered again after March 20. Figure 7 shows the results of tensile forces of bars for Panel No. 2 after the end of construction. The term of measurement is divided into 5 sections. Section I is early winter. Snowfall is not deep, and frost heaving begins. This increases the tension of bars. Section I1 is deep winter, and the wall is covered with about 30cm of snow. Section I11 is early spring. The temperature rises, frost melts, and soil becomes loose. Then, snow water percolates into soil. This causes sudden drop of tensile forces of bars. Bar tension in terms of the coefficient of horizontal earth pressure at the end of construction was K=O.13, and the apparent angle of internal friction was +=50". K and @ values at the end of construction was 0.05 and 60°, respectively. Figure 8 shows the relationship between horizontal displacement and earth pressure on Panel 2. The earth pressure is deduced from bar tension.
Figure 5. Displacement of the fill top. (Nov. 8, 1999-May 1, 2000)
' uiiit;kN
pi
Panel1-1
.M
4
Figure 7. Tension of bars after the end of construction of Panel 2 . (Nov. 8,1999)
0.29
Numbcr of b a d cell
End of construction Nov,8,199
End of observation
Horizontal displacement(mm)
May,5,2000
Figure 8. Horizontal displacement versus earth pressure for Panel 2 during construction
Figure 6. Tension of bars at the end of construction and observation.
348
The coefficient of linear deformation E=3.3 MPa was obtained by taking the following data.
6.6 Earth pressure gages
Horizontal earth pressure oh=8.5kN/m2, Vertical earth pressure ov=18.47x3.5=57 .6kN/m2, Horizontal displacement A 1=0.034m, Horizontal length of the soil layer 1=6m, Poisson's ratio l/m=1/3. Figure 9 shows a soil layer, which is confined by the upper and lower horizontal planes, and a vertical wall at the end of the bars. Figure (a) is the situation at the end of construction. The earth pressure acting on the vertical wall is equal to the friction along the lower surface minus that along the upper surface. The earth pressure acting on the back of the front panel is balanced by the friction on the perimeter of bars. The maximum tensile force is at the back of the panel. The tensile force decreases gradually in the backward direction, and it becomes zero at the end of the bars. Figure (b) show the situation at a certain time after the end of construction. The panel moves a little bit forward, and the earth pressure against the back of the panel become smaller. As the front part of the soil is extended a little, the frictional force of the soil to the bars acts in the forward direction. The rest of the bars act as anchor. Figure (c) shows the situation of the elapsed time during or after winter. The upper bar projects out of the panel, because no rigid connection exists between the panel and the bar. Tensile force of the bar tip is zero. Friction of the front part of the bar becomes larger and acts in the forward direction, and that of the rest of the bar acts in the backward direction. The panel has 4 bars, but those bars are not in the same situation. As to this wall, frictional force depends on the compactness of soil around the bar. The bars near the front wall were bent down due to settlement of the panels, and this part of the bars moved out from the soft layer to the compacted soil layer. Therefore, the pull-out force of the bars near the front wall became larger than the rest of the bars, which were embedded in the soft layer. This is why some bars projected out of the panels.
Four load cells were installed under the front panel No. 1, to measure load from this panel. Three earth pressure gages were installed on the ground, to measure the pressure from the backfill. Table 3 gives the results of measurement. Despite that the influence of the back slope of the backfill on earth pressure of a retaining wall is great, the condition of the back slope has not been studied well yet. To obtain a case record,5 panel-type earth pressure gages were installed on the back slope, which is 45' in inclination. A part of the records is given in Table 4. Maximum frictional resistance of the back slope is computed by using the result of the triaxial compression test, which is cohesion c=23.7 kN/m2,and angle of international friction $=9' .Then, maximum frictional resistance is about 180kN/m, and the sliding safety factor is more than 3. 6.7 Excavating the test wall The wall was excavated in August 2000. The settlement of bars was measured during the excavation. Figure 10 shows the levels of the bars. The bars were bent down near the front wall. The backfill did not settle uniformly. A cone penetration test was conducted, to check the compactness of soil. Alternation of soft and hard layers, and loose soil about 2m from the front panels, were found by the test. 7 EFFECT OF THE DISTURBED ZONE ON THE SAFETY OF THE WALL ITSELF As a result of the test, it was found that a part of the fill about 2m back from the wall face was disturbed. The effect of disturbance on the safety of the wall itself was examined. Figure 10 shows the wall with the disturbed zone. Weight of (oabd), W1=350 kN/m, (acd), W2=206 kN/m. From Table 3, the earth pressure on (oa) at the end of construction is 384 kN/m. Friction on (ad) is 384-350=34 kN/m. Earth pressure on (ad) is obtained, using the force triangle, 37 kN/m (K=O.18). Frictional resistance on (oa) is 180 kN/m. The safety factor along (oa) is
Figure 10. The test wall with the disturbed zone.
Figure 9. A model showing relationship between earth pressure and friction.
349
(2) The backfill about 2m behind the front face was disturbed by low temperature, and the front face moved forward about 46mm, which is 1% of the wall height of 5m. (3) The maximum bar tension was about 10 kN, which was observed in winter. Screwed face steel bars had high pull-out resistance. (4) Mechanism of reinforcing the bench-type wall with steel bars was partly clarified by this test. (5) It will be necessary to replace a cohesive soil with frost-free materials near the front face, if horizontal movement is to be avoided. (6) The method of estimating deformation should be improved in the future.
Table 3. Loads and earth pressure. Date
load (kN)
Nov. 8, 1999 Jan. 8,2000 June 12,2000
48.82 26.06 20.62
Earth pressure
(kN/m2) 75 88 78
Table 4. Earth pressure on the back slope.
Date Nov. 8 Feb. 28 Jun. 8
Normal component, N, kN/m 70.4 79.5 78.6
Frictional compoment, S, kN/m 46.8 57.5 51.2
tan-'(SIN) S/N
(' )
0.66 0.72 0.75
33 36 33
180+37=4.9. On the other hand, at the end of observation, earth pressure on (oa) is 400 kN/m. On (ad), friction is 400-350=50 kN/m. and earth pressure on (ad) is 33 kN/m (K=0.16). Frictional resistance on (oa) is larger than the cohesive shearing resistance of the undisturbed portion of the fill 23.7x3=7 1.1 kN/m. The safety factor is 7 1.1+33=2.1. Assuming the earth pressure on the front wall is equal to the tension of bars, the amount of tension is calculated from Figure 6 as 14.76 kN/m. Earth pressure on (ef) is about (33+14.76)+2=24 kN/m. As a result of the pull-out test of bars, pull-out resistance is more than 1.6 kN/m. The length of bars that remained in the undisturbed zone is 3m.Total pull-out resistance is 1.6x3x10=48 kN/m. The safety factor is 48+24=2.0. In conclusion, the wall with the disturbed zone is still safe.
8 CONCLUSION (1) A bench-type wall with flat slabs and steel bars was constructed with concrete panels; steel bars, which are used for reinforced concrete and cohesive soil with a high moisture content. It could stand safe in severe conditions of heavy snow and low temperature.
350
REFERENCES Fukuoka, M., Imamura ,Y. 1982; Fabric Retaining Walls, Proceed-ings of the 2nd International Conference on Geotextiles, Las Vegas, Vo1.3,575-580. Fukuoka, M., Imamura ,Y. 1984a; Earth Pressure on Retaining Walls and Buried Pipes, Proceedings on Case Histories in Geotechnical Engineering, Rolla, USA, .355-362. Fukuoka, M, Imamura,Y. 1984b; Research on Retaining Walls During Earthquakes, Proceedings of the 8th World Conference of Earthquake Engineering, San Francisco, USA, Vol. 3, 501-508. Fukuoka, M., 1986; Fabric Retaining Wall with Multiple Anchors, Proceedings of the 3rd International Conference on Geotextiles, Vienna, 435-440. Fukuoka, M., Imamura, Y., Morita, S.,1987 Proceedings of the International Conference on Structural Failure, Singapore, 99. 20-38. Fukuoka, M., Imamura , Y., Nishirnura, J.,1986 Geotextiles and Geomembranes, 4,207-221. Fukuoka, M., Okedoi, K., Ozaki, K., Nakayama, K., Ihara, K.,; Stability of Fiber Retaining Walls,1991. Proceedings of the 2nd International Conference on Recent Advances in Geotechnical Earthquake Engineering, Saint Louis, pp.Vo1. 1, 643-648. Fukuoka, M., 1998; Long-Term Deformation of Reiforced Cohesive Soil Fills and Walls, Proceedings of the 6th Intemationalconference on Geotextiles, Geomembranes, and Related Products, Atlanta, Vo1.2,811-814. Fukuoka, M., Earth Pressure (in Japanese), 1999; Kindai-Tosho Publishing C0.,99-140.
Landmarks in Earth Reinforcement, Ochiai ef al. (eds), 0 2001 Swefs & Zeiflinger, ISBN 90 2651 863 3
Dynamic behavior of multi-anchored reinforced soil wall in large-scale shear box M. Futaki B.R.I. Ministry of Land, Infrastructure and Transport, Japan
N. Aoyama P. W.R.I.Ministry of Land, Infrastructure and Transport, Japan
K. Misawa & T. Konami Okasan Livic Co., Ltd., Japan
M. Sato, T. Tatsui & K. Mikami Techno-sol Co., Ltd., Japan ABSTRACT: Based on a lot of achievement, the excellent stability of reinforced wall during the earthquake has been suggested. On the other hand, it is important to confirm the performance when a soil structure is designed. Actual size vibration test with large scale shear box was done to research the dynamic behavior of the "Multi-Anchored Reinforced Wall(MARW)". The acceleration of each part, the tension of the reinforcement material, the vertical load of the wall and the amount of deformation were measured at MARW of the height 5m during the vibration. According to the test results, following dynamic behavior of MARW were confirmed. l)A reinforced field is unified action. 2)Increase in tension of reinforcement(Tie-bar) during the vibration is big in the connection point of the Tie-bar and the wall. 3)Increase in tension of Tie-bar during the vibration is as big as it is deep. 4)Increase in vertical load of reinforced field is big around the toe. 5)Residual stress and deformation after acceleration over 400gal are extremely small. 1 INTRODUCTION
2 TEST APPARATUS AND TEST METHOD
Many retaining walls which were designed to withstand earth pressures by self-weight were severely damaged in the Hyogoken-Nanbu Earthquake in 1995. On the other hand, damage to retaining walls made of reinforced soil was light, which indicates that walls of this type have a high seismic resistance(The Japanese Geotechnical Society 1996, Kobayashi,K., Tanabe,H. and Boyd,M. 1996, Nishimura,J., Hirai,T., Iwasaki,K. et al. 1996). Although many tests using small specimens to investigate the seismic resistance of reinforced soil retaining walls have been carried out so far (Matsuo,O., Tsutsumi,T. and Saito,Y. 1999), the number of studies using fullscale specimens(Tateyama,M., MuratqO. and Tatsuoka,F. 1990) is not many. Particularly, although reinforced soil retaining walls with multi-layer anchor have been proved to be stable even during actual earthquakes, their stability mechanism has not been fully revealed. In this study, tests using a large-scale shear box and a vibrating table were carried out to investigate earth pressures acting on a reinforced soil wall with multi-layer anchor; stresses imposed on reinforcement materials; contact pressures at the bottom of backfill area; and the deformation behaviors of the fi-ont panel during earthquakes. Furthermore, dynamic responses of a tall reinforced soil retaining wall with multi-layer anchor were analyzed to obtain data to establish a rational design method.
The large-scale shear box in shown Fig.1 can produce simple shear deformation. The shear box is 3.6m wide, 1Om long, 5m high, and it has the multi story structure that piled up 17 square frames of Hsteels of 300mmx200mm. The table is vibrated with hydraulic actuators capable of giving a maximum vibration velocity of 20 kine, total weight of 4.4 MN, and amplitude of rtlOOmm. A separate displacement is to occur between the two adjacent frames smoothly, and many rollers work to achieve it.
Figure 1. Schema of large-scale shear box
35 1
As shown in Fig.2, a full-scale reinforced soil retaining wall specimen measuring 5.0 m high, 3.6 m wide, and 9.0 m long was constructed in the shear box. The 3.5 m long tie bars for the reinforced area were placed 0.6 m from the active failure surface which is derived from the conventional design method, to simplify the preparation of models for dynamic response analysis and numerical simulation. Measurement devices were placed in such a way that the ratio of forces shared by the panel and reinforcement materials and the behaviors related to bed soil stability during earthquakes (such as contact pressures) are correctly evaluated. Measurement items included horizontal and vertical forces at the bottom of the panel; distribution of contact pressures at the bottom of backfill area (measured with earth pressure cell); tension in the tie bars; response acceleration of the panel, reinforced area, backfill area behind the reinforced area, and vibration table; deformation of the shear box and ground surface; and the location of potential active failure surface (measured with strip-type tension meters). In order to investigate the relationship between the behaviors of a reinforced soil retaining wall during earthquakes and the wall height, sweep vibration tests and step vibration tests were carried out for three different backfill heights: 3.0 m, 4.0 m, and 5.0 m. In the step vibration tests, the amplitudes of input waves were increased in stages: sine waves and actual earthquake waves (TAFT EW) were used as the input waves. The maximum response acceleration at the ground surface, obtained under these conditions, was as large as 420 gal. 3 BACKFILL MATER AL AND BACKF LL METHOD Nikko Silica Sand No. 6 was used as the backfill material. With a fine fraction content of 7%, a mean particle diameter of 0.25 mm, and a uniformity coefficient of 2.3, this material is classified as sand with fine fractions {S-F}. The physical and mechanical characteristics of this material are shown in Table- I.
Figure 2. Full-scale reinforced soil retaining wall specimen.
352
After uniformly spreading in layers with a thickness of 50 cm, the backfill material was compacted to a designated density using a small vibratory tamper. The density of the backfill material was controlled by measuring each layer's density by the core cutter method and by measuring each layer's uniformity by the dynamic plate loading tests. After the completion of backfilling, the Swedish sounding test and dynamic penetration test were conducted to check the strength of the backfill material. The average degree of compaction of the backfill material was about 83%, which is close to the lower limit of the compaction control criterion (Dc2 85%) for this type of reinforced soil. In addition, the variation in density is relatively small. Table-1 . Physical and mechanical characteristics.
4 TEST RESULTS 4.1 Frequency characteristics The results of sweep vibration tests are shown in Fig. 3: the left diagram gives the ratio between the response acceleration spectrum at the ground surface of the fi-ont section of the reinforced area and acceleration spectrum of the vibration table, while the right diagram shows the ratio between the response acceleration spectrum at the ground surface of four areas (wall height: 5 m) and the acceleration spectrum of the vibration table. The resonance frequency of a specimen with a wall height of 3.0 m, 4.0 m, and 5.0 m was 7.5 Hz, 5.5 Hz, and 4.5 Hz, respectively. This means that the more higher the wall (therefore the heavier), the smaller the resonance
Figure 5 . Distribution of response acceleration
place surface, because the almost equal response characteristic is shown. Fig.5 shows the distribution of response acceleration of the front panel, reinforced area and the backfill area behind the reinforced area. This figure also indicates that the front panel, reinforced area, and backfill area behind the reinforced area behave like a composite mass, since their response accelerations have similar profile. However, the shape and magnitude of the curve for response acceleration change depending on the maximum acceleration and shape of input ground motion.
Figure 3. Results of sweep vibration tests
frequency. The responses of ground surfaces (within reinforced area), the front panel, and the backfill area behind the reinforced area to the vibration of the table are almost the same, which indicates that the reinforced area and the backfill material behind it behaves like a composite mass.
4.3 Actual tension in tie bars
4.2 Response acceleration characteristics Fig.4 shows the response acceleration of the panel, reinforced area, and backfill area behind the reinforced area during vibration. Although there are some variation depending on the magnitude of the input acceleration, the accelerations at the ground surface level of the front panel, reinforced area, and backfill area behind the reinforced area have a phase difference to the vibration table's acceleration of between n/4 to 2/3n. However, the behavior in which reinforced area and backfill area were almost united is proven on response acceleration in each these
Strain gauges were placed on the surface of steel tie bars to measure the tension (corresponding to earth pressures acting on the panel surface) imposed on them and verify the magnitude and distributions of forces acting on the front panel which occur during vibration. Fig.6 shows the relationship of the depth with the tensile forces in tie bars during vibration and with the maximum tension(increment) at the time of the maximum average acceleration. Since the tension given in this figure were measured near the panel, they are considered to be close to the earth pressures. As shown in this figure, the tensile and
Figure 4. Response acceleration during vibration.
Figure 6 . Maximum increment tension of tie bars
353
compressive forces in tie bars during vibration are caused by active and passive earth pressures, with tension dominant, and the directions of the forces are the same regardless of the depths. In addition, the maximum increment of the tension in tie bars is nearly proportional to the depth, but smaller than the values calculated using the current design method. In addition, as shown in Fig.7, the increment of tension in tie bars during vibration is small near the anchor plate due to the surface friction of the bars, but large near the panel.
4.5 Deformation behavior of the panel The left diagram of Fig.9 shows the distribution of panel displacement at the time of maximum average acceleration, when the input ground motion has a frequency of 1.5 Hz and an acceleration of 420 gal, while the right diagram compares the maximum displacement amplitude for input ground motions with various accelerations. The displacement distribution of the 5-m height during vibration for each input ground motion was in the primary mode. The maximum displacement at the top of the front panel was about 50 mm for a ground acceleration of about 400 gal. In addition, the residual displacement after vibration was relatively small.
-
D is p lacem ent ( m m ) Back
50
M ax. disp lacement(mm)
F r o I? t
0
Back
-50
~
Front
50
-50
5 J h
E3 0
1
2
c
3
Tn
Distance from panel ( m )
G 2
Figure 7 . Distribution of increment tension of tie bars.
1
0
4.4 Contact pressures at the bottom of the b a c v l l In order to obtain information for checking the bearing capacity during earthquakes, the magnitude and distributions of the contact pressures at the bottom of both the front panel and backfill area were investigated. Fig.8 shows the distributions of the contact pressures at the bottom of the backfill area. As shown in this figure, the contact pressures working at the bottom of the backfill area during earthquakes increase and decrease from those during stationary state depending on the direction of vibration. The increase in contact pressures tends to be larger at the location immediately behind the panel than at the locations away from the panel. Contact pressures after vibration were almost the same as those during stationary conditions, and residual contact pressures due to vibration were negligible.
Figure 9. Distribution of panel displacement
4.6 Active failure surface during earthquakes In order to identify the location of active failure surface for both during vibration and during stationary conditions, strip type tensile meters were embedded in the reinforced area. Fig. 10 shows the strains in the strips. As shown in this figure, locations for the peak
Figure 8. Distribution of contact pressure.
Figure 10. Strain in the strip-type tension meters.
354
values are almost the same. This means that the active failure surface does not move backward and the soil behind the front panel behaves like a composite mass because of the effect of the reinforcement materia1s.
Table-2. Input parameters Material Embankment Wall(Pane1s) Joint filler fee Anchor rods
4.7 Ground deformation after vibration
Unit weight (kN/m')
Modulus of deforma-
15
26 5 79
(m/m')
Poisson ratio 0 45 0 20 0.10 0.29
12 1-42 2 2,100 5 2,100
Fig. 1 1 shows the locations of vibration-induced cracks in the ground behind the front panel and in the ground surface. The locations of cracks in the ground behind the front panel were identified when the role paper placed in the ground was removed. As shown in the figure, cracks are observed only in the ground behind the reinforced area. It should be noted that the cracks appear near and behind the Coulomb's active failure surface.
Figure 12. Finite element mesh.
5.2 Results of simulation for the vibration test
Figure 1 I . Cracks in the behind the panel and ground surface.
5 DYNAMIC ANALYSIS 5.1 Method of analysis The analysis is called "Super FLUSH" was used as the two-dimensional plane strain condition. Finite element mesh is shown in Fig. 12. This model is used 1197 joint nodes and 1120 elements. A boundary condition set up free at the bottom part, fixed at the wall, and free only the horizontal direction in shear box boundary. Moreover, backfill materials, panel materials, side-glance area materials, and anchor plates were made a plane element. Then equivalent line shape wasn't taken into consideration, and it was made elasticity. And, the mass of shear box are ignored because it is small in comparison with the backfill. Input parameters are shown in Table-2. Modulus of the backfill material is considered as confining pressure dependence using the result of soil test.
The depth distribution of the response acceleration at the wall which are measurement value by the shaking table test and the dynamic analytic result is shown in Fig.13. This figure shows that it can confirm that it almost corresponds as for the analytic result and the experiment value, when the input acceleration of the shaking table is small. When input acceleration is big (sin-wave 400gal), the deformation mode of the wall becomes the multiple modes, and some difference is admitted in the experiment result and the analytic result about the response acceleration of the wall. But, as for the average response acceleration in the depth direction, it is understood that it has good adjustment. The depth distribution of wall displacement got from vibration experiment and dynamic analysis is shown in Fig. 14. When input acceleration is small, good correlation is presumed an experiment result and analytic result in the same way as the depth distribution of the response acceleration. But, the difference around the ground surface grows big when input acceleration grows big. And it is understood that an analytic re-
I
i I
Y
i
l
I
,
'
Figure 13. Distribution of response acceleration of panel
355
Figure 14. Distribution of response displacement of panel
sult is beyond the measurement value. The basic performance of the analytic program was verified from these things as a result of comparing measurement value with analytic value in the deformation nature of the wall and the response acceleration. Therefore, it was judged that to simulate the behavior in of reinforced soil wall using the dynamic analysis was precisely possible. 5.3 Results of dynamic analysis for the wall height I Sm-class
maximum 20kN/m2 appears in the neighborhood of the ground surface or the bottom part in tie bars. And, contact pressure is changing about 20% to customary pressure at the foundation under the panels. The result of arranging the predictive value in each of large and medium-scale ground vibration level is shown in Table-3. Earth pressure against wall in the large-scale vibration level is the degree that it is the same as in the medium-scale, as an absolute value. However, sufficient consideration seems to be necessary, when the appearance depth reaches near the ground level, and when that the withdrawal resistance of the anchor plate is the small range is considered. It is considered that the stability of the wall height 15m-class can be verified fully as a result examined about the stability during the earthquake of the multi-anchored reinforced soil wall by the dynamic analysis. And, it can be expected that rational design method of the multi-anchored reinforced soil wall becomes possible these result by the reflection of the seismic designing method. Table-3. Results of analysis for 15m-height class Earthquake Level
This object is valuation of the resistance to earthquake of the multi-anchored reinforced soil wall, and the dynamic analysis are done by using the verified analytic program about the wall height 15mclass. Considering large and medium-scale ground vibration level, the shape of seismic wave adopted the shaking wave which compressed maximum acceleration amplitude in the place in 150ga1, 400gal (KOBE N-S direction). An example of the analytic result is shown in Fig. 15. The various modes exist in the response acceleration distribution toward input acceleration by earthquake. But, as for the acceleration inputted from the base, it is understood that it doesn’t amplify in the backfill and around ground surface too much. Moreover, the deformation of the crest of panels is about 20cm during the large-scale ground vibration level, and this value is about equivalent on the displacement angle of 1.5/100. At this time, it is understood that tension of about the
Figure 15. An example of the analytic results.
356
Mediumscale Largescale
R~~~~~~~act,
Di;tgtF
magnification
(half-amp,)
13
09
04 -
100 15
100
Tension of tie bars (earth pressure) 2OkN/m’ (bottom) 20kN/mL (bottom,sur-
Contact pressure (ratio) 15%
20%
C
6 CONCLUSION The following findings were obtained from the vibration tests using full-scale specimens of multianchored reinforced soil wall: 1) The panel, reinforced area behind the panel, and the backfill area behind the reinforced area behave like a composite mass. The distributions of the response acceleration, however, vary depending on the acceleration and shape of the input waves. 2) Although the increments of the tensile forces in tie bars are generally proportional to the depth,
about the stability during the earthquake of the multi-anchored reinforced soil wall by the dynamic analysis.
their values are smaller than those derived from the current calculation method. 3) Contact pressures at the bottom of the backfill area during vibration tend to be large at the location immediately behind the front panel. Their values after vibration, however, are almost the same as those before vibration: residual contact pressures are negligible. 4) The displacement at the top of the fiont panel during vibration is about 50 mm for a ground acceleration of 400 gal. The residual displacement after vibration is relatively small. 5) The active failure surface does not move backward during earthquakes. This indicates that the reinforced area forms a composite mass. 6 ) No cracks in the reinforced area were observed during visual inspection after vibration. Cracks were found only in the area behind the reinforced area. 7) It is considered that the stability of the wall height 15m-class can be verified fully as a result examined
REFERENCES The Japanese Geotechnical Society 1996 Report of damagefor the Great Hanshin Awaji earthquake, pp 327-341 Kobayashi,K ,Tanabe,H and Boyd,M 1996 The performance of reinforced earth structures in the vicinity of Kobe during the Great Hanshin Earthquake Proc of the International symposrum on earth reinforcement(IS Kyirshu’96),pp 395-400 Nishimura,J , Hirai,T , Iwasaki,K et a1 1996 Earthquake resistance of geogrid-reinforced soil walls based on study conducted following the southern Hyogo earthquake Proc of the International symposium on earth reinforcement(IS Kyltshu’96), pp 439-444 Matsuo,O , Tsutsumi,T and Saito,Y 1999 Shaking table tests and stability analysis on seismic performance of geosynthetic-reinforced soil retaining walls Civil engineering journal, Vol41, No 1 p 32-37
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Stability test of the multi-anchored reinforced soil wall constructed on soft ground H. Hashimoto, N. Aoyama, H. Miyatake & M. Hirasawa Public Works Research Institute, Ministry of Land, Infrastructure and Transport, Tsukuba, Japan
ABSTRACT: Because the degree of ground settlement that will not destabilize a reinforced soil wall with such following capability is not understood, it has been designed based on a bearing capacity safety factor identical to that for concrete retaining walls. Multi-Anchored Reinforced Soil walls are normally designed with a safety of 3, but large-scale laboratory testing has been performed in order to propose a design method that can lower the bearing capacity safety factor for this design method.
1 INTRODUCTION Because a reinforced soil wall is a flexible structure, it can stably follow a moderate degree of ground set tlement. because the degree of ground settlemen t that will not destabilize a reinforced soil wall wit11 such following capability is not understood, it h a s been designed based on a bearing capacity safetJi factor identical to that for concrete walls. The cur rent designing way of this reinforced soil wall is de scribed at the first of this paper. The bearing capac ity safety factor has been applied to 3 safety factor:s the current way, however, large-scale laborator! r' testing has been performed in order to propose a de sign method that can lower the bearing strengtki safety factor for this design method.
2 STABLE MECHANISM OF THE REINFORCED SOIL WALL
In the equation (Fig. l), if the cohesion is assumed to be constant, the pull-out resistance of anchor-plate is shown as a simple equation of confining pressure of the plate circumference that depends on the ground depth of anchor-plate and the depth dependence is shown as the earth pressure on the wall. The stability of the reinforced soil wall is univocally determined by the relationship between earth pressure on the wall and tensile strength of anchor-plate in the banking of reinforcing area with appropriate internal friction angle cp of the fill material, when a wall becomes higher, not only the earth pressure, but also the pull-out resistance increase. It is important to control the parameters c and cp suitably under this reinforcement mechanism and it is necessary that the stability of the Multi-anchored reinforced soil wall is considered in internal stability and external stability (Fig. 2).
Both of the facing walls are connected to the anchor plate by the tie-bar, and the earth pressure on tht: wall is supported by the pull out resistance of an chor-plate, in this way, the stability is kept. The: pullout resistance of anchor-plate is based on tht: theory of bearing capacity for the horizontal fora : inside of the banking; it is shown according to tht: following equation.
QpIt:
Ultimate pull-out resistance of anchor-plate. c : The cohesion of fill material. qp : The confining pressure of around the anchor-plate. N,,Nq : Coefficient of bearing capacity, for the pull-out resistance of anchor-plate.
Figure 1. Examination of internal stability.
359
4 EXPERIMENT 4.1 Outline The test performs by constructing a clay ground with a depth of 1.5 m inside a large soil tank with a height of 4.5 my width of 10.0 my and depth of 4.0 m installed at the Public Works Research Institute. An anchor type reinforced soil wall with a height of 3.0 m was constructed on top of this ground (Fig3). Kanto loam used to form the foundation ground and sandy soil uses as the banking material. The foundation round fills loosely (wet density: pt = 1.1 g/cm ) so that the embankment and the loading would cause substantial settlement. A loading device capable of applying an overburden load q of 200 kN/m2 that is installed on top of the embankment used to perform loading in steps equivalent to the load of 1.0 m of banking (q = 17 kN/m2) up to the load equivalent of 8.0 m of banking (q = 136 kN/m2) (Fig 4 ) .
F
Figure 2. Examination of external stability.
In the internal stability, the reinforcing area constituted by block wall and anchor-plate group examines force for keeping the stability as reinforced soil wall and balance of resistance force. In the external stability, sliding of reinforcing area, bearing capacity of foundation ground, and circular slip failure including reinforced area. 3 BASIC POLICY OF THE DESIGN The Multi-anchored reinforced soil wall is examined to satisfy a two-point demand of safety factor.
(1) The examination for the internal stability 1) The examination for the rupture of tie-bar 2) The examination on the pull-out resistance of anchor-plate (2) The examination for the external stability. 1 ) The examination for the stability of the reinforced soil wall structure. 2) The examination for the whole stability including the reinforced soil wall structure. In the design, it is necessary to consider sufficiently internal stability and external stability. The situation of the field is suited, and it must be considered in order to be also excellent in work- ability and cconomical efficiency. Multi-anchored reinforced soil wall supports the earth pressure that affects block wall by the pull-out resistance of anchor-plate, and it is the method that constructs the banking with stabilized vertical wall. Moreover, banking region held by block wall and anchor-plate group seems to become the reinforced soil structure, because the deformation is restricted. There are two phases examination, internal stability and external stability, in the design of this method.
360
Figure 3. Outline of the experiment (cross sectional view)
Figure 4 Test sol1 tank (front view>
4-2 The prior In configuration position i= 1, the constructed length is insufficient. Allowable tensile stress of tie-bar and Allowable resistance force of anchor-plate have en-
Table 1. Test soil tank (Front view)
subgrade reaction increases accordingly as the loading increased.
2) Lateral displacement of the wall surface. Figure 7 shows the lateral displacement of the wall surface. Deformation and a large lateral is placement of the wall surface occurred as it followed the settlement of the foundation ground and the overburden load. But it is not deformed to the degree that its
sured the sufficient stability. It was judged that the internal stability had been stabilized (Fig. 3 and Table 1).
4.3 Design of the subgrade reaction The distribution of the subgrade reaction, when the loading is small, the part of wall base is bigger than the banking part, and it almost uniformly increases with the increase in the loading. q=68kN/m2 loading stage, banking central increased excellently. The sub grade reaction compares measured value with designed value. The wall part of measured value is bigger a little than the designed value, and the banking part of measured value is a little smaller than the designed value (Fig. 5). 200 180
-<
160
N
140
5
: 120 * z2 100 0
I -
{
80
Figure 6. Distribution of settlement and subgrade reaction of the foundation ground
t l
n 7 60 v)
40
20 0 0
17 34 51
(H=3rn)
68
85 102 119 136 153
LoadingWeight q(kN/m2)
Figure 5 . Designed value and measured value of the subgrade reaction.
4.4 Results of the test
1) Settlement and subgrade reaction of the foundation ground. Figure 6 shows the settlement and subgrade reaction of the foundation ground. The loading causes a relatively large settlement of the foundation ground. But in the reinforced soil wall, the foundation of the wall surface work do not settle very much and the area behind the reinforced section settle unequally generating the maximum deformation. The
0 50 100 150 200 250 300 Quantity of lateral displacement of wall surface(mm)
Figure 7. Distribution of lateral displacement of the wall surface.
361
functions are obstructed by the falling of the wall surface panel and the breakage of the reinforcement. 3) Tie-bar tensile strength and horizon earth pressure. H=3.0m the banking end, earth pressure before the stage load testing agrees approximately with the design calculation value in the each loading stage at measurement position of H=0.5m and 2.5m wall heights, however, the H=l.5m wall height shows small value from them (Fi 8). but, Large change do not appear in q=68kN/m loading stage of which subsidence of the ground and wall surface horizontal displacement are greatly observed. In the experimental result, tie-bar tension strength and wall surface earth pressure become a distribution that differs from the design theory in the each loading stage. However, the excessive value is not shown. Subsidence deformation for the ground and wall surface deformation do not show the behavior which internal stability causes the adverse effect.
5
wall surface and the reinforced range should not settle unequally. It is necessary to conduct further studies of the foundation treatment method. 2. Even though the reinforced range settles and the wall surface is deformed, serious damage causing the failure of the structure does not occur. 3. The safety factor of the ground bearing strength can be reduced below the conventional level as long as it is at a level that can allow a certain degree of settlement of the reinforced soil wall and lateral displacement of the wall. This conclusion is based on the fact that although the quantity of settlement following the loading test is a maximum of 27cm at the wall surface foundation and 37cm at the center of the reinforced range. The structure continues to function properly. But further study is necessary to determine standard values from the perspective of effects on deformation of the reinforced range when it is applied at the in-situ level (Fig. 9).
Figure 9 The relationship of structure between internal stability and external stability
REFERENCES Public Works Research Center 1998 Multi-anchored soil retaining wall method design and constructing manual ver 2 10-12, 50-100 Japan road association 1999 Road construction The retaining wall guideline 132-149 M Hirasawa, N Aoyama, H Miyatake, H Hashimoto 2000 The stable experiment of the reinforced soil wall constructed on soft ground (No 1) 55th, year scientific lecture association lecture outline collection third part, Japan SOCof Civil Engineers 578-579 H Hashimoto, N Aoyama, H Miyatake, M Hirasawa 2000 The stable experiment of the reinforced soil wall constructed on soft ground (No 2) 55th, year scientific lecture association lecture outline collection third part, Japan SOCof Civil Engineers 580-58 1
Figure 8. Tie-bar tensile strength and horizon earth pressure
5 CONCLUSIONS 1. Under the effects of the settlement of the ground, the reinforced range behaves as an integrated mass and the wall surface is deformed. The reinforced range settles more than the wall surface foundation. It is assumed that while the foundation work execute at the wall surface foundation restricted settlement, the reinforced range is untreated. For this reason the
362
Landmarks in Earth Reinforcemenf, Ochiai et al. (eds), 0 2001 Swets & Zeiflinger, ISBN 90 2651 863 3
Dimension analysis on reinforced soil walls by finite element method Huang Guangjun, Yu Xijian & Yang Canwen Geotechnical Engineering Division, China Academy of Railway Sciences, Beijing, 100081, PRC
ABSTRACT: There are lots of factors that influence the properties of reinforced soil. But in most of former studies, only partial factors were considered and these factors were analyzed separately. In this paper, the relationship between the main factors is revealed by dimension analysis, and they are summarized into three groups of dimensionless parameter: relative rigidity of reinforced soil, relative strength of plain soil, and relative load. When the influence of foundation condition and compaction-induced stresses are neglected, different reinforced soil walls have similar stress field and displacement field only if their dimensionless parameters are equal.
3. There is no slippage between the reinforcement and the soil; 4. Uniform spacing between any two neighboring layers of reinforcement, uniform length of reinforcement, as shown in Figure (1); 5. Only the gravity of soil is considered, all other load that might act on the reinforced soil wall is neglected here; 6. Plain strain condition; 7. The influence of foundation and compactioninduced stresses are all neglected. In the following dimension analysis, soil element, facing element, and reinforcement element are studied respectively.
1 INTRODUCTION There are lots of factors that influence the properties of reinforced soil. These factors mainly include properties of soil, reinforcement, facing system, and the load acting on the reinforced soil, etc. (Huang Guangjun, 1999). But in most of former studies, only partial factors were considered and these factors usually were analyzed separately. (Wong, K.S. and Broms, B.B. 1994; Matichard, Y. et al, 1994; Juran, I. et al, 1989) Since these factors interact each other, it is more reasonable that they be considered in a whole. In this paper, the relationship between the main factors is revealed by dimension analysis, and they are summarized into three groups of dimensionless parameter: relative rigidity of reinforced soil, relative strength of plain soil, and relative load. When the influence of foundation condition and compaction-induced stresses are neglected, different reinforced soil walls have similar stress field and displacement field only if their dimensionless parameters are equal. This conclusion is useful in model tests and engineering practice.
2.1 For soil element For one soil element, the relationship between element node displacement and node force can be expressed as below:
For rectangular element, the stiffness matrix is
2 DIMENSION ANALYSIS OF REINFORCED SOIL WALL In order to simpliiji the analysis course, it is assumed that: 1. Soil is perfect elastic-plastic material, and it comply the yield criterion of Mohr-Coulomb; 2. Both of the facing system and the reinforcement are still in elastic state;
3 63
Let
kb,=
1-2p L1-p x 7 , +-2(1-ppCJ
then Equation (1 0) can be turned into:
I-2p
P 1-P
x 7, XI-A
----'I,&+-
~~
h -the thickness of soil element Es - Young's modulus of soil p - Poisson's ratio 7 , 5 - local coordinate of vertex in the element
P)'
= CFxl
Fxl
4
1 4
1
E l
E1
=-abhy[O 1 0 1 0 1 0 1B'
F x l clr
(5)
Let (F}: =[0 1 0 1 0 1 0 1]T, and have both sides of Equation ( I ) be divided by E, Hh , then
1 1 E,A [K];,,(8)"= --{F}e E,yHh 1, E,,Hh
where E, - Young's modulus of reinforcement of reinforcement element, in Figure (1) , lg = 2a t, - equivalent thickness of reinforcement,
1,
- length
ng - number of reinforcement element in one layer, n, - number of reinforcement layer in the reinforced soil wall
For Mohr-Coulomb, the soil strength is expressed as
z= C + CT tan4
(8)
This equation can be turned into a dimensionless one, that is
2.2 For reinforcement element Figure 1. A simplified FEM mesh.
Reinforcement is treated as bar element. Since it is assumed that no slippage between the reinforcement and the soil, interface element does not need here. For one reinforcement element, the relationship between element node displacement and node force can be expressed as below:
2.3 For facing system element
Facing system is treated as beam element here. For one beam element, the relationship between element node displacement and node force can be expressed as below:
where where
3 64
So Equation (19) is turned into
E, A 1, 0
-
12E,I
Symmetry
1;
6E,I
4E,I
1;
I,
0
o z . . F. - A.
- -
0
E, A
.-
1,
12I<,I
0
6f<, I
I,
0 If 1; 6E,I 2E,I - - 0 1; 1,
0
-
12E, I If
6E,I I;
--
4E , I
__
1,
Syniinetry
--I:’,i,h 1,
o /6t’ /
412
121, -1 0
12 0
2.4 For the whole structure The stiffness of the whole structure is assembled with stiffness of all elements. The diniensionless balance equation of the whole structure is
1
6r2
t’
0 - 1 - 1 0 1; 121, 61’ 21’ 0 1 L 0 121, 12
I’
1 . 1; 6i2 4t2 -/ L 121, 12
If error of numeric calculation is neglected,
5, v,n, ,n,
Ef -Young’s modulus of facing system l f - length of facing system element, in Figure (S), I f = 2b tf -thickness of facing system,
(6)” = [U,
v,
{F)“ = [h
8,
UJ
FNJ
MI
FQl
FQ/
I”
M,
Equation (14)can be turned into
S E , Hh
[ K ] ; {6)”
=
S ~
E,sHh
{F
Symmetfy
15 J
/
12,H - 1 0
E,Hh
is named relative rigidity of reinforcement
y
/
1W’ 0
E At,
E,H’h 1
is usually small, the
tS
In this paper, E’ is named relaESt, E ,t , tive rigidity of facing system and denoted bv E j ; mined by
E,Hh -t;
=.
t
stiffness of facing system is approximately deter-
F,,
0 Q
Since the change range of
>@
1
by have no influence on the stiffness of the
whole structure [K],,. Furthermore, ns can be determined by H and ts. So [K],, is determined by the following dimensionless parameters named relative rigidity of reinforced soil:
8Jr
VJ
2
5’ E,Hh -
L, is named relative length of and denoted by E,r ; H reinforcement.
E,Hh
Let H = n,t, = n / I , - , tem element, then
M,
In Equation (21), (F}” is mainly determined by
E,H’h,
YH The relative strength of plain the relative load -. E,
nj is the number facing sys-
c # . So it is concluded that:
soil is -,
YH
for two different reinforced soil walls, if their dimensionless parameters (relative rigidity of rein365
forced soil, relative load and relative strength of plain soil) are equal respectively, their displacement field and stress field must be similar respectively. In another way, if
Figure 2. FEM mesh
that: for different structure, the parameters of soil, reinforcement and facing system are different, but the dimensionless parameters are same. The result is shown in Figure (3) Figure (5). It is obvious that the relative tension in reinforcement, the relative horizontal pressure acting on the back of facing sys-
-
Table 1. Main parameters that are used in FEM c33.01
No.
Dimen-
E,
0.2
0.2
0.2
0.2
0.2
Efr
37.5
37.5
37.5
37.5
37.5
0.8
0.8
0.8
0.8
L,/H
sionless C/(yH)
4
ters
(”)
(rH)’Es (1 0-7
where T, is tension in reinforcement. Though the conclusion is deduced from a simplified model (Figure I), it is also valid in other more complicated condition (Figure 2). This conclusion will be verified by finite element method in the following part. 3 VERIFY BY FINITE ELEMENT METHOD
30
30
30
4.85
4.85
P
0.35
0.35
0.35
0.35
0.35
C (KPa)
10.0
4.6
6.5
10.0
5.0
19.4
18.0
18.0
19.4
19.4
5.0
2.5
3.5
5.0
2.5
ts (m)
1.0
0.5
0.7
1.0
0.5
E,(MPa)
20.0
9.3
13.0
20.0
10.0
L, (m)
4.0
2.0
2.8
4.0
2.0
1.0
0.4
0.4
0.5
0.5
2000
H (m)
tf (m)
system E I(MPa)
366
30
0.103 0.103
4.85
E,(MPa) Facing
30
0.103 0.103
4.85
Reinfort, cement (mm)
In order to verify the conclusion, a group of reinforced soil walls are analyzed here. The FEM net is shown in Figure (2). It must be noted that the facing system is treated as elastic solid element, this is different from that in Figure (I). It is shown in Table 1
0.8
0.103
4.85
y(KN/m’)
Soil
C3-3.02 C3-3.03 C3-3.04 C3-3.05
4000
2320 4550
8000
0.25
0.125 0.175
0.25 0.125
3000
1400 1950
3000
1500
Figure 5. Relative horizontal displacement of facing system.
tem, and the relative horizontal displacement of facing system are accordant for different reinforced soil walls. It can be deduced that the stress field and displacement field of different walls are similar respectively. 4 PARAMETER STUDY In those dimensionless parameters, relative rigidity Egt, of reinforcement -, relative length of rein-
E\4 L, , and relative rigidity of facing system forcement -
H
4t ,
have the most remarkable influence on the
properties of reinforced soil walls. In Figure (6), it is shown that the relative tension in reinforcement increases with the relative rigidity of reinforcement. rH and p strongly affect the displacement Though E, field, they have little influence on the stress field, as shown in Figure (7).
Figure 3. Relative tension in reinforcement.
Figure 6. Influence of reinforcement relative rigidity on relative tension in reinforcement,
Figure 4.Relative horizontal pressure on back of facing systern.
367
ters are equal. This conclusion is useh1 in model tests and engineering practice. In those dimensionless parameters, relative rigidity of reinforcement, relative length of reinforcement, and relative rigidity of facing system have the most remarkable influence on the properties of reinforced soil walls.
YH and p strongly affect the displaceThough E, ment field, they have little influence on the stress field. Figure 7. Influence of relative load and p on relative tension in reinforcement.
5 CONCLUSIONS In this paper, the relationship between the main factors that influence the properties of reinforced soil walls is revealed by dimension analysis, and they are summarized into three groups of dimensionless parameter: relative rigidity of reinforced soil, relative strength of plain soil, and relative load. If the influence of foundation condition and compactioninduced stresses can be neglected, different reinforced soil walls have similar stress field and displacement field only if their dimensionless parame-
368
REFERENCES Huang Guangjun, 1999, Study on the property and design method of reinforced soil, dissertation presented to China Academy of Railway Sciences, in Beijing, P.R.China, in Partial fulfillment of the requirements for the degree of Ph.D. Wong,K.S. and Broms,B.B., 1994, Failure Modes at Model Test of a Geotextile Reinforced Wall, Geotextiles and Geomembrances, Vol. 13, pp475-493. Matichard,Y. et al,l994, Behaviour of a Geotextile Reinforced Earthwork under Surface Loading, Recent Case Histories of Permanent Geosynthetic-reinforced Soil retaining Walls, Tatsuoka & Leshchinsky (eds), 1994 Balkema,Rotterdam, pp.1 17-130. Juran,I. et a1 ,1989, Laboratory Model Study on Geosynthetic Reinforced Soil Retaining Walls, Journal of Geotechnical engineering, ASCE, Vol. 1 15, N0.7, July 1989, pp905-927.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets 6: Zeitlinger, ISBN 90 2651 863 3
Analyses of a near-fault geosynthetic-reinforced modular block wall damaged during the 1999 Chi-Chi earthquake Ching-Chuan Huang Professor, Department of Civil Engineering, Cheng Kung University, Taiwan
Li-Hwei Chou & Yu-Hung Chen Research students, ditto.
ABSTRACT: A near-fault geosynthetic-reinforced modular block wall (RMBW) damaged during the 1999 Chi-Chi earthquake (M~=7.3)is analyzed using a new ‘three-wedge’ method. The new method calculates the stability of the facing and soil wedges interactively. As a result, a safety margin against shear failure on the potential failure surface is the only output. Therefore, irrelevant safety criteria can be eliminated. The new method gives not only an accurate estimation on the contribution of block facing to the seismic stability of RMBW’s but also accurate failure patterns under seismic conditions. A new method for calculating seismic displacement of RMBW’s based on the ‘three-wedge’ mechanism and the ‘displacement diagram’ is developed. It is demonstrated that this method gives more realistic displacements of RMBW’s than any other existing pseudo-static methods.
1 INTRODUCTION
200mm long) which were used as shear keys between the stacked blocks. (2) Failures at the junction of the geogrids were found behind the collapsed facing. The locations of the junction failures coincided with those for
The Chi-Chi earthquake (M~=7.3) occurred at 1:47AM, 1999 in central Taiwan. Rupture of the Chelungpu fault (about 100km-long and 30km-wide) was reported to be the cause of this earthquake. Intensive damage to the near-fault highway facilities, such as: bridges, embankments, and soil retaining structures were reported (Huang, 2000a). In the severely shaken area, no failure of geosyntheticreinforced soil retaining walls was observed except an inadequately designed and constructed wraparound reinforced slope and two dry stacked modular block reinforced walls. The present study investigated one of the failure sites of the modular block reinforced wall located about 4km east from the surface scarp of the ruptured fault (Site 1 in Figure 1). Geosynthetic-reinforced modular block walls have been increasingly used in Taiwan. However, no authorized design guideline for modular block reinforced walls is available. The investigated RMBW is currently used in a highway-widening project for a secondary earth retaining purpose. At Site 1, a 3.2m-high RMBW collapsed, see Figure 2. Two typical sections of the collapsed walls are shown in Figures 3(a) and 3(b). See Huang (2000a, 2000b) for other RMBW sites. Post-failure investigations at site 1 showed that: (1) At the section where the wall was on the verge of collapse (Figure 2, Figure 3b), the wall buckled and /or bulged between the lower 1/3 and 1/2 of the wall height and caused large openings between the blocks. These openings in turn disabled the function of the FRP rods ( 1 5 m diameter and
Figure 1. Location of the investigated site.
369
Figure 2. The geosynthetic-reinforced modular block wall damaged during the 1995 Chi-Chi earthquake.
the (FRP) rods used to connect the blocks. Ultimate shear strength of FRP rods is 48000 kPa according to ASTM D4475. Most of the rods found at the site were intact. The knitted polyester net, or PET geogrid, had ultimate tensile strength of 75kN/m and junction strength of 0.3kN/junction. (3) In-lab direct shear tests were performed on the soil samples remolded under the in-situ density and natural water content. Strength parameters obtained from 60mm-diameter, 35mm-thick samples are c=O, cp =30.4" , those from 200mm-diameter, 100mm-thick samples are c=O, cp =29.2" .
A preliminary investigation on the failure mechanism of this wall was performed using the seismic design guideline proposed by NCMA (1998). Safety factors against various failure modes obtained from the analysis are summarized in Table 1. In this calculation, the experimental value of connecting force at block-backfill interface, T,,, and a possible range of the block-block friction angle, y , were used (see Table 3). Table 1 indicates that: (1) For kh=O , namely, static condition, F, for toppling and connections are 0.62 and 0.65-0.94, respectively. However, none of these failure modes were observed in the field. Over-estimation on the lateral earth pressure may account for this disparity. (2) For kh=0.22 condition (= 0.5.ama,/g, see Figure l l a ) , in addition to the under-estimated F, for toppling and connections of reinforcement, as discussed in (I), a critical value of F,=0.99 was obtained for pull-out failure. However, pull-out of the reinforcement has not been observed (see Figures 3a and 3b).
Figure 3(a). A cross section of the geosynthetic-reinforced modular block wall damaged during the 1999 Chi-Chi earthquake.
Figure 3(b). A cross section of RMBW on the verge of collapse in the Chi-Chi earthquake.
(3) For kh=0.22 condition, F, for interface shear and internal sliding are 11.61-15.29 and 2.48-3.24, respectively. These values are the highest ones, among all values of F,. Unfortunately, the failure shown in Figure 2 and Figure 3(b) inferred that internal sliding and interface shear failures may occur to some extent. The over-estimated F, against these failures may be due to the inappropriate failure mechanisms used. In the design guideline, horizontal failure surfaces along the re-
Table 1. Safety factors calculated using the NCMA design guidelines. External stability analyses Internal stability analyses Base sliding Overturning Over-stress of reinf. Pull-out Internal sliding 0.00 6.93 3.55 7.83 1.85 7.59-9.90 0.22 1.30 5.00 0.99 1.81 2.48-3.24"
Facing stability analyses Interface shear Toppling Commections 19.39-25.53 0.62 0.65-0.94* 11.61-15.29" 0.06 0.43-0.61 * * Ranges shown in this table indicate the effect of block-block friction angle, p =30"-45".
kh
370
inforcement layer were used to evaluate the seismic stability of RMBW’s. In addition, buckling and/or bulging at the lower 1/3-1/2 of the modular block wall cannot be analyzed using the existing methods (e.g., Cai and Bathurst, 1996, Ling and Leshchinsky, 1998) in which only horizontal displacement along the base of the reinforced zone is taken into account. An inconsistency between the seismic design method and displacement calculation methods is that the displacement calculations are performed using the base sliding failure mode which may not be the dominant failure mode (see F, for base sliding in Table 1). It is also noted that the ‘three-wedge’ mechanism as shown in Figures 4 and 7 is more appropriate for describing the observed failure patterns. Consequently, the present study focuses on the following: (1) To develop a straightforward ‘three-wedge’ method for pseudo-static analysis of the RMBW, in which, only a safety factor against shear failure along the critical surface is the final result. This may eliminate the use of too many, sometimes, irrelevant safety criteria, as used in the current design guidelines. (2) To perform a new deformation analysis, based on the ‘three-wedge’ mechanism, to describe the failure pattern observed in the 1999 Chi-Chi earthquake, and also to avoid the inconsistency discussed above.
thod is a simplified case of the ‘three-wedge’ model shown in Figure 4. When the stability of facing is considered in the analysis, the connecting reinforcement force (TwJ at the facing-backfill interface, and the block-block shear resistance may influence the result of the analysis. Therefore, four types of facing and connecting conditions were used in the present study. They are summarized in Table 2 and Figure 5. In which, Type 1: Stability of facing is not considered. This type of analysis is equivalent to the conventional ‘two-wedge’ analysis. The mobilized tensile force at the failure surface, T,, is equal to the smaller one of Tpb and Tpf, (Tpb: the pull-out resistance for the back of the potential slip surface; Tpr: the pull-out resistance for the front of potential slip surface) Type 2: Stability of facing is calculated interactively with the stability of the two soil wedges. T, = min. of {TPb, Tpf +Tc, Ttensllle}, and TWI= min. of { Tc7 TPft.TPh, Ttensile), Tlenslle: tensile strength of geogrid, T,: junction strength of geogrid at FRP rod-geogrid connection. Type 3: Stability of facing is calculated interactively with the stability of the two soil wedges. TI = min. of { T P ~TPft.Tb+Tc, , Ttensille), and TWI= min. of {Tb+Tc,TPft.TPb, Tten\ile). Type 4: Stability of facing is calculated interactively with the stability of the two soil wedges. In addition to similar calculations of T, and T,, as used for type 3, shear resistance of the FRP rods was added to the block-block shear resistance.
2 PSEUDO-STATIC ANALYSIS A schematic figure of the three-wedge method used in the present study is shown in Figure 4. The socalled “two-wedge” or “bi-linear failure line” me-
R,
Wedge W
Wedge F
Wedge B
Figure 4. Force equilibrium system in the ‘three-wedge’ method.
37 1
2.5
!
Site 1 M . 2 , CP = 30.40, c= 0 "lm' 2
VBf =o
-1
- -
* - - Type4 (upperbound)
----c
-
- 8 -
- -
1.5
-
-
-
+
- Type4 (lower bound) - Type3 (upper bound) - Type3 (lowerbound)
Type2 Type1
- -
-4-
FS
1
Figure 5. Schematic figure of the pull-out and connecting strength of the geogrid. 0.5
Table 2. Facing and connection types considered in the pseudostatic analysis.
Facing stability Geogrid juction strength Block-block friction Shear strength of FRP
Type1 Type2 Type3 Type4 No Yes Yes Yes No Yes Yes Yes No No Yes Yes No No No Yes
I
I
I
I
I
Figure 6. Results of analysis using various types of facing and connection.
rods"
* The shear strength of FRP rods was converted to an equivalent cohesion of 48.3 kN/m at block-block interface in the stability analysis. Table 3. Upper and lower bound values of block-block interface strength parameters used in the present study.
Upper bound Lower bound
Cohesion, c (Wa) 0 0
Friction angle, U (")
45
30
For types 3 and 4, lower bound and upper bound values of the block-block interface friction angle, ,U, were used (Table 3). In the present analysis, h =0.2 was used ( h =kh/k,, kh, k,: horizontal and vertical ground acceleration, respectively) because the main pulse composed from the N-S and U-D ground accelerations (seismograph TCU052) showed that h =0.2.The F, vs. kh relationships using four types of facing and connection are shown in Figure 6. The following points can be seen: (1) The 'two-wedge' method rendered a largely under-estimated value of khcr (khcr is the value of kh when F,=1.0). This infers that facing elements perform a positive role to the seismic stability of RMBW. (2) The values of khcr increase with increasing facing-reinforcement connecting force, TWi. (3) The values of khcr increase with the increase of shear resistance at the block-block interface. (4) The post-failure curves for types 2, 3 and 4 all joined the post-failure curve for type 1 (two- wedge) because beyond the critical failure condition (F,=l .0), the facing loses its soil-retaining function.
Figure 7. Failure lines calculated using the three-wedge method.
The value of (PBF = 0 ((PBF : friction angle at the interface of blocks B and F) is used throughout the present study because of a mainly-open crack close to the back of the reinforced zone, see Figure 7. This figure also shows the failure patterns that were calculated using various types of facing. The failure lines obtained from types 2 and 3 come close to the failure pattern observed in the field. The predicted failure lines, using facing type 4, deviated considerably from the observed ones. For the following deformation analysis of Geosynthetic-reinforced modular block walls, only facing type 3 is used.
372
I
(b) Figure 8(a). Displacement mode considered in the present study. Figure 8(b). Displacement mode considered in the existing methods.
cp = 30.4", c= 0 kNlm' cp,, = 0 ,kk,= 0.134 (lower bound
0
Downward traction
Y
20
Figure 1 l(a). Ground acceleration record and constant values of khcrused in the displacement calculation. 0""
+kl,<,= 0.1 34 (&,lower bound) .......
.............................. +
-r
Block movement
__ -/A
16
Time(s)
i/"
Tension
12
8
4
//
g
25
Figure 9. A possible mechanism for seismic displacement of facing and soil wedges.
4"0-..
k b = 0.15 (&,upper bound)
1
..........................................
kk = 0.134 (&,lower bound) &(= . ..3oOmm) ....... ------.~..'.----.----.....-------...........................
kh,= 0.15 (&,upper bound) 12
20
Time(s)
Figure 1 1(b). Calculated displacement of facing and backfill soil using the three-wedge failure mechanism.
3 DISPLACEMENT ANALYSIS
65 Figure 10(a). Schematic figure for the seismic displacement of geosynthetic-reinforced modular block walls.
643
....................
Figure lO(b). Displacement diagram for geosynthetic - reinforced modular block walls.
Figures 8 (a) and 8(b) compares the different failure mechanisms used in the three-wedge and the existing methods (e.g. Cai and Bathurst, 1996, Ling and Leshchinsky, 1998) for calculating seismic displacement of RMBW. The new method calculates not only the horizontal displacement but also the vertical displacement of the reinforced zone. Figure 9 schematically shows that the buckling andor bulging of the stacked block facing may be primarily due to the traction force induced by the settlement of the soil behind the facing. Therefore, the calculation of vertical settlement of soil wedge behind the wall may facilitate the seismic design of RMBW's. Figures 10(a) and 10(b) show the failure mechanism and associated displacement diagram used in the displacement analysis for the failed Geosyntheticreinforced modular block wall at site 1. The displacement analysis is based on the 'sliding block' concept, proposed by Newmark (1965) and the 'allowable displacement field' which is used in the limit analysis (e.g., Atkinson, 1981). The new method proposed herewith calculates not only the horizontal displacement but also the vertical dis-
373
placement for all components of RMBW’s. Figure ll(a) shows a major portion of the ground acceleration and the calculated values of khcr ,based on facing type 3. Figure ll(b) shows the calculated horizontal displacements for the facing ( 65 ) and the vertical displacements of soil wedge ‘F’ ( 6 2). In the present study, the angle of dilatancyv =O” is assumed. The calculated horizontal displacement ( 6 5 ) for the facing was 300-7OOmm. The calculated vertical displacement (62)for block ‘F’ were 50200mm. These calculated results are comparable with the measured ones (see Figure llb). Larger measured value of 6 2 than the calculated ones may be due to the compression of soil wedge during the earthquake. It is noted, however, the present calculation was based on a constant khcr using peak value of cp . Further displacement calculations using other possible post-failure values of khcr , e.g., khcr based on residual value of y, , or khcrbased on two-wedge mechanism, as shown in Figure 12, should be performed in the near future.
of the facing structure and the backfill soil wedges in an interactive way, resulting in a safety factor relating to the shear strength along the potential failure surface. This can eliminate safety criteria that may be irrelevant to the seismic stability of RMBW. For site 1, the new method results in safety evaluations and failure mechanisms comparable with those observed in the post-earthquake site investigations. A new method for displacement calculation of RMBW based on the ‘three-wedge’ mechanism and the ‘displacement diagram’ is developed. The new method gives more realistic and accurate displacements of RMBW’s than any other existing ‘sliding block’ methods. 5 ACKNOWLEDGEMENT The present study is supported by the National Science Council, Taiwan, ROC under the contract Nos. NSC 89-292 1-2-319-005-05 and NSC 89-22 18-E006-144. The authors acknowledge Prof. F. Tatsuoka, University of Tokyo, Japan, Dr. M. Tateyama, Japan Railway Research Institute, and Dr. Y. Iwasaki, Geo-Research Institute, Osaka, Japan, for their financial and technical supports in the post-earthquake investigations. REFEENCES
Figure 12. Possible variations of k,,,, in the post-failure deformation calculation.
4 CONCLUSIONS Stability and displacement analyses were performed on an RMBW damaged during the 1999 Chi-Chi earthquake (M~=7.3). Analysis of the damaged RMBW, following the seismic design guideline proposed by NCMA (1998), indicated that the failure mode resulted from the use of the design guideline deviated largely from the observed one. A new ‘three-wedge’analysis method is developed. A major advantage of this method is to analyze the stability
374
Atkinson, J.H. 198I Foundations and slopes, an introduction to applications of critical state soil mechanics, McGraw-Hill, London, Ch.4, 6. Cai, Z. and Bathurst, R.J. 1996 Seismic-induced permanent displacement of Geosynthetic-reinforced segmental retaining walls Can. Geothech. J., Vol. 33, pp. 937-955. Huang, C.C. 2000a Investigations of soil retaining structures damaged during the Chi-Chi (Taiwan) earthquake, Journal of the Chinese Institute of Engineers, Vol. 23, No.4, pp. 417-428. Huang, C.C. 2000b Investigations of the reinforced block facing walls damaged during the Chi-Chi earthquake, Geosynthetic Technical Information, Japan Chapter of the International Geosynthetics Society, Vol. 16, pp. 40-45. (in Japanese) Ling, H.I. and Leshchinsky, D. 1998 Effects of vertical acceleration on seismic design of Geosynthetic-reinforced structures Geotechnique, Vol. 48, No.3, pp. 347-373. National Concrete Masonry Association 1998 Segmental retaining walls- Seismic design manual 1 “ edition, authored by Bathurst, R.J. Newmark, N. M. 1965 Effect of earthquakes on dams and embankments Geotechnique, Vol. 15, No. 2, pp. 139-159.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets 6: Zeitlinger, ISBN 90 2651 863 3
A contribution to the design of flexible wire mesh facing J. -M. Jailloux Profractal Engineering & Consulting, Clichy, France
N. Freitag & P. Segrestin Freyssinet International, Vt!lizy, France
ABSTRACT: The current mesh facing technology for MSE structures is validated mainly by observation of actual structures, but the structural design of the mesh remains unsolved so far. Classical approach gives deformations that are much larger than the ones observed on site. This paper presents a new methodology that leads to realistic results due to a better understanding of the actual interaction between the soil and the flexible facing. The communication includes: - the observations on actual structures - the principle of the modeling of the soil, with a behavioral law (stress vs. deformation of the backfill material) which constitutes an innovative feature in the design of such structures - the description of the algorithm which is used to calculate the stresses and deflections of the wire mesh facing units. Other applications that use a mesh facing, noticeably soil nailing, can take advantage of the method.
1 INTRODUCTION Facing panels constituted of steel mesh have been widely used in Mechanically Stabilized Earth structures under various versions. Sometimes bent or flat, with or without hooks, connected to reinforcements made of steel or synthetic strip or mesh, or geotextile, the different facing elements have been justified only by observation on actual structures and occasionally by experimentation. So far no comprehensive approach of the mechanisms that govern the deformation and the stress in the steel wires has been presented. The subject of this project was to analyze the situation and to derive some guidance as regard the justification of the facing technology. A new methodology is now proposed. It introduces an innovative feature by taking account of a behavioral law stress vs. deformation for the backfill material combined with the non-linear calculation of the deformation of the mesh.
Figure 1. Components of a steel mesh facing panel.
helps to support the earth pressure during backfilling operation. So that it is possible to backfill a course of panel up to its top, and to compact the soil properly before installing the next course. We observe that for each reinforcement, four bearings are created on the flexible facing panel, plus two bearings due to the overlapping of the next row of panels. In this resides the principle of this type of mesh facing: transferring the earth pressure applied on a flexible panel to the localized reinforcing strip by means of several separated connections so that the deformation of the facing is limited. Several versions have already appeared on this principle: diameter of wires (@)8 or 10 mm) and openings in the mesh (100 or 150 mm). Also the overall length (2.4 to 6 m) and height may vary. Stone filling or topsoil for vegetation is currently
2 PRESENTATION OF THE TECHNOLOGY For the purpose of this study we have considered one type of facing currently used for green MSE slopes. Its different components are presented in Figure 1. A flat steel mesh is supported at the proper angle of inclination of the facing by means of prebent hooks at the bottom. The reinforcing strip is connected to this main hook with a flat tie-strip. A secondary hook further maintains the facing and 375
used to back the mesh. These different versions finally are answers to different local manufacturing constraints and marketing requirements.
mesh panel used is the one described in Figure 1. The overall length of one panel is 3 m. The mesh is constituted by wires 4 8 mm, spacing 100x100 mm. Three steel reinforcing strips per panel are connected to hooks, which create 18 bearings per panel. Figure 2 presents the technique to measure the deformation: a frame made of steel square tubes is placed more or less parallel to the facing. A Tshaped rule slides on the upper side of the frame. The distance between the vertical sliding rule and each node of the mesh is measured and recorded (300 nodes per panel). The exact position of hooks and surrounding panels were also noted. Three similar panels with small apparent deformation were measured on that particular project. A Reinforced Earth design program for sloping wall was used to estimate the loads. The total force applied by the earth on that panel is 30 kN (10 kN per reinforcing strip). The raw results (readings of the distance between a node and the frame) are transferred in a spreadsheet. The position of the reference plane is modi-
3 THE DESIGN PROBLEM Anyone who built a MSE wall with a mesh facing knows that the deformation of the facing is very dependent on the way the workers place and compact the soil behind the facing. When the pressure of a foot behind the mesh is released, a permanent localized deformation is possible; it does not however question the stability of the wall. Our concern is to assess the theoretical deformation and the normal tensile stress in the wires as if the wall was perfectly built (as it could be done in a laboratory under controlled conditions), without taking account of construction defects or damages to the facing. The problem is complex because the large number of elements brought into play and also the unknown response of the soil to deformation. However we can anticipate the benefit of a suitable method to justify and improve the existing technique, or test new solutions. When one tries to calculate the deformation by means of classical formulas of a beam on supports, the deformation is large and much larger than what is actually observed on sites. In fact the wires act as cables as well and carry a load that reduces the deformation. A better approach is to use a more sophisticated calculation tool: a computer program for structural design with a non-linear method. The entire mesh is modeled with elementary bars connected to nodes (the cross welds). In that case the neutral axis in one element can elongate and generate a normal tensile load, which better represents the actual behavior of a wire in the facing. For purpose of simplification, the earth pressure is applied like a local force at each node proportionally to the surface area it represents (usually one mesh opening). A connection to one hook is modeled as a simple support, fixed in space. The program increases the load by small steps (method P-Delta for example). For the real wall described below, under the most probable earth pressure (without any load factor), the result provides a too high deformation (84 mm) instead of some 20 mm observed on the wall. Although rather sophisticated, the approach is not enough adapted to the reality. We need something else. Before going into details, let's review how was measured the deformation on a real structure.
Figure 2. Measurement of facing deformation
4 MEASUREMENTS ON SITE
The structure is a Reinforced Earth wall in Italy. The facing about 5 m high is at an angle of 80". The
Figure 3. Analyze of actual deformation.
376
fied by calculation to find the best adjustment i.e. the plane that gives the minimum standard deviation. This is shown in Figure 3. Of course it is diffcult to have on a construction site all connection points in the same plane within a millimeter and there is no need to be so accurate. An ideal geometry after construction could be approached only in a laboratory where all parameters could be better controlled. Finally we can estimate that the maximum deformation due to earth pressure is around 20 mm. No need to try to obtain a greater precision. We just want an order of magnitude.
5 MECHANISM OF DEFORMATION Now, let’s consider the nature of the mesh panel and the parameters that play a role in the mechanism of deformation: - A very flexible beam, or more precisely a plate, constituted of wires, multi-supported; its inertia module is very small for such small witres, - A network of cables in which a normal tension is developed as the deformation increases; it behaves somehow like a membrane, - A soil for which we have little information particularly on the way it behaves behind such flexible facing The two first mechanical points could be dealt with, thanks to powerful calculation programs. Taking account of the soil behavior is a new feature in the design of a MSE facing. Although it is obvious that the earth pressure decreases when the deformation increases, we must describe this behavior and quantify it. Terrasol made a study (1988) on the subject of modeling the behavior of a facing constituted by rigid vertical posts or soldiers, between which are placed a thin concrete facing panel. Since the facing panel can deform much more relatively to the soldiers, an arch of soil is built up that span over the panel and rest on the rigid components: the soldiers. The Figure 4 shows the stress concentration effect on soldiers as well as the rotation of stress direction. The report concludes: “a small deformation of the facing (of some mm) is sufficient to activate the mechanism of arching effect”. The earth pressure behind the facing is thus reduced drastically (for example to one tenth of the average pressure) depending mainly on the length between the supports, while the pressure increases behind the stiff elements. Let’s remind some basic concept in soil mechanics, noticeably as regards retaining sttuctures. The movement of a rigid screen toward or outward the soil from the position at rest gives a certain force F called respectively passive or active earth pressure. The result of that experimentation can be plotted as F in relationship with the deformation (see Figure 5).
Figure 4. Arch effect on stiff elements (Terrasol).
Force
t
Passive Pressure
Active Pressure
Deformation Figure 5. Force on a screen.
Also a more fundamental test, the triaxial test, allows drawing a similar curve for the intrinsic properties of the soil (Figure 6). Under the earth pressure, the steel mesh deforms outward hence the earth pressure reduces considerably. For negative movement (like it can happen locally on the mesh) the pressure increases quickly. Since the decrease of the earth pressure is due to the build up of arches, we must also consider the length of the span: we assume that the deformation is proportional to the span. Therefore, the general behavior law takes account of the relative deformation instead of the absolute deformation. After several attempts, we propose an exponential law, see Figure 7, where: qi is the load (or earth pressure) at a given node i and relative deformation di/LV q0 is the load at the supports (supposed to be fixed: no deformation) a is the coefficient for the soil (no dimension); greater is a, stiffer is the soil.
377
Oats preparationand analysis
Densesand
(P-Delta method) Create the mesh
Triaxial test
Define and locatethe supports
5
0"
% /a
Calculate deformations
End if convergence
Put a same load on each node
h l o d i i loads on the nodes accordingto the behavioral law of the soii
Read and pfot deformations, tension, stress, etc.
__-
I
- 1 0
5
0
.
I
Figure 8. Flowchart of operations.
-
10
5
A i, Figure 6. Triaxial response of a sand: K vs. deformation. qi: Local Applied Force at a node
qi = q0 * exp(-a * di/lLV)
\I
1
a : coefficient for the soil a = 15foraverysofisoif ' a = 120 for stones LV : a reference distance for the arch span i.e. greatest distance between tie-point9 on the facing
I
Figure 9. Modification of loads at the nodes.
dinV :Relative Deformation
0
done with a new distribution of load: higher load at the supports, lower load in span. The process is repeated till no significant change is observed. Practically we have created a small spreadsheet program to deal with the setting of the proper loads used by the calculation program. The flow chart of operations is shown in Figure 8. Figure 9 presents the way the new set of loads is defined after the first run (tl). The horizontal line represents the deformation caleulated by the program after all 300 nodes were loaded at qi(t1) = 167 N.The total load is 50 kN. Now the qi'(t1) points represent the direct application of the behavioral law: the load increases for the negative deformation and decreases as the positive deformation increases. But the total is now down to 37.6 kN. A "damping" coefficient p can be used to take a part only of the difference between the previous load value qi(t1) and the next one qi'(t1). This precaution was found to be useful to better control the convergence. Once done, each value is multiplied by an overall factor to reach again the total load of 50 kN and thus obtain the final qi(t2) values ready for the next run.
Figure 7. Earth pressure vs. facing deformation.
di is the deformation at node i LV is the longest distance between two consecutive supports. The advantages of this law is two fold: - A quick convergence, - A unique parameter for the soil: "a".
6 METHODOLOGY The problem is now to implement the above behavioral law in the calculation. The idea is to run first the calculation of a panel with a uniformly distributed load using a non-linear static calculation program. The method of analysis is P-Delta for geometrically non-linear beam element (Gachon and Galka 1978). The best results were obtained by using the algorithm BFGS (BreydonFletcher-Goldfard-Shanno). Then the deformation at each node is analyzed, input in the exponential law equation and the load is modified accordingly. Then again a new run will be
378
Figure 10. Presentation of results (loads and deformations) after convergence; the 18 tie-points are at di = 0.
7 RESULTS 7.1 Soft soil
A new set of results was obtained for a typical panel made of wires @ l O mm, mesh size 100x100 mm. Three reinforcing strips are connected to one panel 3 m long; each strip provides four tie-points (the main and secondary double-hooks) plus two due to the overlapping of the next panel. The Figure 10 presents the maximum deformation at 13.6 mm obtained after only 3 runs, under a total load of 50 kN with a soft soil (a = 30). On this view of the load distribution and mesh deformation, we see that the mesh is equally supported by the 18 tiepoints. Tension in the wires is an information provided by the calculation program. It is thus possible to set a specification for the maximum stress in the steel wires. After scanning different wire diameters, mesh size and number of tie points, a graph was drawn, as shown in Figure 11, in which the field of application of the various mesh facing is clearly defined. Note that the value of q0 is greater now: the load increases on the bearing points. This was expected. The new set of loads is then sent to the calculation program and processed, and so on. Usually convergence is obtained after a few iterations only: 3 or 4 for a soft soil to 6 or 8 for a stiff soil. Different values for "a" were tried to fit with the observations on the actual wall. For the wall described in 94 above (with an estimated total load on one facing panel of 30 kN), a calculated deformation of 23.7 mrn was obtained (compared to 20 mm observed on site) with a coefficient a = 30, which can be considered typical for a soft soil as the one generally used as backfill for MSE structures. A similar graph can be drawn for another criteria like the allowable deformation for example.
Figure 11. Tensile stress in the wires - Mesh 100 x 100 mm.
7.2 Stiffsoil In Figure 12 are presented the results obtained on exactly the same panel and load as described in figure 10, but the soil is now much stiffer with a "a" coefficient of 120. This case would correspond to a stone filling behind the panels. The loads are extremely concentrated at the tie-points. The ratio between the maximum and minimum value of the load at one node reaches 6.3. The calculation is a little longer and a damping ratio of 1.5 (it means that only 67% of the difference between qi (tl) and qi' (tl) was used to define the next qi (t2)) helps for convergence. The deformation after the first run was 17 mm; it is the value we would have without considering the soil behavior. After 5 or 6 runs, the maximum deformation was stabilized at 9 mm, but it took some more runs to balance the distribution of loads. 379
Similar technologies in which a flexible mesh receives the earth pressure can also take advantage of this calculation technique to estimate the behavior of the structure; this is noticeably the case for soilnailed embankment where a steel mesh is used to retain the soil between the heads of the nails.
8 CONCLUSION More and more Mechanically Stabilized Structures use light steel mesh panels as facing. So far no comprehensive approach of the mechanisms that govern the deformation and the stress in the steel wires has been presented. The use of non-linear analysis combined with a behavioral law stress vs. deformation for the backfill material constitutes an innovative feature in the design of mesh facing technology. The deformation is now realistic and all tensile stress in the wires can be estimated. Although used first for MSE facing technology, the methodology presented in this paper is valid for other kind of applications.
Figure 12. Same mesh and loading as above in Fig. 1 I ; a = 120. All 18 peaks correspond to tie-points.
7.3 Limitations We have considered that the tie-points are fixed. If in the model we allowed them to slide in the plane of the panel, the tension in the wires due to the cable effect would be reduced and the deformation increased. In fact, the stress decreases much quicker than the deformation increases due to this modification. Further work would be needed to refine the model, but we admit that the results obtained with this method are quite satisfactory and certainly conservative as regards the tensile stress in the wires. Finally this calculation tool is very useful to define new geometry of panels in accordance with the desired specifications on load, deformation and stress.
REmRENCES Gachon, H. & Y.GalCa 1978. ModCle d'analyse non linkaire des structures 5 barres - mtthodes d'approche du seuil de bifurcation. Construction Me'tullique. N"2, 7-34. Terrasol, 1988. Simulation de l'effet de volite dans le soutenkment par paroi berlinoise en vue d'une application aux parements Terre ArmCe. Rapport TA 87.59/02.
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Landmarks in Earth Reinforcemenf, Ochiai et al. (eds), 0 2001 Swefs & Zeiilinger, ISBN 90 2651 863 3
A field instrumentation and FEM analysis for an isolated-reinforced earth wall Y.Y. Kim, K.J. Han & K.M. Kim E&S Engineering Co., Ltd., Seoul, Korea
ABSTRACT: During the construction of reinforced earth wall, an amount of accumulative horizontal deformation induced by backfilling compaction is inevitable. To reduce the horizontal displacement, an isolatedreinforced earth wall method was developed by isolating the facing blocks from the reinforced embankment. Field tests were performed by sequential construction of a 7.2 m high trial embankment in order to veri@ the effect of the method. Firstly, a reinforced embankment composed of geogrids, geotextiles and backfills was constructed. As a second construction stage, the facing blocks were installed in front of the embankment. During the construction, displacement pins and strain gauges were used to measure the horizontal deformations of the wall and the tensile strains of geogrids, respectively. Based on the field instrumentations of trial embankment, it is found that the reduction in horizontal displacement of facing blocks is significant; the horizontal deflection in the first construction stage was 70 mm and the corresponding value of second construction was approximately O(zero). It is also found that the prediction by FEM analysis simulates well the general trend observed from field instrumentation.
1 INTRODUCTION
2 DESIGN METHOD OF KOESWALL
Construction of reinforced earth retaining walls has been continuously increased due to economical considerations, the effectiveness and practicality of the construction technique, and excellent appearance (Cho and Shin, 1999). However some amount of horizontal deformation is induced during the construction stages of reinforced earth walls. As the reinforced earth wall is constructed with compacting the backfill step by step, the accumulative horizontal deformation is inevitable. To reduce the horizontal displacement, an isolated-reinforced earth wall method (named KOESWall System) was developed by isolating the facing blocks from the reinforced embankment(E & S Engineering, 1998). The main purpose of this study is to verify the usefulness of this method and the reduction effect of horizontal displacement. A 7.2 m high trial embankment was tentatively constructed in order to verify the effect of KOESWall. During the construction, displacement pins and strain gauges were used to measure the horizontal deformations of wall and tensile forces of geogrids, respectively. The result from the field instrumentation was compared with a FEM analysis.
The design of KOESWall system is based on tieback analysis and slope stability analysis. The tieback analysis is a sort of limit equilibrium analysis(Jewel1, R.A., 1985). It has been assumed that the thrust of reinforced backfill is equilibrium with the resistant force induced from the soil-reinforcement interaction. In the slope stability analysis, we can calculate the equilibrium of forces and of moments on the potential failure line enhanced with reinforcement. The potential failure line of geosynthetic reinforced earth wall is supposed to be a log-spiral, in which safety factor for moment about pole is minimum. In the geosynthetic reinforced earth wall without facing blocks, the factor of safety for moment equilibrium(FS,,,)about point 0 in Figure 1 (a) is defined by
381
wall by Kim et aL(1996) was adopted in the design program for the KOESWall system. 3 FIELD INSTRUMENTATION The KOESWall was constructed on October, 1998 in Whasung, Kyunggi-Do, about 30 km south of Seoul. This wall was constructed to extend the site for the precast concrete manufacturing facilities. The cross-section of the wall is schematically drawn in Figure 2. The major elements of the wall are categorized into three parts; backfill soil, geogrid as a reinforcement, and modular facing block. The decomposed granite soil as a backfill was obtained around the construction site. Its specific gravity is about 2.72 and contents of sand is about 82.6%. The silt and clay contents are 17.4%. Kim and Lee( 1996) presented the effectiveness of decomposed granite soils as a backfilling material in reinforced earth structures. Total height and length of the constructed reinforced wall are 7.2 m and 53.2 my respectively. The surcharge height of 1.0 m was constructed by soil Table 1. Details of compacted soil and geogrids (after Kim & Lee, 1996). Compacted Soil 21.34 kN/m’
ydrnny
Material
ES-grid PET/PVC
wOp,
8.3 %
Tensile strength (Wide width)
113.01 kN/m
PI
N.P
Aperture size
25mm x 50mm
USUC
SM
Rib thickness
1.37 mm
Figure 1. Potential failure surface of KOESWall
In which, MA,,, Mdq and Mdtr is driving moment by self weight of wedge, surcharge load (4) and seepage force, respectively. And M,, and M,., is resisting moment by cohesion and the maximum tension force of reinforcement(T,,), respectively. As the facing blocks are placed in the front of reinforced earth (see Figure 1 (b)), an additional moment induced from facing blocks is developed. In this case, the additional moments may be considered as follows.
In which, MdL, and MrL, are driving and resisting moment by facing blocks, respectively. The method which was proposed for the prediction and internal stability analysis of horizontal deformation of geosynthetic reinforced earth retaining
382
Figure 2. Cross section used for field instrumentation.
over the top of the wall as shown in Figure 2. The nonwoven geotextile was used for rapping part of soil to restrain the backfill soil. The compaction of backfill soil was proceeded by 0.7 ton (KD85, 7HP) hand-guide roller. Geogrid from the main body of reinforced wall was connected to the modular facing block by placing it on the top of the block without using any special gear. The front top part of block is a little bit higher than the rest part of block in order to hold the geogrid with self-weight of stacked blocks. The shear test was performed to evaluate the required shear resistance in this type of connection. Major measurement devices such as horizontal displacement pins and strain gauges were installed at various levels of the wall. Five displacement pins were fixed on the various levels of facing blocks to measure the wall displacement due to the compaction-induced earth pressure. The strain gauges were attached on the geogrid at 9-12 places along the geogrid layer to measure the mobilized tensile strength and strain due to the horizontal earth pressure. The strain gauges are set at the same levels of displacement pins. Figures 3 and 4 shows the measured horizontal deformation at the two instrumentation points, respectly. The maximum horizontal deformation developed at a height of about 2/3H in both cases. The reduction effect of horizontal displacement and the mobilized tensile strain of the reinforcement of the wall will be compared with the results of FEM analysis. 4 FEM ANALYSIS PENTAGON-2D developed in Korea was used for the FEM analysis of test wall. PENTAGON-2D is the comprehensive finite element analysis program to calculate the variables in the continuum model subject to the external load and/or seepage boundary conditions. The displacement method is implemented to solve the equilibrium equation. Here, the primary variable is the displacement vector and the secondary variable is the stress. To limit the stresses the structure may retain, the plasticity theory is used as a constitutive model. If the elastic stress violates the limit, the stress-strain relationship is assumed to follow the flow rule. Figure 5 shows the finite element mesh used for the analysis. It is composed of 2873 nodes, 912 Quad8 elements (eight nodes Quad elements), 534 Truss2 elements (two nodes truss elements) for geosynthetic reinforcements. The soil is modeled by Quad8 elements. The fill was assumed to be an elastic-perfectly plastic material with a Mohr-Coulomb failure criterion. The foundation of the wall was to exhibit an elastic behaviour. 383
Table 2. Material properties.
Bulk modulus, K(MN/m’) Shear modulus, G (MN/m’)
Fill
Foundation
Crushed rock
Facing
24.53
735.75
57.22
98.10
11.32
339.57
26.41
81.75
Cohesion, c (kN/m2j
9.81
9.81
9.81
98.10
Internal friction angle, 4 (”)
30
35
35
40
Unit weight, y (kN/m3)
20.6
19.6
20.6
23.5
Young’s modulus, E (MN/m’) Sectional area, A (m’) Tensile strength, (MN/m2) Compressional strength, (kN/m2j
Grid Reinforcement
Nonwoven geotextile
4563
98.10
0.00032
0.03
14.60
9.81
0
0
The facing block is modeled using Quad8 elements and was assumed to be elastic. The geotextile sheets were modeled using elasticperfectly plastic truss elements with negligible compressive strength and no bending stiffness. And the reinforcements is modeled with perfect interface adherence to the adjacent soil at the point of maximum tension. This means that there is no slip between the soil and the reinforcements; the soil and reinforcement strains are the same at this interface. In the FEM analysis, by only the applying body force, the deformation behavior of geosynthetic reinforced earth wall under the compaction can not be fully simulated. Thus, we attempt to simulate the compacting effect by loading and unloading of a uniform surcharge(0, 49.05, 68.67, 98.10 kPa) at the top of every soil layer. To reflect upon backfilling compaction efforts, Figure 6 shows the calculated horizontal displacement of reinforced embankment with incremental uniform surcharge loads. In case of only considering the selfweight of embankment, the horizontal displacement is relatively small and the location of maximum horizontal displacement is H/3 from bottom of embankment. But the horizontal displacement is increased with increasing the applied load and the location of maximum horizontal displacement changed to 2H/3 from bottom of embankment, which seems to be similar to the actual behavior of geosynthetic reinforced earth wall. So the case of the surcharge load q = 9.81 kPa is recommended as a modeling of compaction effects in the FEM analysis.
Figures 7 and 8 show the horizontal displacement contour with respect to before and after block installation. In these figures, the trends and values of horizontal displacement are nearly same. The horizontal displacement of backfilling embankment does not affect on the front blocks. Figure 9 shows the horizontal deformation of geosynthetic reinforced embankment with the construction stages in the FEM analysis. In this figure, with the construction stages, the horizontal deformation of upper part of the wall became greater than that of the lower part, which is similar to the field measurements. Figure 10 shows the distribution of tensile force developed in the reinforcement. The location of maximum tensile forces of each reinforcement is near the wall face.
0
30
60
90
120
150
Horizontal Deformation (mm) Figure 6. Variation of horizontal deformation with applied compaction load.
Figure 7. x-displacement contour after the construction of geosynthetic reinforced earth wall (q = 9.81 kPa).
384
5 COMPARISON BETWEEN MEASURED AND COMPUTED RESULTS 5.1 Comparison of horizontal deformation
In Figure 11, the horizontal deformation measured in the trial embankment is compared with the result of FEM analysis and design program. In the most cases, the maximum horizontal deformations are developed at about 2H/3. However, in the FEM analysis, there were relatively large deformations in the lower part of the wall height.
Figure 1 1. Horizontal deformation of geogrid reinforced embankments.
5.2 Comparison of the horizontal displacement of facing block Figure 12 shows the horizontal displacement of front blocks after their installation. The measured values were greater than the calculated values. It may be caused by the instrumental error. But the values are much smaller than the mobilized horizontal deformation of the reinforced embankment. This shows the effectiveness of discrete construction method in KOESWall system.
Figure 10. Distribution of tensile force in the reinforcements.
Figure 12. Horizontal deformation of facing blocks
385
5.3 Comparison of the strain in the reinforcements
6 CONCLUSIONS
Tensile strain distribution on the reinforcement is shown in Figure 13. In case of FEM, there were large strains in the reinforcement located near the wall. But in case of the design program, the upper part reinforcements(# 14, # 16, # 17) mobilize large strain. So the design program should be modified properly.
Field Instrumentation and FEM analysis for the isolated-reinforced earth retaining wall constructed with geogrids and concrete modular facing block is performed. Based on the results of the field measurement and FEM analysis, the following conclusions are drawn. 1) In the FEM analysis, only with the applying body force, the compaction effect can not be fully considered. Thus, an attempt is made to simulate the compacting effect by loading and unloading of an uniform surcharge(0, 49.05, 68.67, 98.10 kPa) at the top of every soil layer. Using a surcharge load of q = 98.1 kPa is recommended in modeling the compaction effects in FEM analysis. 2) In the case of FEM analysis employing the uniform surcharge load q = 98.1 kPa, the maximum horizontal deformation of geosynthetic reinforced embankments developed at a height of about 2/3H, which is the same trend of field measurements. 5)
In both cases, when the isolated-reinforced earth wall method were adopted, over 80 ?40 of horizontal displacement can be reduced. The displacement of facing block is very small compared to the geosynthetic-reinforced earth body.
REFERENCES Cho, S D and Shin, E C (1 999), "Application of geosynthetics and earth reinforcement technique in Korea", Special Volume for Proc of the 1lth Asian Regional Conference on SMGE, pp 43-49, Seoul, Korea E & S Engineering (1 998), Technical Report of KOESWall System, E & S Engineering CO, Ltd , Seoul, Korea Jewell, R A (1985), "Limit Equilibrium Analysis Reinforced Soil Walls", Pro of 11th ICSMFE, San Francisco, Vol 3, p 1705 Kim, S K and Lee, E S , (1996), "Use of decomposed granite soils as backfill for reinforced earth structures", Proc of the Synzposizimon Earth Reinforcement, Kyushu, Japan, Voi 1, pp 51-56 Kim, H T , Kang, I K , Lee, E S and Bang, Y K (1 996), "Internal Stability analysis and Deformation Prediction for Fabric Reinforced Earth Structure", Earth Reinforcement, Proc of the International Symposriinz on Earth Reinforcenient, Fukuoka, Kyushu, Japan, November, pp 389-394
Figure 13. Distribution of tensile strain on the reinforcements.
386
Landmarks in Earth Reinforcement, Ochiai et al. (eds), @ 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Limit analysis of soil structures subjected to constraints by reinforcement S. Kobayashi, A. Tanaka & T. Tamura Kyoto Universily, Japan
ABSTRACT: Many reinforcement methods of soil structures have been proposed and used in engineering practices. This paper aims to present a theoretical methodology to evaluate the safety of soil structures with reinforcement members from the point of limit analysis. The upper bound method is extended to deal with rigid-work hardening materials. Its mathematical properties are also discussed.
1 INTRODUCTION Many reinforcement methods of soil structures have been presented. They have great advantages in engineering practices. Then, what are key issues of reinforcement mechanisms? This is a simple motivation of this paper. Stability problems of soil structures with reinforcement members are discussed from the point of limit analysis. In many engineering practices, the limit equilibrium method is used to estimate the stability of reinforced soil structures. In the previous study, theoretical properties of the limit equilibrium method was discussed and concluded as an approximate method of the upper bound analysis (Kobayashi 2000b). The effect of reinforcement members are usually replaced as member forces in engineering design. Then, according to the limit equilibrium method, equilibrium of forces or moments of the region of interest are calculated to estimated the stability of the system. However, it must be noticed that member forces are not known prior to the solution. In other words, a collapse mechanism of a system depends on the distributions and the strengths of reinforcement members via deformation restriction effect. Therefore, actual member forces essentially depend on the collapse mechanism (= solution of the analysis). In this sense, a kinematic approach to the rigid plastic boundary value problems - the upper bound method - is one of the promising methods. As a simple and a typical example, a stability problem of self-standing type sheet pile walls has been investigated and discussed by the author (Kobayashi 2000a). Interactions between soil structures and reinforcement members must be considered carefully. From the point of limit analysis, deformation mechanisms
of soil structures with reinforcement members can be characterised by the following two features; e
0
Rigidities of geo materials and reinforcement members are very different. Some reinforcement materials are very ductile. Geo-textiles and geomembranes are typical examples. On the other hand, all most all deformation of soil materials can be treated as plastic deformation. Therefore, deformation (or strain) levels of the reinforcement members at their peak strengths and deformation levels of the geo materials at their peak may be mismatched. Many geo materials show apparent strain softening behaviours. For example, stress-strain relations of over consolidated clays or dense sands show their peak strengths and decrease the strengths to their residual values with the increase of shear deformation. Then, it must be noted that deformation levels should be kept within a certain magnitude to utilise the potential strength of the geo material.
It is impossible to deal with these two features by the ordinary upper bound analysis. Therefore, an extended method of the upper bound method is proposed and discussed in this paper. 2 PROPOSED METHOD 2.1 Fundamentals of upper bound method Let us consider a stress field Q satisfying a yield condition, i.e. f ((2) = 0. Let be plastic strain rates correspondingto the stress Keld Q. Let us also consider another stress field Q* which never violates the
2
387
yield condition at same material points. Hill assumed the following inequality called "principle of maximum plastic work" (Hill 1948).
Geometrical interpretation of equation (1) is well known, as Yielding surface is convex in stress space. Figure 1. Piecewisely linearised yielding locus.
Plastic strain rate 4 P is always parallel to the outward normal direction of yielding surface
( a f /ag>.
respectively, and
D@) =
Thus, the associated flow rule is derived from Hill's assumption,
=
Equation (1) can be represented in the large as,
lT
s,
Q(k)
idS
+
qPdV
-
1
SD
+
s,
@)
E(') . a d s +
. ["dS
s,
(6)
b('). &dV. (7)
This inequality insists that internal dissipation D@) by a certain trial kinematically admissible velocity field is never less than external plastic work D(') for the actual limit load.
for a region V and on a velocity discontinuity plane I?, where $ and [i] denote the surface traction and the amount of velocity discontinuity on I?, respectively. Let us consider any kinematically and plastically admissible velocity fields zi,which satisfy velocity strain rate relation, boundary conditions on the velocity boundaries and the associated flow rule. According to the constitutive relations (2), a corresponding stress field Q@)to the assumed velocity field can be found. It should be noted that stress field Q ( k )is not always determined uniquely, nor staticallyadmissible. On the other hand, let the exact solution for a stress field be Q"). Principle of-maximum plastic work is applied for this problem;
-
After some arrangements, the following inequality can be derived
2.2 Linear programming formulation of upper bound method Linearisation of yielding functions. Linear programming formulation is convenient to the theoretical consideration of the proposed method. By introducing piecewise linearisation of yielding function, limit analysis can be formulated in the form of linear programming (LP). This formulation was firstly introduced by Maier (Maier 1970), and developed in the field of structural plasticity (Maier & Munro 1982; Maier & Lloyd Smith 1986; Lloyd Smith 1990). Yielding function can be expressed as a closed convex locus containing the origin in the stress space, according to Hill's maximum plastic work principle. This locus is then piecewisely linearised by intersectional polyhedron, as shown in figure 1. This piecewisely linearised yielding locus also contains the convexity. Let rzi be the outward normal unit vector of the ith plane and hi be the distance from the origin to the i-th plane. Then, a condition that an arbitrary stress state Q exists within or on the yielding locus can be expressed as
3 .Q 5 h.j.
where the superscripts k and t implies the terms of the kinematically admissible field and the exact field,
388
(8) Let be a tensor arranged each outward unit normal vectors giand & be a vector arranged each distances h;, a stress state Q which is not violating the yielding condition can beexpressed as
N T Q I h , w h e r e-N = ( r z , , r z ~ , . . . , r z ~ ) .
(9)
Then, let us consider a targeted boundary value problem in a weak form. According to virtual work rate principle, a following equation in weak form can be obtained for any arbitrary kinematically admissible 4P velocity field &, its corresponding strain rate field and for any arbitrary stress field __ Q.
This weak form equation can be transformed into a following equation by introducing spatial discretization,
Figure 2. Targeted rigid-plastic boundary value problem.
Associated flow rule for the piecewise linear yielding locus is expressed as follows.
where h is a vector arranged non-negative plastic multipliers of each planes. Compatibility condition of stresses and plastic strain rates can be expressed as
Equation (11) insists the complementary condition between stresses and plastic strain rates. This inequality insists the following two possibilities. 0
0
A stress state is on the yielding locus, i.e. hi -
For the concise notation, volume integration in equation (16) is omitted in this paper. Velocities on the fixed velocity boundaries S,f can be expressed as follows,
where tensor D has zero or unity components. Its =f component [ D f ] j j is unity, if the i-th fixed velocity condition is corresponding to the j-th component of else, [ D f ] i jis zero. nodal velocity vector iN, Finally, weak form equation can be expressed as follows,
ni - Q = 0, or
A stress state is within the yielding locus and no plastic strain rates are occurred, i.e., hi = 0.
where scalar in is a load factor. The internal dissipation energy rate @jilt for a trial kinematically admissible velocity field is expressed in a spatially discretized form as follows,
Targeted boundary value problems and spatial discretization. Targeted rigid-plastic boundary value problems are schematically shown in figure 2. For the simplicity of the discussion, only fixed boundaries and free surface boundaries (no traction on stress boundaries) are considered.
zi = Qon S,f,
_.
t = Qon S,.
(12)
QP are stresses corresponding to a given velocwhere ity field. Plastic multipliers vector are evaluated at integral points in numerical calculations. According to Hill’s maximum plastic work principle, a following inequality holds,
External loads are only body forces in&, where scalar m is a load factor. Velocity field i is spatially discretized by shape functions __ and nodal velocities tir\l, as
For rigid-plastic materials, plastic strain rates q p are associated with nodal velocities i and plastic multipliers via associated flow rule in the following manner,
where Q* is the exact stress solution. This inequality can betransformed as follows by spatial discretization,
389
Linear programming formulation. Let us consider upper bound analysis for this boundary value problem. After some calculation of equation (21), the upper bound of a loading factor m is evaluated as
By noting that the external plastic work rate Pexr and the internal dissipation energy rate Witit are homogeneous in the first order of velocities, a loading' factor is independent of the amount of the velocities themselves. Therefore, the additional condition of GN . f = 1 doesn't loose generality in the calculation of a loading factor. Thus, LP formulation of the upper bound method is derived as follows, h
. i -+min
zi
zi
{. & }
=
Winr ( t ) d t
Finally, the upper bound theorem is integrated by time to estimate the upper bound value of a load factor m,
i
--+
j", T=T
Win,
The same assumption is applied for the calculation of external plastic work We,, . Consequently,
subject to
Moreover, as nodal velocity vector & is formally di= -ziG, vided into two non-negative variables; iN where 3 0and z i 2 0, a standard form LP can be derived.
[@,9'. 0'$1
-
Then, internal dissipation Wint done during t = T is considered. It is assumed that kinematically admissible velocities @ are constant with time to avoid the time domain integration and save the amount of calculations. This procedure is different from the ordinary elasto-plastic analysis, where step by step integration is used in the time domain.
0
min
subject to
Additionally, it is convenient to normalise the external plastic work We,, = 1 in the calculation, as game as the ordinal upper bound analysis. Moreover, if a hardening coefficient tensor is positive definite, equation (29) is concluded tothe quadratic programming.
h 0,L i ; 20,bi;
(24) 30 where vectors noted here mean column vectors, and superscript t means transpose of column vectors, i.e., row vectors.
-3
2.4 Proposed calculation procedure
2.3 Extention to work hardening materials Extention of limit analysis to rigid-work hardening materials is discussed in this section. For the simplicity of the discussion, a linearly hardening model with piecewisely linearised yield functions is considered. According to the linearly hardening model, strengths 12 are linear to the magnitude of plastic deformation, i.e., a plastic multiplier h (Konig 1976). where tensor is hardening coefficient tensor and vector myveclzois an initial strength. By noting that a plastic multiplier h is a monotonous increasing function with time, a strength vector & at time t can be obtained by the following integration. OT
*T
-11,
+ho = E
&dt
+ho
(26) 390
No limit of plastic deformation for the linearly hardening model is given in the previous section. Accordingly, a strength can be increased infinitely. This is far from the real behaviour of materials. Therefore, selection of a time parameter T is important to evaluate the stability rationally. Physical meaning of a time parameter T is explained here. Though parameter T holds a time dimension, it is not necessarily a physical time itself. However, a product of T and a corresponding velocity field must be a displacement. It should be reminded that absolute values of a velocity field are meaningless due to the first order homogeneity of velocities in both the internal dissipation rate Wit,[ and the external plastic work rate Similarly, the absolute value of time parameter T is meaningless. It should be emphasised again that a product of a time parameter T and its corresponding velocity field must be a physical displacement. As pointed out previously, the additional stability criteria of a deformation, which ensures that no
Sheetpile
work-softening behaviour is occurred in anywhere of a domain, is applied to estimate the stability of soil structure-reinforcement members system, as follows.
R
/--Spring
A
1 Assume initial time parameter 7;:. 2 Evaluate the optimum velocity field based on the quadratic programming equation (29).
3 Calculate strains and displacements in all the region by the product of the obtained velocity field and the assumed time parameter.
4 Check the strains and displacements in all the region to ensure that they &e within threshold values. a) If strains or displacements are more than the threshold values, modify a time parameter 7;.+1 = Ti - 6T to return to step 1. If b) strains and displacements are all with the threshold values, modify a time parameter 7;.+1 = 7;: - 6T and check the convergence of the calculation.
*
A load factor in is converged. Stop the
*
calculation. A load factor m is not increased, Return to step 1.
It should be noted that strains and displacements are checked after the optimisation of the quadratic programming. Only kinematically and plastically admissible velocity fields are required for the upper bound analysis. If additional restrictions of the deformation were applied in the optimisation process, its solution might be overestimated.
2.5
Practical interpretation of the proposed method
Practical interpretation of the proposed method is discussed in this section. The proposed method does not require a large amount of calculations in comparison with an evolutional (step by step) method. This is a big advantage for numerical calculations. An assumed velocity field is constant with time to avoid the time integral calculations. This assumption might be applicable, if a collapse mechanism is not changed drastically with time. If the assumption of the positive definiteness of hardening coefficient tensor is applicable, numerical calculations are stable dueto the convexity of the analysis. A load factor, a time parameter and its conesponding optimised velocity field are obtained by this method. A distribution of displacements can be estimated from these results. A distribution of displacements is good information on the practical management, as displacements can be measured easily in the
Sub sheet pile
Figure 3. Example boundary value problem.
practical sites. Member forces of reinforcements can be also estimated. A distribution of these member forces is good information to consider the locations of the reinforcement members. The proposed method is also applicable to the elasto-plastic problems, if only loading processes are considered. Elasto-plastic problems under monotonic loading conditions are essentially same to the rigidplastic problems with zero initial strengths. It is observed in some practical cases that reinforcement members are still under the elastic state, although surrounding soils are in the plastic state. However, the proposed method is applicable to this situation.
3 NUMERICAL EXAMPLE A very simple example is considered to demonstrate the proposed method. The stability problem of a selfstanding sheet-pile wall is considered. A sheet pile with plastic bending moment M , is installed sufficiently to the supporting layer. Soils behind the wall is c , @ = 0 material with a density p. An elastic spring with the elastic modulus k is pin-connected to the top of the sheet pile. The other end of the spring is firmly supported and its influences are negligible. Frictional effects between soils, a sheet-pile and a spring are also negligible. It should be calculated for all the kinematically admissible velocity field to obtained the best optimised solution. However, only one deformation mode shown in figure 3 is considered here for the demonstration. A velocity field of the soil for this mode is
vx = -@(y - tana . x),
Then, strain rates are calculated as
i., = -@ tana, i., = @ tana,
9 = 2iXy= -e(l
+tan")
(31)
The internal dissipation is occurred at the soils, the sheet pile and the spring. On the other hand, the external plastic work is done by the self weight of the soils.
39 1
I
l
l
I \ \
'
\
\
'
'
a
\
\
'
'
. . I '
2
1
of limit analysis is discussed in this paper. The conclusions of this paper is summarised as follows. The extention of the upper bound analysis to rigid-work hardening materials is presented. The formulation is concluded to the quadratic programming.
0'
0
"
0.05
0.1
*
"
"
0.15 0.2 0.25 0.3 0.35 Magnitude of deformahon: T6
"
0.4
0.45
The evaluation method of the stability of soil structures and reinforcement members system is proposed based on the quadratic programming and the additional deformation conditions of the threshold values for strains and displacements.
'
0.5
Figure 4. Numerical result.
The proposed method does not require a large amount of calculations in comparison with an evolutional (step by step) method.
By optimising about the parameter a which governs the area of plastic deformation, the stational point is at the angle of tan a = 1. Therefore, non-dimensional critical height of the sheet pile is a solution of the following cubic equation.
A load factor, a time parameter and its corresponding optimised velocity field are obtained by this method.
where 1 = c / ( p g ) is a specific length, h = H c r / l is a non-dimensional sheet pile height, m = M Y / ( c l 2 )is a non-dimensional plastic bending moment of a sheet pile and K = k / c is a representative elastic modulus of a spring. A numerical result in case of a non-dimensional plastic bending moment m = 10 is shown in figure 4. The horizontal axis means the magnitude of plastic deformation TO. This value coincides with the magnitude of plastic (engineering) shear strain of the soil behind the wall. Parameter 6 shown as thin lines in figure 4 is non-dimensional horizontal displacement at the top of the sheet pile normalised by a specific length 1. It is clearly observed that non-dimensional height magnitude of deformation relations depend on the elastic moduli of springs. If additional conditions for the threshold strains or displacements are given, the critical height of the sheet pile for this mode can be evaluated from figure 4.
-
4 CONCLUSIONS Theoretical methodology to evaluate the safety of soil structures with reinforcement members from the point
392
The proposed method is also applicable to the elasto-plastic problems, if only loading processes are considered. REFERENCES Hill, R. (1948). A variational principle of maximum plastic work in classical plasticity. Quart. J. Mech. Applied Math. 1 : 1828. Kobayashi, S. (2000a). Limit analysis of sheet pile type retaining walls. In Proc. of the Int. Symp. IS-Yokohama 2000, Volume 1: 315-320. JGS. Kobayashi, S. (2000b). Teaching earth pressure problems and stability problems from the point of applied mechanics. In Proc. 1st Int. Con$ Geotechnical Engineering Education and Training, 289-296. Konig, J. A. & Maier, G. (1976). Adaptation of rigid-workhardening discrete structures subjected to load and temperature cycles and second-order geometric effects. Computer Methods in Applied Mechanics and Engineering 8,37-50. Lloyd Smith, D. (Ed.) (1990). Mathematical programming methods in structural plasticity. Springer-Verlag. Maier, G. (1970). A matrix structural theory of piecewise-linear plasticity with interacting yield planes. Meccunica 5( I), 5566. Maier, G. & Lloyd Smith, D. (1986). Update to mathematical programming applications in engineering plastic analysis. In Applied Mechanics Update 1986, pp. 377-383. ASME. Maier, G. & Munro, J. (1982). Mathematical programming applications to engineering plastic analysis. Applied Mechanics Reviews 35,1631-1643.
Landmarks in Earth Reinforcemenf, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Seismic earth pressures acting on reinforced-soil and conventional type retaining walls Junichi Koseki Institute of Industrial Science, University of Tokyo, Japan
Kenji Watanabe, Masaru Tateyama and Kenichi Kojima Railway Technical Research Institute, Japan
ABSTRACT: Comparisons are made on the resultant force of normal earth pressures that were measured in a series of model tests on six different types of soil retaining walls with a height of 530 or 500 mm, backfilled with dense dry sand. They consist of three reinforced-soil retaining walls with a full-height rigid facing and three conventional (cantilever, gravity and leaning) type retaining walls. The resultant forces of normal earth pressures measured in tilting tests were, in a broad sense, comparable with theoretical ones based on the Mononobe-Okabe or its equivalent method. On the other hand, the resultant forces measured in sinusoidal and irregular shaking tests were smaller than those measured in the tilting tests. However, by making corrections for the horizontal and vertical response accelerations of soil wedge in the backfill, the measured values became much close to the theoretical ones, in particular, in the region at high seismic loads.
2, the length of the top and fourth reinforcement layers was increased to 800 mm and 450 mm, respectively. The length of all the reinforcement layers was increased to 350 mm for the reinforced-soil wall type 3 . A grid of phosphor-bronze strips was used as the model reinforcement. To form a model grid reinforcement layer, strips having a thickness of 0.1 mm and a width of 3 mm were soldered to each other at an interval of 50 mm in the longitudinal direction, in parallel with the side wall, and 100 mm in the transverse direction, in parallel with the facing, as shown in Fig. I . To effectively mobilize friction between the reinforcement and the backfill, sand particles were glued on the surface of the strips. The details of the model wall and reinforcement configurations are described in Koseki et al. ( I 998a).
1 INTRODUCTION In order to establish practical design procedures to evaluate seismic stability of different types of retaining walls (RWs) against high seismic loads, a series of shaking table tests with irregular wave were conducted by Watanabe et al. (2001) on RW models consisting of six different types. This series of tests followed the previous series of tilting tests and sinusoidal shaking tests conducted on the same types of RW models, as reported by Koseki et al. (19984 1999). In the present paper, comparisons are made on the resultant force of normal earth pressures that were measured in these model tests. 2 TEST PROCEDURES 2.1 Model walls
Phosphor-bronze strips (0.1 mm thick) with glued-on sand particles (soldered at each overlap)
The cross-sections of RW models are shown in Fig. 2 of Watanabe et al. (2001). All the models were 600 mm in width. The cantilever type and gravity type RWs were 530 mm high with vertical back face and base footing width of 230 mm. The leaning type RW was also 530 mm high, while it had inclined back face and base footing width of 180 mm. The three types of reinforced-soil RWs were 500 mm high with a full height rigid facing having different arrangements of reinforcement layers. For the reinforced-soil wall type I, ten layers of reinforcement strips having a length of 200 mm were horizontally placed in the backfill sand at a vertical spacing of 50 mm. For the reinforced-soil wall type
pb
_,i __________
-r ~
I
200mm I
i Figure I . Plan of model grid reinforcement layer
393
..... 7. ..
h
E
Tilting Test -a- C, Cantilever -0- G, Gravity -A- L, Leaning R I , Reinforced-I -0-- R2, Reinforced-2 -v-R3, Reinforced-3
90
E 80
v
a
-v-
70
F" 60 Sinusoidal Shaking Test U C, Cantilever 0 G, Gravity A L, Leaning R I , Reinforced-I 0 R2, Reinforced-2 7 R3, Reinforced-3
8
v
Ecn
40
5 30
Irregular Shaking Test E8 C, Cantilever EI G, Gravity $- L, Leaning 'iQ R I , Reinforced-I d R2, Reinforced-2 Y$ R3, Reinforced-3 I
50
a.
9 -
2 I
20
10 0 0.0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0 1.1 1.2 1.3
Seismic coefficient, k,=tan 8 =amax/g Figure 2 Accumulation of residual horizontal displacement near the top of the wall.
Shear load in the vertical direction and normal lateral load acting on the backface of wall were measured with a number of small two-component loadcells that were set on the back of wall along its center line, as shown in Fig. 3 of Watanabe et al. (2001). By using a piece of sponge covered by Teflon-sheet and smeared with silicone grease, the friction between the side of the wall and the side wall of the sand box was reduced, and the sealing between them was also achieved. 2.2 Backfill and subsoil material Air-dried Toyoura sand, having enlux= 0.977; en7,,,= 0.605; G,$= 2.64; D ~ =o 0.1 1 mm and D ~ o 0.23 = mm, was pluviated through air to form the backfill and subsoil layers at a void ratio of about 0.650 (Dr=S80/,). A series of plane strain compression (PSC) tests under low confining pressure of 9.8 kPa, which were performed on the same batch of Toyoura sand prepared at the same density, revealed that its peak and residual angles of internal friction b e a k and heswere 5 1 and 43 degrees, respectively. After finishing the pluviation of sand, the surface of the backfill was trimmed to the prescribed geometry, and a surcharge of 1 kPa was applied by placing lead shots on the surface of the backfill to simulate such a structure as the railway ballast fill. To separate sand from the lead shots, 0.2 mm-thick rubber membranes were placed between them.
2001). About 50 cycles of sinusoidal waves at a frequency of 5Hz were employed as the sinusoidal base acceleration. Each model was subjected to several shaking steps, where the maximum amplitude of the base acceleration was increased stepwise at a prescribed increment. Results from these shaking tests were compared with those from tiliting tests, where seismic loads were applied by tilting the whole sand box at a rate of approximately 1.O degredminute to simulate pseudo-static loading conditions. Based on the pseudo-static approach, the observed seismic coefficient kJfin the tilting tests was defined as;
where 6 is the tilting angle of the sand box. On the other hand, the observed seismic coefficient k17in the shaking table tests was defined as;
where a,7luxis the single amplitude of maximum base acceleration at the active state (i.e., when the inertia force of the backfill is acting outward) for each shaking step, and g is the gravitational acceleration.
3 TEST RESULTS AND DISCUSSIONS
2.3 Application of seismic loads
3.1 Residual displacement of wall
Seismic loads were applied by shaking the sand box horizontally with an irregular or a sinusoidal base acceleration. The irregular base acceleration was prepared based on a strong motion that was recorded as N-S component at Kobe Marine Meteorological Observation Station during the 1995 HyogokenNanbu earthquake (refer to Fig. 4 of Watanabe et al.,
Relationships between the seismic coefficient kh and the horizontal displacement dropmeasured at a distance of 5 cm below the top of the wall are shown in Fig. 2. For the shaking tests, the values of dlot at the end of each shaking step are plotted. In the sinusoidal shaking tests as well as the tilting tests, after exceeding about 25 mm, which corresponds to about 394
where khl x W is the horizontal inertia of the soil block located above the wall base and separated by the vertical failure plane from the remaining part of the backfill (i.e., W is the weight of this soil block, and kjll is the measured horizontal response acceleration ab of this soil block divided by the gravitational acceleration g for the shaking table tests); T and khl x Ware defined positive when they act in the direction toward the facing (i.e., at the active state). In this case, theoretical relationships with b = S= 4 e a k and 4 = S= hes are added to the figure, since the frictional angle S a t the vertical failure plane can be assumed equal to qi
5% of the total wall height, the dropvalue increased very rapidly, soon resulting into the ultimate overall wall failure. In the early steps of irregular shaking tests up to the kh value of about 0.5, the dropvalue accumulated in a similar manner among different types of RWs. When the kh value exceeded about 0.5, however, the rate of increase in the dto1,value became larger for the three conventional type RWs than for the three reinforced-soil type RWs. Such different extents of ductility that depend on the RW type are discussed elsewhere (Watanabe et al., 2001). 3.2 Resultant force of normal earth pressures Relationships between the resultant force P O acting normally on the facing from the backfill and the seismic coefficient ki, are shown in Fig. 3. The P a values are evaluated by integrating normal stresses measured with loadcells along depth of the facing, which include initial values measured before the start of shaking or tilting. For each irregular or sinusoidal shaking step, the P, value was defined under three different conditions; i.e., when either one of the P, value itself, the wall top displacement drol,,or base acceleration (on the negative side, inducing outward inertia force) becomes respective peak state. The kiz values are evaluated based on Eqs. (1) and (2). Note that for the tilting tests, the measured values of the normal stresses at tilted conditions were corrected for the effects of the sand box inclination by a factor of l/(cos 0), where 8 is the tilting angle. In Figs. 3a through 3c, theoretical relationships based on the Mononobe-Okabe method are shown, while in Figs. 3d through 3f, those based on limitequilibrium stability analysis assuming a two-wedge failure mechanism (Horii et al., 1994) are presented. In obtaining these relationships, the shear resistance angle q5 of the backfill was set equal to 4 e a k (=51 degrees), and the frictional angle 6 at the interface between the backfill and the wall facing with sand paper was set equal to 3/4q$,& (Koseki et al., 1998a). For comparison, the residual condition of 4=hes(=43 degrees) and S = 3/46.,, was also employed in the calculation. For the cantilever type RW, the resultant forces measured at the backface of the wall cannot be directly compared to the calculated values, because the calculated resultant forces are those acting on the vertical failure plane in the backfill, which was actually observed to develop from the heel of the wall base. Therefore, the resultant force P a acting on this vertical failure plane was estimated from the measured values of the normal force P,l acting on the backface of the facing and the shear force T acting on the top of the wall base from the backfill as:
00
02
04
06
10
08
Seismic coefficient, k,
I Foimation offailure /I Sinusoidal11 Irregularshaking I I Irregular shaking 1 J
-
2 , ' o a t '0 0
'
02
5
'
'
04
L-m-4-c06
atpaktaseacc 08 10
I
U
Seismic Coefficient, k,
1
Fcimirior, or failwe
/S;nusoldail I irregular shaking I I
-D-l--5-
a,
00
02
04
06
Irregular shakir
at peak base acc
08
10
Seismic coefficient, k, P O
=Pal
+ T - khl X W
(3)
Figure 3 Relationships between resultant normal force and seismic coefficient
395
I
1
Formation of failure
lSlnusoldal1
Jlrregularl
h
z
1.6
1.4
.-@
1.2
2
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-m
I
s
0.8
0.6
a,
2 0.4
,o w c
U -0- peak value of P,
0.2 -m-f
$
00
02
at peak base acc
-0-
06
04
08
10
Seismic coefficient, k,
I I
Forniation of failure
. $
I llrregularl
Sinusoidai shaking
h
E
1.6
Q"
1.4
6,
.g
1.2
Q 0
1.0
g c
b
0.8 0.6
a,
$
0.4
$J
0.2
vc
3
d . .
-D-/ I
I
00
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c
I
I
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,
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-n-
at peak base acc.
,
I
I
I
, ' I 12
10
08
Seismic coefficient, k,
1
I
Foimation offailure
could be defined as the state either when the failure plane is about to develop (for q$ = beak), or after the failure plane has developed in the backfill, where the q$valueshas dropped to bes. Phase difference in the shaking tests in the vertical distribution of horizontal response accelerations of backfill would be one of the reasons for the different P, values from the tilting tests. In addition, as discussed by Tatsuoka et al. (1998), different from the case of tilting tests, the earth pressure acting on the back of the facing in the shaking tests is controlled largely by an interaction between dynamic response of the backfill and the wall structure. In fact, the P, values defined under the three different conditions as mentioned above were, in general, different from each other. In Fig. 4, the peak horizontal response acceleration (ah)nl,xin the backfill inducing outward inertia force was compared with the peak base acceleration anluxin the irregular shaking tests. The values were evaluated based on the records of an accelerometer located near the mass center of the soil wedge (in the unreinforced zone for reinforced-soil RWs) above the failure plane. With the increase in the shaking level, the (a17)m,x value became gradually smaller than the arrlUx value. In particular, after the failure plane was formed in the backfill, the rate of increase in the (ah)n2,xvalue was significantly reduced, or even the increase stopped temporarily, due possibly to sliding of the soil wedge along the failure plane.
lSilusoidal1 oil wedge in backfill
9.
U.
00
02
04
06
08
10
12
1
Formation of failure plane
h
14
Seismic coefficient, k,
Figure 3. (continued) Relationships between resultant normal force and seismic coefficient. V
$ 1 2 1
It can be seen from Fig. 3 that, in general, the P, values measured in the tilting tests were larger than those measured in the sinusoidal or irregular shaking tests. In a broad sense, the results from tilting tests were comparable with the theoretical ones, except for the leaning type RW. It should be noted, however, that the direct comparison of the measured values with those calculated by the Mononobe-Okabe or its equivalent method, be valid at the active failure state in the backfill. The active failure state
14 12 10 08 06
2
0
5
0
2 0
04
02 00
J = O
m
Y
R
O
Peak base acceleration, amax (G)
Figure 4. Relationships between peak horizontal response acceleration in backfill and base acceleration.
3 96
For gravity type and reinforced-soil type 1 RWs, results from sinusoidal shaking tests are also shown in Fig. 4. Note that, using 20 cycles of sinusoidal waves, these shaking tests were additionally conducted on limited types of RWs. In these tests, a noticeable amplification of the response acceleration took place before the formation of the failure plane, in contrast to the attenuation in the response observed in the irregular shaking tests. On the other hand, after the formation of the failure plane, a sudden reduction in the response acceleration took place in the sinusoidal shaking tests, which may also be due to sliding of the soil wedge along the failure plane. It should also be noted that, the horizontal response of the soil wedge above the failure plane was accompanied by its vertical response, as typically shown in Fig. 5. In this case, the first failure plane had been already formed in the backfill during the previous shaking steps, and several large cycles of horizontal base acceleration induced relatively large response of the soil wedge not only in the horizontal but also in the vertical directions. The peak horizontal response acceleration was mobilized after a certain phase lag after the peak base acceleration, and, as mentioned above, the value was smaller than the ammvalue. When outward inertia force was acting on the soil wedge, it was also subjected to vertical upward inertia force (i.e., downward acceleration) in the beginning, which was reversed into the downward inertia force (i.e., upward acceleration) in the later stage.
This peculiar behavior could be explained qualitatively by considering the sliding of the soil wedge along the failure plane as follows: The broken curve in Fig. 5b is the base acceleration. When the soil wedge started sliding (after point A in Fig. 5), its horizontal response acceleration became smaller than the base acceleration. At the same time, it slid down along the failure plane with negative (downward) vertical acceleration (between points A and B in Fig. 5a). Since reversal of the base acceleration took place, the sliding of the soil wedge was terminated eventually (at point C in Fig. 5a). Before the termination, the sliding movement was decelerated with positive (upward) vertical acceleration (between points B and C in Fig. 5a). The point B’ in Fig. 5b is the point after which the horizontal response acceleration of the soil wedge became larger than the base acceleration (i.e., when the relative horizontal acceleration of the soil wedge to the base was reversed). It was slightly different fiom point B (when the vertical acceleration of the soil wedge was reversed) in Fig. 5a, due possibly to that the horizontal response acceleration in the underlying non-sliding soil mass was not equal to the base acceleration. Similarly, the point that corresponds to point C in Fig. 5a (after which the horizontal response acceleration of the soil wedge became equal to the base acceleration) could not be clearly defined in Fig. 5b. In Fig. 5, the peak horizontal response acceleration was mobilized while the sliding movement was decelerated (between points B and C). In some of the other cases, however, the peak horizontal response acceleration was mobilized while the sliding movement was accelerated (between points A and B). In Fig. 6, correction for the effects of horizontal and vertical responses of the soil wedge during the irregular shaking was made for some RW models on the seismic coefficient kh and the measured resultant force P,, respectively. The kh value was evaluated from the (a&,m value. The P, value was obtained at the moment when (a~t)rTlux was mobilized, and it was corrected by dividing with a factor of “1 +a,/g”, where a,,is the vertical acceleration of the soil wedge obtained at the same moment as above (defined as positive when it induces downward inertia force). The corrected relationships are represented by using open symbols in Fig. 6. For reference, measured relationships between uncorrected kh and P, values that were obtained at the moment when the base acceleration became its peak (i.e., when arllux was mobilized) are plotted by using solid symbols, and the aforementioned theoretical relationships are also shown. It can be seen that, by making a correc-
Leaning type RW during irregular
a, UI U -
%$-0.8 -0.6
Broken base acc
b
I
.&
4.0
4.2
4.4
4.6
4.8
5.0
Time (sec)
Figure 5 Typical response acceleration of soil wedge above failure plane for leaning type RW.
397
.
h
a
-
2 04-
2Z
K
ace , kh=amJg
O O '
00
'
02
'
'
04
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06
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08
'
.
I
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Seismic coefficient, k,
Seismic coefficient, k, h
;--
rination of
a,
00
02
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04 06 Seismic coefficient,
08
2
10
00
02
k,
04 06 08 10 Seismic coefficient, k,
12
14
Figure 6 Effects of correction for response of soil wedge on relationships between resultant normal force and seismic coefficient
making corrections for the horizontal and vertical response accelerations of soil wedge in the backfill, the measured values became much closed to the theoretical ones, in particular, in the region at high seismic loads.
tion to the response of the soil wedge, the measured relationships became much closer to the theoretical ones, in particular, in the region at high seismic loads. In summary, the experimental data supports the overall trend of the Mononobe-Okabe method. In addition, sudden increase in the corrected P, values after formation of the first failure plane as shown in Figs. 6% 6b and 6d can be explained by the modified Mononobe-Okabe method as proposed by Koseki et al. (1998b), considering the effects of strain softening and strain localization. However, the detailed quantitative evaluation of the original or modified Mononobe-Okabe method was not possible, because of the delicate nature of dynamic earth pressures. It is readily seen that the reinforced-soil RWs could stand without exhibiting ultimate failure against earth pressures that were much higher than those acting on the conventional type RWs.
REFERENCES I-Iorii,K, Kishida,H , Tateyama,M and Tatsuoka,F 1994 Computerized design method for geosynthetic-reinforced soil retaining walls for railway embankments, Recent Case Histones of Permanent Geosynthetic-Reinforced Soil Retaining Walls, Balkema 205-21 8 Koseki,J , Munaf,Y , Tatsuoka,F , Tateyama,M , Kojima,K and Sato,T 1998a Shaking table and tilt table tests of geosynthetic-reinforced soil and conventional retaining wall, Geosynthetrcs International, Vol 5, Nos 1-2 73-96 Koseki,J , Tatsuoka,F , Munaf,Y , Tateyama,M , and Kojima,K 1998b A modified procedure to evaluate active earth pressure at high seismic loads, Soils and Foundatrons, Special Issue, No 2 209-2 16 Koseki,J , Munaf,Y, Tateyama,M , Kojima,K and Horii,K 1999 Back analyses of case histories and model tests on seismic stability of retaining walls, 11th Asian Regional Conf on Soil Mechanics and Geotechnical Engineering, VOI 1 399-402 Tatsuoka,F , Koseki,J , Tateyama,M , Munaf,Y , Horii,K 1998 Seismic stability against high seismic loads of geosynthetics reinforced soil retaining structures, 6th International Conference on Geosynthetics, Vol 1 103-142 Watanabe,K , Tateyama,M , Kojima,K and Koseki,J 2001 Irregular shaking table tests on seismic stability of reinforced-soil retaining walls, this symposium
4 CONCLUSIONS The resultant forces of normal earth pressures measured in the tilting tests were, in a broad sense, comparable with theoretical ones based on the Mononobe-Okabe or its equivalent method. On the other hand, the resultant forces measured in the sinusoidal and irregular shaking tests were smaller than those measured in the tilting tests. However, by 398
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Influence of reinforcement's inclination on bearing capacity of RS wall M. Kulczykowski Institute of Hydroengineering Polish Academy of Sciences, Gdansk, Poland
ABSTRACT: The paper presents the experimental results obtained on the small scale models of RS retaining wall loaded on its crest with the footing. The influence of reinforcement's inclination on a bearing capacity and a orientation of failure surface is tested. The data are compared with the theoretical prediction based on limit states theorems. It is shown that the reinforcement's inclination reduces the bearing capacity of the structure.
1 INTRODUCTION
limit states techniques is shortly described. It is shown that the structure reinforced horizontally is the strongest, and that the inclination of reinforcement reduces the bearing capacity of RS retaining wall.
In the most of reinforced soil structures the reinforcement is placed in horizontal layers. However, in soil nailing construction the inclination of the reinforcement is always recommended. There is a lack of the papers analyzing an influence of reinforcement inclination on a bearing capacity and failure mechanism of RS (Reinforced Soil) structures in a complex manner. A few publications on that subject present contradictory results. Kitamura et al. (1988) have been describing the model tests of RS retaining wall reinforced horizontally and with reinforcement placed at the angle of 20' to the horizontal. They noted that the bearing capacity of construction is slightly decreased for inclined reinforcement in comparison with horizontal one. Juran et al. (1990) have presented the results of the similar model tests. They did not state the relevant influence of reinforcement inclination on critical height of RS construction. Huang et al. (1990) have considered the effect of reinforcement direction on bearing capacity of RS slope. They pointed out that the most effective direction for reinforcing members was the one which coincided with the direction of the minor principal strain in the failure zone. In numerical analysis Bang et al. (1992) confirmed that reinforcement inclination at the angle of 5+20' was the most effective. On the other hand, Lesniewska (1992) using the method of characteristic showed that the horizontal placement of reinforcement in RS wall was optimal. The similar conclusion was drawn by Sabhahit et al. (1995) and Sawicki (2000). Herein are presented results of experiments in which the influence of reinforcement's inclination on bearing capacity of RS retaining walls is tested. In the next step the theoretical analysis based on the
2 MODEL TESTS 2.1 Experimental method The model walls were constructed in strong box with an inner dimension of 66 cm long, 50 cm high and 26 cm wide. To reduce the friction effect the frontage sidewall was made of glass and the opposite sidewall was covered with a smooth aluminum plate. Each model wall was 32 cm in height and was reinforced with ten layers of reinforcement, which were placed at equal vertical spacing of 3.2 cm. The configuration of the model is shown in Figure 1 .
Figure 1. Configuration of the model wall.
399
The reinforcement used in the models consisted of aluminum strips 2.5 cm wide, 18 pm thick and 30 crn long. In each layer three strips were connected to the cardboard panels used as the wall facing (see Figure 2). The backfill and the foundation were constructed from the sand that was rained through air by using the hopper kept at 100 cm high from the sand surfaces. The models were built on a 15 cm thick foundation. A temporary support in the form of wooden plate and platform was positioned on the top of the foundation soil in front of the wall face to keep the facing in place during construction. To obtain the required inclination of reinforcement, each of sand layer was flattened with the grader. Then the layer of reinforcement was placed on the exposed portion of sand. Next layer of soil was placed, in turn, on it, and this process was repeated for successive layers, until the model wall reached the desired height. The temporary support was then removed. The position of reinforcement was marked near by the glass sidewall using the thin layer of colored sand. It was useful to detect the failure surface in the model wall.
2.2 Material properties The soil used as a backfill and a foundation was silica sand. Results of triaxial compression tests indicated that the soil exhibited a friction angle 4 of 3 1' at confining stresses within the range expected in the models. The sand was rained through air under controlled condition to a dry density yof 17.3 kN/m3. Results of tensile tests of reinforcing strips indicated that the plastic limit of reinforcement R was
400
41.6 x 103 kN/m2 and the elongation at break was 4.1%. No test was performed on the interfaces between the strips and the model soil. 2.3 Results of the model tests Five tests on uniformly reinforced model walls were conducted using different angle a of reinforcement inclination with respect to the horizontal: a=O', 5', loo, 15' and 20' (see Figure 3). All walls were loaded at their crests using the 9 cm wide smooth rigid strip footing. The footing base was loaded at a constant vertical displacement rate of 3.3 mm/min, until model failure occurs. The applied load and vertical displacement of footing were measured using data acquisition system. The deformation of the wall during loading was observed through the glass sidewall. The failure mechanism and the orientation of failure surfaces, detected by means of colored sand were recorded by photo camera. The failure zones in model constructions are presented in Figure 4. For all tests the same failure mechanisms were observed. The failure was developed along the surface from the footing edge to an intermediate point along the wall height. This mechanism was associated with tensile failure of the reinforcement. It was shown that the reinforceinent inclination strongly influenced the height of failure zones. The experimental results of the critical load, footing displacement measured at failure and the height of the failure zones are presented in Table 1, Table 2 and Table 3, respectively. These results indicate that the structure reinforced horizontally is the strongest. The increase of the angle of reinforcement inclination reduces the bearing capacity of RS model and enhanccs the footing displacement at failure and the height of the failure zones. It should be noted that for little value of angle a the decrease of critical load is rather small, but for the angle a > 10' the reduction is significant. The bearing capacity of model wall reinforced horizontally is nearly two times greater than of the similar structure with the angle of reinforcement inclination a=20°. Simultaneously the height of the failure zones is more than two times higher.
3 THEORETICAL PREDICTION In this paper the influence of reinforcement's inclination on bearing capacity of RS retaining structures is analysed using limit states techniques. The homogenised rigid-plastic theory of RS proposed by Sawicki (1983) is applied. It is assumed that the reinforced soil consists of the soil matrix and unidirectional reinforcement, and that the perfect bonding between these constituents exists. The limit states theorems enable estimating so called upper-bounds of critical load of RS retaining wall. The estimate of the critical load can be obtained from the analysis of the kinematically admissible mechanism of failure, shown in Figure 5. The mechanism depends on slippage of the rigid wedge ABC along the planar failure surface AB. The velocity Y of uniform translation is inclined at angle Cp with respect to AB. This inclination results from the assumption about the associated flow rule. The expression for the upper-bound estimate of the critical load can be derived by equating the energy dissipated along the failure surface to the work done by external forces and self-weight forces of RS structure. The following formula represents the upperbound estimate of critical load:
where Cp= angle of internal friction, a= angle between the direction of reinforcement and the horizontal direction, a= width of loading area. The strengthen behaviour of reinforcement is described by the coefficient 00, which indicates the
Figure 4. Failure zones in the model tests
Figure 5. Kinematically admissible mechanism of failure.
401
4 COMPARISON BETWEEN EXPERIMENTAL RESULTS AND THEORETICAL PREDICTION
tensile strength of reinforcement per unit cross section of entire structure:
The comparison between observed and predicted results are presented in Table 1, Table 3 and in Figures 6-7. In Table 1 and in Figure 6 the experimental results and theoretical prediction of the critical load of RS model walls with different inclination of reinforcement are compared.
where T = tensile limit force for single element of reinforcement, A, = cross section area of construction normal to the reinforcement direction, n = number of reinforcement elements. For sheets of geotextile or geogrid GO can be defined as Go=-,
Table 1. Comparison between experimental results and theoretical prediction of the critical load of RS wall.
TG
Inclination of reinforcement
M
deg 0 5 10 15 20
where TG = limit force per unit length of reinforcement and AH = the vertical spacing. For reinforcing bar or strips, GO can be taken as
Experimental data kN/rn2 27.5 26.8 26.1 21.9 15.2
Theoretical prediction kN/m’ 20.1 19.6 18.5 16.6 13.9
Table 2. Experimental results of the footing displacement measured at failure.
where TB= tensile limit force of a single bar or strip and A7 and AV denote horizontal and vertical spacing, respectively. The value of unknown angle p can be obtained from the expression:
Inclination of reinforcement deg 0 5 10 15 20
(5)
Experimental data mm 6.7 9.1 9.4 13.5 11.1
which leads to the following result: 1
1
co sin2p - - tan p sin(p + 4) cos2a [ 2 +Go{ I tan p sin 2cr[sin’ p + cos2(p +
01)
1
+--yacos2(p++)tanp= 0. 2 From the analysis of Eqs.(l) and Eqs.(6) it follows that there should be:
Due to the complicated form of the expression (6) the Of p can be Obtained for the experimental data.
402
Figure 6. Comparison between experimental data and theoretical results of the critical load of RS wall.
Moreover the analysis based on the limit states theorems is presented. With regard to the experimental and theoretical results it is interest to note that: 1. The RS retaining wall reinforced by the horizontally placed layers is the strongest and that the reinforcement inclination reduces the bearing capacity of the structure. 2. The inclination of reinforcement strongly influences the orientation of the failure surfaces. 3 . The kinematical limit approach with the failure mechanism depends on slippage of the rigid wedge along the planar failure surface gives a good estimation of the experimental results. These conclusions support the theoretical results of Lesniewska (1992), Sabhabit et al. (1995) and Sawicki (2000). On the contrary to the other authors' suggestions about the positive effect of the reinforcement inclination on the strength of RS structures, the present study confirms the advantage of the horizontal placement of reinforcement in RS retaining walls.
Figure 7. Comparison between observed and predicted height of the failure zone.
It is shown that the horizontal placement of reinforcement makes the structure the strongest and that the inclination of reinforcement reduces the bearing capacity of RS retaining walls. The difference between measured and predicted critical load (of about 25 9'0) is probably due to boundary friction effect on the sidewall. In Table 3 and in Figure 7 the experimental results and theoretical prediction of the height of failure zone of RS model walls with different inclination of reinforcement are compared. It is shown that the lowest value of the failure zone height corresponds to the horizontal reinforcement and that the inclination of reinforcement strongly influences the orientation of the failure surfaces. The experimental heights of the failure zone are approximately equal to that obtained from the theoretical prediction.
REFERENCES Bang, S., Kroetch, P.P. & Shen, C.K. 1992. Analysis of soil nailing system, In Proc. Earth Reinforcement Practice: 457-462. Rotterdam: Balkema. Bridle, R.J. 1989. Soil nailing - Analysis, Ground Eng.: 52-56. Elias, V. & Juran I. 1985. Manual of Practice for Soil Nailing, USA Dip, FHA Contract DTFH-61-85-C-00142. Gassler, G. & Gudehus, G. 1989. Soil nailing - some aspects of a new technique, In Proc. Tenth ICSMFE: 665-670. Stockholm. Gray, D.H. &Ohashi, H. 1983. Mechanics of fiber reinforcement in sand, J Geotech. Eng. ASCE 109 : 335-353. Huang, C., Tatsuoka, F. & Sato, Y. 1990. Reinforcing a sand slope supporting a footing using steel bars, In Performance of Reinforced Soil Structures: 323-329. London: Thomas Telford,. Jewell, R.A. & Wroth, C.P. 1987. Direct shear tests on reinforced sand, Geotechnigue 37 (1): 53-68. Jewell, R.A. & Padley, R.J. 1990. Soil nailing design - The role of bending stiffness, Ground Engineering, March: 30-36. Juran, I., Baudrand, G., Farrag K. & Elias V. 1990. Kinematical limit analysis for design of soil-nailed structures, J. Geot. Eng. ASCE, 16 (1): 55-72. Kitamura, T., Nagao A. & Uehara S. 1988. Model loading tests of reinforced slope with bars, In Proc. Earth Reinforcement Practice: 3 1 1-316. Rotterdam: Balkema. Lesniewska, D. 1992. Influence of reinforcement's inclination on slip line and bearing capacity of reinforced soil steep slopes, In Proc. Earth Reinforcement Practice: 493-496. Rotterdam: Balkema. Sabhahit, N., Basudhar, P.K. & Madhav, M. R. 1995. A generalized procedure for optimum design nailed soil slopes, Int. Jnl for Numerical and Analytical Methods in Geoniechanics Vol. 19: 437-452. Sawicki, A. 1983. Plastic limit behavior of reinforced earth, J. Geotech. Eng. ASCE 109 (7): 1000-1005. Sawicki, A. 2000. Mechanics of Reinforced Soil. Rotterdam: Balkema.
Table 3. Comparison between observed and predicted height of the failure zone. Inclination of reinforcement
Experimental data
deg 0 5 I0 15 20
cm 17 19 22 27 35
Theoretical prediction cm 18 19 24 27 35
5 CONCLUSIONS In the paper the laboratory tests performed on the models of RS retaining walls with different reinforcement inclination are described.
403
Sawicki, A. & Kulczykowski M. 2000. Influence of reinforcement's inclination on bearing capacity of RS retaining walls, In Geotechnics in Civil and Environmental Eng.: 1000-1005. Gdansk: WBWiS PG. Schlosser, F. 1983. Analogies et differences dans le comportement et le calcul des ouvrages de soutenement en terre ar-
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mee et par clouage du sol, Annales ITBTP, Soles et FondationsNo.418: 26-38. Stocker, M.F., Korber, G.W., Gassier, G. 8~(hdehus G. 1979. Soil nailing, In Int. Con$ on Soil Reinforcement: 469-474. Paris: Presses ENPC.
Landmarks in Earth Reinforcernenf, Ochiai ef al. (eds), 0 2001 Swefs & Zeiflinger, ISBN 90 2651 863 3
Effect of facing and construction sequence on the stability of reinforced soil wall Y . Lirn Department of Civil and Geotechnical Engineering, Paichai University, S. Korea
J. Jung & Y . Park Geotechnical Engineering Division, Highway Research Center, Korea Highway Corporation, S. Korea
Y . Suh Department of Transportation Engineering, Hanyang University, S. Korea ABSTRACT: A small-scale reinforced soil wall was constructed in a laboratory to investigate role of the wall facing and the effect of construction sequence on the wall. A fkll continuous facing wall and a blocktype facing wall were introduced for test. These two different facing systems adapted different construction procedures. The model wall was built with geogrid reinforcement, sand, and facings on rigid surface. The model wall was instrumented with earth pressure gages, LVDTs, and strain gages. The experimental results have shown differences in wall behavior related to construction sequence and types of wall facing. It is found in this study that the reinforced soil wall system built with full continuous facing be the safest reinforced soil wall ever compared to the block-type facing wall. Thus, it is recommended that study for the wall system be necessary for further wide usage for the future. removal of the props. Different wall behavior can provide different pattern and amount of wall deformation. In this study, a small-scale model wall reinforced with geogrid adapting different construction sequences was investigated in order to validate usage of full continuous facing wall for reducing wall deformation effectively compared to conventional reinforced soil wall.
1 INTRODUCTION Recently, many types of reinforced soil wall with geogrid reinforcement and block facing have been proposed due to its simplicity of construction and economic construction procedure. However, wall deformations generated during the construction procedure are comparatively larger than those of conventional gravity retaining wall and excavated cut wall such as soil nailed wall (Jones, 1994). This is due to use of different construction procedure and construction materials. The reinforced soil wall is built usually from bottom to top. The soil nailed wall is built from top to bottom. Typically, the reinforced soil wall is vulnerable to deformations. In this study, a reinforced soil wall built with full continuous facing was investigated in order to check possibility of reducing wall deformation. Cardoso and Lopes ( 1996) divided construction procedures of the reinforced soil structures into two typical types. The first one is the so-called common type of reinforced soil wall built from bottom to top with block-type facing, reinforcement and backfill soil. In this type of the wall, tension in the reinforcement is generated from the beginning of construction. On the other hand, construction of the other type of the wall may start with use of paneltype facing with aid of props in front of the wall first. After setting up the propped facing, the wall is backfilled and reinforced from bottom to top. Then the props are to be removed from the wall after backfilling. Therefore, the tension in the reinforcement is to be generated when the wall moves due to
2 DESIGN OF MODEL TEST 2.1 Test apparatus and instrumentation A schematic diagram of the testing apparatus used for the tests is shown in Figure I . The model wall was constructed by adapting two different construction sequences. The testing apparatus consisted of external steel frames and internal soil retainer. The external steel frames include vertical and horizontal loading machines attached on them. The size of the external frames is 2.0m long, 3.0m high, and 0.8m wide. The soil retainer used for making the model wall to be filled with the Jumunjin sand was 1.2m long, 0.8m high and 0.8m wide. The bottom of the soil retainer was rigid steel slab. The vertical load is applied to the wall by a 196kN capacity of linear servo motor and screw gear. Either a constant loading rate or a constant pressure system is available in this loading machine. Total displacements of the wall were monitored by five LVDTs as shown in Figure 2.
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change in earth pressure in process of loading and wall deformation. Strain gauges were also attached on the surface of the reinforcement in order to monitor the generated tensile strains. The measured signals were transmitted to readout box and personal computer for analyzing and saving the data. The model wall was loaded vertically to failure. The failure of the wall was determined when the wall was deformed horizontally to 10% of the wall height (H=O.8m). 2.2 Properties of backfill soil and reinforcement The soil used for the backfill of the model wall was selected as S. Korea’s standard Jumunjin sand. Properties of the Jumunjin sand and of the reinforcement are shown in Table 1. The used reinforcement for all tests was obtained from commercial production (geogrid type). Dry Jumunjin sand was compacted to 77% of relative density by free falling method. Table I . Material properties of Jumunjin sand and geogrid reinforcement.
* T= tensile strength at 5% elongation of reinforcement; ** €=elongation at failure of the reinforcement 2.3 Design of model tests
Flgure 2 Sect~onalview of model walls (a> block facing, (b) continuous full facing
The load that was to the was measured by a load Earth pressure gauges were also embedded behind the wall to monitor 406
All tests were focused on investigation of effects of reinforcement length, facing types, and load intensity on the wall behavior, especially the wall deformation. The vertical load was applied to the wall after the wall completion. The variable parameters for loading were width of loading plate, location of loading plate, and loading intensities. The red data from earth pressure gauges were too irregular to be valuable data. Thus it was ousted. The used facings (0.05m thick) were made of recycled and pressed polyurethane board and have 5% of glass fiber. The size of the full continuous facing was 0.8m wide, 0.8m high and 0.05m thick. The size of the block facing was 0.8m wide, 0.2m high and 0.05m thick. The block facings were inter-connected using small rail and long hole made on the top and bottom of each facing board. However, each block facing moves freely by making the rail to have half-circular section. The reinforcements were connected to the facing using small steel hooks. Material characteristics ofthe facing are presented in Table 2. All tests were performed following the design parameters as shown in Table 3.
3 ANALYSIS OF TEST RESULTS
Table 2. Material characteristics of facing.
Full cont. facing Young's modulus (kN/m2)
Block facing
6.27E6
6.27E6
5.88
5.88
3.1 Wall deflections Figure 3. illustrates how the horizontal wall deflections are different with changing the types of wall facing and the construction sequences. The horizontal wall deflections generated in the block-facing wall were 10 to 17 times those of the full continuous facing wall immediately after the wall construction was completed. Furthermore, the horizontal wall deflections of block facing wall were 4 to 5.7 times those of the full continuous facing wall when the
Section area
Table 3. Design parameters for test.
Facing and reinforcement conditions Test
Facing
Length
no
type
(L)
1
FHP
0.9H
FHP
0.7H
FHP
7H
FHP
0.5H
0.7H
8
1 1 I BP
Loading conditions
width (m) 0.6 (0.75H) 0.6 (0.75H) 0.3 (0.371-1) 0.6 (0.75H)
(0.75H) 0.3
10.0
19.6
39.2
10.0
19.6
39.2
10.0
19.6
39.2
10.0
19.6
39.2
10.0
19.6
39.2
10.0
19.6
39.2
I
I
10.0 19.6 39.2 O'' (0.75H) N.B) FHP: full continuous facing, BP: Block facing, H: wall height (H=O.Sm) 0.5H
The model wall was constructed by following two different construction sequences independently. The block-facing wall was constructed built from bottom to top with block-facing, extensible reinforcement and backfill soil. On the other hand, construction of the continuous full facing wall started with setting full facings and props up first. After setting the propped facing up, the wall is backfilled and reinforced from bottom to top. Then the props are to be removed from the wall after backfilling. Thus the construction sequences adapted in this test were drastically different in these two wall types. In the continuous full facing wall, the facing was not hinged at the bottom of the wall resting on a rigid foundation. Therefore the facing at the bottom of the wall.
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Figure 3. Comparison of horizontal wall deflections. (a) L=0.9H, (b) L=0.5H
vertical load was applied up to 39.2 kN/m2. Thus, difference in wall deflections between the two types of wall facing decreased when the vertical load applied. Nevertheless, the horizontal wall deflections of the full continuous facing were still smaller than those of the block-facing wall. Generated additional net wall deflections measured after the vertical loading applied increased similar proportion in both facing walls. However, the net increase of wall deflection in the block-facing wall tended to increase relatively smaller than those of the full continuous facing wall. This may be due to enough generation of wall deflection during wall construction when in case of the block facing wall. 3.2 Tensilforce distribution in reinforcement The typical generated tensile forces in the reinforcement are shown in Figure 4. to Figure 7. for block facing and full continuous facing, respectively. Intensities of vertical loads were varied from 9.8 kN/m2 to 39.2 kN/m2 and were applied to top of the wall. As shown in the figures, the generated tensile forces in the reinforcement of the block facing wall were generally greater than those in the reinforcement of the full continuous facing wall. In the full continuous facing wall, the generated tensile forces in upper (two) reinforcement were greater than those in lower (two) reinforcement and tended to decrease with depth. In addition, in case of the full continuous facing wall, the maximum tension was found in reinforcement length (L) of 0.7H (not shown here). The generated tension in the reinforcement of the full continuous facing wall increased relatively very small when the wall was vertically loaded after the pros were removed, compared to the block-facing wall. On the other hand, the generated tensile forces in the block-facing wall werc larger in lower reinforcement than those of upper reinforcement. In addition the maximum tensile forces in each reinforcements of the blockfacing wall represented greater values in middle row of reinforcements than the upper most and lower most reinforcement. This may be due to the differences in mode of wall movement: the block facing wall moves in horizontal translation mode, but the full facing wall moves in rotational mode mostly. In case of the full continuous facing wall, the locus of maximum tensile force in the reinforcements tended to be generated close to the facing rather than the Rankine failure surface when the reinforcement length (L) was longer than 0.7H. 4 SUMMARY AND CONCLUSIONS This study has presented positive possibilities Of reducing wall deflection when the reinforced soil wall is equipped with full continuous facing. In this study,
Figure 4. Typical generated tensile force distribution in Case of block facing: (a) row 1 (top) to (b) row 4 (bottom),
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Figure 6 . Comparison of maximum tensile force generated in reinforcement after completion of wall construction (without loading): (a) full continuous facing, (b) block facing.
Figure 5 . Typical generated tensile force distribution in case of continuous full facing: (a) row 1 to (b) row 4.
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small-scale reinforced soil wall was constructed in a laboratory to investigate role of the wall facing and the effect of construction sequence on the wall. A full continuous facing wall and a block-type facing wall were introduced for test. These two different facing systems adapted different construction procedures. The model wall was built with geogrid reinforcement, sand, and facings on rigid surface. The model wall was instrumented with earth pressure gages, LVDTs, and strain gages. It is found in this study that the reinforced soil wall system built with full continuous facing be the safest reinforced soil wall ever compared to the block type facing wall since it can reduce wall deflection drastically compared to the conventional block-facing wall. Thus, it is recommended that study for the wall system, for example, full size of the full continuous facing wall, be necessary for further wide usage for the future. 5 ACKNOWLEDGEMENTS This research has been performed as a part of Advanced Highway Research Center Project funded by Korea Ministry of Science and Technology, Korea Science and Engineering Foundation. REFERENCES Cardoso, A.S. and Lopes, M.L. (1996) The influence of the construction method on the behavior of geosynthetic reinforced walls-a numerical study, Proc. The Int. Symp. On Earth reinforcement, Kyushu, Japan, 35 1-355 Jones, C.J.F.P. (1994) Economic construction of reinforced soil structures, Proc. Recent Case Histories of Permanent Geosynthetic-Reinfroced Soil Retaining Walls, Ed. Tatsuoka and Leshchinsky, Balkema, Rotterdam, 103-1 15 Rowe, R.K. and Ho, S.K.(1997) Continuous panel reinforced soil walls on rigid founations, J. Geotech. and Geoenv. Eng., Vol. 123, NO. 10, ASCE, 912-920 Simac, P.E. et al. (1997) Segmental retaining walls, NCMA Design Manual, 2nd. Ed. Tatsuoka, F. (1992) Roles of facing rigidity in soil reinforcement, Key Note Lectiire, Proc. The Int. Symp. On Earth Reinforcement, Kyushu, Japan.
Figure 7. Comparison Of maximum tensile force generated in reinforcement after vertical loading: (a) full continuous facing, (b) block facing.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Interface friction coefficient of extensible reinforcement and its influence on designing of retaining structures P. Michalski & K.M. Skarzyliska Department of Soil Mechanics and Earth Structures, Agricultural University of Krakdw, Poland
ABSTRACT: The interface friction coefficient is the basic factor influencing the frictional resistance of reinforcement and as a result the internal stability of an earth structure. Field and laboratory investigations on interface friction coefficient between materials of extensible reinforcement (geotextiles, flexible straps) and fills (industrial wastes) were carried out. Two methods were used to determine interface friction coefficient i.e. shear box test and pull-out test. The test results obtained from both methods demonstrated that values of interface friction strongly depend on vertical stress (with its increase are diminishing) and derived from shear box test were greater than those obtained from pull-out test. Detailed calculation procedures for designing two experimental retaining structures using geotextiles and flexible straps are given and results of calculations are presented. Strong dependence of reinforcement length on values of friction coefficient was evidently revealed.
1 INTRODUCTION In the Upper Silesian Region of Poland great concentration of underground coal mining industry causes a lot damage to the earth surface infrastructure due to mining subsidences. Another problem is connected with the great amounts of colliery spoils and fuel ashes as the area for depositing them is strongly limited and they also affect the human environment. One of the basic ways of utilization of great amount of these wastes is applying them for civil engineering structures as man made building soil. As the reinforced soil techniques have become quite popular nowadays all over the world in numerous application areas, the use of colliery spoils and fuel ashes as a fill in retaining structures using reinforced soil technique can give considerable technical and economic benefits. Since there is no experience in erecting retaining structures from industrial wastes in Poland, the research project comprising two experimental reinforced soil structures 3 m high has been initiated and performed on fuel ash lagoon and colliery spoils dumping ground in Przezchlebie next to Gliwice: First of them was constructed from exhausted railway timber ties as facing units, straps from cut exhausted conveyor belts as reinforcement and colliery spoils as a fill. The second one was constructed in a different way with the use of geogrids and geotextiles as reinforcement, forming a facing wall by wrapping around the particular fill layers. Colliery spoils and fuel ashes were used as a fill. Both walls consisted of a number
of sections 3-m high, independent from static point of view. These structures were designed according to two design procedures, based on the coherent gravity hypothesis and tie-back hypothesis (Jones, 1988). The retaining wall constructed with the use of reinforcing strips from exhausted conveyor belts was designed basing on coherent gravity hypothesis, following the rules given by the relevant Polish Standard. The embankment reinforced with geotextiles and geogrids was designed according to the tieback hypothesis basing on the principles given by LOTRAK designer's handbook (DON & LOW LTD. 1996). In both cases only the internal stability of structures was considered. The interface friction coefficient is the basic factor influencing the frictional resistance of reinforcement and as a result the internal stability of a reinforced earth structure. Length and spacing of reinforcing elements strongly depend on this factor, so determining of the interface friction coefficient is of essential importance. Taking that into consideration two methods for determining the friction characteristics of reinforcing materials were applied, i.e. laboratory shear box test and field pull-out test in full scale.
2 DESCRIPTION OF METHODS USED The shear box test was performed using a large (300 x 300mm) shear box apparatus, adapted specially for the test by fitting a timber block into the bottom of
41 1
Table 1. Results of shear box determining of interface fiction characteristics. Fill
Colliery spoils R.C. = 0.99 4 = 41" C=41kPa Colliery spoils R.C. = 0.98 4 = 38" C = 51 kPa Pulverised fuel ash R.C. = 0.88 4 = 28" C = 29 kPa
Material of reinforcement
Angle of Adhesion friction 4, C, 1"1 Wal
exhausted conveyor belts 36.8 new conveyor belts 34.5 "Paraweb" straps 32.6
35.4 3 1.8 24.2
Geotextiles F 650M Geotextiles Ha Te 600 AW (geocomposit)
37.8 34.1
19.2 29.7
Geotextiles F 650M Geotextiles Ha Te 600 AW (geokompozyt)
33.1 32.5
20.5 19.1
Note: R.C. = Relative Compaction = Degree of Compaction obtained from Standard Proctor Test
Figure 1. Determining the interface friction coefficient: a) shear box test b) pull-out test.
the lower half of the box. The reinforcing material was attached to the block on its one edge so it could move on the block towards the acting shearing force to include the material flexibility during the test. The upper half of the box was then filled with compacted fill material (degrees of compaction of the particular fill materials are given in Table 1) and a normal load was applied. The shear force required to cause sliding was then measured using a constant rate of strain of 1.0 mm/min at normal stresses of 50, 100, 150,200,250 kPa (Fig. la). The friction coefficient p obtained from the shear box test was calculated by the formula:
where C, = adhesion [@a]; crU = vertical stress applied, or overburden pressure [Wa]; @, = angle of 412
interface friction ["I. The full scale pull-out test was carried out at Barony colliery (Scotland) using a large open-topped box with one open end, constructed from rigid steel frame and timber railway sleepers. The test the box was 5 m long by 4.60 m high being 3.1 m wide. During test the box was filled with compacted test material and at specific levels reinforcing straps were laid passing through the front wall of the box. A specially designed pull-out jack delivered by British Coal Minestone Services was used to pullout individual straps with continuous record of load and displacement up to achieving the maximum pull-out force just before sliding of the strap (Fig. 1 b). The apparent friction coefficient p was obtained from this test using the formula:
where T, = maximum pull-out force [kN]; B = width of strap [m]; L = length of strap [m]; y = unit weight of fill [kN/m3]; H = height of fill above reinforcement [m]. 3 TESTRESULTS Results obtained from the shear box tests carried out on reinforcing straps made from exhausted conveyor belts and new Paraweb straps (British make) in colliery spoils and pulverized fuel ash are presented in the Table 1. Having these data and the data obtained from pullout tests (Michalski 1989) the relationships between values of interface friction coefficient p, developed
Figure 2. Values of interface friction coefficient p obtained for colliery spoils versus vertical stress oU.
Figure 4. Influence of backfill parameters: 4 (a) and c (b) on the angle of interface friction @,y. Results obtained from shear box tests for geotextiles.
Figure 3. Values of interface friction coefficient obtained for the British straps in colliery spoils and fuel ash.
using formulas 1 and 2, and vertical stress ou were made. They are demonstrated in a graphical form in Figure 2. For comparison the analogous graphs were prepared basing on British test results, performed in the same way given in (Bouazza et al. 1990). They are presented in Figure 3. The test results obtained from both methods demonstrated that values of interface friction p strongly depend on vertical stress and with its increase are diminishing however are discrepant. Values p coefficient derived from shear box test were much greater than those obtained from pull-out test, especially at low vertical stresses. These results confirm the results obtained by other authors (Finlay et al. 1988) from the tests carried out in similar fills. In the authors’ opinion p values obtained from shear box tests are overestimated, particularly for small values of vertical stress ou but values p obtained from pullout test are underestimated. It can be explained by the fact that the probable distribution of the frictional force along the pulled-out strap is uniform only towards the front end and is dropping off to a low value at the free end. It reduces the value of friction coefficient (Finlay et al. 1988). So, the basic question arises
413
which values should be used for design calculations. Test results also demonstrated the influence of the fill shearing strength parameters on the interface friction characteristics. These relationships between values of angle of internal friction and cohesion and the angle of interface friction obtained from shear box tests are presented in Figure 4. It is evidently visible from the graphs that the grater shearing strength of a fill material gives the higher values of interface friction between the reinforcement and the fill.
4 INFLUENCE OF INTERFACE FRICTION PARAMETERS ON THE PREDICTED LENGTH OF REINFORCEMENT
To facilitate the analysis in what way the values of interface friction coefficient influence predicted lengths of reinforcement some calculations were made according to two hypotheses, i.e. coherent gravity hypothesis and tie-back hypothesis (Jones 1988) (Fig. 5). Calculation procedure basing on the coherent gravity hypothesis and rules given in PN-83B-03010 were
Tensile force at the facing panel: T,
-
= 0.75 Z, max [kN]
(8)
Working length of the reinforcement strip:
where z,- max, a1 = as above; B - width of reinforcement Em]; ,u = apparent coefficient of adherence between soil and reinforcement [-]; ml = safety factor = 0.75. Total length of the strip: L = L,
+ AL Em]
(10)
where AL = extent of active zone [m]. The embankment with a vertical facing slope reinforced with geotextiles and geogrids was designed according to the tie-back hypothesis and specifications given in LOTRAK designer's handbook (DON a LOW LTD. 1996). The applied calculation procedure was as below (see Fig. 5b). Tensile force pulling out the reinforcement at given level was obtained from: Figure 5 . a) coherent gravity hypothesis, b) tie-back hypothesis.
7;. = a,, . s [kN/m]
applied for the retaining structure reinforced with Paraweb straps and the straps made of exhausted conveyor belts. The procedure was as it follows (see Fig. 5a). Tensile stress in fill acting on reinforcement:
where s = vertical spacing of reinforcement (in all cases s = 0.5 m);
K=(l-sin@).
= angle of internal friction of for z 5 6 m, where the fill, z = depth below soil surface [m]; a1 = vertical stress on reinforcement strip:
where cry = overburden pressure:
where y = unit weight of soil [kN/m3], h - fill height above reinforcement [m]; q = extra load on the soil surface [kPa]. Maximum tensile force in reinforcement was obtained from:
L n
*
03
[kPa]
(I - - 2 +-.tan2 :
45 - - (13) (
O
3
where K = soil pressure coefficient;
where K = soil pressure coefficient:
trmz =
(11)
where T, = as above; Fb = safety coefficient = 1.2; y = unit weight of fill [kN/m3]; z = as above; 6 = angle of friction between fill and reinforcement material, obtained from shear box test ["]. Lo value at given level was calculated from the formula: Lo=(H-z).tan
(45
O
3
--
[m]
(15)
where H = total height of the embankment [m] ( H = 3.0 m); z , @ = as above. Having calculated Lb and LO values, the L value was obtained as:
(7)
where 0 3 = as above; Y = vertical spacing of reinforcement [m]; n = number of strips per running meter of the wall [ 1/m] .
414
Table 3. Predicted lengths o f geotextile reinforcement calculated for the constant value o f the angle of the interface friction 6 obtained from shear box tests. H
2.5 2.0 1.5 1.0 0.5
Figure 6. Influence of p values on predicted strap length.
Results of strap length calculations are presented in Table 2. Standard design procedure assumes the constant value of interface friction coefficient irrespective of reinforcement level. The value of 0.4 was developed from shear box tests as the equal half of stabilised value p (= 0.8) and it approximately equals the value p obtained from pull-out tests (Fig. 2). It results from Table 2 that predicted lengths of reinforcing straps are diminishing with the depth when the constant g values are assumed, but they are changing in a different way when the variable p values are used, relevant to the actual vertical stress crU. Generally, lengths of straps calculated at p = 0.4 are 3-4 times greater then those determined for variable p values whereas the variable p values determined for particular reinforcement levels are 3-18 times greater then value of 0.4. The most interesting is the analysis including an extra load imposed on the embankment surface. Results of the calculations in graphical form are illustrated in Figure 6. These graphs demonstrate the dependence of predicted strap lengths on extra vertical load acting on the surface of the structure. It results from these Table 2. Total length of reinforcing straps in colliery spoil predicted for various and constant values of p. H
[m] 2.75 2.25 1.75 1.25 0.75 0.25
Collierv sDoils 6 = 35" Pulverised fuel ash 6 = 32"
1.22 0.98 0.73 0.49 0.24
0.16 0.15 0.15 0.14 0.14
1.38 1.13 0.88 0.63 0.38
1.50 1.20 0.90 0.60 0.30
0.24 0.24 0.23 0.22 0.22
1.74 1.44 1.13 0.82 OS2
graphs that when the constant p value is assumed, the predicted length of reinforcing straps does not depend on an extra load 40. This apparent paradox results from the fact that the greater horizontal forces caused by the greater load qo are compensated in full by the greater frictional resistance produced by the greater vertical stress acting on the reinforcing straps. In the case when variable p values relevant to the actual vertical stress are assumed the calculated length of straps are increasing with the increase of vertical stresses, because greater horizontal forces are not compensated in full by frictional resistance due to diminishing p values, therefore the lengths of straps have to be greater. Generally the predicted lengths are diminishing with the depth below structure surface because the extent of acting zone is getting smaller (Fig. 5). The similar analysis was made for geotextiles where the length of reinforcement was calculated according to the tie-back hypothesis (Fig. 5 ) and (DON & LOW LTD. 1996), assuming the constant values of the angle of the interface friction obtained from the shear box tests. Results of these calculations are presented in the Table 3. As previously, where the constant value of the friction coefficient was assumed the predicted lengths of reinforcement are diminishing with the depth below the embankment surface. The calculations were also made including an extra load qo and the results obtained revealed that there is no influence of an extra load on the predicted length of the reinforcement due to the same reason as in the previous case because the 8 value was assumed as con stant. The limitation of an extra load is only due to the tensile strength of reinforcement itself (ultimate limit state) or its extensibility (serviceability limit state).
Overburden p determined from shear box tests pressure C T ~ (variable) p = + tg4s p = constant = 0.40 Conveyor Paraweb Conveyor belts belts straps and Paraweb [kPa] p L[m] p L[m] strapsL[ml 5 EXPERIMENTAL RETAINING STRUCTURES 5 7.050 1.04 5.50 1.11 3.87 15 2.810 1.23 2.26 1.39 3.77 Taking into consideration the calculation results pre25 1.960 1.36 1.61 1.56 3.66 sented above two experimental structures in full scale 35 1.600 1.29 1.33 1.51 3.43 (3 m high) were erected but due to technical reasons 45 1.395 1.04 1.17 1.27 3.11 55 1.260 0.77 1.08 1.01 2.68 in practical design the constant lengths of reinforce-
2
ment on all levels have been applied. The retaining
415
5. The design procedures described in the paper can be recommended to use for temporary structures however designing should be yet conservative up to establishing the optimum method for determining the interface friction coefficient because this problem remains still open.
wall was erected with the use of 4 m long straps and in the reinforced embankment the reinforcement 3S O m long from geotextiles and geogrids was applied both in colliery spoils and in pulverized fuel ashes. The horizontal displacement of these two structures in their lifetime and caused by dynamic load (produced by numerous passes of heavy building plants) have been systematically recorded and the results (Michalski 2000) show that the behaviour of the structures is fully satisfactory and the internal stability has not been affected.
REFERENCES DON & LOW LTD. 1996. Lotrak Geotextiles. Designer’s handbook (Polish translation). Jones, C.J.F.P. 1988. Earth Reinforcement and Soil Structures. Butterworths, London. Michalski, P. 1989. Study on friction characteristics between straps from conveyor belts and various fills in reinforced earth. British Coal Corporation. Minestone Services Philadelphia, Houghton-le-spring. Research Report ms. Polish Standard PN-83B-03010. Retaining walls. Static calculation and design. Michalski P. 2000. Retaining structures constructed using reinforced soil technique from industrial wastes. Proc. of All Colloque Franco-Polonais de Mkcanique des Sols et des Roches Appliquke. Paris, France. Finlay, T. W., Wei, M. J., Hytiris, N. 1988. Friction characteristics of polypropylene straps in reinforced minestone. Proc. of the Int. Geotech. Symp. on Theory and Practice of Earth Reinforcement (Fukuoka Japan). Balkema, Rotterdam, pp. 87-92. Bouazza, A., Finlay, T. W., Hytiris, N., Wei, M. J. 1990. Friction characteristics of polypropylene reinforcing straps in various fills. Proc. of the lnt. Symp. on Reclamation, Treatment and Utilization of Coal Mining Wastes (Glasgow UK). Balkema, Rotterdam, pp. 349-354.
6 CONCLUSIONS 1. The interface friction coefficient strongly depends on the method of its determination and values obtained from shear box tests are greater than those derived from pull-out tests. 2. The value of normal stress 0, acting on the reinforcement influences the value of friction coefficient - when 0, is increasing, the values of friction coefficient are diminishing irrespectively of the determination methods. 3. The predicted length of reinforcement strongly depends on the friction coefficient and at its lower values the designed length of reinforcement should be greater. 4. If the constant value of interface friction coefficient is assumed in both design procedures described in the paper the calculated length of reinforcement does not depend on an extra vertical load.
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Reliability analysis of geosynthetics reinforced soil wall Y. Miyata, S. Shigehisa & K. Kogure National Defense Academy, Yokosuka,Japan
ABSTRACT: This paper examines reliability analysis for GRSW (Geosynthetics Reinforced Soil Wall). Proposed analysis method evaluates stability of GRSW against six failure modes with uncertainty of the design parameter. Performance functions needed for the analysis are derived fiom the design manual for GRSW of Japanese Public Works Research Institute. Compared failure probability calculated by proposed method, the most critical failure mode can be evaluated. In this paper, outline of the analysis method is explained, and advantage of reliability analysis is discussed on the basis of numerical results for simple condition. shown in Fig. 1. In this study, these failure modes are considered for reliability analysis of GRSW.
1 INTRODUCTION The shear strength of compacted soils changes with compaction condition or strain level, and the tensile strength of geosynthetics changes with temperature or strain rate. It is difficult to determine design parameters for compacted soils and geosynthetics. In the design of GRSW (Geosynthetics reinforced soil wall), uncertainty of the design parameter should be considered. GRSW is a kind of hybrid structure. Hybrid structure has generally some failure mode. Safety against every predictable failure mode should be evaluated with suitable index to compare safety in each failure mode. In reliability analysis, uncertainty of the design parameter is considered by assuming the design parameter such as random variable, and safety of structure is evaluated with failure probability, which is suitable index to compare safety in each failure mode. This analysis may be a useful tool to determine partial safety factor in the limit state design. In this paper, basic concept of reliability analysis method is proposed for GRSW, and some advantage of reliability analysis is discussed on the basis of numerical results for simple condition.
2.2 Failure probability and performance function In reliability analysis, a performance function Z(X) is used to describe that structure is in a “safe state” (Z(X)>O) or in a “failure state” (Z(X)
=
P~[z(x) < 01
2 ANALYSIS METHOD 2.1 Failure mode In Japan, PWRI (Public Works Research Institute) has established the design manual for GRSW (PWRI, 1992). The PWRI manual recommends to check safeties against six failure modes, which are
Figure 1. Failure mode considered in reliability analysis
417
Based on the assumption that Z follows a standard normal distribution, Pf is expressed as follows.
2.3 Uncertainty of design parameter Design parameters in the PWRI manual are shown in Table 1. Many researchers have investigated uncertainty of soil parameters [Vanmarcke (1977), Matsuo (1984), etc]. Uncertainty of design parameters of geosynthetics seems to be larger than that of soils, however it have not been investigated enough. In this study, all design parameters but p are assumed to be random variables to follow a standard normal distribution. Probabilistic assumption for each design parameter is shown in Table l .
in which @ is the distribution function of the standardized normal distribution, ,8 is reliability index, pz or oz is first or second statistical moment of Z(X). Expressed Z, ,8 can be calculated. In this study six performance functions, on failure modes shown in Fig.1, are derived from the PWRI manual. Derived performance functions are as follows.
Table 1. Considered design parameters Design parameters Deterministic ( 0 ) or random variable(o)
A-1 (Rupture of geosynthetics);
ZAi I$,, ( T = ?<) -1 A-2 (Pull out of geosynthetics); ZA2=FIN(T=Tp)-l ?/I unit weight density, c cohesion, 4 angle of shear resistance, c* or @ cohesion or friction angle between fill and geosyn-
in which M,<+U(T) I-& (1’) = , M, =liC{ci+~cos~tan@), Ms AM(1’)= + sin0 tanh) ,M , = cos0 ,
thetics, q,L ultimate bearing capacity
RCW
c cos^
reinforcing force, TI<: Of geosynthetics, TI’: Pull out strength of geosYnthetics, R: radius of circular arc, I: Arc length of sliding surface split with a slice, Weight of soil in a slice, 8: angle between a sliding surface split with a slice and geosynthetics.
2.4 Calculation of reliabiliv index The failure probability of GRSW is calculated with reliability index; p proposed by Hasofer and Lind (1974). The p is defined as the minimum distance from surface Z(X)=O to the origin of the uncorrelated random variables. Formulation to calculate p can be written as follows.
B-1 (Sliding); Lc + W,?tan4
-1 P H B-2 (Overturning); W&i+ &a2 282 = PH H/3 B-3 (Failure of foundation) 281
=
(7)
(8) in which L: length of geosynthetics, WR:weight of reinforced zone, PH or horizontal or vertical earth pressure to the reinforced zone, qu: Ukimate bearing capacity.
in which (aZ/aX). is the gradient vector at the most probable failure point X*=(Xl*, X2*, *, X,*) and % is the vector of mean value of the basic input random variables respectively. The calculation of p is performed on the surface of minimum safety factor defined in the PWRl manual.
3 RESULTS AND CONSIDERATION
P p r :
C-l (Overall stability); Z,, FAs- 1
(9)
in which F,, = M R / 4 .
418
3.1 Deterministic and Reliability Analysis In order to investigate the differences between deterministic and reliability analysis, these analyses were performed for same condition. Input parameters and cross section of analyzed condition are shown in Table 2 and Fig.2 respectively. Mean values of input parameters were determined on suppos-
Table 2. Input parameters ___
Design parameters Mean value, Coefficient of variant c (kNlmL) tanL-
Figure 3. Reliability index and Safety factors
Figure 2. Cross section of analysis condition
ing that GRSW was constructed with sandy soils on clay deposit. Coefficients of variant for design parameters were assumed to be ten percentage for soils and to be twenty percentage for geosynthetics respectively. Layout of geosynthetics in analyzed condition was determined to make Fc. larger than FT against each failure mode, in which Fc. is a calculated safety factor and FT is a target value of safety factor recommended in the PWRI manual. F(, from deterministic analysis was transformed to “safety ratio index; 2’ defined as follows.
‘I
The referenced FT is shown in Table 3. The Fc. on internal stability was calculated by following equation.
dex as p from the results of deterministic analysis with mean value. In the case of comparing safety in each failure mode, reliability analysis should be performed. 3.2 Effects of Uncertainty of Design Parameter Design parameter for GRSW consists of soil parameters such as c, tan4 and geosynthetics parameters such as TR, c*, tan@. Comparative analysis was performed in order to investigate the effect of uncertainty on soil parameters i+and on geosynthetics parameters VR. Assumed layout condition of geosynthetics is shown in Fig.2 and input mean values are shown in Table 2. Fig.4 shows calculated relations between Pf (A- 1) and P’s or V I values. ~ When VSvalue is below fifteen percentage, Pj (A-1) is larger with increasing of VR value. When Vs value is over twenty percentage, the effect of bt is too small to neglect. In the case of calculating Pj (A-I), it is unnecessary to consider uncertainty of geosynthetics parameters when soil parameters is over a value, which is twenty percent-
Reliability index p on internal stability was derived from product Pf (A-1) and Pf (A-2), in which Pf (A1) and (A-2) is failure probability on failure mode A- 1 and A-2 respectively. Fig.3 shows comparison between safety ratio index x and reliability index p on five failure modes. The most serious or safe mode depends on analysis method. It is impossible to derive an equivalent in-
ef
Table 3. Target of safety factors; FT Mode Fl
A-1&2 1.2
B-1 1.5
B-2 1.2
B-3 2.0
C-1 1.2
Figure 4. Pf(A-1) vs. VS and VR
419
age in this study. Fig.5 shows calculated relations between Pf(A-2) and VSor VZ< value. At same condition, Pf(A-2) is larger than Pf(A-I). When V,y value is over twenty percentage, Pf(A-2) is larger with increasing of VR value. In the calculation of Pf(A-2), uncertainty of geosynthetics parameters should be always considered whether uncertainty of soil parameters is large or small. Coefficient of variant for the design parameter should be determined on the basis of the results of reliability analysis for various cases and of accident investigation.
Figure 7. Effect of number of layer, N
is smaller with increasing of geosynthetics length. When length of geosynthetics is below four meters, external stability is the most serious. When length of geosynthetics is over six meters, internal stability is the most serious. The most serious failure mode changes according to length of geosynthetics. Calculated relations between failure probability on internal stability and number of layer are shown in Fig.7. Pf(A-2) is always larger than Pj(A-1) regardless of number of geosynthetics. Pf (A-1) and Pf (A-2) changes smaller value with decreasing of number of layer. Conducted reliability analysis, layout of geosynthetics can be determined according to acceptable risk.
4 CONCLUSIONS Main conclusions of this paper are as follows. Reliability analysis method was proposed for geosynthetics reinforced soil wall. Compared failure probability calculated by proposed method, the most critical failure mode can be evaluated. The effect of uncertainty of the design parameter depends on failure mode. (3 1 Conducted reliability analysis, layout of geosynthetics can be determined according to acceptable risk.
Figure 6. Effect of geosynthetics length, L
3.3 Effects of Layout Condition of Geosynthetics Stability of GRSW changes according to layout condition of geosynthetics. Relations between failure probability and length or spacing of geosynthetics were investigated by parametric study. In the investigation, failure probability was calculated by changi% the length or number of layer for the condition shown in Fig.2. Input parameters were shown in Table 2. Calculated relations between failure probability in each mode and length Of geosynthetics are shown in Fig*6. probability Of mode but mode A-1
REFERENCES Hosofar, A M and Lind N C 1974 Exact and invariant second Of Engineering Mechanics Dzv, moment ‘Ode Proc ofASCE, Vol 100, No EM1 11 1-121 Matsuo, M 1994 Geotechnzcal Engineering - 7heory and Practice of Reliability Deszgn, Gihodoh Publisher (in Japanese) Public Works Research Institute 1995 Design Manual for Reinforced Earth with Geotextzle (in Japanese) Vanmarcke, E H 1977 Probabilistic modeling of soil profiles, J of Geotechnzcal Engineerzng Div, Proc. of ASCE, Vol.103, No.GTI 1: 1227-1249. -
420
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Experimental research of reinforced soil wall for rock-fall protection T. Nomura & S. Inoue PROTEC ENGINEERING Co., Ltd., Japan
M. Fuchigami & Y. Obata A C D Co., Ltd., Japan
Y. Yokota & T. Kubo MAEDA KOSEN Co., Ltd., Japan
K. Arai Fukui University, Japan ABSTRACT: The retaining reinforced soil wall for rock-fall protection has been developed recent years. It needs appropriate spaces to construct the wall due to its structural shape. However, there is often insufficient space to construct the wall where needs such a protection structure. Therefore the pocket type reinforced soil wall for rock-fall protection was developed. It can be constructed in narrow space at a roadway side and it can catch a rock by it's top. This paper describes the experimental research of the reinforced soil wall for rock-fall protection. The test was performed two models of prototypes, the miniature size model (wall height 1S m ) and actual size model (wall height 6.0m). The experiment of miniature size model confirmed the effect of geosynthetics. The experiment of actual size model demonstrated behavior of the wall, when at the moment of clashing of the rock and confirmed it's safety against a huge scale rock fall.
1 INTRODUCTION Recent years, rock-fall protection techniques have been being researched and developed in U.S.A. (e.g., Hearn et al. 1995). Rock-fall countermeasures have been improved technologically and the rock-fall protection engineering has been diversified far more than ever. Under such circumstances, the first author, et al. developed the "bank type rock-fall protection retaining wall", which was constructed into a bank out of a geosynthetics-applied reinfoW3ment embankment as shown in Figure 1. Such retaining wall has been put into some practical uses. On the route where a rock-fall is likely to take place, however, it is often impossible to secure the space enough to install a protection facility on the roadside. As a solution to this problem, a "pocket type rock-fall protection retaining wall" was designed, which has a function of preventing a falling rock from reaching the road, with a flat place located on the roadside by the reinforced earth engineering method as shown in Figure 2. The bank type wall is designed to catch a rock at the lateral side, while the pocket type wall is used to stop a rock at the upper surface. This paper, deals with the results of the actual size experiment conducted for a pocket type wall. The impact force acting upon the pocket type wall is to be discussed to propose a simplified design method.
42 I
Figure 1 Bank type rock-fall protection retaining wall (conventional m e )
Figure 2 Pocket type rock-fall protection retaining wall
2 ACTUAL SIZE EXPERIMENT 2.1 Experiment method A actual size model shown in Figure 3. was fabricat& and a weight falling experiment was conducted. The embankment material applied had parameters as shown in Table 1. On the model, geosynthetics were laid out height wise in intervals of 500 millimeters as a geosynthetics with a tensile strength of 32 kN/m (strain 5%). For the wall surface material, a wall block was used, with a 300 millimeters thick buffer layer (one-size crushed stone) provided on the back. In addition, a 1.4 meters thick non-rolled layer of sand was provided at the ceiling end of the wall as rock-fall buffer material. To carry out the experiment, a cylindrical weight, which had a diameter of 1.54 meters, weighing at 51.7 kN, was dropped from a height of 20 meters onto the model at the center as shown in Figure 3. Measurement items included weight acceleration, a vertical earth pressure, a wall surface displacement and a penetration of the weight. To determine an intra-earth value of the vertical earth pressure, however, a load meter was buried after secured onto a 300 x 300 millimeters steel plate.
Table 2 shows the displacements by block on the wall surface. A maximum wall surface displacement of 97 millimeters was recorded in front of the weight drop point, with a penetration of 950 millimeters observed. The weight stopped in the non-rolled layer the geosynthetics at the uppermost stage broke down*
2.2 Results Figure 4 shows a time series of changes in the weight impact force obtained by multiplying the measured weight acceleration by the weight mass. It shows a weight impact force of 2,417 kN at the maximum in 17 msec. Figure 5 shows a time series of changes in the vertical earth pressure measured with a load meter. Load Meter 2, located just under the weight drop point, read a maximum vertical earth pressure of 355.6 kN/m2 in 67 msec. Table 1. Properties of soil Unit weight p
, (kN/m3)
Angle of shear resistance
16.0
36.0
@I(')
0.0
Cohesion c'(kN/m2) Table 2. Wall surface displacement.
1 2 3
4
6
3
4
b-)
1
2
5
6
7
8
am am aoii am
am am am am
am a074 aw a076 am am am am am am am am aces am a a am am am am am am a063 am am
QCEO
0061 0 0 a61 aC60
am 002
Figure3. Shapeofactualsizeexperiment.
422
3 SIMPLIFIED DESIGN METHOD Based on the weight impact force obtained from the actual size experiment results, a design impact force was assumed while an attempt was made to check out the model by examining the internal stability of the reinforced earth retaining wall. Figure6 is a schematic of simplified calculations. To examine the internal stability of the pocket type rock-fall protection reinforced-earth retaining wall, it was assumed that the design impact force would act as an additional load on the reinforced earth retaining wall at the ceiling end. For a possible distribution of the impact earth pressure is concerned, the vertical earth pressure was eventually found to distribute by 1: 0.5 or more as measured during the actual size experiment. The rock-fall impact force, therefore, was made to act as a distributed load on the top surface of the reinforced earth retaining wall where the rock-fall impact force had was distributed at 1 : 0.5 in the non-rolled layer as shown in Figure 6. For the earth quality constant used in the calculations, refer to the conditions given in Table 1. It was assumed, moreover, that the geosynthetics would not have its strength affected by its own creep and that it had a tensile strength of 32
Figure 4. Weight impact force vs. time
Figure 5. Vertical earth pressure vs. time
The actual size experiment recorded a inaximum weight impact force of 2,417 kN. Based on this value, the rock-fall impact force assumption formula given in Expression 2.1 was used to calculate the Lame's constant that expresses the rigidity of an impact recipient. As a result, the impact recipient was found to have approximately 700 kN/m'. The buffer material used in the design by Lockshed, et al. reportedly Lame's constant have 1,000 kN/m2in general. The impact force measured in the experiment reported herein may be deemed to fall nearly within a range of reasonable values. A weight penetration relative to the rock-fall impact force assumption formula, moreover, may be expressed in Expression (2.2).
where, P is a weight impact force, Ya penetration, W a weight load, H a drop height, hlame's constant and yl a converted radius of the weight. From Formula 2.2, a penetration of 929 millimeters was obtained subject to the actual size experiment conditions. And it was found to agree nearly with the experiment result.
Figure
SIn,pllfiedcalcuiatIons
Figure 7. Slip surface and minimum safety factor
423
kN/m. The weight, moreover, had an impact force of 2,4 17 kN experimentally determined. And it was really loaded in the experiment. In the experiment reported herein, the pocket type rock-fall protection retaining wall did not come to break down. It was assumed, therefore, that the retaining wall had a safety factor of 1 .O% or more relative to a rock-fall impact force. And the reinforced earth retaining wall with a safety factor of more than 1.O had its shared width obtained on a trial and error basis. The term, load's shared width, as used herein, means an extension of the width by which the geosynthetics would resist an impact load. The examination resulted in a load's shared width of 6.0 meters and a minimum safety factor of 1.030. Figure 7 shows the sliding surface with the safety factor minimized. And Figure 8 shows the shared width. As shown in Figure 8., moreover, a distribution gradient of approximately 1: 0.7 was obtained by connecting the distributed load at the end with the center
Figure 8. Stress distribution Table 3 . Tensile force acting on geosynthetics NO.
Tmsile strength JwIgtmt
Tensile force to act (IdVd 51. @XI 28.22 24.753 22.609 21.282 20.492 20.075
T,(M/d
10 9 8 7 6 5 4
Dism fim the top@ 0. 5 1. 0 1. 5 20 25 3.0 3. 5
2
45
20.452
32
32 32 32 32 32 32 32
X
0 0
0 0 0 0
of the wall height at the end of the shared width. The sliding surface, moreover, appeared from the loadacting point at the end to a position of 1.O meter above the reinforced earth retaining wall at the bottom end. During the experiment, moreover, the wall surface displacement was found to increase upwards from a position of approximately 1.0 meter at the bottom end of the wall. From this, it may be gathered that the sliding surface was located almost reasonably. Table 3 shows the results of checking out each geosynthetics for tensile strength. Consequently, the geosynthetics at the uppermost stage only was found to have an acting tensile strength exceed the geosynthetics tensile strength. During the experiment, the uppermost geosynthetics broke down, showing an agreement with the experimental results. 4 SUMMARY The results obtained in the study reported herein could be summarized as under. (1) It is possible to enhance the safety of a retaining wall against a vertical load by installing a geosynthetics on the retaining wall body. (2) A actual size experiment allowed the author et al. to make certain that a pocket type rock-fall protection retaining wall would be safe enough as tested under the conditions of 51.7 kN in weight and 20 meters in drop height. (3) An assumption formula would permit us to obtain weight impact force acting upon the pocket type rock-fall protection retaining wall at the ceiling end, including a penetration. (4) Replacing the impact force with a static additional load would permit us to make a design, using the reinforced earth retaining wall stability checkout method commonly used. From the above, the author et al. think that the pocket type rock-fall protection retaining wall is satisfactorily applicable as a rock-fall protection work unless a flat space is available alongside the road. The author et al. would like to establish a simplified calculation technique in the future. To this end, it is planned that the experiment reported herein will be reproduced analytically to grasp the transfer and distribution behaviors of internal stresses, thereby making certain of the load's shared width and deformation behaviors of a reinforced earth retaining wall. REFERENCE
0
George. H., Robert K.B., & Henrie. H. (1995) Transportation research record 1504.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Numerical analysis of a sheet pile mooring wharf having several tie rods S. Ohmaki & K. Saeki National Research Institute of Fisheries Engineering, Fisheries Agency, Japan
M. Kiyozumi Nagaoka University of Technology, Japan
ABSTRACT: When an anchor pile-type sheet pile mooring wharf is constructed on soft ground, it sometimes undergoes tremendous lateral deformation due to the lateral flow of the ground that accompanies embankment construction behind the sheet pile. In the present study, various measures were considered to counteract this lateral flow of the ground. The results of the analysis led to the following conclusions: 1) By replacing the clay in front of the sheet pile with rubble mound, the horizontal displacement of the sheet pile could be reduced. 2) By increasing the horizontal distance between the sheet pile and anchor piles, the horizontal displacement could be greatly diminished. 3) When the horizontal distance between the sheet pile and anchor piles was unchanged, there was almost no reduction in horizontal displacement at the top of the sheet pile even if the number of steps of tie rods was increased.
1 INTRODUCTION
In addition, joint elements (Goodman, 1976) were inserted between the sheet or anchor piles, and the foundation. The finite element analysis (Kobayashi, 1984) was conducted under plane strain conditions. Figure 1 is the ground profile used in the analysis. The upper part of the figure shows the entire profile used in the analysis, while the lower part depicts the area where the anchor piles and rubble mound were constructed. Table 1 lists the material constants that were used to calculate the ground for the model. The upper and lower clay layers and the rubble mound were modeled as elasto-viscoplasticity following Sekiguchi and Ohta (1977) and the foundation layer was considered to be a linear elastic body.
When an anchor pile-type sheet pile mooring wharf is constructed on soft ground, sometimes greater than expected horizontal displacement occurs on the top of the sheet pile, hindering the use of the wharf. This is probably the result of lateral deformation of too closely spaced sheet and anchor piles that occurs when the ground moves laterally during and after the construction of an embankment behind the sheet pile. At least three different reinforcement measures have been considered to prevent or suppress the lateral displacement of piles and make them resist the lateral flow of soft ground: 1 ) Replacing the soft soil in front of the sheet pile with sand, gravel, etc., 2) increasing the horizontal distance between the sheet and anchor piles, and 3 ) increasing the number of steps of tie rods. To investigate the results of these measures, we subjected the ground-structure system composed of soft ground, and sheet pile, anchor piles and rubble mounds, to two-dimensional consolidation deformation analysis.
2 GROUND MODEL AND MATERIAL CONSTANTS USED IN THE ANALYSES The analysis considered the ground to be a twophase material composed of an elasto-viscoplastic structural skeleton of Sekiguchi and Ohta (1977) and incompressible pore water, and the sheet pile, anchor piles and tie rods to be linear elastic beam elements.
Figure 1. Ground profile used in the calculations
425
Table 1. Soil parameters used in analysis. Upper clay layer
Lower clay layer 0.0 1.600 1.484 0.173 0.027
770
0.0 1.600 1.484 0.173 0.027 1.200 0.315
1.200 0.315
0.480 0.407
(l/day)
10-5
10-j
10-j
Parameters Young's modulus Poisson's ratio Density Stress ratio at failure Compression index Swelling index
E(MPa) V
p (tlmj)
A4 A U
Initial void ratio Anisotropic parameter
eo
Initial volumetric strain rate
v
Secondary compression index Coefficient of permeability
a k (m/day)
10-3 10-3 4 . 3 2 ~ 1 0 - ~ 8.64~10.'
Table 2. Material constants of beam elements Properties
Sheet pile
Bending rigidity H(MNom2)
46,46
Axial rigidity EA (GN)
3,83
Rubble mound
Anchor pile
Tie rod
22.05
6.21
1.88
0.126
Table 2 shows the material constants for the sheet pile, anchor piles and tie rods used in the analysis. The material constants of the joint elements were joint rigidity in the vertical direction E, of 107 kN/m3, and in the tangential direction Es of 10 kN/m3. The boundary conditions of displacement were as follows: Both the vertical and horizontal displacement of the upper boundary (ground surface) of the analytical profile in Figure 1 were unrestricted, the boundaries of both left and right sides were unrestricted in the vertical direction and restrained in the horizontal, and the lower boundary (bottom) was restrained both vertically and horizontally. As the hydrological boundary conditions in the analysis, drainage was allowed only in the upper boundary (ground surface) of the analytical profile in Figure 1; there was no drainage at any other boundary. Assuming that the load acted uniformly on the ground surface behind the sheet pile and increased at a constant rate, the analysis considered various combinations of the following conditions: 1) Presence or absence of rubble mound in front of the sheet pile: If there was no rubble mound, that area became the upper clay layer. 2) Horizontal distance between anchor piles and sheet pile: As we can see in the lower part of Figure 1, the calculated horizontal distances from the sheet pile were 1Om for anchor pile A, 20 m for anchor pile B and 40m for anchor pile C.
426
0.30 1.600 1.735 8.315x10" 2.810~10-~
Base stiff layer
9.80 0.33 1.800
10" 8.64~10~
3) Number of tie rods: The positions of tie rods used in the calculations are shown in the lower part of Figure 1. Here, tie rod 1 signifies just one tie rod; two are denoted by tie rod 1 and tie rod 2; and three rods are shown as tie rods 1 , 2 and 3. 4) The velocity of a uniformly distributed load acting on the ground surface behind the sheet pile: Calculated for 1.O kPa/day. Calculations were stopped when the finite element mesh in the ground was destroyed. This load is called the "critical load" for convenience. In addition, the conditions for the following calculations, particularly in sections where there were no calculation conditions, were 1) no rubble mound, and 2) anchor pile B. 3 ANALYTICAL RESULTS 3.1 Effect of rubble Figure 2 shows the relation between the horizontal displacement of the top of the sheet pile and the load for the case of no anchor piles and load velocity of
Figure 2. Effect of rubble mound on the relation between load and horizontal displacement of the top of the sheet pile.
Figure 3 . Effect of rubble mound on the depth distribution of the bending moment of the sheet pile.
1.0 kPdday, and for both with and without a rubble mound. According to the figure, when there was no rubble mound, there was great horizontal displacement of the sheet pile, and its form was nonlinear. Figure 3 shows, for the same calculations, the depth distribution of the bending moment of the sheet pile at that point in time when the load was 40 kPa. When the load was large, the rubble mound greatly reduced the maximum bending moment. 3.2 Effect of anchor piles Figures 4(a) and 4(b), which show the cases of no anchor piles, and anchor piles A, B and C, illustrate the relation between load for one and three tie rods, respectively, and the horizontal displacement of the top of the sheet pile. Here we can see that as the distance between the anchor piles and sheet pile increased, the lateral displacement of the sheet pile decreased, and the "critical load" increased. A likely explanation is that as the lateral distance between the anchor piles and the sheet pile increased, the length of the anchor piles that were within the lateral flow region of the ground behind the sheet pile decreased. Figures 5(a) and 5(b) show the results of calculations for depth distribution of bending moment of the sheet pile, made under the same conditions as in Figure 4. It is clear that the maximum bending moment was largest when there were no anchor piles, but when there were anchor piles, it decreased as the distance between the sheet and anchor piles increased. In addition, Figure 5(a) shows that for one tie rod, the bending moment near the ground surface was close to zero or negative at shallow depths, while at deeper depths it tended to be positive.
Figure 4 Effect of anchor piles on the relation between load and horizontal displacement of the top of the sheet pile
3.3 Effect of tie rods Figures 6(a) and 6(b) show, for the cases of no rubble mound and rubble mound respectively, the relation between the load for one, two and three tie rods and the horizontal displacement of the top of the sheet pile. As we can see, this relation was only slightly affected by the number of tie rods. This that even when the number of tie rods was in-
Figure 5 Effect of anchor piles on the depth distribution of the bending moment
427
Figure 7, which is based on calculations made under the same conditions as Figure 6, depicts the depth distribution of the bending moment of the sheet pile at a load of 20 @a. According to Figure 7, a negative bending moment occurred near the ground surface when there was one tie rod, and the greatest positive bending moment occurred when there were three tie rods. 4 CONCLUSIONS A sheet pile mooring wharf built on soft ground was the object of this study. Numerical simulations were used to investigate the effects of rubble mounds on its bending moment and horizontal displacement, as well as the effects of the positioning of anchor piles, and the number of tie rods. The results led to the following conclusions: 1. The rubble mound helped to increase lateralsupport effect of the sheet pile, reducing the horizontal displacement and maximum bending moment. 2. Increasing the horizontal distance between the anchor piles and sheet pile helped to decrease both the maximum bending moment and horizontal displacement of the sheet pile, for the same load* This was due to the fact that there was little effect of lateral flow in the ground on the anchor piles. 3 . The relation between load and horizontal displacement of the top of a sheet pile was nearly constant, regardless of the number of steps of tie rods. However, increasing the number of steps of tie rods helped to increase the lateral support effect of the sheet pile. In addition, different numbers of steps of tie rods had different effects on the depth distribution of the bending moment of the sheet pile.
Figure 6 Effect of number of tie rods on the relation between load and horizontal displacement of the top
REFERENCES Figure 7 Effect of number of tie rods on the depth distribution of the bending moment of the sheet pile
creased, the horizontal displacement of the sheet pile could not be suppressed. However, increasing the number Of tie rods did tend to increase the to increase the load" Of the ground, that is, it lateral support effect of the sheet pile.
428
Sekiguchi, H & H Ohta 1977 Induced anisotropy and time dependency in clays, Proc of SpeciaIP Session 9, 9th ICSMFE 229-238 Goodman, R E 1976 Methods of Geological engineering in discontinuous Rocks (translated to Japanese by Akal, K , C Kawamoto 8~ Ohnishi)~ shuppan LTD 250258 Kobayashi, M 1984 Stability analysis of geotechnical stmctures by finite elements, Report of the Port and Harbor Research Institute, 23-1 482-499 (in Japanese)
Landmarks in Earth Reinforcemenf, Ochiai et a/. (eds), 0 2001 Swefs & Zeiflinger, ISBN 90 2651 863 3
Relation between wall displacement and reinforcement for reinforced retaining wall K. Okabayashi & K. Tagaya Kochi National College of Technology, Nankoku, Japan
M. Kawamura Toyohashi University of Technology, Toyohashi, Japan
ABSTRACT: From previous studies and results of field tests by the authors, the allowable wall displacement for design of reinforced retaining walls was defined. In addition, two dimensional elasto-plastic FEM analyses for the prototype retaining wall with stiff reinforcement were carried out to determine the optimum reinforcement considering the allowable wall displacements. 1 INTRODUCTION
are shown in Figure 1. The results of centrifuge model tests and the simulations are the values at failure and the others are the values when the walls are stable. The data at the failure state were plotted for the centrifuge model tests and simulation analyses, in Figure 1, and the data at the construction works and immediately after the construction were plotted in the same figure. It is recommended that the wall displacement is limited within the wall displacement immediately after the construction to maintain the small stress level and the wall stability. The maximum wall displacement at failure is approximately H/60 and the wall displacement after construction is approximately H/150. Therefore, it seems that the safety factor is about 2.5. H/150 is reasonable as the allowable wall displacement, as this number is a little bigger than the values by field measurement in I12 fields by Ogawa (Ogawa1993) and the other measured values are also within this range.
The current design methods for reinforced retaining wall being employed are based on the theory of rigid-plasticity that takes no account of wall displacement and deformation of reinforcing material in backfill. It does not correspond with a real phenomenon. For instance, Rowe et al. collected measured tensile forces of reinforcements of actual reinforced retaining walls and showed that the measured values of tensile forces in reinforcement are smaller than those by the current design when the wall is stable (Rowe & Ho, 1992). To obtain a rational solution to this structure, it is needed to clarify the relations among the displacement of wall, the tensile force in reinforcement, the earth pressure acting on the wall, and the frictional force between reinforcement and soil. These relations were observed in the series of centrifuge model tests (Kawamura & Okabayashi, 1998a) and its simulation analysis (Kawamura & Okabayashi, 1998b). In these papers, it was recommended that the strain level of the backfill material had to be restricted, and that the wall displacement should be prescribed as the design parameter, to be able to determine the amount of the reinforcement rationally. In this study, two dimensional FEM analyses were carried out for models by applying gravitational force. The relation between the wall displacement and the amount of optimum reinforcement will be discussed. 2 ALLOWABLE WALL DISPLACEMENT FOR THE RETAINING WALL REINFORCEMENT The relationship between the wall height and the wall displacement of the reinforced retaining wall by the centrifuge model tests and actual measurements,
Figure 1 Wall height and wall displacement
429
3 FEM ANALYSES FOR THE RITEINING WALL Figure 2 shows an example of analytical model for model 4. In these analyses, it is assumed that the reinforcement and the facing are elastic, and the backfill is Elasto-plastic. In the elasto-plastic constitutive equation, plastic softening was considered. In plane strain analyses, the reinforcement was simulated as a continuous sheet with an equivalent thickness (rigidity equivalent). The self weight analyses by the backfill are performed. The material constants of the reinforced retaining wall are shown in Table 1 . Joint elements shown in Figure 3, between facing and soil, and between reinforcement and soil were employed for the discontinuity. Material constants of the joint elements are shown in Table 2. In Table 2, $s dilatancy angle, Ks is the shear modulus of rigidity (kPa), Kn is the normal modulus of rigidity (kPa), and 4 is the angle of shear resistance. These
properties were obtained by the FEM inverse analysis in which the Centrifuge model tests were simulated (Kawamura & Okabayashi, 1998). Figure 4 shows the cases to be studied, in which model 0 is unreinforced one. The heights of retaining wall, H, for each case are 6.0 and 12.0m. The length, L, of the reinforcement laid in the backfill varies in L/H, which are 0.1875, 0.375, 0.5625, and 0.75. The spacing, h, varies in h/H which are 0.125,0.25,0.33 and 0.5. Figure 5 shows the calculated lateral displacement of the wall for each model. The lateral displacement of the wall decreases as length of the reinforcement becomes larger, and as the spacing of the reinforcement becomes smaller. The calculated maximum lateral displacement of the wall occurred at the middle height of the wall. Table 2. Properties of joint element Ks Faciiidsoil Reinforcemendsoil
Figure 2. FEM model for model 4 Table 1. Material properties. Elastic
Poisson's
Unit
Modulus
Ratio
weight
E
V
&Pal
Y
C
b
(kpa)
("1
(kNh')
Wall face
2107000
0. 2
23.52
-
Reinforcement
2 1600000
0. 3
77.03
-
__
Backfill
19600
0.3
15.5
0
35
-
Figure 4. Models Figure 3. Place ofjoint element.
430
100000 100000
KII
(kP4 10000 I00000
6
(i,
("1
("1
10
10
10
10
ratio of the earth pressure, Rp, is expressed by the following equation
in which PO is resultant force of Rankin’s active earth pressure acting on the wall in the unreinforced earth structure, and Pg is that of earth pressure resulted from the analysis with reinforced soil backfill, as shown in Figure 8.
Figure 5 . Wall displacement.
3.1 Relation between the wall displacement and the length of the reinforcement. The maximum wall displacements resulted from the reinforcement in the case of h/H=O. 125 are plotted in Figure 6, for the length of reinforcement in the backfill. The maximum wall displacement ,&, and the length of reinforcement L, are nondimensionalized by the wall height. In this figure the dotted line indicates the allowable wall displacement. The relation between the maximum wall displacement and the length of reinforcement is like a hyperbola and the maximum wall displacement decrease, as the length of reinforcement becomes larger. And the values of maximum wall displacement vary by the difference of the wall height. From Figure 6 all calculated values are within the allowable displacement incase of H=6m, and it is understandable that all cases require the longer reinforcement than Model 2 in case of H=12m.
Figure
drspiacementand length Of
3.2 Relation between the wall displacement and spacing of reinforcement Figure 7 shows the relation between the dimensionless maximum wall displacement, 6,,,/H, and the dimensionless spacing of reinforcement, h/H, in the case of L/H=0.75. The dimensionless maximum wall displacement by the reinforcement Of backfill becomes larger as the spacing becomes larger. And the values of maximum wall displacement changed according to the wall height. From Figure 7, all calculated values are within the allowable displacemcnt incase of H=6m, and it is understandable that making the spacing of reinforcement closer than Modcl 6 in case of H=12m.
Figure 7 Wall displacement and spacing of reinforcement
3.3 Relation between the reduction ratio of the earth pressure and the length of the reinforcement. The reduction of earth pressure resulted from the reinforcement are evaluated by reduction ratio, which is proposed by (Ogisako et al., 1988). The mi’ction
Figure 8. Reduction of earth pressure (h/H=0.125)
43 I
of wall displacement is larger than that of the earth pressure reduced by the reinforcement. From these results the earth pressure acting on the wall is changed according to the condition of the wall deformation, and tensile force acting on the reinforcement. 4 CONCLUSION
Figure 9. Reduction ratio and L/H (h/H=0.125)
Figure 10. Reduction ratio and h/H (L/H=0.75)
Reduction of earth pressures resulted from the reinforcement in the case of h/H=O.125 are plotted against the dimensionless length of the reinforcement in the backfill in Figure 9. The reduction ratio of the earth pressures decreases gradually as the length become larger, and the reduction ratio of the earth pressure has the almost same values of the wall height independently. 3.4 Relation between the reduclion ratio of the earth pressure and spacing of reinforcement Figure 10 shows the relation between the reduction ratio of earth pressure and dimensionless spacing of the reinforcement in the case of L/H=0.75. The reduction ratio of the earth pressure decreases rapidly as the spacing becomes larger, and the reduction ratio of the earth pressure has the almost same values independently of the wall height. In case of dimensionless spacing of the reinforcement h/ H=0.25, the reduction ratio of earth pressure becomes negative. It can be supposed that the increment of earth pressure due to the reduction
432
As the results, the followings were made clear. 1. From the results of author’s model tests, FEM analyses, field tests and prototype experiments with regard to the displacement of a reinforced retaining wall, H/150 is considered as an allowable displacement for a stable state. 2. The dimensionless maximum wall displacement by the reinforcement of backfill decrease, as the length of reinforcement becomes larger. And it becomes larger as the spacing becomes larger. 3. The values of maximum wall displacement changes different due to the difference of the wall height. 4. The reduction ratio of the earth pressure decreases gradually as the Length becomes larger. And it decreases as the spacing becomes larger. 5. The reduction ratio of the earth pressure has the almost same values independently of the wall height. 6 . In this paper, the deformation due to self weight of the backfill was analyzed. The consideration of the surcharge and other external lodes and the verification by experiment are necessary. REFERENCES Ogawa, N 1993 Relationship between filling material and wall deformation in TERRE ARMEE METHOD Journal of Geotechnical Engineerzng lll-27 1 19-125, in Japanese Ogisako, E Ochiai, H Hayashi, S and Sakai, A 1988 FEM analysis of polymer grid reinforced lsoil retaining walls and its application to the design method Int Synzpo Theory and Practice of Earth Reinforcement 329-334 Kawamura, M and Okabayasi, K 1998a Centrifugal model test on interaction between earth pressures against reinforced retaining walls and tensile force of reinforcement Journal of Geotechnical Engineering D-42 263-273, in Japanese Kawamura, M and Okabayasi, K 1998b Finite element analyses on interaction between earth pressures against reinforced retaining walls and tensile force of reinforcements Journal of Geotechnical Engneering D-43 239-247, in Japanese Kerry Rowe, R and Ho, S K 1992 A review of the behavior of reinforced soil walls, Keynote Lecture Proc of Znt Symp on Earth Reinforcement Practice Vol2 801-830
Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
A fully synthetic connection of polyester based strip reinforcement to concrete panels. Development, tests and first application. Pierre Orsat Freyssinet International Group, MENARD division, France
Nicolas Freitag Ecole Nationale des Ponts et Chausse'es. France
ABSTRACT: In order to be able to offer a coherent ground reinforcement system based on vertical panels and synthetic reinforcement elements, a totally synthetic corresponding connection had to be developed. This publication describes the pre-dimensioning steps, experimental justification and the first applications undertaken. It emphasizes the advantages of such a development for all cases where the use of chemically aggressive backfill (in relation to reinforcement and metal attachments and in combination with vertical facing) was excluded. 3 INTRODUCTION
3 SOLUTION RETAINED
The FREYSSINET/TERRE ARMEE INTERNATIONALE Group must be able to offer two complementary and coherent families of strip-reinforced flexible retaining walls and vertical facing panels, thus a totally synthetic system, and for which the development forms the subject of this publication.
After briefly studying the matter, it immediately appeared that an attachment system using synthetic loops anchored partly in the concrete and equipped with two external hoops and synthetic pin located inside these two hoops would be the best adapted to fulfil1 the specifications (figure 1)
2 SPECIFICATIONS The new attachment system to be development had to meet the following specifications: 2.1 Mechanical characteristics The tensile strength of the assembly had to be 65 kN for the attachments corresponding to 30 and 50 kN grade reinforcements and 130 kN for the other grades. It was agreed that this new attachment should be able to be used with the set of existing molds thereby requiring only minor modifications to be made.
Figure 1 . An overall view of synthetic connection
3.1 Synthetic loops (1) Overall design The synthetic loops are manufactured from reinforcement strips which are identical to the main reinforcements, although not as wide (33 or 44 mm instead of the normal 90 mm). The idea consists in forming continuous loops wrapped several times around themselves using a successfully developed technology. These strips are made of polyester fibers grouped into 10 separate bundles; the assembly is protected by a common polyethylene sheath. From a geometric point of view, the idea consists in making "straight, flat mother loops" which will then be deformed to obtain a "gondola" shape in or-
2.2 Physical and chemical characteristics This new attachment had to make it possible to use the main strip/panel reinforcement assemblies for ground of the same type as those for synthetic reinforcements are preferably used, that is: - ground having a pH level outside the 5/9 range -risks of sneak currents, chloride ion or high sulfate content (due to the use of de-icing salts, for examp1e) -for use as sea walls, for reinforcements without electromagnetic interference (airports,. ..). 433
The pins alone, in order to determine their premature breaking criterion as well as their resistance over time. The assembly consisting of a loop/pin attachment anchored in the concrete.
der to fit the two "interior" hoops corresponding to the anchoring in the concrete and the two external hoops corresponding to the pin's bearing area. (2) Characteristics sought The straps used are of the following type:
4. I Tests conducted on loops Table 1 - Loops characteristics Individual "grid-web
33 mm/l 5kN
strans
Strap width Breaking strain guaranteed on a single strip Theoretical breaking strain of a straight mother loop (4strands)
( I ) Prior testing Prior testing has been conducted concerning the alkaline hydrolysis resistance of the main reinforcements used. (2) Resistance testing on "mother" loops Resistance tests on the straight mother loops were conducted. These tests allowed us to fine tune the number of strands necessary as well as the amount of overlap to be respected. (3) Sheathbiber slippage tests Slippage tests were also conducted in order to highlight the possible existence of this phenomenon and to measure its speed in order to simulate its evolution for the expected service life of the work. These measurements showed that significant slippage occurred during loading, then extremely slow slippage which would amount to only 4mm over a period of 70 years. (4) Accelerated ageing tests In order to validate the ageing coefficient used, accelerated ageing tests must be conducted in order to fine tune, as the case may be, the coefficient used which is currently at a maximum.
44 mm/25kN
33 m m
44 mm/25 kN
15kN
25kN
15kNx4=60kN
25kNx4=100kN
(3) Identification The two loop models developed were: Loops "50 kN/500-600" for reinforcement grades 30 and 50 kN (or 500MM represents their flat measurement and 600MM represents the coverage length). Loops 100 kN/600-900" for primary reinforcement grades 75 and 100 kN. 3.2 Synthetic pins ( 1 ) Overall design The pin which is, strictly speaking, a pin functioning as a short beam resting on two simple bearing surfaces which are formed by the two hoops outside the concrete, described above. In this manner, this pin is subjected to flexion and shearing traction forces. (2) Investigations to be conducted While the technology of pultruded glass fiber components is being increasingly used in the field of civil engineering, two specific axes of research have appeared: Knowledge of the rupture mode for this specific type of stress, Physico- chemical behavior for the foreseen chemical environments. (3) Initial choices In light of these specificities, the initial choices were based on the use of "vinyl-ester" resin-based pultruded glass fiber pins.
4.2 Pin tests
(1) Mechanical strength tests The mechanical strength tests showed that the designing characteristic is the interlamilinar shear strength (ILSS). All of the measurements conducted showed an interlaminar shear strength of 60 Mpa (60 N/mm2). (2) Accelerated ageing tests These pins are being more widely used, although are relatively recent; as a result, their long-term behavior is not totally known. (3) Dimensioning retained A reduction/safety coefficient of 3 was retained relative to the long-term dimensioning of these pins, such that the following diameters were retained: - Reinforcements, grade 30 kN: 30 mm - Reinforcements, grade 50 kN: 36.8 mm
4 TESTS CONDUCTED Complete tests were conducted on: The isolated loops, either straight or gondola shaped, concerning their intrinsic mechanical resistance as well as the appreciation of possible fiber/sheath slippage risks, and their possible measurement.
4.3 Testing of the loop + pin assembly anchored in the concrete
-
Tests of the loop/pin assemblies anchored in the concrete were conducted, simulating the operating mode of the panels as much as possible.
434
5 FABRICATION 5.1 Loop fabrication
The loops are fabricated in simultaneous pairs (strips then loops) and are separated during cooling. The overlap weld is performed hot while maintaining strict control of the polyethylene softening temperature.
-
a seaside quay wall in Ireland, subjected to the effects of tidal fall. The use of reinforcements and metal attachments for this structure was excluded (figures 2 , 3 , 4 and 5 ) longitudinal canal walls, in the United Arab Emirates (figure 6)
5.2 Pin fabrication The pins are fabricated by continuous pultrusion. They are cut into 200 mm bars and the ends are dipped in the same resin to seal them thereby increasing their resistance to alkaline hydrolysis. The price of the pins depends directly on their diameter and the quantity manufactured during one fabrication series. This is why the following combinations (the most economical) were retained: Table 2 - Toggles combination Reinforcement grade Proposed
30
50
75
100
1x30
1x36,s
2x30
2x36,s
6 PRACTICAL APPLICATIONS Two initial practical applications were successfully performed :
Figure 5.A global view of DUN LAOGHAIRE quay wall Figure 2. A PARAWEB V strip connected to synthetic connection.
Figure 6. A night view of AL MAJAS CANAL once completed
Figure 3 . Loops in place before casting
43 5
7 DISCUSSION
8 CONCLUSION
Except for more precise accelerated ageing tests which remain to be conducted, the development of this entirely synthetic attachment may be considered as complete. It has the interest of being "universal" in that it does not require significant mold modification and functions with panels of varying thickness, regardless of the layout of reinforcement bars. This attachment has formed the subject of a patent application which is currently under examination. This connection works also with totally unreinforced panels; its strength depends directly of the handle spacement inside concrete: the larger the distance is, the strongest the connection will be. The ultimate strength of various components of this connection ie PARAWEB, loops, toggle, concrete are coherent. Moreover, whatever can be the factor of safety applied to main reinforcing strips, the same value shall be applied to connecting loops which means that this system remains always coherent.
The first true applications conducted allowed the pertinence of the choices made to be validated and the interest in this type of technology for construction structures in delicate physico-chemical conditions. This innovation extends the scope of use of concrete walls erected onto reinforced soils to uses or fill types that were previously discard, the latter corresponding to an inescapable evolution, as immediately useable backfill tends to be progressively unavailable
43 6
REFERENCES Orsat, P , Khay, M , McCreath, M 1999, reinforcements of polyester5bre housed in polyethylene ducts study of alkaline hydrolysis in the presence of high ph value "soil reinforcements by geosynthetics and related techniques" October 12-13 October 1999, Bordeaux France
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeiflinger, ISBN 90 2651 863 3
Mechanical behaviour of soil reinforced by geocells N. Racana LERMES, Blaise Pascal University, Clermont-Ferrand & Sol Solution S.A., Riom, France
R. Gourvits & M. Grediac LERMES, Blaise Pascal University, Clermont-Ferrand, France
ABSTRACT : Some techniques of earth reinforcement are based on geocells. This paper deals with the interaction between the geocells and the mechanical properties induced by the reinforcement. After the stage of soil densification, geocells can be roughly considered as blocks. Since the different cells are overlapped, the function of the mechanical connections between the cells must be clarified. To answer this question, an experimental and numerical analysis has been performed. The mechanical behaviour of the cellular layers will be presented in this paper and illustrated with the main results of the experimental and numerical studies.
1 INTRODUCTION A reinforced soil is a composite material made of inclusions resisting to the traction forces and placed into a compacted soil. The interaction between the soil and the reinforcement is essential for a good transmission of the stress. This interaction is usually created by the use of granular materials which permits to generate a rubbing interface between the soil and the reinforcement. The concept of soil reinforcement with geocells gives some more complex interactive mechanisms and features confined soil. Among the different techniques of cellular reinforcement, the Sol Solution walls made of threedimensional geocells is studied. This geocell reinforcement technique has contributed to realize more than thirty walls in France. For example, Sol Solution has developed an innovative technique consisting in tank confinement based on a complete stability study carried out with earthquake resistant data (see Figures 1, 2). The present new research consists in an analysis of the mechanical behaviours of the geocell wall. The structural behaviours of geocell reinforcement,
Figure 2. Building site scheme.
the interaction between the geocells and the influence of the textile modulus are studied to optimise the type and the repartition of the inclusions. It is expected to better understand the mechanisms of this process. 2 BEHAVIOUR OF ONE GEOCELL LAYER 2.1 Experimentation This first experiment allows to analyse the interaction between nonconnected geocells. The filling material is a two-dimensional granular medium (Schneebeli 1956, Mezghani 1987). The layer rests onto an inclined plane so that it is mainly subject to the force of gravity (Fig. 3). This experiment is useful to study the compaction effect of the structural behaviour. When the force of gravity is applied to the geocell layer, the lack of density of the granular medium creates voids inside the geocells. After the stage of densification, geocells can be roughly considered as blocks. The compaction effect is clearly pointed out by the small deformation (less than l 'Yo) of the layer with an important densification.
Figure 1. Confining transverse section.
43 7
The interaction between the different cells is emphasized by the repartition of horizontal stress. The stress concentration in the middle of each cell is presented in Fig. 5. This stress repartition shows that the geocells create a rubble mechanism. 2.3 Conclusion Those experimental and numerical approaches allow to analyze the interaction between the geocells. This interaction depends on the density of the filling material as well as on its compaction. With an important density, geocells can be roughly considered as blocks. Their interaction creates a self-blocking balance. Figure 3 Experimental setup
3 STRUCTURAL STUDY OF REDUCEDSCALE WALLS
2.2 Numerical approach The numerical model uses the finite difference method (Billaux 1992). This model built with the FLAC package allows to study qualitatively of the nonconnected geocells' behaviour. Geocells are modeled with the cable elements (Fig. 4). This numerical approach permits to better understand the interaction between the different cells. The gravity force is applied to the geocell layer which is fixed at the top by beam elements. The mesh and boundary conditions are shown in Figure 4.
3.1 Experimental approach Many experimental analysis of geocell reinforced walls show that the reinforcement leads to a shear strength increase which can be assimilated to an apparent cohesion (Breuil 1993, Gourves 1996, Reiffsteck 1996). The object of these new experiments is to study the effect of geocells without connections. Two reinforced reduced-scale walls have been built up with several nonconnected geocell layers. The only difference between those reduced-scale walls is the mechanical properties of the reinforcement. The filling sand has a friction angle of 38" and has no cohesion. 3.2 Paper reinforcement The wall to be tested is reinforced by paper nonconnected geocell layers. The young's modulus of the paper is 3GPa. The geocell layers dimensions are hight 1.7cm, depth 70cm and length 83cm. The diameter of each cell is roughly 5cm. The loading is applied to the wall through a metal load repartition plate. A l0OkN jack and its hydraulic power station are used.
Figure 4. Geometry and boundary conditions.
Figure 6. Wall instrumentation.
Figure 5. Horizontal stress repartition
438
The measurements of the external behaviour are based on the analysis of the facing strain and the sheet displacement. The measurement device helps us to control nine displacement sensors (from A to 1) and one force sensor (sensor J) (Figures 6,7). Friction occurs between the soil and the lateral plates. To avoid this effect, the lateral plates are lubricated (the soil is protected by a thin plastic membrane and grease). Sensors D and E permit to study the lateral effects. The loading increases at a fairly linear rate, as can be observed in Figure 8. No boundary effect is obtained (see Fig. %a).
The stress-strain relation applied to the loading distribution plate permits to determine the stiffness defined by the ratio between applied loading and displacement : 7kWmm. At a load of SOkN, a sudden collapse of the wall occurred. The failure line (Mohr-Coulomb type) underlines that the soil-geocells composite exhibits the same friction angle as the filling sand. The cells have an effective strength which can be interpreted as an apparent cohesion (Fig. 9). With the limit equilibrium method of the Coulomb corner, the tension sum of the reinforcement can be determined. cos(n/4- p/2) - tan(p).cos(n/4+ (012) T=(F+W) . Sln(Z/4 - p/2) + tan(p).sin(n/4 + p/2) where W is the weight of the Coulomb corner, F the vertical loading and tp the friction angle of the filling soil. The block method gives the average tension in each reinforcement. A comparison between the average tension and the strength can be performed with the tensile test. The reinforcement of the wall broke with the traction forces and did not with the shearing strength. It is important to notice that the passive anchorage has not been deformed. 3.3 Polyan reinforcement The Young’s modulus of the polymer used in the second experiment is thirty times less important than the modulus of the paper (Fig.10).
t 0
10 20 DISPLACEMENT [mrn] a
90
0
- Boundary effect
Figure 9. Shear strength increase
,
-2
4.E+04 3.E4-04 2.€+04 _.
80
N
E
2 30
O.E+OO 0.1
20 10
0 0
5 10 15 DISPLACEMENT [mm]
IQ0
Figure 10. Stress/strain curve ; Paper and polyan specimens.
20
b - Facing displacements.
1 10 Strain %
439
In this second experiment the usual displacement sensors could not be used to measure the facing strain. The dimensions and the design of the wall are the same as the paper reinforcement’s one. Only the distribution plate is equipped with displacement and force sensors (Fig. 1I). The facing displacement is captured and taped by a video system (Peuchot 2000). Following the observation of the gravity center of some marks (Fig. 12) the three-dimensional measurements of displacement is obtained. The distribution plate is also determined by the cameras. The accuracy of this technique is 1/10000 of the distance between the instrumented object and the video camera. In this experiment, the accuracy is therefore of 0.lmm (the camera is 1 m far from the wall). The measurements of the facing obtained with the forty-two marks show that the global displacement of the facing is homogeneous (Fig. 13). This second structure is softer than the paper geocell reinforcement. An homogeneous displacement of the plate and an initial stiffness (ratio between applied loading and displacement) of 1 . 4 k N / m (Fig. 14) are observed.
The shearing zone is observed for each layer (Fig. 15). The passive anchorage has not been subject to any strain. The failure surface and the distortion of the reinforced soil are presented in Figure 16. The shearing zone is bounded by the passive anchorage and the active block. The failure of MohrCoulomb type shows that the composite reinforced soil exhibits the same friction angle as the filling soil.
Figure 13. Displacements evolution.
Figure 1 1. Experimental shape Figure 14. Plate displacement.
Figure 12. Experimental setup. Figure 15. Layer’s distortion.
440
geocell reinforced wall. Cells’ interactions are analysed with a geocell layer without connections. This approach is useful to observe the rubble mechanism. The reduced-scale wall experiments show that geocell reinforcements can be assimilated as an apparent cohesion. Failure mechanisms are not different if a soft or a rigid structure is used. Some previous tests carried out on reinforced soils with connected geocells gave similar results in terms of displacement and strength. Consequently, the geocells without connections exhibit the same properties as the geocells with connections in this type of test.
Figure 16. The failure surface.
3.4 Comparison between the two reinforced walls REFERENCES
The only difference between the paper and the plastic reduced-scale walls is the nature of the reinforcing material. The ratio between their elastic modulus is about thirty. The behaviour of a reinforced soil depends on the mechanical characteristics of the reinforcement. In this case, the initial stiffness of the composite soil-reinforcement is 7kN/mm for the rigid structure and 1.4kN/mm for the soft wall. The magnitude of the displacement is very sensitive to the Young’s modulus of the geotextile. Nevertheless, the failure mechanism is the same: no passive anchorage movement is detected, the same phenomenon is observed. We have an active block corner.
Billaux, D. & P. Cundall. 1993. Simulation of geomaterial with a lagrangian elements method. Revue Franqaise Geotechniqzie. N 63:296-303. Breuil, D. 1993. Etude du comportement de mur de soutenement realises a l’aide d’un geotextile tridimensionnel alveolaire. technical report. Blaise Pascal University. Gourves, R., P. Reiffsteck. & J.F. Vignon 1996. Design of geocells reinforced soil structure through a homogenization method and a finite difference method: Comparison and charts. International symposium on Earth Reinforcement, Kyushu. 357-362. Mezghani, A. 1987. Milieu granulaire : Analyse statistique de I’etat de contrainte macroscopique au sein d’un milieu analogique. doctoral dissertation. Blaise Pascal University. Peuchot, B. 2000. 3 D video measurement material brochure. Vidkonzetric. Reiffsteck, P. 1996. Etude du comportement mecanique du geotextile tridimensionnel alveolaire ARMATERB. doctoral dissertation. Blaise Pascal University. Schneebeli, G. 1956. Une analogie mecanique pour les terres sans cohesion. Compte rendu des sckances de 1’Academie des Sciences, Tome 243 : 125.
4 CONCLUSION The mechanism of a geocell layer under loading is analysed to better understand the behaviour of a
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Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lS5N 90 2651 863 3
An analytical approach to compute design loads of reinforced earth embankments M.V. Ratnam Consultant, Hyderabad, India Cformerly: Dean, Kakatiya University)
M.B. Rao Professor of Civ. Eng., Arbaninch Water Technology Inst., Ethiopia Cformerly: Research Scholar, Kakatiya University)
ABSTRACT: This paper presents the details of a large scale fully instrumented reinforced earth (RE) model wall (provided with teak wood facing panels and hand cut metallic strips for reinforcement) tested to failure under surface area (rectangular) load with varying edge distances. All the instrumentation used in this study, such as strain gauges, lateral and vertical earth pressure cells, an eighty six point light emitting diode (LED) tell tale system for sequential record of strip breakages etc.(designed, fabricated, calibrated and tested by the authors) was used to measure various values including lateral pressure on facing elements and tensile stress variation along reinforcing strips. For correlation studies, these values are compared with the corresponding values computed from classical earth pressure theories modified to simulate the test conditions. The load intensity causing first strip breakage is taken as the ultimate load on the wall and this value divided by a factor of safety 3 considered appropriate in view of the various constraints affecting field construction procedures and quality assurance, is taken as the allowable load.
1 INTRODUCTION
The model box of size 1.03 x 1.12 x 1.08 m has three sides made up of wooden planks stiffened with steel sections and the fourth side is open and kept free from the adjacent sides to facilitate fabrication of RE wall comprising of hand cut steel strips 800 x 10 x 0.137 mm for reinforcing elements provided with a hole at either end to receive a 3mm steel bolt with washers to facilitate connection to the skin element at the facing end and LED circuit connection at either end. The horizontal and vertical spacing of these strips are 27 cm and 6 cm respectively. The skin elements (86 nos.) of teak wood, cruciform in shape and 180 x 120 x 20mm in size are made up of two pieces 13 mm and 7 mm thick laid one over the other and nailed such that edge projections are available for necessary interlocking. The facing end of each strip is connected to the facing element at its center through an aluminium T bracket connector of size 20 x 18 x 18 mm screwed into the panel at its center. Soil used for the back fill is a medium sized sand (sw - sp) obtained from a nearby river source. The soil is compacted in layers such that the finished thickness of each layer equals the vertical spacing of strips. The strips are laid horizontally on the surface of the compacted layer and properly positioned and the next layer of soil is spread uniformly to needed thickness and compacted. This process is repeated till the model is completed. All the relevant properties of the materials used for facing elements, joints, reinforcing strips and the soil used, are determined and checked to satisfy the specifications and for use in the design. The arrangement of facing elements,
Reinforced earth walls with galvanized steel strips for reinforcement and R.C.C. cruciform shaped facing panels have come to stay as a credible innovative construction technique for approach embankments of flyovers, abutments etc. Considerable research and performance studies on laboratory model and field prototypes have resulted in better analysis, design methods and construction techniques as well as specifications for materials to be used with its attendant improved performance and quality assurance. Nevertheless, correlation studies between lateral earth pressures actually measured in tests and those obtained from classical earth pressure theories duly modified to simulate test conditions have been scarce. This paper details an attempt to modify the expressions (based upon classical theories) to evaluate the load intensity that causes the first strip breakage and the variation in lateral pressures on facing elements to obtain reasonable correlation with those actually measured in laboratory tests.
2 DESCRIPTION OF THE MODEL The assembly of the RE wall model was built in a model box that fits into a self straining loading steel frame fitted with an appropriate loading mechanism and load measuring device comprising of a hydraulic jack and a proving ring each of 100 tons capacity. 443
Figure 3. Layout of instrumentation - L.E.D. display unit (common to all tests).
pressure cells to record i) toe pressures at either end and center of RE wall, ii) lateral pressure along the height of one rigid vertical side of the model box, iii) vertical pressure in the compacted soil fill at three different depths vertically below the center of the loaded area, d) strain gauged (half bridge) reinforcing strips to record the tensile stress variation along their length. These are located at the relevant and vulnerable positions (usually along the central section of the loaded area) to obtain the peak values of various parameters. Also, included is the apparatus to conduct pull out test on the dummy additional strips provided, equipment for calibration of pressure cells, dial gauges and datum bars to measure lateral deflection of RE wall, and dial gauges to measure surface settlement at the four corners of the loaded surface area in all the tests conducted up to failure.
SURFACE SETTLEMENT GAUGES, @ : STAINLESS STEEL EAI?TH PRESSURECELL EMBEDDED IN FACING PANEL, 52: FACING PANEL NUMBER. - : LOCATION OF REINFORCING STRIP ATTACHED TO PANEL, :DIAL GUAGE TO RECORD LATERAL DEFORMATION OF PANEL, @ :PANEL WITH STRAIN GAUAGED STRIP
V 1 ,V2.V3.V4:
a
Figure 1. Arrangement of facing panels, and location details of instrumentation.
Figure 2. Plan of loaded surface area for different tests showing edge distance from r.e.wal1.
and the details of instrumentation provided, common to all tests, are presented in Figure 1. Each strip is connected to the LED assembly that is provided with a bulb projecting out at the center of each skin element which blows out when the strip breaks and thus breaks the circuit. The actual lay out of loaded surface area and edge distances adopted for all the tests conducted are shown in Figure 2. The circuit details of the LED assembly are shown in Figure 3.
3 INSTRUMENTATION USED a) Multi-channel switching and balancing units for measurement of strain gauged bridge output simultaneously from about 60 locations. b) 50mm diameter stainless steel diaphragm type strain gauged earth pressure cells to record lateral pressure on skin elements. c) 100 m diameter brass diaphragm type
444
4 TEST PROCEDURE The construction of model is completed after ensuring that the instrumentation has been duly installed at each stage as planned and ensuring its satisfactory working. The plates one over the other decreasing in size from bottom to top plate for transfer of load applied, are arranged over which the load application device and measuring device are properly positioned. All initial readings from the instrumentation are also recorded. The load is applied gradually in increments of 50 kN/m2. After each load increment, all the instruments are checked, read properly and the values recorded. This is repeated till the first strip breakage occurred, as indicated by the blow out of the glowing bulb at the center of the skin element connected to the broken strip, and the test is continued till failure, as indicated by either excessive lateral deformation of the wall leading to considerable gap between adjacent skin elements with attendant leakage of soil fill or by breaking of reinforcing strips accompanied by noise and excessive outward
movement of skin elements or a possible combination of both. Later, the set up is carefully dismantled, and the sand filling is carefully removed in stages to facilitate complete and correct record of location and mode of breakage of all torn strips. Quite a few strips failed at the bolt hole connection with the facing elements. The same procedure is repeated for all tests conducted.
5 COMPUTATION OF RESULTS 5.1 Load intensity causingfirst strip breakage 5.1.1 Experimental approach For any test, the load intensity corresponding to the first breakage of strip(s) is obtained from the LED display board. The total load applied as indicated by the proving ring divided by the loading area gives the value.
Figure 4. Boussinesq’s approach (for area loading).
B= x2ylz / (XI2+z2) R x2, C= tan
-’ (xlyl / zR ,I),
D=(xi ylZ)/(X12+z2)R, 1, Rxl= ( X I
5.1.2 Analytical approach After a discreet study of the various classical theories, Modified Spangler’s (1982) approach is found to give values closest to the experimental values in respect of the surface area load intensity causing the first strip breakage. The approach to derive the expression is described below briefly. For a point load P, Boussinesq’s formula for the lateral pressure s xx on a flexible wall imagined to be a simple vertical plane in an elastic half space is given by
2+
y1
+z
andR,2=(x22+y I 2 + z 2 ) ” Lateral pressure distribution on the RE wall facing elements due to an area surcharge will be maximum on the central vertical section of the loaded area. The values for each symbol used in the equations are obtained for each test conducted. The value of the reduction factor R* is evaluated from the relationship R*’= 1 / (1+ k”)
s xx = (3P.x2z)/ (2 ER’)
(1)
(5)
Where k* is the relative stiffness factor (Relative flexural rigidity) of the facing or skin of a RE structure.
taking poisson’s ratio p for back fill as 0.5 and R = ( x2 + y2 + z2 )”.. The Spangler’s formula, a modification of Boussinesq’s point load formula for a rigid retaining wall, considered by Terzaghi (1943) and Spangler (1982) etc. is given by
s ,, = 2 (3P .x * z ) / ( 2 n R’)
)”
For a perfectly rigid facing, k* -+ 00 while for a perfectly flexible facing, k*+ 0. If k*=O then
(2)
Spangler’s modified formula intuitively arrived at for computing the lateral pressure on a RE wall facing having partial rigidity is given by the expression s xx = (2 - R”) (3P.x’ z) / (2 n R5 )
(3) where ps and pp are the Poisson’s ratios of the reinforced soil (assumed as 0.35 ) and wall facing material (taken as 0.25 for teak wood ) respectively. For steel and reinforced concrete these values are 0.3 and 0.15 respectively. E, and E, are the stress-strain moduli for facing material and reinforced soil. The latter is taken approximately equal to that of unreinforced back fill soil, which depends upon a number of factors such as in-situ density, rate of loading, vertical stress, confining pressure, stress history etc.
where R* is a reduction factor explained below. This equation is integrated for a rectangular surcharge load intensity q (t/m2), shown in Figure 4, and the lateral pressure on the wall facing due to superimposed area load is given by s xx = (2 - R*) (2) q (A-B-C+D) / 27~ Where A = tan
(x2 y1/ z R
(4)
2),
445
and Rf is a failure ratio equal to (0, - oh) at failure divided by the asymptotic value of (S, - Sh), k is a modulus number, n is an exponent determining the rate of variation of initial tangent modulus with Oh, G~ is atmospheric pressure (10.13 t/m2), c is cohesion and 4is angle of internal friction of the backfill E is taken as zero i the value of M 21. Values of Rf, k, n, c, and Cp are to be determined experimentally. However, in the absence of tests the following values can be assumed in respect of a cohesion less soil fill. R ~ 0 0 . 7 0 ,k=400, n=0.5, c=O and Cp=36'. The major principal stress o,= (01) is due to the self weight of backfill and the surface load above. The vertical stress component due to the surcharge is assumed to disperse uniformly on an area bounded by 2: 1 dispersion. This assumption is justified as shown by Bowles (1982) that there is insignificant difference between the vertical pressure under a rectangular area load predicted by Boussenesq's theory and that derived from 2 to 1 stress distribution at critical depth equal to or greater than the width of the area load. The lateral pressure on the RE wall facing ox, can be computed as the sum of the lateral pressures caused due to compacted soil fill and the surcharge load applied on the surface.
6 CORRELATION OF DATA AND DESIGN 6.1 Lateral earth pressure on RE wall facing
Figure 5. Lateral pressures (recorded and computed).
This is to be determined from the slope of the tangent drawn at the relevant point on the stress strain curve obtained from triaxial compression test conducted in the laboratory on back fill sample (CD test). Alternatively this may be obtained as E, (Tangent modulus for the backfill) as proposed by Kondener and Zelasko (1963) and modified by Duncan and Chang (1970) from the equation:
(7) where M = Rf (1-sin Cp)
(0, - o h )
(2ccos @2oh sin Cp) 446
Figure 5 presents the correlation study in respect of lateral pressure variation along the height of RE wall facing for its full height as obtained from (a) modified Spangler's approach and (b) the recorded values from instrumentation installed in the model for the surface area load intensity causing first strip breakage. The results are presented for four of the tests conducted with varying edge distance and size of loaded area as detailed in Figure 2. The analytical values fall within a small band width while the measured values show wide variation. A notable factor is the edge distance variation from test to test. In some tests it is less than minimum value usually specified. This could not be monitored because of limitations imposed by the model dimensions chosen. However, the edge distance is taken as the minimum specified or the actual value (whenever it is higher) for computation purpose. The analytical values are slightly on the higher side of observed values, thus ensuring safety. A factor of safety 3 is recommended to make good for the constraints that affect the quality assurance on the field. As such it is recommended that for any given RE wall either built or to be built, the area load intensity that causes the first strip breakage can be computed as the ultimate load. The permissible load can be arrived at with a factor of safety subject to a minimum edge distance of 1 m or one tenth height of wall (higher of the two
Figure 6 (a), (b), (c). Measured values of tension compared with computed values.
6.2 Maximum tension occurring in the strips
values). The design of the reinforcing strip against tension mode of failure is as follows:
T = oXx AH AS = o,,b t / F.S.,
(8)
where b and t are the breadth and thickness of the strip and AH and AS are the vertical and horizontal and o,, are the stresses. spacing of the strips, and oxx Against pull out mode of failure.
T =oxx AH AS= o,, (max)2b tan (pL, / F.Sp
(9)
where F.S., and F.S., are factors of safety against tension and pull out modes of failure of strips, cTzz(max) for a given load intensity q, taken as the maximurn of all the values computed for different depths up to the bottom of the wall, is adopted in the design. Leis the effective grip length of the strip.
447
The model wall is subjected to incremental load intensity over the total loading area in each of the tests. Figure 6 presents the values of maximum tension occurring in each of the strain gauged strips extending for the full height, at the center section of the loaded area for the load intensities of 10, 20 and 30 tons per square meter, marked (a), (b) and (c) in the figure, in respect of 4 tests conducted. The measured values shown are the corresponding maxi-mum tensions measured in each of the strips strain gauged at five locations along the length. As can be seen, the computed values as per modified Spangler’s approach, form a band width and nearly envelop the recorded maximum tensions. This again indicates the validity of the use of modified Spangler’s analytical approach.
7 CONCLUSIONS
REFERENCES
(a) A light emitting diode tell tale system that can be connected to an innumerable number of reinforcing strips, has been designed and its satisfying performance has been established, for obtaining a record of surface area loading intensity that causes the first strip breakage, followed by a record of sequential strip breakages that occur upto the total failure of the RE wall.
Bowles, J.E.( 1982) Foundation analysis and design, McGrawHill International Book company,New York Carothers, S.D. (1920). Direct determination of stresses.Proc. Royal Society, London.Ser.A.Vol.97 Duncan, J. M. & C.Y. Chang, (1970). Non-linear analysis of stress and strain in soils. ASCE Journal of Geotech. Eng.,Vol. 96(5):1629-1653. Hausmann, M.R.( 1976) Strength of reinforced soil. Proc. 8th Aust. Road Research Conf.8, Sect.13:l-8 Ingold, T.S. (1982). Reinforced earth. Thomas Telford: London. Jones, C.J.F.P.( 1985) Earth reinforcement and soil structures. Butterworths, London Kondener, R.L. & J.S. Zelasko, (1963). A hyperbolic stressstrain formulation for sands. Proc. 2"d Pan American conf. on S.M.F.E..Brazil Vol. 1: 289-324. Rao, M.B. (1987) Design, construction and performance. Studies of a fully instrumented large scale model of reinforced earth wall under surface area loading. Ph..D. thesis, Kakatiya Univ., India. Spangler, M.G. & R.L. Handy, (1982) Soil Engineering, Harper and Rev. publishers, New york. Terzaghi, K. (1943) Theoretical Soil Mechanics, John Wiley & sons, New York. Venkata Ratnam, M. et al. (1987) Stability of a large scale model R.E. retaining wall. Proc. 81h ARCSMFE, Kyoto, Japan. Vidal, H. (1978) The development and future of reinforced earth. Keynote address: Proc. ASCE symposium on Earth reinforcement, Pittsburg, USA.
(b) The analytical approach using modified Spangler's theoretical equation, so far as the limited data available from this study, holds promise to predict the ultimate load as well as design load intensity for any proposed or built up RE wall for which the data in respect of all the materials either used or proposed to be used is made available. (c) It is considered prudent to recommend a factor of safety 3 for the load that causes the first trip breakage, to compute the safe design load for any given RE wall with details of design load, construction materials and construction technique as well as site characteristics. (d) The limited data available from this analytical approach, tallies reasonably with the criteria for internal stability against tension failure and pull out failure usually adopted in the design of RE walls.
448
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Evaluation of seismic performance in Mechanically Stabilized Earth structures J.E. Sankey The Reinforced Earth Company, Vienna, Virginia - USA
P. Segrestin Freyssinet International, Velizy, France
ABSTRACT: Recent earthquake events have brought about renewed interest in the response of a variety of structures to seismic loads. In the case of mechanically stabilized earth structures, such as Reinforced Earth@, current seismic design codes do not appear to fully incorporate their inherent flexibility. A brief catalogue of major earthquakes and corresponding descriptions regarding the condition of local Reinforced Earth structures is provided to demonstrate the realistic flexibility of the structures. A call for better consideration of the ductile response of Reinforced Earth is recommended based on its flexible composition of discrete steel reinforcements and select soil matrix. will then be compared to the criteria used in design for the walls. Even in cases where the seismic accelerations exceeded the design accelerations, it will be shown that little if any distress resulted. The rationale shall be presented that the ductility of the Reinforced Earth may allow minor permanent deflections to occur without distress that would affect service life. Although methods of calculation could be presented to predict the allowable deflections, it is instead suggested that a monitoring program be established to better assess stability versus deflection in seismic events. Suggestions for the future monitoring program will be discussed at the conclusion of this paper with an eye toward establishing a displacement criterion in design of Reinforced Earth.
1 BACKGROUND In the last decade there have been major earthquake events in the United States (Northridge, California, 1994, 6.7 Richter magnitude), Japan (Great Hanshin, Kobe, 1995, 7.2 Richter magnitude), and Turkey (North Anatolian, Izmit, 1999, 7.4 Richter magnitude). The Northridge Earthquake was responsible for 57 deaths, 11,000 injuries and $20 billion US in damages. The Kobe Earthquake was a terrible tragedy that killed more than 5,000 people, injured 27,000 more and destroyed over 150,000 structures (houses, buildings, bridges, elevated roads, port works and utility services). The even more tragic Izmit Earthquake resulted in 16,000 deaths, 30,000 injuries and over $16 billion US in damage. In the three earthquakes cited, there were numerous Reinforced Earth structures constructed near the respective epicenters of the seismic events. The relative flexibility of Reinforced Earth walls and their ability to withstand distress in the face of large horizontal and vertical accelerations appears to set these structures apart from the more rigid structures where significant damage occurred under seismic action. Yet by most building codes, Reinforced Earth structures are routinely required to be designed using the same quasi-static design loads as those given for rigid structures. The use of quasi-static analysis, though simple, neglects the fact that Reinforced Earth can displace to a certain extent without showing significant damage. Resulting designs may extend the reinforcements to unreasonable lengths. The purpose of this paper is to briefly catalogue the condition of Reinforced Earth structures subjected to seismic events in the Northridge, Kobe and Izmit Earthquakes. The actual physical condition
2 REINFORCED EARTH SEISMIC DESIGN French engineer and architect Henri Vidal invented Reinforced Earth over 30 years ago. The most common system is a composite material formed by the placement of granular soil and steel reinforcements as shown on the section in Figure 1. The linear steel reinforcements are connected in turn to individual concrete facing panels. The panels are held into place by the interfriction that results between the reinforcements and granular soil. Reinforced Earth is simply a coherent gravity mass engineered to be internally stable while at the same time resisting externally applied loads. Under seismic conditions, Reinforced Earth is usually designed considering quasi-static design loads. The design method is intended to proportion structures on the basis of horizontal accelerations given in seismic areas with appropriate factors of 449
Figure 1. Typical section of a Reinforced Earth Structure
safety for internal and external stability. A summary of the seismic design procedure for Reinforced Earth follows. 2.1 Internal stability The flexibility of Reinforced Earth is fairly well addressed in internationally recognized design codes, which mirror pseudo-static numerical or reducedscale models. The reinforcements found in Reinforced Earth are designed to withstand a combination of static forces and horizontal dynamic forces generated by the inertia of the retained soil in the active zone. The inertia force is distributed to the individual reinforcements proportionally to the available resistance of the reinforcements (resistant zone) at each level within the structure. Under seismic loading conditions, it is understood that pullout resistance of the earth reinforcements can be reduced as much as 20% for accelerations as high as 0.4 g. Therefore, pullout resistance is conservatively taken at 80% of the resistance used in static only conditions. Factors of safety against pullout and tensile rupture of the reinforcements under combined static and seismic loading may be reduced to 75% of the factors of safety required in static only conditions. 2.2 External stability The Mononobe-Okabe pseudo-static method is used in external stability computations of Reinforced Earth. However, the static force is only combined with !A of the dynamic earth pressure and 50% of the full inertial force of the wall. Factors of safety against sliding and overturning may be reduced to 75% of the factors of safety required in static only conditions. 450
3 PERFORMANCE OF REINFORCED EARTH STRUCTURES A review of the Reinforced Earth structures near the epicenters of the Northridge, Kobe and Izmit earthquakes indicated very little damage occurred to any of the structures. Of significant interest are the design-versus-actual horizontal accelerations and resulting permanent wall deflections recorded. 3.1 Northridge earthquake A total of 23 Reinforced Earth structures were located within the affected area of the earthquake. Of these, over 65% were higher than 5 m and over 25% are higher than 10 m (Frankenberger, Bloomfield & Anderson 1997). The distance of the Reinforced Earth structures from the earthquake epicenter ranged from 13 to 83 km.The estimated ground accelerations varied horizontally from 0.07 g to 0.91 g and varied vertically from 0.04 g to 0.62 g. The only damage that appeared was minor spalling of concrete panels in some of the walls. It is interesting to note that adjacent structures to the Reinforced Earth structures, such as buildings, suffered much more severe damage and in some instances were posted as unsafe. Of even more interest is the fact that over 75% of the Reinforced Earth structures were designed using lesser horizontal ground accelerations than actually occurred, and over 50% of the Reinforced Earth structures were designed using no consideration for horizontal ground accelerations at all. 3.2 Kobe earthquake Of the 120 Reinforced Earth structures inspected after the earthquake, 70% were over 5 m high and 15% were over 10 m high. The structures were de-
signed using estimated ground accelerations of 0.15 to 0.2 g. The actual ground acceleration was 0.27 g. Ground movement was evident above or next to 22 of the structures, with 10 walls showing minor cracking of isolated concrete panels and 3 walls exhibiting significant lateral movement (Tatsuoka 1995 and Kobayashi, et.al., 1997). Deformations recorded in walls at Awaji Island and Hosiga-oka Park varied between 4 mm to 113 mm (displacement relative to bottom of the panel at mid-height and top of walls). All of the walls remained functional after the earthquake. 3.3 Izmit earthquake
A full evaluation of the Reinforced Earth structures for this particular earthquake area has not yet been completed. However, one bridge and ramp structure was surveyed at Arifiye, almost immediately adjacent to the epicenter (Segrestin 2000 and Asheim & Mander, 2000). Although the bridge itself collapsed, the Reinforced Earth ramp walls sustained only nominal damage and remained stable (Figures 2). Shear deformations from differential settlements propagated upward through the panels, separating some by as much as 75 mm. The Reinforced Earth walls were designed for a ground acceleration of 0.10 g. This resulted in only a minor increase in the amount of reinforcing strips compared to static design. Yet the actual ground acceleration was measured at 0.4 g. It is interesting to note that if the full effect of ground acceleration were considered in design under current practice, then at least 40% more reinforcement would have been added. The fact that the increased reinforcement did not prove to be needed is a good indication of the safety of the technology and conservative nature of current design principles. 4 MODIFICATIONS TO SEISMIC DESIGN PRACTICE Design of Reinforced Earth structures under seismic conditions is based upon the concept of a rigidplastic mechanism. However, reports of favorable deformation behavior of Reinforced Earth during recent seismic events in the United States, Japan and Turkey suggest that plastic deformation evaluation of the soil-reinforcement system should be admissible. The extent of deformation is dependent somewhat on the height of the structure and length of reinforcement. The Reinforced Earth Company is contemplating a survey of its structures in seismic areas to further the concept of plastic deformation and its evaluation. It is noted that the plastic deformation model is relevant only as far as the soil-reinforcement interaction is concerned. It may not be enough in itself to take into account the potential brittle failure of corroded strips, which would require separate consideration. 45 1
Figure 2. Arifiye Reinforced Earth ramp wall. Photo Asheim & Mander, EERI.
Finite element analyses (Segrestin & Bastick 1988) indicate that the zone of maximum stress moves out only slightly with the addition of dynamic forces and may essentially be ignored. Shake table studies (Bathurst, et. al., 1996, and Sakaguchi, et. al., 1992 and 1996) suggest that lateral wall displacements only tended to increase with decreasing reinforcement length if the reinforcement length was less than 0.7H, where “H” is the total height of wall. The reinforcement length of 0.7H is the minimum basis for design of Reinforced Earth structures. Adding to the minimum length for purposes of seismic external stability alone should not be necessary and the requirements of the static mode of design (sliding and overturning) should generally govern. The height of a Reinforced Earth structure affects the amplification of acceleration found in the structure, i.e., fundamental frequency of the wall versus predominant frequency of the earthquake motion. It is noted that strong motion earthquakes typically have predominant frequencies of 3 Hz or less. Compare this predominant frequency to the fundamental frequencies of short walls at 10 Hz and tall walls at 3 Hz. It is not surprising that short walls, even if not designed for horizontal ground accelerations, are more or less unaffected by seismic events. It is recommended that seismic design consider the height of the Reinforced Earth structure when making evaluations, and de-emphasize the reliance on ground acceleration. Finally, design codes should introduce a plastic deformation model in the evaluation of Reinforced Earth structures. Displacement-based design is suggested here primarily as a means to justify the reduction of reinforcement length (Michalowski & You 2000). Current design relies on no deflections to occur, as would be calculated in a steel frame structure. Instead, a deformation-based design is suggested for Reinforced earth structures that applies a safety fac-
tor to the true soil-reinforcement strength parameters, and by indirect means to the displacements calculated. 5 CONCLUSIONS Reinforced Earth structures have proven to be safe and flexible in the presence of seismic events throughout the world. Current design codes apply a very conservative approach, especially in the determination of external stability. This paper suggests that a plastic deformation approach be taken instead in the design of Reinforced Earth structures; whereby consideration for seismic design will be based on wall height, ground acceleration, and allowable deformation. The wall height will determine how much reliance needs to be paid to seismic design, with lower height walls being less restrictive than moderate to tall walls. Recognizing that Reinforced Earth walls can deflect and remain stable means that establishing an inventory of wall deflections after seismic events and corresponding wall heights will be important. It is recommended that a survey of Reinforced Earth structures be undertaken. To be reliable, the location of the relationship of the base of the walls with respect to the upper portions of the walls needs to be established, preferably in seismically active cities where a number of these walls would be concentrated. When significant seismic events occur in the cities where base line surveys have been completed, then follow up measurements should be undertaken. It is anticipated that actual deformation readings may be used to tailor better design models and establish more realistic and economical reinforcement lengths in safe design.
452
REFERENCES Bathurst, R J & Alfaro, M C 1996 Review of Seismic Design, Analysis, and Performance of Geosynthetic Reinforced Walls, Slopes, and Embankments Keynote Paper, IS-Kyztshu '96 32 pp Fukuoka, Japan Frankenberger, P C , Bloomfield, R A & Anderson, P L 1997 Reinforced earth walls withstand Northridge Earthquake Earth Reinforcement, Technical papers prepared by Groupe TA1for The International Symposium on Earth Reinforcement, Firkuoka, Kyushu, Japan, 12 - 14 November 1996 47 - 52 Rotterdam A A Balkema Kobayashi, K , Tabata, H & Boyd, m> 1997 The performance of Reinforced Earth structures during the Great Hanshin Earthquake Earth Reinforcement, Technical paper prepared by Grozipe TA1for The International Symposiiini on Earth Reinforcement, Fukiioka, Kyshii, Japan, 12-14 November 1996 41 -46 Rotterda A A Balkema Michalowski, R L & You, L 2000 Displacements of Reinforced Slopes Subjected to Seismic Loads ASCE Journal of Geotechnical and Environniental Engineering 685 - 694 Reston, Virginia ASCE Production Services Department Sakaguchi, M 1996 A Study of the Seismic Behavior of Geosynthetic Walls in Japan Geosynthetic International 13 30 Volume3,No 1 Sakaguchi, M, Muramatsu, M & Nagura, K 1992 A Discussion on Reinforced Embankment Structures Having High Earthquake Resistance Earth Reinforcenient Practice, Proceedings of the International Symposium on Earth Reinforcement Practice 287 -292 Fukuoka, Japan Segrestin, P 2000 Performance of Reinforced Earth Retaining Walls Near the Epicentre of the Izmit Earthquake (Turkey, August 17, 1999) Soiltech Monograph 12 pp (TA 2000) A 901 Freyssinet, Velizy, France Segrestin, P & Bastick, M S 1988 Seismic Design of Reinforced Earth Retaining Walls - The Contribution of Finite Element Analysis International Geotechnical Symposium on Theory and Practice of Earth Reinforcement Fukuoka, Japan Tatsuoka, F , Koseki, S & Tateyama, M 1995 Performance of Geogrid-Reinforced Soil Retaining Walls During the Great Hanshin-Awaji Earthquake, January 17,1995 Proceedings, 1' I International Syniposiiini on Earthquake Geotechnical Engineering 55 - 62
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Use of a l g geogrid reinforced-wall model to evaluate the effectiveness of a FE numerical code M. Schiavo & P. Simonini IMAGE, University of Padova, Italy
G. Gottardi & L. Tonni DISTART, University of Bologna, Italy
ABSTRACT: The paper presents the results of the numerical analyses carried out with the well-known FE code PLAXISto simulate the response of a 1g small-scale model wall reinforced with polypropylene geogrids, under loads applied on top of the sand bed behind the wall. The capability and reliability of the numerical code to describe the measured behaviour is discussed, particularly with reference to the most suitable material model and parameters. A good general agreement between experimental data and numerical results is shown, especially when considering the complexity of the physical system, characterized by an initially low stress level combined with a highly confined soil-geogrids-wall interaction. published and discussed elsewhere (Gottardi & Simonini, 1997, 2000; Simonini et al. 2000). In this research - besides a brief presentation of the physical model wall - the results of the experimental study were used to check the capability and the effectiveness of the well-known numerical code PLAXIS(1998) to describe the response of such a rather complex system.
1 INTRODUCTION The ever increasing use of commercial numerical codes in geotechnical engineering practice urgently demands for new and accurate experimental work, in order to evaluate the applicability and reliability of these codes to describe the behaviour of structures interacting with the soil such as, for instance, shallow and deep foundations, embankments, tunnels or various types of walls. A special case of wall is represented by the geosyntetic-reinforced walls, whose use is rapidly increasing due to its construction simplicity and flexibility coupled with the low cost of the reinforcement technique. In fact, several types of geotextiles and geogrids, made up of various polymers, are today produced and used as reinforcements. Small-scale l g physical models represent a useful tool to reproduce the behaviour of such reinforced structures and already provided a rational basis for most calculation methods (e.g. Juran & Christopher, 1989; Palmeira & Lanz, 1994; Helwany & Wu, 1995; Karpurapu & Bathurst, 1995), but their direct application to design purposes is still rather limited by the fact that some important similarity requirements are not satisfied. However, the results of l g models can be of great benefit for the calibration of the relevant analytical and numerical models (Simonini, 1996), due to their controlled and repeatable test conditions. A rather new experimental research was recently carried out at the University of Padova (Italy), concerning the behaviour of walls reinforced with polypropylene geogrids; some experimental results - especially focused on time and temperature effects - were already
2 THE PHYSICAL MODEL The physical model wall (1200 mm long, 400 mm wide and 600 mm high) intends to reproduce a plain strain state within the reinforced soil mass. Figure 1 shows a general view of the small-scale model, where the main components - lateral walls, facing elements, reinforced soil and loading plate - can be observed. The retaining wall is made up of a set of rigid metallic strips, hinged each other and kept vertically only by the interposition of the geogrid layers. The geogrid (1200 mm long and 400 mm wide) are locked into the facing strips and spaced 70 mm. The wall is constructed from bottom to top by anchoring the metallic strips to a provisional vertical track, which is removed after the wall construction is completed. The sand layers are prepared by raining technique. All mechanisms for sand deposition are fully automatic and allow for the achievement of homogeneous and highly reproducible layers, the standard deviation of relative density (85%) being less than 1%. The reinforced retaining wall is loaded, through a rigid steel plate (200 mm x 400 mm) resting on top of the sand surface, by an electrical stepper motor. 453
Figure 1. Model wall lay-out
The load or displacement path generation and the data acquisition from all measurement devices are fully automatic via a personal computer and an A/D interface. The selected position of the displacement transducers allows for the continuous monitoring of the horizontal movements of the wall and the vertical displacement and rotation of the plate. Figure 2. Triaxial tests on large diameter sand samples
3 SAND AND GEOGRID PROPERTIES The soil used for layers preparation is a medium-fine quartz river sand with mean particle size D50 = 0.42 mm; non-uniformity coefficient C, = 2.0; specific gravity G, = 2.71, minimum and maximum dry unit weight 13.6 e 16.5 kN/m3. In order to evaluate the stress-strain behaviour of the sand, isotropically consolidated and drained (CID) triaxial compression tests were carried out on large diameter samples reconstructed by raining technique at the same relative density of 85%. The deviatoric stress ( o a - s) vs. axial strain &a and the volumetric strain E" vs. cacurves are plotted in Figure 2. As expected, the dense sand showed a strongly dilatant response under shear. Due to the presence of dilation, the peak-strength envelope is slightly curved and characterized by friction angles decreasing from 42.6" to 41.2" in the range of the investigated stresses. The critical shear strength angle of the sand, estimated from CIU triaxial compression tests, is 33". As reinforcing material, a suitably scaled polypropylene geogrid was used (grid size: 12 mm x 14 mm; mass per unit area: 63 g/m2; tensile strength: 4.5 kN/m). Results of elongation tests, carried out. with different strain rates, are reported in Figure 3.
454
Figure 3 Results ofelongatlon tests on the geogrlds
4 EXPERIMENTAL RESULTS In the test used to calibrate the numerical analyses the load was applied on top of the sand bed behind the wall, with a constant loading rate of 5.10-2kN/s, up to failure. Unloading-reloading cycles with an
5 PLAXIS ANALYSES
amplitude of 5 kN were interposed at increasing load levels. Failure was characterised by the collapse of the reinforced wall due to the progressive breakage of the polymeric grids. Figure 4 provides the results of the relationship between the applied load Q and the vertical displacement of the plate w. A similar kind of information, related to the wall movements, is provided in Figure 5 , where the measurements from all the horizontal transducers are plotted together, in order to provide the deformed configuration at various load levels. Dashed lines represent the position of the reinforcements.
The reinforced wall was schematised with the 2D mesh of Figure 6, with six-node triangular elements for the soil, geogrid elements and hinged beams as wall facing on the right-hand side. No interface was inserted at the beam-soil contact, since it turned out to be not particularly important in this case. The geogrid elements were fixed to the relevant beam hinge, in order to simulate the effect of an anchor to the facing elements. The loading plate was composed of rigid beam elements with a fully rough interface. Horizontal and vertical displacements were restrained at the base of the wall, whereas vertical displacements were free at the left boundary of the mesh. A KO initial stress state due to a self-weight y = 16.0 kN/m3 was assumed in all the simulations. The most suitable selection of the material model and related parameters for the various elements forming the reinforced wall was first analysed.
Figure 4. Load-displacement curve of the loading plate Figure 6. FEM mesh used in the numerical analyses
5.1 Materkd models The behaviour of the dense sand was modelled by using both available elasto-plastic constitutive laws implemented in the code, namely the bilinear elasticperfectly plastic model and an advanced constitutive law, referred as hardening-soil model. At failure, both matcrial models are based on the classic MohrCoulomb strength criterion, with the possibility of having a non-associated flow. with dilatancy y/ different from the shear strength angle 4. The response of geogrids is linearly elastic: the PLAXISgeosynthetic elements in the basic formulation do not allow to take into account any tensile strength limit. In order to estimate the strength parameters of the sand, the results of triaxial compression tests shown in Figure 2 were used, whereas the geogrid properties were determined from the elongation tests reported in Figure 3.
Figure 5 Measured wall deformation at different load levels
455
Table 1 . Material parameters SAND STRENGTH ' ( O ) 42 GEOGRID STIFFNESS EA (kN/m)
" 9 55
WALL FACING STIFFNESS E(GPa) 2 0 6
v 030
A summary of the adopted material parameters is reported in Table 1. A small cohesion was introduced in order to avoid possible numerical instabilities and excessive computational times. The sand constant stiffness modulus required by the elasticperfectly plastic material model was obtained via back-analyses of the experimental data. The numerical output of the overall response of the wall proved, as expected, to be very sensitive to such parameter, assumed equal for all the FE elements. An elastic modulus of 6.5 MPa was found to provide a good agreement in terms of applied load-vertical displacement curve (see Figure 8). The basic feature of the advanced hardening-soil model, based on the hyperbolic approximation of the soil response under shear, is the stress-dependency of the soil stiffness. Therefore, no back-analysis was required in this case and the relevant stiffness parameters could be independently determined on the basis of the results of the triaxial compression tests already described in Section 3. In addition, plastic strains due to both the deviatoric and the hydrostatic stress increments can be taken into account. The complete description of the hardening-soil model can be found in the PLAXISuser manual. Table 2 summarizes the stiffness parameters used with the two material models (p',,f=lOO kPa).
Figure 7. Triaxial tests on sand: comparison between experimental data and numerical simulations.
q&,. = 33"), the volumetric strains turn out to be underestimated. 5.3 Model wall sirridation In order to reproduce the wall response, vertical displacements were applied at the nodes of the top plate. The final displacement of 20 mm (equivalent to the physical modcl failure) was given in 416 increments, with a relative error tolerance in each computational step of 0.003. The main outcome of the analyses, to be compared with the measured curve of the top plate applied load-vertical displacement, is plotted in Figure 8.
Table 2. Stiffness parameters of the soil material models ELASTIC-PERFECTLY PLASTIC MODEL E (MPa) 6 50 V HARDENING-SOIL MODEL \
E$ E,:;
I
(MPa) (MPa)
m
0 20
l 6 35 20.67
Elf'' (MPa)
65 40
Vur
0.20
0.62
Xi
09
5.2 Validation of the hardening-soil model
A preliminary check of the effectiveness of the hardening-soil model to reproduce the behaviour of the dense sand was first carried out, simulating with PLAXISthe CID triaxial compression tests. Figure 7 compares the experimental and the simulated (dashed lines) response in both (0, - s) vs. E, and vs. E, planes. Note the good agreement between experimental curves and numerical simulations, especially in the deviatoric stress - axial strain plane. Assuming a dilatancy angle of only 9" (and
456
Figure 8. Load vs. vertical displacement of the top plate: comparison between experimental data and numerical simulations.
Apart from the final, fragile failure of the physical system, shown by the peak load in the experimental curve and by the formation of a clear mechanism inside the reinforced soil, which cannot be reproduced in such numerical analyses, there is a substantial good agreement between the experimental data and both numerical models used, especially when considering the hardening-soil model, which seems to be able to reproduce the softer wall behaviour at lower load levels. It is interesting to observe that, whereas the results of the MC bilinear model are essentially a good fit of the experimental data, the prediction with the hardening-soil model was obtained using the material parameters independently determined with actual element tests (i.e. triaxial tests for sand and elongation tests for geogrids). Such results comparison is well summarised in Figure 9, which shows the wall deformation at the maximum load level and at half-way to it. The numerical model is not able to reproduce the top reinforcement layer slippage, which is in fact a clear feature of the physical model. However, the hardeningsoil model seems to better capture the larger horizontal displacements at higher stress levels. Displacement vectors, computed at 20 mm of vertical displacement of the top plate, are plotted in Figures 10a and lob. Note the general trend inside the reinforced soil mass and behind the wall, that confirms the trend of the horizontal wall displacements of Figure 9. No strain localisation along a well defined sliding surface, as clearly noticed in the physical model, is possible in these numerical analyses.
Figures 1 l a and 1l b show the distribution of the tensile forces along the geosynthetics at the same final situation. The greatest values are observed in the third and fourth layer (from top), in correspondence of the maximum lateral wall movements, the maximum tensile force being equal to 2.30 kN/m. The general pattern corresponds to what observed in the physical model, except for the already noticed slippage of the shallowest reinforcement layer.
~i~~~~9, wall djsplacements: compar,son between experimental data and numerical simulations.
Figure 11. Tensile forces in the reinforcements: a) MC bilinear model; b) Hardening-soil model.
Figure 10. Displacement vectors: a) MC bilinear model; b) Harden ing-soi I model.
457
parameters as deduced from specifically performed standard compression triaxial tests. Both numerical simulations showed a remarkable good agreement with the experimental trend of the vertically loaded top plate. As regards the wall deformation, neither type of analysis can reproduce the slippage of the top reinforcement layer, which is in fact a constant feature of the physical model. Furthermore, in both cases it was observed that the tensile forces in the geogrids at the maximum applied load remained relatively distant from the ultimate strength and thus cannot explain the actual occurrence of the sudden and catastrophic collapse of the model wall, caused by the progressive breakage of the reinforcements. This difference could be partially explained when we consider that the FE code and the constitutive models used do not take into account the material softening and the strain localisation - and the consequent stress concentration along a well defined sliding surface.
Figure 12. Loading-unloading-reloading cycle at Q = 20 kN.
Finally, with respect to the classic MC bilinear model, the hardening-soil model seems to be more able to reproduce thc actual unloading-reloading response. The typical behaviour is presented with reference to the unloading-reloading cycle displayed in the upper-left part of Figure 12 (between 15 and 20 kN). Note the very good agreement of the numerical results. The deformed shape of the wall is also reported at the end of each loading cycle.
6 CONCLUSIONS The present rcsearch was aimcd at testing the effecwhcn tiveness of the well-known FE code PLAX~S reproducing the behaviour of a small-scale reinforced wall. loaded up to a stress levcl corrcsponding to thc failure onset in thc physical model. The reinforced system is made up of a very dense dilating sand, initially at a very low stress lcvcl. and heavily confined gcogrids conncctcd to stiff facing elements. Two suitable constitutive models implemented in the code were used to describe the soil behaviour. namely the classic Mohr-Coulomb. bilinear model and an advanced hardening-soil model. With the latter it was possible to introduce the stress-dcpendency of soil stiffness and the actual unloading-reloading sand response. Most important, only with the hardcning model it was possible to introduce the soil
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REFERENCES Gottardi, G , Simonini, P 1997 Long term behaviour of reinforced walls model results Proc Itit Conf Geosynthetics Asia ’97, Vol 2 43-50 Bangalore, India, November 1997 Gottardi, G , Simonini, P 2000 Time and temperature effects on the behavior of geogrid reinforced walls Proc 2”’ Eiir Geosyiitlietics Conference, EUROGEO 2000, Bologna, Vol 1 175-180 Patron Editore, Bologna Helwany, M B , Wu, J T H 3995 A numerical model for analyzing long-term performance of geosynthetic-reinforced soil structures GeoJynthetics Iiiternationul, Vol 2, N 2 429-453 Juran, I , Christopher, B 1989 Laboratory model study on geosynthetic reinforced soil retaining walls .Join.iial of Geot Eiig, ASCE, 115, N 7 905-926 Karpurapu, R , Bathurst, R J 1995 Behaviour of geosynthetic reinforced soil retaining walls using the finite element method Coinputer und Geotechnics, 1 7 279-299 I’almeira, E, M , Lanz, D 1994 Stresses and deformations i n geotevtile reinforced model walls Geotextiles mid Geonienibraiia, 13 33 1-348 PLAXIS - Finite Element Code for Soil and Rock Analyses 1998 Version 7 Netherlands PLAXIS B V Siinonini, P 1996 A finite element approach to the strength of granular soils reinforced with geosynthetics Proc of die Internntioiia1Sviiiposiirni on Earth Reiiifoix enieiit 675-680 Fukuoha, Kiushu, Japan, 12-14 November 1996 Simonini, P . Schiavo, M Gottardi, G & Tonni, L (2000) Numerical analysis of a model wall reinforced with polypropylene geogids Proc 2”’ Ein GeosyiifheticsCoi7ference, ElJROGEO 2000, Bologna, Vol I 231-236 Patron Editore, Bologna
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Seismic stability of preloaded and prestressed reinforced soil structure against strong shaking M. Shinoda, T. Uchimura & F. Tatsuoka Department of Civil Engineering, University of Tokyo, Bunkyo-ku, Tokyo, Japan
M. Tateyama Japan Railway Technical Research Institute, Kokubunji-shi, Tokyo, Japan
T. Natsuki A raigumi CO., L TD, Nish inomiya-shi, Hyogo, Japan ABSTRACT: To substantially increase the vertical stiffness and reduce the vertical residual compression of geotextile-reinforced soil (GRS) structures subjected to long-term traffic load, the preloading and prestressing (PLPS) method has been proposed. The seismic stability of PLPS GRS structures was investigated by pcrforming shaking table model tcsts using sinusoidal waves with horizontal acceleration of 700 gals. The use of a newly developed device, called the ratchet system, is very effective in increasing the seismic stability of the structures by maintaining high prestress when the backfill tends to contract and preventing the expansion of the backfill. These functions effectively prevent the occurrence of resonant state while restraining the bending deformation of the structurc, which are essential for the high seismic stability of the structure. 1 INTRODUCTION A new construction technology, called the preloading and prestressing method, has been proposed so as to substantially increase the vertical stiffness and decrease the vertical rcsidual compression of geosynthetic-reinforced soil (GRS) structurcs against long-term traffic load (Figure 1; Tatsuoka et al., 1996; Uchimura et al., 1996, 1998). That is, the deformation of the backfill is made essentially elastic by applying sufficiently large vertical preload to the backfill and, while the structure is in service, the stiffness of the backfill is kept sufficiently high by not fully unloading the preload, but maintaining sufficiently high prestress. Shinoda et al. (1999) performed laboratory model tests and showed that the transient and residual deformation of the backfill against cyclic load, such as traffic load, could be made very small by means of the PLPS method.
Figure 1 Stress-strain behaviour of PL/PS soil structure (not to scale)
In the summer of 1996, the first prototype PLPS GRS bridge pier was constructed in Fukuoka City, Japan, to support a pair of railway bridge girders (Figure 2). The backfill of the pier is densely compacted well-graded crushed gravel, reinforced with geogrid layers with a vertical spacing of 15 cm. Since having been opened to service the summer of 1997 until now (March 2001), the pier has shown nearly zero residual settlement against about 1 20 train passing per day (Uchimura et al., 2001). The maximum transient compression of the backfill by each train passing is only about 0.02 mm, which is equivalent to a vertical strain of as small as
Figure 2 First prototype PLPS GRS bridge pier (Uchimura et al ,2001)
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about 0.001 ?40 in the backfill. This strain value is within the elastic limit strain of the material (Tatsuoka et al., 1997). It is one of the keys for the success of the PLPS method that the transient strain has been kept as small as above by attaining a high stiffness of the backfill, which was realized by the PLPS procedure. One end of the bridge is supported by an abutment, which is a GRS structure that is similar to the pier, constructed using the same backfill material and reinforcement as the pier, but without PLPS. In comparison to the PLPS GRS pier, the GRS abutment exhibited a total residual settlement of 12 mm until now, despite that the load by the same train passing applied to the abutment is nearly a half of that applied to the pier. This relatively large residual settlement was due to a much larger maximum transient settlement of about 0.2 mm, which is equivalent to a strain of about 0.01 %, exceeding the elastic limit strain. These different behaviours of the pier and abutment indicate that the PLPS method is very effective in reducing the vertical transient and residual settlement of GRS structures. Another important required property of such PLPS GRS structures as this pier is the capability of surviving high seismic load. To have a better insight into the seismic stability for PLPS GRS structures and to develop the relevant seismic design methodology, a series of shaking table model tests were performed (Shinoda et al., 2000a, b). The functions of a new device, called the ratchet system, which was developed to substantially increase the seismic stability of GRS structures were evaluated. 2 TEST METHOD The models of GRS structures were 55 cm-high and 35 cm timcs 35 cm in cross-section (Figure 3). This relatively slender dimension was selected to investigate into the behavior of PLPS GRS structures in a rather critical use. The backfill of the models was a compacted well-graded gravel of crushed sandstone ( D ~ o =2.52 mm; Uc= 5.41; FC= 0 To; emax=0.986; and emln=0.481). The backfill was compacted to a relative density of 90 %, reinforced with 12 grid layers with a vertical spacing of 5 cm. Each grid layer consisted of 34 phosphor bronze strips (3.5 mmwide, 0.2 mm-thick and 350 mm-long), 17 in each perpendicular direction, with an aperture of 8 mm. The periphery of each sub-layer of the backfill was protected with gravel bag models with a diameter of about 5 cm. A square steel platen of 5 cm in thickness and 45 cm times 45 cm in cross-section with a weight of 282 N was placed on the top of the completed reinforced backfill. Horizontal accelerometers and displacement transducers were set at the top platen and the mid-heights of the second, sixth and eleventh sub-layers from the bottom of the model. 460
Figure 3. GRS structure model placed on a shaking table
Vertical preload 30kPa was applied to the backfill by using four steel tie rods fixed to the top and bottom reaction steel platens, which was then decreased to the initial prestress of 15 kPa. These preload and prestress values were determined considering the model similitude. Each tie rod was equipped with a load cell to measure the tension. To examine the deformation of PLPS reinforced structure during strong shaking, a series of shaking table tests were performed using sinusoidal waves with horizontal acceleration of 700 gals and a frequency of 5 Hz at the table. The results from the following two tests, among many others that were performed in the present study, will herein be reported. In the first test, a pier model was not equipped with a ratchet system on the top of each tie rod (Figure 4; the ratchet system is explained later), but the tie rods were fixed to the top reaction platen with nuts. Due to a relatively high stiffness of the tie rods, the tie rod tension decreases at a high rate when the backfill exhibits vertical compression by creep and cyclic loading. The consequence of the reduction of tie rod tension could be serious for the seismic stability of structure. That is, the initial natural frequency M structures under undamaged conditions of a PLPS GRS should be designed to be sufficiently higher than the predominant frequency w of considered seismic load to avoid the resonant state during a seismic event. The resonant state can be avoided also by making M of the structure sufficiently lower than the w value of the input motion. This method will, however, be penalized by too large cyclic deformation, which may result into a large compression of the backfill. With an initial ratio w/n smaller than unity, a rapid and significant decrease in n during cyclic loading may result into the transient resonant state, which may result into excessive deformation or even the total collapse of the structure.
of the backfill tends to increase by whatever cause, such as bending deformation of the structure or dilatancy during monotonic or cyclic loading. The high stiffness of the system is attained by locking the displacement of the top end of each tie rod relative to the top reaction platen (Figure 4c). it is to be noted that the bending deformation of the backfill, as illustrated in Figure 5a, is one of the most dangerous causes for the failure of such slender GRS structures. By these two functions of the ratchet system described above, the bending deformation of the structure can be effectively restrained, as illustrated in Figure 5b and shown below.
In the second test, a ratchet system was used to fix the top end of each tie rod to the top reaction platen (Figure 4). This system was developed to alleviate the above-mentioned problem. In the present study, after the prestressed condition was reached, a ratchet system was fixed to the top end of each tie rod before the start of shaking test. 3 RATCHET SYSTEM The ratchet system was designed to show a low stiffness, under high prestress conditions, when the backfill tends to vertically contract by whatever cause, such as creep deformation and shakinginduced compression. This low stiffness is attributed to a low stiffness of a relatively long spring attached between the top end of each tie rod and the top reaction platen (Figure 4b). in this case, the tie rod tension decreases only slightly even when the backfill contracts relatively largely. On the other hand, the ratchet system can exhibit a very high stiffness, increasing largely the tie rod tension, when the height
4 TEST RESULTS Figure 6 shows the time histories of horizontal acceleration at the top platen and the shaking table, the average prestress acting at the top of the backfill (equal to the sum of the tension acting in the four tie rods and the weight of the top reaction platen divided by the cross-section of the backfill), the rotation angle of the top platen and the averaged vertical compressive strain of the backfill. The averaged vertical prestress in the test not using ratchet systems decreased to nearly zero already by an elapsed time of five seconds from the start of shaking (Figure 6b). A large rotation at the top plate was then induced (Figure 6b), indicating a large bending deformation of the backfill. This behavior was due to a large reduction in the stiffness of the backfill caused by a substantial decrease in the prestress. It may be seen from Figure 6b, on the other hand, that the ratchet system functioned very well in keeping the prestress and restraining the bending deformation of the model despite very strong shaking. Importantly, the averaged prestress did not become lower than the initial value at any moment during the shaking test, but it became very large transiently in each cycle by restraining the backfill expansion, resulting into a very small rotation of the top platen and a small settlemcnt of the backfill.
Figure 4. Behaviour of the ratchet system when the backfill tends to vertically contract and expand.
Figure 6. Time histories of; a) horizontal acceleration at the top platen and the table,
Figure 5. Schematic diagrams of the deformation of GRS structure; a) without a ratchet system; and b) with a ratchet system.
46 1
1 + 4h2(w/n)’
+ 4h2(o/n)z
= /{I - (w/n)2)2
(la>
where h is the damping ratio (unknown), w is the given input frequency (equal to 5.0 Hz) and n is the non-constant natural frequency of the structure (unknown). Figures 7a and 7b show the theoretical relationships between L and o / n and between p and o / n for diffcrcnt h values obtained based on Eqs. l a & 1 b. The transient values of the damping ratio h and the frequency ratio w/n for respective measured values of L and p were obtained by solving Eqs. l a & 1b with a process of iteration. The relationships betwecn the measured value of L and the estimated value of w/n and between the measured value of p and the estimated value of w/n arc presented in Figures 7a and 7b. It can be seen that the model without using the ratchet system exhibited the transient resonance while the value of w/n was increasing, which was associated with a substantial decrease in the tic rod tension due to shaking-induced vertical deformation of the backfill (Figure 6). The increase in the W / M value was enhanced by the decrease in the stiffness ofthe back’iill due to the non-linearity of stiffness. This observation is well consistent with the fact that the single amplitude of shear strain y,,, at the top of the model showed noticeably a high value at the resonant state and the y,, value was very high, close to that during resonance. even when the ratio w/n exceeded the value at resonance (about unity) (Figure 7c). On the other hand, the model using the ratchet systems did not exhibit thc resonance, because the ratio w/n was kept always lower than unity. The value of y , , was therefore was kept much smaller (Figurc 7c). This behaviour was due to the positive functions of the ratchet system (i.e., maintenance of high prestress and restraining of the expansion of backfill).
Figure 6 (continued) Time hlstorles Of, b) average Prestress, C) rotation of the top plate, and d) averaged vertical strain
5 DYNAMIC RESPONSE CHARACTERISTICS The dynamic response characteristics of the models were very complicated, mainly because they were not stationary due to continuing changes in the stiffness of the backfill caused by shaking-induced changes in the tie rod tension and associated deformation of the backfill. For the first approximation, the response characteristics of the models wcre analyzed based on the theory of a single degree of freedom. The time histories of the response ratio L and the phase difference p between the top platen Of the structure and the shaking table were measured in the respective test. Thcse quantities arc givcn theoretically as:
Figure 7 Theoretical and measured response curves, a) response ratio plotted to the ratio of the input frequency to the natural frequency (to be continued)
462
2) the major deformation mode of the backfill is bending and shearing as a simple beam; and 3) the rotation of the top platen is zero due to the full restriction of tie rods (this assumption is relevant before the tie rod tension becomes very small). From the condition that the maximum kinematic energy is equal to the maximum strain energy, an equation is derived to obtain the natural frequency M of the model. Figure 9 shows the ranges of the theoretical relationship between the natural frequency n and the shear modulus G of the backfill for the two tests obtained by this procedure. The shear modulus G of the backfill at each cycle during shaking was back-calculated by substituting the respective natural frequency M that was estimated by assuming that the PLPS reinforced pier model deformed as a single degree of freedom system (Figure 7a). Figure 10 shows the relationships between the shear modulus G back-calculated as above and the corresponding measured single amplitude shear strain y,, for the two tests (with and without using the ratchet systems). The shear stress z that is indicated in Figure 10 is equal to the shear modulus G multiplied by y,, . Figure 11 shows the relationships between the back-calculated shear modulus G and the corresPondh3 measured ~ O l - m dstress (i.e. aver-
Figure 7 (continued) Theoretical and measured response curves, b) phase difference and c) dynamic shear strain of backfill, plotted to the ratio of the input frequency to the natural frequency
Figure 9. Theoretical relationship between the natural frequency and the shear modulus of the backfill with the ranges estimated for the two tests (with and without using ratchet systems).
Figure 8. Model configurations to compute the natural frequency of the model.
6 STIFFNESS OF THE BACKFILL DURING SHAKING In the seismic design of PLPS GRS structure, the initial natural frequency and those during shaking should be evaluated with a reasonable accuracy so as to avoid the occurrence of resonance during a given design seismic load. To that end, a simplified method was developed based on the following assumptions: 1 ) the backfill is a uniform isotropic linear elastic modulus E, material having a constant shear G with a maSSOf s);
Figure 10 Relationship between the back-calculated shear modulus of the backfill and the measured shear strain in the test with and without ratchet systems
463
Figure 1 1. Relationship between the back-calculated shear modulus of the backfill and the measured vertical stress in the test with and without ratchet systems.
aged vertical stress on the backfill)o,. The maximum and minimum values of o,,in each cycle are plotted against the respective shear modulus G . It is clearly seen that in the test without using the ratchet systems, the increase in the shear strain is associated with a large decrease in the shear modulus, which is associated with a substantial decrease in the vertical stress 0” and a large increase in the shear strain due to large shear stresses z as a result of resonance. On the other hand, in the test using the ratchet systems, the shear strain y,>, was kept relatively small, which was associated with high shear modulus values and relatively low shear strain values due to the non-occurrence of resonance (Figure IO), which was associated with high vertical stress values kept high during strong shaking (Figure 1 I). These data clearly show that the ratchet system can maintain high prestress and thereby can restrain low shear strains by keeping high shear modulus values while avoiding the resonance state.
7 CONCLUSIONS The following conclusions can be derived from the test results: 1) The initial natural frequency n, under undamaged conditions of a given PLPS GRS structure should be designed to be sufficiently higher than the predominant frequency U,, of design seismic load so as to avoid the resonant state. 2) With an initial ratio U,/., smaller than unity, a rapid and significant decrease in the natural frequency n during cyclic loading may result into the occurrence of transient resonant state, which may result into excessive deformation or even the total collapse of the structure.
464
3) Therefore, the natural frequency n should not decrease largely from the initial value n, (which is larger than U , ) so that the n value does not approach U,. To this end, the tie rod tension should be maintained to a sufficiently large value during shaking. The use of ratchet systems is effective for this purpose. 4) The ratchet system also does not allow the height of the backfill to increase, which is effective in restraining the bending deformation of the backfill, which is very dangerous deformation mode to be avoided. It will be also important to ensure that the structure can survive the resonant state, in case it takes place. For the above, the tie rod tension should not decrease largely to a near-zero value even at the resonant state. PLPS GRS structures equipped with ratchet systems described in this paper can be constructed as permanent important structures having a high stiffness for long-term repeated load, such as traffic load as well as a seismic stability. REFERENCES Shinoda,M , Uchimura,T , Maruyama,N and Tatsuoka,F (1 999) “Effects of preloading and prestressing on the vertical stiffness of GRS”, Proc of the 11th ARC on Soil Mechanics and Geothecnical Engineering, Vol 1 4 19-422 Shinoda,M , Uchimura,T , Tatsuoka,F and Tateyama,M (2000a) “Seismic stability of preloaded and prestressed reinforced soil structures”, Proc of the 2nd Asian Geosynthetics Conference, V o l 2 43-48 Shinoda,M , Uchimura,T , Sugimura,Y, Tatsuoka,F and Tateyama,M (2000b) “High seismic performance of preloaded and prestressed geotextile-reinforced soil structures”, CD-ROM Proc of an International Conference on Geotechnical & Geological Engineering, GIGS0447 PDF Tatsuoka,F , Uchimura,T and Tateyama,M (1 996) “Preloaded and prestressed reinforced soil”, Soils and Foimdutions, JGS, Vol 37, No 3 79-94 Tatsuoka,F , Jardine,R J , Lo Presti,D , Di Benedetto,H and Kodaka,T ( 1997) “Characterising the Pre-Failure Deformation Properties of Geomaterials”, Theme Lecture for the Plenaiy Session No I , Proc of XIVlC on SMFE, Hamburg, September 1997 Uchimura,T , Tatsuoka,F , Sato,T , Tateyama,M and Tamura,Y ( 1996), “Performance of preloaded and prestressed geosynthetic-reinforced soil”, Proc Int Symposium Earth Reinforcement, Fukiioka, Balkerna (Ochiai et a1 , eds), VOI 1 537-542 Uchimura,T , Tatsuoka,F , Tateyama,M and Koga,T (1998) “Preloaded-Prestressed Geogrid-reinforced Soil Bridge Pier”, Proc the 6th Int Conf on Geosynthetics, Atlanta, VOI 2 565-572 Uchimura,T , Shinoda,M , Tatsuoka,F and Tateyama,M (200 1) “Performance of PLPS geosynthetic-reinforced soil structure against working and seismic loads”, Proc of the 15th International Conference on Soil Mechanics and Geotechnical Engineering, Istanbul (to appear)
Landmarks in Earth Reinforcement, Ochiai ef al. (eds), 0 2001 Swefs & Zeiflinger, ISBN 90 2651 863 3
Seismic behaviour of earth reinforcement walls R.A. Sofronie UNESCO Chair Ecoland, University of Bucharest, Romania
C.A. Taylor Earthquake Engineering Research Centre, University of Bristol, UK
F. Iosif Technical University of Civil Engineering, Bucharest, Romania
ABSTRACT: The paper presents the experimental results of a test programme performed on the shaking table of Bristol University, UK. Two similar models of reinforced soil retaining walls, one of conventional type and another self-confined, have been investigated. Both models supported the same elastic structure and were submitted to inputs of increasing intensities. The seismic behaviour of the models was comparatively analysed with the aid of cumulative diagrams of settlements, horizontal displacements and response accelerations. The ultimate limit state occurred by tilting of the facing, but the two models reached this state differently. The numerical models revealed the local phenomena that developed in the reinforced soil as well as the large influence of the elastic structure. Both conceptual design and the development of numerical analysis are supported by the results. more advanced. First the facings were fixed with hinges at their bottom and in this way only a single DOF was allowed, that of tilting. Secondly the steel cables were anchored in the backfill allowing in this way a self-confining mechanism of the model. Finally a numerical distinct element model based on the UDEC 3.10 program was developed. The significance of the research is to assess the behaviour of the two types of reinforced earth structures in seismic conditions, to determine some practical recommendations for safe seismic design and to draw out some structural measures for mitigation of their seismic vulnerability.
1 INTRODUCTION Eurocode 8 does not contain provisions for seismic design of earth reinforcement. Reinforced soil structures for retaining walls, abutments and slopes are currently designed according to BS 8006-95 and DIN 4017. There are two criteria of design: 1) external stability; 2) internal stability. Confining earth reinforcement improves the stability of both types (Sofronie & Feodorov 1995, 1998). The buildings supported by soil structures are currently considered only as static surcharges. There are no provisions for soil-structure interaction under seismic actions. The former tests carried out on two pairs of models, for conventional and confined retaining walls, one in Bristol, UK in 1998 and another to Iasi, Romania in 1999, offered some information of practical interest (Sofronie, Taylor & Greening 2000). Indeed, earth reinforcement structures tend to settle continuously without collapse. Obviously, this stable behaviour is mainly due to the granular structure of the soil but also to its degree of compactness. Similarly, horizontal displacements occur as a consequence of facing tilt. The response accelerations characterise the dynamic behaviour of earth reinforcement and emphasise the amplification or attenuation phenomena developing in models during earthquakes. Now the problem was to investigate the influence of elastic structures on earth reinforcement retaining walls. By conducting the testing program according to the same principles used for the former program and comparing the results one could assess the influencing factors. However, the current research was
2 MODEL s E r - u P For the two models, an existing shear stack was used. This is a flexible parallelepiped box, composed of aluminium rings alternately connected with rubber elements, specially designed to reproduce the field boundary conditions experienced by soil structures during earthquakes. The philosophy of modelling was to compare three phenomena developed under seismic actions, namely settlements, horizontal displacements and response accelerations, for the two types of retaining walls. Leighton Buzzard Sand was used for both infill and backfill. This is a siliceous, uniform and cohesionless sand with a unit weight (as placed) of 15.4 kN/m3 and the internal friction angle cp7= 45". As reinforcement Tensar geogrid SR55, with quality control strength of 55 kN/m and specific weight of 0.5 kg/m2, was adopted. The grids were fixed to the facing by nails at five 465
levels equally spaced at 150 mm. For the conventional wall they were carefully cut at L=540 mm, corresponding to the required aspect ratio L/H=0.6 (Fig. 1). The self-confined wall was derived from the first model by shortening the grids to L=360 mm for an aspect ratio L/H = 0.4 and by adding the two steel cables of @ = 6 mm anchored in backfill. The idea of self-confinement is not new, it has been already tested on the shaking table of INCERC Iasi in 1999, but the results were not reported yet (Fig. 2).
The innovation of this test was the elastic structure with one DOF. It was installed on the backfill in the most disadvantageous position for reinforced soil. Four loading cases were considered for the structure, with natural frequencies of 1 1 Hz, 6.1 Hz, 4.7 Hz and 3.4 Hz, respectively. Since this range of frequencies is much lower than the natural frequency of the earth reinforcement model of 30 Hz, there was no danger of resonance occurring in the oscillating system (Sofronie, Taylor & Crewe 2000).
3 RECORDING DEVICES Twenty-eight channels of instrumentation monitored the actual motion of the shaking table and the behaviour of the model incorporated in the shear stack. Of these, 14 devices were used for the model itself, 10 for the elastic structure, 3 for the shaking table and 1 for the shear stack. The infill settlements and horizontal displacements were recorded with non-contacting transducers. Two wire-potentiometer transducers recorded the facing displacements. Response accelerations were recorded at six positions in the soil and at four positions on the facing. The settlements of the elastic structure were measured at the four corners and at the centre of the roof mass. Orthogonal horizontal response accelerations of the structure were measured at the foundation and roof. The shaking table accelerations were measured in the three principal directions. The horizontal displacement of the shear stack was monitored at one location. For the accelerations induced in the sand, 6 unidirectional piezoelectric accelerometers were placed at different levels: 3 in the infill between the geogrid reinforcements, and 3 in the backfill (Figs. I and 2). These miniature accelerometers, with 0.7 mm in diameter and 4.5 g in weight, were used so as to reduce the disturbance to the material behaviour. They were mounted inside small, light perspex boxes size so that the whole unit had the same mass density of the sand, and were positioned during sand deposition. A thin layer of sand was glued to each side of the boxes to guarantee good contact with the adjacent granular material. All accelerometers were connected by thin flexible coaxial cables. The cables were protected during the experimental tests to prevent potential damage induced by the sand movements. Before starting the testing program, all recording devices were carefully calibrated. In addition, the initial level of sand was manually mapped in 13 vertical cross sections. After the collapse of each model, the final surfaces were mapped again. By difference of measured heights the controlling values of settlements in both models were been obtained. Indeed, these values have been compared with the recorded values for settlements and a good agreement was found. The aspect ratio of area affected by collapse was L/H=2.45 - 2.80.
Figure 2. Model of the self-confined walls.
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4 TESTING PROGRAM The six degrees of freedom shaking table at Bristol University was opened in 1987. It has a cast aluminium platform measuring 3 m by 3 m, weighing 30 kN, and has a maximum payload of 150 kN. The platform is driven by eight 50 kN servo-hydraulic actuators, four acting horizontally around the perimeter and four acting vertically at the corners. The platform has peak displacements of h150 mm in all translation axes, peak velocities of 0.5 m/s and peak bare platform accelerations of 4.8g horizontally and over 7.5g vertically. A PC-based real-time digital control system allows a wide range of recorded and synthetic earthquake and other motions to be reproduced by the shaking table. Up to 64 data acquisition channels are available (Taylor 1997). For fulfilling the three objectives of the comparative test programme, a single degree of freedom shake was applied in the longitudinal direction of the models. Three types of input excitations were successively induced: the El Centro’40 acceleration (Fig. 3), a Eurocode 8 compatible artificial acceleration (Fig. 4) and a sine dwell function with a frequency of 5 Hz (Fig. 5). Since the dominant input frequencies were much lower than model natural frequency of 30 Hz, there was no danger of resonance occurring in the soil deposit at low levels of table acceleration. All inputs were alternately induced with gradually increasing intensities beginning with 0. l g up to over lg. For each of four weights of the elastic structure, namely 50 kg, 100 kg, 150 kg and 250 kg, the three types of inputs were successively induced. The first model of a conventional wall was shaken with
52 quakes, and the second one with 57. The El Centro’40 input, as the weakest of the three inputs, was induced 13 times in each model. The Eurocode 8 input was induced 19 times in the model of conventional walls and 22 times in the model of selfconfined walls. The sine dwell quasi-harmonic function has a constant length with increasing amplitude from zero to the maximum value over the first ten cycles, constant amplitude over the subsequent twenty cycles, with the amplitude decreasing down to zero over the last ten cycles. This kind of input motion is more severe than the former two but it allows clear investigation of the kinematics of soil structure models and their behaviour at accelcration values close to the ultimate limit state. It was induced with gradually increasing intensities 20 times in the model of conventional walls and 22 times in the model of self-confined walls. At the end of the test program, strong sine dwell inputs were induced in each model. The conventional wall model yielded at 1.25g by facing tilt, while the self-confining wall failed similarly, but at 1.69g and only after one of the two steel anchorage pulled out. Both cases of collapse were strongly influenced by the elastic structure under its maximum loading (Figs. 6&7). From the 109 tests performed 3,052 recordings in real time were obtained.
Figure 7. Testing program for self-confining wall model.
5 SETTLEMENTS The settlements of the two models were recorded at the top of the infill (Sl at the level +900 mm) and at the top of the elastic structure (S2 at the level of +1,220 mm), both located on the longitudinal axis of symmetry. They were measured from one test to the subsequent one in a cumulative manner. For each model the specific behaviour was observed of the in-
Figure 5 Sine dwell acceleration time history
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fill and backfill under the three types of inputs with increasing intensities, and then the settlements of the two models were compared. In the case of the conventional wall, no significant settlements were recorded during the first succession of 23 quakes. After the sine dwell quake, no.24, with an intensity of 0.3g, the infill settled by 8 mm and the backfill by 4.4 mm. During the next quakes the settlements gradually progressed. Some larger values occurred after quake no. 47 but they were not critical. Collapse occurred during quake no.52, under a sine d well input of 1.25g, when the settlements reached 58 mm in the infill and 48 mm in the backfill. Then the facing tilt and the elastic structure inclined without overturning. It was a controlled yielding, similar to a fail-safe condition (Fig.8). The self-confining wall behaved slightly different in settling. From the first quakes the anchoring cables relaxed little by little and the infill started to settle so that after quake no. 7 a cumulative value of 7 mm was recorded. The whole irregular shape of infill settlement diagram reflects the influence of both, on one side of the facing and on the other side of the elastic structure. At the same time the backfill continuously and uniformly settled. The collapse occurred during quake no. 57, also under a sine dwell, but at 1.69g and only after one of the two anchoring cables was pulled out. By coincidence the settlements reached 59 mm in the infill and 48 mm in backfill. The mechanism of yielding was similar, and during the collapse the synthetic reinforcement was not damaged (Fig.9). By reporting the settlements to the increasing values of induced accelerations it appears that in both infill and backfill the main phenomena occurred when the intensities assumed values between 0.3g and 0.7g (Figs. 10&11). The only difference between conventional and self-confining walls was that the later resisted longer and to intensities 35% higher.
Figure 9. Settlements in the model of self-confined walls.
The shapes of the sand surface for both before and after collapse were drawn from measurements. It is evident that due to ar steel cables the self-confined wall behave better and therefore was safer (Figs. 12& 13).
6 HORIZONTAL DISPLACEMENTS All measured displacements were in the don direction of the models and were influenced facing tilt and the elastic structure. In the conventional wall model the infill displact than the backfill and the influence of thc structure was not felt before quake no. 40. displacements before collapse were of 77 mi infill and 68 mm in backfill (Fig. 14). In 1 confining wall, the infill and backfill remain
Figure 14. Displacements in conventional wall model
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Figure 18. Dynamic response of the elastic structure.
Figure 15. Displacements in self-confined wall model.
However, the excitation time being too short, significant did not occur. A selection of maximum amplitudes developed during each for both models are comparatively presented in two diagrams (Fig. Surprisingly, the dynamic response of elastic structure located on the conventional wall was much stronger than in the case of the self-confined wall. A rigorous explanation of this behaviour could be given only aRer a analysis of the dynamic phenomena.
and behaved during 46 quakes’ Only after the anchoring cables started to yield (by pulling Out of the anchorage) did the infill displacements increase and finally reach 99 mm, while in backfill the last recorded was Of 33 mm (Fig. 15). By plotting the settlements with respect to the increasing values of induced accelerations, it appears that the infill of the conventional wall started to displace from the first quakes, while that of the selfconfined wall only started after the input intensity reached about 0.4g. However, for intensities higher than 0.6g the infills in both models no longer displaced (Fig. 16). The backfill displaced similarly in both models for inputs under 0.4g. Further, for intensities between 0.5g and 0.6g, the backfill of the conventional wall model displaced by 77 mm and remained at this value until the collapse occurred at 1.25g. In the other model, the same phenomenon occurred, but in a larger range of intensities, namely between 0.4g and 0.Q what indicates a better behaviour (Fig. 17). Horizontal displacements were little influenced by the elastic structure; they mainly depended on the facing tilt. As concerns the self-confining procedure, it barely prevented the horizontal displacements, because the anchors were also moving together with the backfill. Only in the case of confining with a fixed cable were the horizontal displacements restrained. The elastic structure was designed and constructed of steel members with only one horizontal DOF. Two of the structure’s natural frequencies, 4.7 Hz and 6.1 Hz obtained for different loading cases, were rather near to the sine dwell excitation and therefore amplification phenomena could arise.
7 RESPONSE ACCELERATIONS The dynamic behaviour of the elastic structure located on the two models was characterised with the aid of response accelerations recorded at its base and top. For practical reasons the ratio between peak the response acceleration (PRA) and peak induced acceleration (PIA) was considered. Values of this ratio larger than one show amplification, and those lower than one show attenuation. The recordings obtained for both models show that at the base of the elastic structure the PRA/PIA ratio remained almost constant and assumed unit values. On the contrary, at the top-level, large amplifications occurred. The amplification ratio reached about 6.5 for the conventional wall model and 5.5 for the self-confining wall model (Figs. 198~20). Again this different dynamic behaviour should be explained by mathematical analysis. It could be of practical interest when mitigation of seismic risk is considered.
Figure 19. Dynamic behaviour of the elastic structure on conventional wall model. Figure 16. Infill displacements in the two tested models.
Figure 20. Dynamic behaviour of the elastic structure on selfconfining wall model. Figure 17. Backfill displacements in the two tested models.
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8 NUMERICAL ANALYSIS With the aid of the UDEC 3.10 distinct element program, the two types of retaining wall were also numerically modelled and submitted to the same actions as the physical models tested on the shaking table (Fig. 21). The elastic structure was also modelled. However, the first results of the analysis brought more questions than answers. Under the seismic action at the bottom of soil structure a strong perturbation occurred. Since the direction of displacement vectors was nearly parallel with the reinforcement a special check would be necessary for that zone (Fig. 22). The upper layers of soil seemed to uplift and nothing prevented them from doing so. This would justify the need of confining soil structures and inducing over those layers some pressure. As concerns the settlements, the analysis showed that the layers of reinforced soil detached and settled separately, mainly after the critical time occurred around second 4 (Fig. 23). This separation of layers is a strong for confining the reinforced soil. The same separation of layers also appeared when
Figure 25 Displacement vectors in conventional wall model
horizontal displacements were analysed (Fig.24). This problem could be of great concern for the behaviour of soil structures in seismic areas. Finally, the dynamic influence of the elastic structure was much larger than expected, and the reinforced zone was mainly affected (Fig. 25). 9 CONCLUSIONS
1. The three parameters: settlements, horizontal displacements and response accelerations satisfactorily characterise the seismic behaviour of earth reinforcement walls. 2. The comparative method adopted in both physical and numerical analyses proved useful by preserving only the common phenomena and allowing their critical review. 3. Even if the confining is still a theoretical concept any efforts to make it true are worthwhile because it brings safety and economies. 10 ACKNOWLEDGEMENTS The financial support of the European Commission DGXII for Science, Research & Development through the Pro-ject INCO Copernicus IC 15-CT960203 EUROQUAKE and ECOLEADER programme is gratefully acknowledged. Thanks are also due to the Company Tensar International, based in UK, for supplying the polymer grids used in testing programs. Figure 23. Infill settlements in conventional wall for 0.1g EIC.
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REFERENCES Sofronie, R , Feodorov, V , 1995 Stability of Reinforced Slopes Proceedings of the 10‘” Danube- Eiaopeaii Conference on Soil A4echanics and Foiindation Engineering 1215 September 1995Mamaia, Romania, ,Vol2, pp 423- 428 Sofronie, R , Feodorov, V., 1998 Confining degree of Reinforced Soil Structures Proceedings of the 11‘” DanubeEiiropean Conference on Soil Mechanics and Foundation Engineering, 25-29 May 1998, Porec, Croatia, pp 279-282
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Sofronie, R A , C A Taylor, & P G Greening 2000 Seismic resistant retaining walls of reinforced soil Proceedings of the 12‘” World Conf on EQ Eng Auckland, N Z , 30 Jan -4 Feb 2000, paper #2029 Sofronie, R A , C A Taylor, & A J Crewe 2000 Mitigation of seismic risk by confining soil structures Proceedings of the Workshop on Mitigation of Seisniic Risk - Support to Recently Affected European Coztntries Hotel Villa Carlotta, Belgirate (VB), Italy 27-28 November 2000, paper #44 Taylor, C A , 1997 Large scale shaking tests of geotechnical structures ECOEST/PREC8, Report No 3
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Landmarks in Earfh Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Numerical analysis of soil nailed retaining wall B .R. Srinivasa Murthy Indian Institute of Science, Bangalore, India
G.L. Sivakurnar Babu Indian Institute of Science, Bangalore, India
A. Srinivas Indian Institute of Science, Bangalore, India
ABSTRACT: Two 5.0m high soil nailed permanent retaining walls as a part of a subway underneath a busy National Highway are designed, constructed and in operation at Indian Institute of Science, Bangalore, India. The walls are designed by conventional design methods and constructed with different types of facing using locally available technology and personnel. The paper examines the behavior of these soil nailed retaining walls using numerical simulations. The behavior of the soil nailed wall is numerically simulated using FLAC (Fast Lagrangian Analysis of Continua). The paper describes the modeling features such as simulation of excavation, installation of nails and construction of facing element and examines aspects such as variation of lateral earth pressure coefficients, tensions mobilized and deformations developed. length of the ramp is 60m. The walls are designed using conventional methods of analysis and constructed. Typical sections are shown in Fig. 1.
1 INTRODUCTION One of the challenging aspects of in-situ earth reinforcement is in understanding the behavior of soil nailed retaining walls. In soil nailed retaining walls, the properties and material behavior of three components namely the native soil, reinforcement (nails) and facing element significantly affect the performance of the structures. The behavior and performance of the wall are influenced by interaction of the components. The behaviour of the reinforced soil walls can be understood to some extent by studying the state of stress within the reinforced zone (Rowe and Ho, 1996). These influences are not adequately addressed in the conventional design procedures, based on limit equilibrium methods. In this paper, the performance of the retaining wall and the role of soil nailing in increasing the stability and in controlling the lateral deformations of the wall has been examined. Two types of facing and placement conditions are simulated. Variation of lateral earth pressure coefficients with depth, maximum tensions mobilized in soil nails have been obtained as a function of percent lateral deformation of the wall and the results are examined.
Figure 1 . Cross sections of soil nailed walls
2.2 Features of the subway One of the walls is vertical and the other wall has a batter of 5' and is adopted by providing 0.15m offset for each 1.5m depth of excavation. The total offset of 0.45m in stages, for 5.0m height is considered to be a better alternative as it enables better concreting of the facing during construction. The length of the nail is 3.5m. The behavior of soil nailed walls is studied from numerical simulations and some results are presented in the following sections.
2 FEATURES OF THE SOIL NAILED WALLS 3 NUMERICAL SIMULATIONS 2.1 Geometrical features
3.1 Problem descritisation in FLAC
Soil nailed walls have been constructed as a part of the subway in Indian Institute of Science campus, Bangalore. The height of the walls varies from zero at one end of the ramp to 5.0m at the other end. The
The complete soil mass is descritized into 6480 zones in plane strain condition. An excavation of 5.0m depth is simulated using null model by deleting 473
the corresponding elements from the mesh. Nails are simulated with pile elements and are introduced at designed spacing. The RCC facing is simulated using beam elements and is connected rigidly to the nails. The size of the mesh is 14m x 12m and sufficiently large enough to avoid boundary effects. Displacement boundary conditions are such that along the vertical boundary, in the sides x-displacements are fixed while y-displacements are free. At the bottom horizontal boundary, both x and y displacements are fixed (restrained) while at the top boundary it is unrestrained. 3.2 Simulation of excavation stages The complete soil mass is initially simulated for gravity stresses before excavation using elastic model and later converted to Mohr-Coulomb material. At the end of the simulation, the unbalanced force is brought to negligible level (up to 1 Newton) and displacements due to gravity are initialized to zero. Simulation of excavation, introduction of nails and providing the RCC facing are carried out as described in the following sections for two types of facings (vertical and offset) following the sequences of construction explained below.
1.5m to 3.0 m is rigidly connected to nails. The above sequence is repeated till the desired depth of excavation is reached. Table 1 shows the soil and nail properties used for both the sequences. Since the properties of soil at the location are highly variable in nature, three representative values of cohesion of soil (10, 15 and 20 kPa) are used for numerical analysis. Simulations are conducted for both sequences of construction. Reducing 3D problems with regularly spaced beams, cables or piles to 2D problems involves averaging the effect in 3D over the distance between the elements. Donovan et al. (1984) suggest that linear scaling of material properties is a simple and convenient way of distributing the discrete effect of elements over the distance between elements in a regularly spaced pattern. The above procedure is used in the present study. Table 1 . Properties of in-situ soil and nails
Sequence I Excavation up to 0.5 m depth and obtaining stability in terms of reduction of unbalanced force to the minimum value (1 Newton) is simulated. Reinforcement modeled as pile element is introduced into the soil at 0.25 m depth from top (corresponding to 0.5 m vertical spacing). Subsequently a beam element representing RCC facing is introduced at 0.25 rn and is rigidly connected to the pile element. Further excavation of 0.5 m depth is simulated and numerical stability is ensured. Reinforcement is introduced at 0.75 m depth from top. Facing element is introduced between 0.5 m to 1.0 m. The above sequence is repeated till the desired depth of excavation is reached. Sequence I1 Excavation up to 1.5 m depth in three steps, each 0.5 m depth and obtaining stability in terms of reduction of unbalanced force to the minimum value (1 Newton) is carried out. Reinforcement modeled as pile element is introduced into the soil at 0.25 m, 0.75 m and 1.25 m depths (corresponding to 0.5 m vertical spacing). RCC facing modeled as beam element is introduced between 0.0 and 1.5 m and is rigidly connected to the facing. Further excavation to next 1.5m depth in three 0.5m steps is conducted and numerical stability is obtained. RCC facing is introduced from 474
Parameter adopted Soil cohesion, c Soil friction angle, 4 Soil unit weight, y Soil elastic modulus, Es Soil Poisson's ratio, v Nail diameter, d Nail length, L Nail spacing, S, x sh
Value 10, 15 and 20 kPa 25" 18 kN/m' 20 MPa 0.3 0.02m 3.5m 0.5m x 0.5m
4 RESULTS AND DISCUSSION
4.1 Lateral deformation of wall Effective performance of reinforced soil structures is dictated by the mobilized strains in soil and reinforcement as a result of force field equilibrium (Jewell, 1987). Strains or deformations mobilized need to be evaluated to establish appropriate serviceability limits for the retaining walls. Normally, horizontal deformations at the top of stable retaining wall are of considerable importance. Figure 2 shows the variation of horizontal deformation of the soil without nailing and nailing using sequences I and 11. It can be observed that the wall is stable without nailing and the deformation is of the order of 25mm (0.5% of H). Soil nailing leads to considerable reduction in percent horizontal deformation as shown in Fig.2. Sequence I gives comparatively lesser horizontal deformations. The maximum horizontal deformation is 0.07% and the corresponding deformation in sequence I1 is 0.12%. The maximum tension mobilized for sequence I is 7 kN and for sequence I1 is 5.75 kN. The results suggest that the mobilization of strains
4.3 Mobilization of maximum tensile force in nails with cohesion and corresponding percent horizontal deformation
Figure 2. Variation of horizontal deformation (%H) with z/H (c = 20 kPa, = 25O, Es = 20 MPa).
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is affected by sequence of construction and the insertion of nails in soil mass before larger deformations are developed in soil mass is advantageous (as in the case of sequence I). The results are in agreement with the observations of Ho and Smith (1993) which suggest that early placement of the nail in the soil is useful.
Figure 4 shows the variation of maximum tensile force in nails versus deformation at the top of wall (%H) for three cohesion values (10, 15 and 20 kPa) which are considered for the present analysis. It can be observed that at the same percent horizontal deformation of the soil, for higher values of cohesion, higher is the maximum tensile force mobilization. This can be attributed to the stiffness of the soil surrounding the nails. Figure 5 shows the variation of percent horizontal deformation of wall as a function of maximum tensile force corresponding to different critical heights of excavation with sequence I type of construction.
4.2 Variation of lateral earth pressure coefficients Figure 3 shows the variation of lateral earth pressure coefficients with depth for c = 20 kPa, 4 = 25" and Es = 20 MPa. As mentioned in section 4.1, the wall is stable and the deformations are 0.5% of H. The corresponding lateral earth pressure coefficients immediately next to the facing are very small. The effect of nailing and sequence of construction on lateral earth pressures is also shown in Fig. 3. As the deformations are restrained, lateral earth pressures are mobilized in the opposite direction. The corresponding lateral earth pressure coefficients are shown and the values are high and in the range of 0.7 to 0.8 immediately at the nail of leveling.
Figure 4. Variation of maximum tensile force in nails with deformation of wall (%H) (sequence I, vertical facing, 4 = 25", Es = 20 MPa).
Figure 3 Variation of lateral earth Pressure coefficients (C = 20 kPa, = 25", Es = 20 MPa)
Figure 5 Maximum tensile forces in nails and horizontal deformation of wall at top (%H) at different depths of excavation for different values of cohesions (1 0,15 and 20 kPa)
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Critical height of excavation in the present study is the height beyond which the soil nailed wall collapses resulting in unacceptably high deformations. It can be observed that as critical height of excavation increases, maximum horizontal deformation increases. From the above results one can estimate maximum horizontal deformation as well as maximum tensile force in nails and critical height of excavation for known value of cohesion. Development of relationships indicated in Fig.5 for a given sequence of construction and range of soil parameters likely to be encountered in the in situ ground state are useful as design guidelines for construction of soil nailed walls.
5 CONCLUDING REMARKS The study examined the behavior of soil nailed walls using numerical simulations. The results suggest that the performance of the walls in terms of maximum tensile force in nails and horizontal deformation at
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the top of walls is dependent on sequence of construction and placement of nails. REFERENCES Donovan et a1 (1984) “Finite element approach to cable bolting in steeply digging VCR slopes“, Geomechanics application in underground hardrock mining, 65-90 Fast Lagrangian Analysis of Continua (1 996), Version 3 3, Reference manuals I, 11, I11 and IV Itasca consulting group, Inc Minnesota Ho, D K I-I and Smith, I M (1993) “Modeling of soil nailing construction by three dimensional finite element analysis”, Mathematical modeling of soil improvement (continuing education course ), Milano Jewell, R A (1988) “Compatibility, serviceability and design factors for reinforced soil walls”, Proceedings of the international symposium on earth reinforcement, Fukuoka, 6 1 1616 Rowe, R K and Ho, S K P (1996) “Some insights into reinforced wall behavior based on finite element analysis”, Proceedings of the international symposium on earth reinforcement, Fukuoka, 485-490
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Ultra-high hybrid wire and concrete-faced Mechanically Stabilized Earth bridge abutments K.M. Truong, M.J. Berkebile & R.A. Gladstone The Reinforced Earth Company, Vienna, Virginia, USA
ABSTRACT: Hybrid Reinforced Earth@abutments support a two-span highway bridge crossing a mine haul road in Arizona, USA. Concrete abutments were physically impractical and economically impossible at this 24m height, while a four-span bridge would have required massive, expensive earthmoving. Bridge loads are supported by spread footings bearing on the Reinforced Earth backfill. The superstructure's adjacent 1.4m deep precast box girders have an asphalt wearing surface, making the bridge easy to disassemble when the highway is relocated after the 10 year life of the mine. These unique abutments have a hybrid facing of Reinforced Earth precast concrete panels and TerraTrel@galvanized wire mesh facing, all connected to galvanized steel strip reinforcements. Exposed to impact, the lower 9m has 180mm thick precast facing. The upper 15m, with TerraTrel wire facing, directly supports the abutment footings. This hybrid Reinforced Earth design using TerraTrel solved an otherwise economically unsolvable problem. Settlement Tolerance - The bridge site is underlain by mine dump (waste rock previously used as backfill) of varying size, density and depth. Since differential settlement was possible between abutments and pier, or within the individual abutments, the modular nature of Reinforced Earth structures and their ability to deform without failure were significant benefits. Economics - Economy is always required, especially for temporary structures. Both the extremely high structural dimensions and the high standards of public safety (as compared to the sometimes lesser requirements for purely industrial-use structures) were economically met by the hybrid Reinforced Earth abutments without compromising either safety or quality. Indeed, no other abutment construction system was seriously considered due to the extreme requirements of this site.
1 BACKGROUND Improvements in mining technology have made recovery of lower-grade ores more economical, resulting in the opening or reopening of sections of many mines. This change in economics required the relocation of a state highway that passes through a major open-pit mine property in Arizona, in the southwestern United States. The nearly 2 km relocation traverses hilly terrain to avoid the new mine pit, requiring the highway to cross a haul road located in a deep ravine. A 24 m high, two-span bridge was required and Reinforced Earth@technology, utilizing a hybrid facing of both precast concrete and wire (Figure l), was selected for the abutments and wing walls. This selection was based on several factors, including: Temporary Structure - The expected economic life of the new mine pit is 10 years, after which the highway will be relocated to its original alignment and the bridge will be torn down. Ease of ConstructionDismantling - Reinforced Earth is a modular mechanically stabilized earth (MSE) retaining wall system that uses a low volume of manufactured components and a high volume of (in this case) readily available on-site backfill. Fabricated materials are shipped and erected easily and inexpensively compared to alternatives like cast-in-place concrete. After the anticipated 10-year service life, abutment dismantling will be almost as easy as was the original construction.
GEOTECHNICAL CONDITIONS AND DESIGN L.
1 Influence of miize dump.foundation material
Borings revealed that granitic bedrock underlay the bridge site at depths varying from 7 m to 26 m. Above the bedrock, almost to the ground surface, was mine dump (generally coarse waste rock) having a relative density ranging from medium to extremely dense. The variations in both density and thickness of the mine dump layer, and the previous 477
Figure 1. Hybrid Reinforced Earth concrete and wire-faced abutments.
observations of 5-10 cm of subsidence at other locations where the existing road is underlain by the same material, suggested the proposed 24 m high Reinforced Earth abutments (and the reinforced concrete pier) could experience some movement. The geotechnical report emphasized the development of stable and reliable foundation support for the complete bridge structure, including the pier and both abutments, by reducing the effect of mine dumpinduced subsidence.
2.2 Designing to accommodate mine dump Due not only to the heavy load to be imposed by the ultra-high Reinforced Earth structures, but also to the anticipated subsidence of the mine dump, the geotechnical engineer required a 6 m thick geogridreinforced MSE mat beneath the abutments and pier. The primary purposes of the mat were to: Provide uniform bearing for the Reinforced Earth abutments and the pier through removal, then replacement as engineered fill, of some of the dump material, Distribute stresses over an area wider than the abutment and pier footprints, reducing contact pressure on the untreated mine dump material below and reducing subsidence, Reduce the chance that any subsidence of the deep, untreated dump material would reach the ground surface, and Provide damping and resist settlement caused by mining activity-induced vibrations. 478
Mat design called for excavation of approximately 7.5 m of mine dump followed by placement of 21 layers of biaxial geogrid alternating with 0.3 m thick compacted backfill. The backfill was crushed and/or screened excavated mine dump, processed to a dense-graded granular consistency. The top of the mat was the foundation elevation for the Reinforced Earth abutment walls.
2.3 Anticipated settlement Settlement was estimated based on immediate elastic compression of both the reinforced fill in the MSE mat and the mine dump beneath the mat. Under each bridge abutment, settlement was estimated at 5-9 cm, except in areas with bedrock directly beneath the MSE mat, where expected settlement was on the order of 2 cm. The pier settlement estimate was 12.5 cm. Estimated differential settlement between abutments and pier was 1-8 cm. Settlement calculations assumed the elastic compression discussed above and did not estimate either long-term subsidence or settlement of reinforced fill within the Reinforced Earth abutments themselves.
3 STRUCTURE ECONOMICS AND DESIGN As indicated above, the abutment design had to meet both the rigorous safety requirements of public highway structures and the equally stringent demands for economy in temporary industrial structures. These combined requirements, plus the remote
project site, made MSE walls and their prefabricated materials the natural choice. The specific choice of Reinforced Earth was made by the low bidder, from among three alternates acceptable to the Arizona Department of Transportation (DOT), in a designbuild contract awarded and paid for by the mine owner. The structure and roadway design met state design and construction standards.
3.2 Design changes due to hybrid facing
3.1 Temporary-structure economics vs. traditional design requirements The initial design concept for the Reinforced Earth abutments used traditional precast concrete facing. When questioning by the owner revealed a $200,000 saving by using the Terratrel wire facing instead, state approval of this approach was sought. The temporary nature of the bridge and its remote location, with no public access to the abutment faces themselves (the haul road under the bridge is private mine property, minimizing risk of accidental damage or vandalism from the public), justified using the wire facing. The owner realized there was still a risk to the physical integrity of the abutment facings, however, from mine truck traffic under the bridge, meaning the full 24 m height could not be wire facing. The haul road vehicle mix would include heavy trucks carrying ore and spoil from the mining pit, with the attendant risk of large boulders being thrown against the wall face. Such impact or abrasion could damage the wire facing, forcing the owner
~i~~~~ 2, ~ ~ of wire facing b and~ precast panel and wire facings
to reconsider using precast panels for the lower portion of the walls. It was decided the precast would rise to an eIevation 7.5 m above the proposed haul road surface (9 m above foundation elevation), with the remaining 15 m height to be constructed of wire facing. Acceptance of this revised plan by the Arizona DOT resulted in the world's first hybrid Reinforced Earth facing system.
dbetween ~
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The decision to use a hybrid facing forced two design changes, even as it provided a net economic benefit. The two changes together lengthened the bridge by 60 cm, using up a small portion of the savings from changing to the hybrid system. The first change was to embed the wire facing 30 cm below the top of the concrete panels to protect the bottom of the wire wall from erosion. To create this embedment, the wire facing was positioned 15 cm behind the back of the concrete facing and 30 cm below the concrete panel's top, putting the wire only 7.5 cm above the topmost reinforcing strip of the concrete panel (Figure 2). Along the wingwalls, where the finished grade on the wall face grade climbed above the top of the and avoid loss of backfill at the transition (Figure 3). The extra 15 cm setback (per abutment) of the wire faces relative to the precast faces lengthened the bridge by 30 cm.
Figure~3. Transition concrete to wire facing. Note ~ from precast ~ finish grading to be required to protect bottom of wire facing.
The second change occurred at the bridge footings on top of the Reinforced Earth walls. To limit bearing pressure forces transmitted into the precast panels, normal clearance between the front of the footing and the back of the panels is at least 15 cm. Because of the flexibility of the wire facing system used on this project, however, a 30 cm bridge footing setback was required at each abutment (Figure 4), adding another 30 cm to the bridge length. 3.3 Reinforced Earth abutment design The bridge design conformed to the requirements of AASHTO (American Association of State Highway and Transportation Officials) and the Arizona DOT. Accordingly, the bridge live load for design was 96.2 kN/m (2-lane roadway, AASHTO HS 20-44 loading with no reduction allowed). The bearing pressure under the bridge seat was 163.7 kPa, well below the allowable of 191.5 kPa. Based on the subsurface conditions and the depth to bedrock, seismic design used an Arizona- and AASHTO-recommended horizontal acceleration of 0.05g. Since all construction would be done with crushed rock from mining operations, backfill properties were very favorable for Reinforced Earth design. Whereas the normal assumed angles of internal friction for select backfill, random backfill and foundation soil are, respectively, 34", 30" and 30", the design benefited substantially because all three angles of internal friction were 36".
The North abutment was 4.5 m higher than tht south abutment and had a maximum bearing pres sure at its foundation of 765.5 kPa. As originally de signed, the maximum horizontal stress within the re. inforced soil at the bottom of the north abutment wa: 191.5 kPa, requiring 14 reinforcing strips per panel Each strip carried a tensile force of 30.3 kN, as closc as possible to the strip's allowable tension of 32 kN Strip length was 18.3 m. (By comparison, soutt abutment loads were slightly lower, with a bearing pressure of 727.7 kPa, resulting in a bottom-layei strip density of 13 per panel and a strip length ol 17.1 m.) North abutment design had to be revised wher excavation for the MSE mat revealed bedrock in the lower 5.5 m of the wall was much closer to the wall face than had been anticipated. This condition would have required substantial blasting and removal of rock to make room for the 18.3 m long reinforcing strips, unless that length could be reduced. A review of the calculations showed that the lower 8 layers of strips could be shortened to 13.1 m without affecting the internal safety factors of the wall, although this increased the applied bearing pressure almost 23% (from 765.5 kPa to 940.7 kPa). The changes in reinforcing strip length and bearing pressure were reviewed and approved by the geotechnical engineer.
Figure 4. Looking down on front of bridge seat. Note clearance behind wire facing, top of precast panel walls below. Bridge deck is in foreground.
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Figure 5. Completed hybrid Reinforced Earth abutment.
4 CONSTRUCTION Manufactured materials were delivered to the project site in late December 1999, and the abutment walls were substantially complete by April 2000. The precast panel facings were installed on a customary 15 x 30 cm unreinforced concrete leveling pad, with the Terratrel wire facings founded on the backfill of the concrete panel walls. When wall construction began, the superstructure box girders were already in fabrication, making wall alignment and plumbness critical to insure proper girder fit. The plumbness issue dominated field quality control during the whole wall erection phase of the project. Each Terratrel wire facing panel sits immediately behind the wire facing panel below, so the thickness of the panel itself automatically creates a nearly 1.5 % batter in the wall unless measures are taken to compensate. For a pair of 15 m high wire walls, 1.5 % batter of both walls potentially puts the tops of the two abutment faces almost 0.5 m farther apart than intended. Indeed, field checks revealed this was happening, forcing crews to outward-batter the walls to preserve clearance behind the facings for the concrete bridge seats. Clearance between the front of the bridge seat and the back of the wire facing (Figure 4) was so important it was also decided to modify the bridge seat design slightly. The centerline of bearing was fixed due to girder fabrication, so the footing was both widened (extending it farther back from the wall face) and thickened. This change ensured con-
tact pressure remained below 191.5 kPa, that stresses were distributed into the reinforced soil at a safe distance behind the wall, and that local overstress of the facing was avoided.
5 CONCLUSION Ultra-high hybrid Reinforced Earth abutments (Figure 5 ) were the clear economic choice for relocating a highway across a mine haul road in a deep ravine. Adverse foundation conditions were overcome by removal of certain subsurface material and replacement with an MSE mat. Structure cost was reduced by using Reinforced Earth precast concrete facing for the lower 9 m and Tenatrel wire facing for the upper 15 m of these 24 m high structures. The change in facing type resulted in lessons learned regarding offset distances and plumbness with regard to the eventual setting of bridge girders. Postconstruction performance, including settlement, has been as expected. After mine economic life ends, demolition will be simplified by the modular nature of these unique Reinforced Earth abutments.
6 ACKNOWLEDGEMENT The authors gratefully acknowledge the owner's permission to write this paper and have willingly accepted his requirement that all identifying information be kept confidential.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2007 Swets & Zeitlinger, ISBN 90 2657 863 3
Combined reinforcement by means of EPS blocks and geogrids for retaining wall structures Y. Tsukamoto, K. Ishihara, H. Nakazawa & H. Kon Department of Civil Engineering, Science University of Tokyo, Japan
T. Masuo & K. Hara Taiyo Kogyo Corporation, Japan
ABSTRACT: This study illustrates that a compressible layer such as EPS blocks installed immediately behind a rigid wall combined with geogrid reinforcement proves to be a good solution to reduce earth pressures on earth retaining wall structures. A series of large-scale model tests were conducted with medium loose sand reinforced as follows; one test with EPS blocks only, another test with layers of geogrid reinforcement only, and the other test with layers of geogrid reinforcement fixed to EPS blocks behind the rigid wall. The roles of controlled yielding due to the inclusion of the EPS blocks and the fixity provided by the EPS blocks to the layers of geogrid reinforcement are discussed with respect to the earth pressures at rest and at active states and the geogrid tensile strains.
1 INTRODUCTION
2 EXPERIMENTAL DETAILS 2.1 Test apparatus
It has become known that the use of a compressible geosynthetic layer behind the rigid wall allows lateral expansion of soil in retaining wall structures, and therefore leads to a reduction in the lateral earth pressure, which enables the retaining wall structures to gain more stability. This technique was successfully employed in the fields and reported by Partos & Kazaniwsky (1987), who used a prefabricated expanded polystyrene bead drainage board for a 10 metres high basement wall. Karpurapu & Bathurst (1992) supported the effectiveness of this technique by employing a FEM analysis, and introduced a controlled yielding concept as follows. A compressible geosynthetic layer behind the rigid wall is subjected to lateral compression due to the action of the earth pressure. The soil behind the wall is on the other hand subjected to lateral expansion, and therefore caused to deform plastically towards an active state in proportion to the lateral compression of the geosynthetic layer, while the rigid wall is kept stationary. This phenomenon is characterized by the tradeoff between lateral compression of the geosynthetic layer and lateral expansion of soil, in other words, the increase in the compressive stress on the geosynthetic layer and the reduction in the lateral earth pressure exerted by soil. As a result, the earth pressure at rest in effect reduces. In this study, the use of compressive expanded polystyrol (EPS) blocks for the controlled yielding of granular materials in retaining wall structures is examined by employing large-scale model tests. The use of geogrid reinforcement together with EPS blocks installed behind the rigid wall is also examined.
Figure 1 shows the large-scale model test apparatus used in this study, which is described in more detail by Tsukamoto et al. (1999). The model soil container is 1.5 m wide, 1.5 m long and 1.05 m high, in which one of the wall with 1.5 m in width and 1.O m in height is a motor-driven mobile rigid model retaining wall and produces a translational mode of wall displacement. The earth pressures are measured with the pressure cells installed at the mobile rigid model retaining wall, and the displacement of the model retaining wall is measured with dial gauges attached to the model retaining wall. The surcharge loading unit is located at the top surface of the model soil specimen. All the voltage outputs from the pressure cells, dial gauges and electrical resistance strain gauges instrumented along layers of geogrid reinforcement embedded in the model soil specimen, where necessary, are stored in the data acquisition system. 2.2 Soil material, geogrid and EPS Toyoura sand is used as a model granular backfill in this study, which is a poorly graded clean fine sand with no fines. The physical properties of this material are as follows ; G,=2.65, D50=0.19mm, U,=1.70, F,=O%, em,,=0.988 and emi,=0.616. One type of the geogrid reinforcement is used in this study. The polymer type is vinylon, and the physical properties are as follows; m a s s h i t area = 200g/m2, aperture size (LxW) = 19mmxl6mm, rup-
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small blocks with 15cm in height, 20cm in length and 100 cm in width, which are used in the model tests described below. 2.3 Test series Figure 2 shows the test series carried out in this study. In all the tests, the dry soil specimens are prepared with a relative density D, of 75%. The soil specimen is air-pluviated to a designed height and the top surface is tamped equally. A layer of geogrid reinforcement is placed where necessary, and this procedure is repeated until the entire soil specimen is ready. Then, the surcharge loading unit is set up on the final top surface of the soil specimen. In test A, there is neither geogrid reinforcement nor EPS blocks in the model soil specimen. In tests B-1 and B-2,4 EPS blocks of 20 cm long and 30 cm long are installed immediately behind the retaining wall, respectively. In test C, 3 layers of geogrid reinforcement are embedded in the model soil specimen and are not fixed to any wall of the model container. In test D, 4 EPS blocks and 3 layers of geogrid reinforcement are introduced and each layer of geogrid reinforcement is fixed between the adjacent EPS blocks. Figuel . Test apparatus.
2.4 Test procedure
ture strength = 29.4kN/m, secant stiffness at 0.5% strain = 560kN/m and strain at rupture = 10%. The physical properties of the EPS used in this study are as follows; mass/unit volume = 25kg/m2, allowable compressive strength = 7OkPa, Young's modulus under triaxial com ression at a confining pressure of lOkPa = 2 . 5 ~ 1Ip0kPa and Poisson's ratio = 0.13. A large block of the EPS material was provided and cut into pieces to produce a dozen of
Figure 2. Test series.
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It is more comprehensive and informative to see the test procedure in the p - q plot of the loading history, where p = (ov+ot,)/2, q=(o,-oh)/2, ovis the overburden pressure at the location of the pressure cells, and o h is the lateral earth pressure measured by the pressure cells. Figure 3 shows the p - q plot for test B-1. First, the surcharge pressure of qo = 100kPa was applied to the top surface of the soil specimen. The mobile model retaining wall was then moved way
the K, value is even less than that for test C. The physical interpretations of these findings are given below. 3.2 Influence of EPS blocks Figure 4 shows the earth pressure K - wall displacement d curves for tests A, B-1 and B-2, in which the K values move from at rest to active state conditions. It is seen that although the comparison of the curves for tests A and B-1 gives erroneous results, in which these curves cross over each other, the KO value for test B-1 is found to be lower than that for test A, but the K, values are similar to each other. This reduction of the KOvalue due to the EPS blocks inclusion may be interpreted with a controlled yielding concept. Figure 5 shows the interpretation of the controlled yielding concept as follows. Koi denotes the earth pressure coefficient at rest against a rigid wall. When a compressible layer such as EPS blocks is introduced behind a rigid wall, the backfill material is caused to expand laterally in relation to the amount of the compression of the EPS blocks, AL = &hxL,where &h and L are the cornpressive strain and length of the EPS blocks, respectively. Subsequently the earth pressure reduces down to KO,. By assuming a plane strain condition, the following formula may be used for the initial estimates of AL ;
Figure 3. Test procedure in p - q plot.
from the soil specimen until the active stress state was achieved, while the surcharge pressure was maintained constant. The surcharge pressure was gradually released. Finally the model wall was moved back to the original position. This procedure was repeated under the different surcharge pressures of qo = 100, 150 and 2OOWa. The KOand K, values can be inferred from the p - q plot by interpolating three data points corresponding to the test conditions under the three different surcharge pressures, as shown in Figure 3, where KO= (l-a)/(l+a), K, = (1p)/(l+p), a and p are the inclinations of the KOand K, lines, respectively.
3 TEST RESULTS 3.1 Earth pressure coeficients, K, and K, Table I summarizes the earth pressure coefficients inferred from the test results, as described above in Section 2. It is seen that in comparison to test A where no reinforcement is introduced, the KO value greatly reduces with the EPS blocks inclusion, but the K, value appears to be the same, as seen in tests B-1 and B-2. On the other hand, the K, value reduces with the geogrid reinforcement, but the KO value appears to be the same, as seen in test C. In test D where the layers of geogrid reinforcement fixed to the EPS blocks are introduced, the KO as well as K, values are found to reduce. In this case, Table 1. Summary of K, and K, values. Test
Test conditions
K, value
K, value
A B-1 B-2 C D
no reinforcement EPS(k0.2m) EPS(k0.3m) geogrid EPS+geogrid
0.395 0.354 0.120 0.403 0.292
0.1 14 0.1 12 0.105 0.076 0.042
Figure 5. Interpretation of controlled yielding concept.
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where E and v may be assumed as described above in Section 2.2. By some manipulation of the test results, the initial estimates of AL thus obtained seem to provide a good evaluation of the reduction of the earth pressure due to controlled yielding. The KO value for test B-2 is extremely lower than those for test A and test B-I. This significant reduction is beyond description from the viewpoint of the controlled yielding concept. This phenomenon might rather be related to the fact that the significant portion of the soil, where plastic flow would be expected to occur due to wall movement if there were no EPS blocks, is replaced by the lightweight EPS blocks in case of test B-2, which in turn greatly reduces the earth pressure at rest. 3.3 Influence of geogrid The K - d curves for tests A and C are included in Figure 6. The layers of geogrid reinforcement embedded in the model backfill have little influence on the earth pressure at rest, where the geogrid tensile stresses are not mobilized enough to counteract against the lateral stress exerted by the model backfill on the retaining wall. It is only when the wall moves and reaches an active state that the geogrid tensile stresses are mobilized, and therefore affect the earth pressure at an active state. 3.4 Influence offixity between EPS and geogrid Figure 6 also includes the comparison of the K - d curves for tests C and D. It is to note that each layer of geogrid reinforcement is not fixed to any wall of the model container and left free in test C, while in test D each layer of geogrid reinforcement is fixed between the adjacent EPS blocks. The earth pressure at rest for test D is found to be lower than that for test C. This can be explained by the controlled yielding of the EPS blocks, as described in Section 3.2. The earth pressure at an active state for test D is found to be lower than that for test C. This difference is considered to stem from the fixity provided by the EPS blocks to the layers of geogrid reinforcement, which affects the magnitudes of tensile strains induced along the layers of geogrid reinforcement. In tests C and D, the layers of geogrid reinforcement were instrumented with electrical resistance strain gauges to observe the distribution of the geogrid tensile strains. Figure 7 shows the distributions of the geogrid tensile strains with wall displacement for tests C and D, in which D is the distance from the original position of the model retaining wall. The maximum geogrid tensile strains are observed at around D = 40 cm for both tests C and D. In this diagram, the geogrid tensile strain at D = 0 for test C is theoretically assumed to be zero, because the layers of geogrid reinforcement are left free. However, the geogrid tensile strain at the point of fixity to the EPS blocks for test D develops with 486
Figure 7. Distribution of geogrid tensile strain with wall displacement.
wall displacement, which most probably allows the maximum geogrid tensile strain for test D to become greater than that for test C, as seen in Figure 7. This increase in the maximum geogrid tensile strain is most likely to contribute to the difference of the lateral earth pressures at active states in tests C and D. 3.5 Summary of experimental findings The experimental findings from the test results can be summarized as follows: (1) The inclusion of the EPS blocks behind the rigid wall reduces the earth pressure at rest, in comparison to the unreinforced rigid wall, due to what is called controlled yielding. (2) The layers of geogrid reinforcement embedded in the model backfill reduce the earth pressure at active states, mainly due to the tensile strains induced along the layers of geogrid reinforcement. (3) The combined reinforcement by means of the EPS blocks and the layers of geogrid reinforcement reduces the earth pressures at rest as well as
5 CONCLUSIONS From the viewpoint of the earth pressures on earth retaining wall structures, the roles of a compressible layer such as EPS blocks installed immediately behind a rigid wall and layers of geogrid reinforcement embedded in a granular backfill were clarified in this study. The role of the compressible EPS blocks was examined by a controlled yielding concept, which contributes to the reduction in the earth pressure at rest. The role of layers of geogrid reinforcement was examined by geogrid tensile strains, which contributes to the reduction in the earth pressure at an active state. The combined use of the EPS blocks and layers of geogrid reinforcement enables the fixity to be provided by the EPS blocks to the layers of geogrid reinforcement, which contributes to the even more reduction in the earth pressure at an active state. To make this technique into practical use, several practical problems need to be overcome, such as fatigue occurring at the fixed positions between fragile EPS blocks and layers of geogrid reinforcement, drainage system, installation damage, and various endurance and degradation characteristics of EPS blocks and geogrid among others.
Figure 8. Schematic of K - d curves (nothing, EPS only, geogrid only, EPS+geogrid).
at active states. The reduction in the earth pressure at rest can be demonstrated by the controlled yielding concept. The reduction in the earth pressure at an active state can be related to the fixity provided by the EPS blocks to the layers of geogrid reinforcement. Based on the above findings, the behaviour of earth pressures with wall displacement under various conditions may be schematically illustrated as shown in Figure 8.
4 IMPLICATION TO FIELDS
6 ACKNOWLEDGEMENTS This study examined the use of EPS blocks and geogrid reinforcement for retaining wall structures. It needs careful considerations to make this reinforcement technique into practical use. Because of a fragile nature of EPS, a fatigue problem at the fixed positions between EPS blocks and layers of geogrid reinforcement and an installation damage problem need to be addressed. Because of a buoyant nature of EPS, a proper drainage system needs to be designed for this particular structure. A design life time of this structure needs to consider various endurance and degradation characteristics of EPS and geogrid materials used. It is noteworthy that from the viewpoint of the experimental findings, any compressible geosynthetic materials, which guarantee the occurrence of controlled yielding, may be able to replace EPS for this type of structures.
The authors would like to acknowledge Mr. T. Tsuruniaki for his efforts in carrying out the test series reported in this paper. REFERENCES Karpurapu, R. & R.J. Bathurst 1992. Numerical investigation of coiitrolled yielding of soil - retaining wall structures. Geotextiles ~ i i dGeonienzbranes, 1 1 : 1 15 - 131. Partos, A.M. & P.M. Kazaniwsky 1987. Geoboard reduces lateral earth pressures. Proceedirzgs ofGeosynthetics’87 Confereiice, New Orleans, USA : 628 - 639. Tsukamoto, Y., Ishihara, K., Higuchi, T. & H. Aoki 1999. Influence of geogrid reinforcement on lateral earth pressures against model retaining walls. Geosynthetics Internutional, 6 , 3 : 195 - 218.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Irregular shaking table tests on seismic stability of reinforced-soil retaining walls Kenji Watanabe, Masaru Tateyama & Kenichi Kojima Railway Technical Research Institute, Japan
Junichi Koseki Institute of Industrial Science, University of Tokyo, Japan
ABSTRACT: In order to establish practical design procedures to evaluate seismic stability of different types of retaining walls against high seismic loads, a series of shaking table tests with irregular wave were conducted on retaining wall models consisting of six different types. In these tests, reinforced-soil retaining wall models with a rigid full-height facing exhibited ductile behavior compared to conventional type retaining wall model such as gravity-type, leaning-type and cantilever type ones. This is because when the conventional type wall started to tilt, the subsoil reaction force at the toe of wall suddenly decreased due to loss of bearing capacity. On the other hand, under similar conditions, tensile force in the reinforcement of reinforced-soil walls could be mobilized effectively to resist against the overturning. It should be also noted that, in the current aseismic design procedures based on the pseudo-static approach, the different extents of ductility and seismic performances of different types of retaining walls are not taken into account properly. In this study, therefore, a series of relatively small-scale model tests was conducted on the different types of RWs to compare their different performances during irregular shaking.
1 INTRODUCTION In recent years, there have been several damages on retaining walls(RWs) due to large earthquake. For example, the Hyogoken-Nanbu earthquake of January 17, 1995, caused serious damages to conventional masonry and concrete gravity-type RWs for railway embankments. On the other hand, the reinforced-soil RWs exhibited ductile behavior and did not reach to critical failure as reported by Tatsuoka et al. (1996). The main loads acting on the retaining wall during earthquake are inertia force, subsoil reaction, seismic earth pressure and tensile force in the reinforcement as schematically shown in Figure 1. The subsoil reaction force and the seismic earth pressure are affected by the dynamic interaction between the wall and the soil, of which the detail mechanisms have not been understood well. Further, for the reinforced-soil wall, the interaction between the soil and the reinforcement, and the seismic behavior of the reinforced and unreinforced zones of the backfill have not been understood in detail, either. For these reasons mentioned above, it is diffi-cult to predict the seismic behavior of the RWs rationally.
2 TESTING PROCEDURES 2.1 Model of retaining walland backJill The model tests were conducted by using the shaking table at Railway Technical Research Institute. A rigid soil container (260 cm long, 60 cm wide, and 140cm high) was fixed to this table. The cross-sections of six different model retaining walls are shown in Figure 2. The models were 600 mm in width (Figure 3). They consists of three conventional RWs (cantilever type, gravity type and leaning type) and three types of reinforced-soil RWs with a full height rigid facing having different arrangements of reinforcement layers (reinforced-soil RWs type 1, type 2 and type 3). The total height of the conventional walls was 530 mm, while that of the reinforced-soil walls was 500 mm. The bottom width at the base of the cantilever and gravity type walls was 230 mm, while it was reduced to 180 mm for the leaning type wall. To adjust the dead load of the gravity and the leaning type walls, extra weights were added nearly at the center of gravity of these walls.
Figure 1. The main forces acting on the wall during earthquake.
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Figure 2. Cross-section of model retaining walls (unit in cm).
To observe the deformation and displacement of sand layers, horizontal layers of black-dyed Toyoura sand having a thickness of 10 mm were prepared at a vertical spacing of 50 mm adjacent to the transparent side wall. The seismic earth pressure acting on the backface of the wall and the subsoil reaction at bottom of base footing were monitored by using two-component load cells which can record both the normal and shear components of earth pressure as shown in Figure 3. The load cells were placed along the center line of the wall surface in order to reduce the effects of the side wall friction of the soil container. To measure the response of each retaining wall and backfill, a number of displacement transducers and accelerometers were installed. The detail of the model is described in Koseki et al. (1998) 2.2 Shaking table tests
Figure 3 Details of typical wall model (Gravity type)
Seismic loads were applied by shaking the sand box horizontally with an irregular base acceleration. A strong motion that was recorded as N-S component at Kobe Marine Meteorological Observation Station during the 1995 Hyogoken-Nanbu earthquake was employed as the base acceleration (Figure 4). Its atn-
For the reinforced-soil RW models, a grid of phosphor-bronze strips was used as the model reinforcement. Ten layers of reinforcement strips having a length of 200 mm were horizontally placed in the backfill sand for reinforced-soil wall type1 . The length of the top and fourth reinforcement layers was increased to 800 rnrn and 450 mm, respectively. for the reinforced-soil wall type 2 in order to increase the stability against overturning failure, as is the common practice in Japan. To study into effects of length of the reinforcement layers, the length of all the rcinforcernent layers was increased to 350 mm for the reinforced-soil wall type 3. Strain gauges were attached to the reinforcement to measure the tensile force. The subsoil and the backfill layers were made of air-dried Toyoura sand (D50=0.23mni, Gs=2.648, e,,,=0.977, ern]"=0.609). The sand layers were prepared by using a sand hopper with keeping the falling height of sand constant. An average relative density of 90% was achieved by this method.
Figure 4. Typical time history of base acceleration.
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This behavior suggests that the horizontally placed short reinforcement layers cannot effectively resist such simple shear deformation of the reinforced backfill. This simple shear deformation of the reinforced backfill should be considered to evaluate the residual displacement of the wall. Figure 6 shows the locations of failure plane and the reinforcements for reinforced-soil RW type 2. The arrows indicate the end of longer reinforcement at the moment when the failure planes were formed. The two failure planes were formed almost simultaneously. The upper one developed from the back of the reinforced zone towards just beside the end of the extended reinforcement (45cm), stopping somewhere below the longest reinforcement. On the other hand, the lower failure plane was formed just beside the end of the longest reinforcement (80cm) and reached the surface of the backfill. This demonstrates that the reinforcement resisted against the formation of the failure plane, and the location of the failure plane was strongly governed by the existence of the extended reinforcement. Accordingly, large tensile force was mobilized in the extended reinforcements which will be discussed later.
plitude and time scale were adjusted so that the base acceleration has a prescribed maximum amplitude with a predominant frequency of 5 Hz. Each model was subjected to several shaking steps, where the maximum amplitude of the base acceleration was initially set to 100 gals and increased at an increment of 100 gals. Shaking was terminated when the wall displacement became considerably large.
3 TEST RESULTS AND DISCUSSION 3.1 Failure pattern of models Figure 5 shows the residual displacement of the wall and the residual deformation of the backfill, which were observed at the end of final shaking step. For all RWs, the major failure pattern of the walls was overturning, which was associated with bearing capacity failure for the cantilever, leaning, and gravity type RWs. For the reinforced-soil walls, no failure plane was observed at the bottom of the front wedge in the reinforced zone. The front wedge did not behave as rigid, but it suffered simple shear deformation along horizontal planes. This is because the resistance against the formation of failure plane penetrating though the reinforcement was larger than that against the simple shear deformation of the reinforced zone.
3.2 Residual displacement of wall Figure 7 shows relationships between the seismic coefficient k17and the horizontal displacement d,,, at
Figure 7. Accumulation of residual horizontal displacement near the top ofthe wall.
Figure 5. Residual displacement of the wall
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the end of each shaking step which was measured at a distance of 5cm below the top of the wall. The seismic coefficient kh was defined as k~~=a,,,,,/g, where alnuxis the maximum base acceleration at the active state (i.e., when the inertia force was oriented towards the active direction) for each shaking step, and g is the gravitational acceleration. In the early steps of irregular shaking tests up to kh value of about 0.5, the d,*[, value accumulated in a similar manner among different types of RWs. On the other hand, when the kh value exceeded about 0.5, the rate of increase in the dropvalue was larger for the three conventional type RWs than for the reinforced-soil RWs. Further, though the total length of reinforcement of type2 was 80% as much as that of type 3, the seismic stability of them was on the same level. Such different extents of ductility in each type of RW agree with the damage observed after Hyogoken-Nanbu earthquake. This is caused by the different resistance mechanism against the external forces acting on the wall such as inertia force and seismic earth pressure. The details will be discussed in the next two sections.
bottom of base footing. When the inertia force was oriented towards the active direction (Point A), the normal force increased at the toe of the base footing, and decreased at the heel in contrast. This experiment clearly revealed that the wall was rotating around the center of the footing, causing stress concentration at the toe of the base footing. Figure 9 shows the relationship between the reaction force from subsoil and the horizontal displacement of the wall d,(),,. Figure 10 shows the relationship between the resultant normal force from subsoil and relative location of its application for gravity type retaining wall. Each experimental data was taken when the inertia force towards the active direction was maximized in each shaking step. In the early shaking steps, the normal stress measured at the toe of the base footing increased rapidly (0-0 in Figure 9). At this moment, the application point of the resultant force gradually moved toward the toe
3.3 Reaction force from subsoil The conventional type RWs resist against the overturning by the reaction force from subsoil. Figure 8 shows the time history of normal stress of the reaction force which includes initial values measured before starting shaking for Gravity type RW. The reaction force was measured by four loadcells at the
Figure 8 Time history of normal pressure at the bottom of base footing for gravity type retaining wall
Figure 10 Relationship between resultant normal reaction force from subsoil and relative location of its application
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of the base footing, accompanied with only a slight increase in the amount of the resultant force (0-8 in Figure 10). After attaining the peak state, the dtoi, value suddenly increased due to loss of bearing cain Figpacity near the toe of the base footing (0-8 ure 9). At this moment, the resultant force decreased suddenly, and its application point moved back toward the heel of the base footing (0-8 in Figure 10). This behavior caused large decrease in the resisting moment against overturning, which led to the low ductility of conventional type RWs. 3.4 Tensile force in reinforcement layer
As mentioned before on Figure 7 , the rate of accumulation of the d,, value did not increase rapidly for three types of reinforced-soil RWs in irregular shaking tests. The Reinforced-soil RWs resist against the overturning moment by the tensile force in the reinforcements in each layer. Figure 11 shows the time history of tensile force in each reinforcements which were measured at a horizontal distance of 2.5 cm from the facing in the uppermost, middle-height and lowest layers. All tensile forces increased simultaneously to resist against the overturning when the inertia force was oriented towards the active direction (Point A in Figure 11). Figure 12 shows the relationship between the tensile force and the horizontal displacement of the wall dlop. The tensile force data was measured when the inertia force towards the active direction was maximized in each shaking step. For all types of rein-
forced-soil RWs, the tensile force increased with the d , , value, not showing such a sudden drop as observed in the reactions from subsoil for gravity type RW (Figure 9). This may explain the ductile behavior of reinforced-soil RWs. It can be also seen from Figure 12 that the tensile force in the uppermost layer was largest for reinforced-soil type 2 with longest reinforcement, while it was smallest for reinforced-soil type 1 with shortest reinforcement. In particular, the former value increased even when the dloi, value was relatively small, while the latter value increased only after the dtoi, value exceeded about 20 mm. These different behaviors may suggest that extension of upper reinforcement layer, such as the case with reinforcedsoil type 2, effectively mobilizes the tensile force in the extended reinforcement. Figures 13a,b and c show the horizontal distribution of tensile force in each reinforcement. As the facing was rigid, almost all the tensile forces were maximized at the nearest point from the facing. For the reinforced-soil RW type 2, however, the tensile force of the uppermost reinforcement (80cm) was largest at a point near of its tip. Such different degree of mobilization of tensile force may be linked to the different locations of these reinforcements relative to the failure planes as typically shown in Figure 6. That is, the tensile force in the uppermost reinforcement for the reinforced-soil RW type 2 was largely mobilized to resist against the formation of the failure plane No.1 (Figure 6) which would have otherwise reached the surface of the backfill.
Figure 12 Tensile forces in reinforcement layers measured at a distance of 2 5cm from facing of reinforced-soil RWs
Figure 1 1 Time history of tensile force in reinforcement for reinforced-soil RW type 2
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Figure 13. Horizontal distribution of tensile forces in reinforcement layers for reinforced-soil retaining walls a) type 1, b) type 2 and c>type 3 .
4 CONCLUSION Shaking table tests in 1-G field were carried out, and the following conclusions were drawn. 1. At high seismic loads, reinforced-soil type RWs showed more ductile behavior than conventional (cantilever, gravity and leaning) type RWs. When the model walls started to tilt, concentration of subsoil reactions at the toe of conventional type RWs resulted into local failure due to loss of bearing capacity. On the other hand, under similar conditions, tensile force in the reinforcement of the reinforced-soil RWs could be mobilized effectively to resist against the wall movement. 2. It was demonstrated that by extending several upper reinforcements, the seismic stability of rein-
494
forced-soil walls could be improved effectively. This is because the tensile force in the extended reinforcement was largely mobilized to resist against the formation of the failure plane. REFERENCES Tatsuoka, F , Tateyama, M & Koseki, J 1996 Performance of Soil Retaining Walls for Railway Embankments Soils ar7d Foundations, Special issue of Soils and Foundations on Geotechnical Aspects of the January 17 1995 HyogokenNanbu Earthquake 3 1 1-324 Koseki, J , Munaf, Y , Tatsuoha, F , Tateyama, M ,Kojima, K & Sato, T 1998 Shaking Table and Tilt Table Tests of Geosynthetic-Reinforced Soil and Conventional Retaining Wall, Geosynthetics International, Vol 5, Nos 1-2 73-96
Landmarks in Earth Reinforcement, Ochiai e f a/. (eds), 0 2001 Swets & Zeiflinger,/SBN 90 2651 863 3
Issues in the use of clay in reinforced earth construction L.D. Wesley Department of Civil and Resource Engineering, University of Auckland, New Zealand
ABSTRACT: The use of clay fill in the construction of reinforced earth walls involved substantial cost savings, and is being increasingly adopted by geotechnical engineers in New Zealand. This paper looks at some of the issues this raises, in the New Zealand context, especially the following: ( I ) The appropriate soil strength parameters to be used in design, ie the choice of the peak value ($'J, the critical state or constant volume value ($'cv), or the residual value (2) The expected deformations which may occur at the wall facing, or elsewhere, either during construction, or after completion, because of the lower stiffness of clay compared to granular material. (3) Possible adverse effects from excessive down-drag on the facing elements. The increased difference in stiffness between the facing and the clay fill could lead to excessive vertical load on the facing. (4) Compaction difficulties and the possible development of pore pressures during construction. (+lr).
the term "4' critical", or $lent is also used to designate the critical state angle, and is the same as 4'cv). The Netlon guide also quotes the UK Dept. of Transport Advice Note HA 68/94 (1 994), which recommends the use of'the residual strength for clays having Plasticity Index values over 25. In New Zealand, the appearance of documents and codes for reinforced earth which recommend the use of $'cv has given rise to some mild unease and "puzzlement" amongst geotechnical engineers. Critical state soil mechanics has not made much impression on the practising geotechnical community in New Zealand, and reinforced earth design in the past, along with all other effective stress analysis, has been carried out using the familiar Mohr Coulomb peak parameters c' and 4'. Together with the issue of the $' parameter for use in design has been an interest in building reinforced earth walls using local clays, because of their ready availability. The choice of 4' parameter becomes more important with clays because the peak value is lower and the difference between peak, critical state, and residual values may be greater than for granular materials. A short research programme has therefore been carried out to investigate the parameter $'cv for some typical local soils (primarily of a cohesive nature), and try to establish its relationship with the standard "peak" Mohr Coulomb parameter 4'.
1 INTRODUCTION The advent of new reinforcing materials such as geotextiles and geogrids has made possible the use of clay fill for reinforced earth (R.E.) walls. The cost reductions associated with the use of clay has made its use increasingly attractive This paper looks at several issues which arise from the use of clay.
($lr)
2 SOIL STRENGTH PARAMETERS 2.1 General There are differing opinions among geotechnical engineers as to whether the peak, critical state, or residual strength should be used for the design of reinforced walls and slopes. For example, Zornberg et a1 (1 998), describe model tests which show that failure is clearly governed by the peak value. and argue strongly (Zornberg et al. 2000) for its use in design. Leshchinsky (2000) proposes a hybrid design procedure in which the peak value is used to determine the critical slip surface, but the residual value is used in determining the grid anchor length. The current Netlon (1996) guide for the design of reinforced earth structures using geogrids suggests that the critical state parameter $'cv is the appropriate parameter. (The suffix cv denotes "constant volume";
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Table 1 . Details of samples and test results
SAMPLE DETAI I S
Atterberg Limits Natural SHEAR STRENGTI-I L.L. P.L P.I. water Parameter Peak Large strain Residual content YO “end’ values 1. CLAY; moderate plas- 65 32 33 49.5 cr ( H a ) 20 0 0 ticity, pale grey $r (deg.) 28.2 30.5 15.0 2. CLAY; high plasticity, 84 27 57 45.5 c’ (kPa) 11.2 0 0 dark grey. Q‘(deg.1 26.8 25.1 1o* 3. SILTY SAND; yellow - N.P 26.2 c’(kPa) 14.5 0 0 $‘(deg.) 35.6 33.3 27.0 4. SANDY CLAY; with 104 60 44 61.5 c‘(kPa) 22 0 0 some coarse material. $‘(deg.) 35.8 36.8 34.1 2.2 Experimental investigation Four samples were obtained from local sites; all were residual soils. Three came from weathered sandstone, and the fourth from weathered volcanic ash. Classification and compaction tests were first carried out, followed by triaxial testing. Consolidated undrained and drained triaxial tests were carried out on each material; they were continued to large axial strains - about 30% in the hope or expectation that this would lead to critical state behaviour and define the critical state parameter $‘cv. In addition to the triaxial tests, ring shear tests were carried out on three of the samples to determine the residual strength friction angle $rr. The results of the tcsts are described fully by Wesley and Davidson (2000), and only a summary is given here. Sample details and results are found in
Figure. 1 Typical triaxial test results (from Sample 1).
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Table 1, and typical triaxial results shown in Figures 1 and 2. Figure 1 shows both the consolidated undrained and drained tests for Sample 1. In the consolidated undrained tests the deviator stress increases rapidly up to strains of 2 to 3%, and then undergoes a very slight but steady increase up to the maximum strain of 20 to 30%. The pore pressure initially rises and then shows a steady decrease. In the drained tests, the deviator stress rises less steeply and reaches a peak at strains between about 10 and 15 ‘-Yoand then shows a slight but steady decrease. The volume initially shows a significant decrease, followed generally by a slight but steady increase. As neither deviator stress nor pore pressure (or volume change) have levelled off to constant values at the end of these tests, the critical state has clearly not been reached, and b‘cvhas not been established.
pragmatic approach could be taken and either the peak 4‘ value be adopted with the c‘ value neglected, or the “end” value at a strain of about 30% be adopted. These values can be expected to be very similar. As mentioned earlier, some codes recommend that when clay is used in reinforced earth, the residual strength should be used in design. This seems a grossly over-conservative approach, without any theoretical justification. There are no other situations in geotechnical engineering where the residual strength is used on an intact material, and there is no reason for it to be used with RE walls. It is rather difficult to see why the issue of using anything other than the peak strength has arisen with RE walls. Safety factors are used in their design, and the deformations they undergo are generally small, so there is no reason to think the soil will be stressed beyond its peak strength. Also, deformations occur in an overall manner, so there is little likelihood of displacement on specific planes leading toward the residual strength.
Figure 2. Typical peak, “large strain”, and residual shear strength parameters (from Sample 1).
This is perhaps not surprising, as there is real doubt as to whether the critical state can be created in clays, especially in standard triaxial tests. Once the peak deviator stress has been reached, clays tend to develop specific failure planes; movement is then concentrated on these planes and deformation is no longer uniform. Instead of moving towards the critical state, the strength may simply decline from peak towards the residual value. Fig. 2 shows the same results plotted as stress paths on a q/p‘ plot where: I
0 1- 0 3
4
=
2
3 DEFORMATIONS WITH CLAY FILL
I
0 1 + 0 3 p f = _____ 2
“Best fit” lines have been fitted to the results as shown, giving peak Mohr-Coulomb parameters. In addition lines have been drawn through the end points of the stress paths to define the “end” or “large strain” values of the parameters. Table 1 shows peak, “end”, and residual strength parameters from all the samples. It is evident that the post failure decrease in strength is quite small, and in none of the tests does it approach the residual value, except for the volcanic ash sample. This material shows the typical behaviour of allophane clays, with a high residual strength not far below the peak value. 2.3 Discussion The difficulty of measuring the critical state parameter 4‘cv for clays from conventional triaxial tests is clearly a strong pragmatic reason for not using it in design. However this is not just an issue of pragmatism; the difficulty of creating the critical state in clays suggests that it is not a useful theoretical concept for such materials. If the material fails in triaxial tests by passing from the peak strength progressively towards the residual strength then the same can be expected in field situations. There are thus both pragmatic and theoretical reasons for not using 4’cv
At the same time, it should be noted that the difference between the peak strength and that at large strains is not great so the question of which of these two parameters to use is not really a major issue. A
497
An argument sometimes raised against the use of clay fill is the likelihood that it will lead to excessive deformations. A theoretical study of this issue has been undertaken using finite element analysis of a hypothetical wall. The wall was 6m high with typical properties and spacing of the geogrid reinforcement. The analysis was carried out for both a “soft” facing with similar modulus to the soil, and a stiff facing with modulus similar to that of concrete. In each case a range of soil modulus values was investigated. The results are summarised in Figures 3 and 4. For clay fill the Young’s Modulus (E,) is likely to be in the range of 10 to 20 MPa, while for a well compacted granular fill it is likely to be in the range of 30 to 50 MPa, or possibly higher. Figure 4 (soft facing) shows that the deformation with clay fill is likely to be about double that for a granular fill. Figure 5 (stiff facing) suggests that the clay fill may result in over twice the deformation for a granular fill. However with the stiff facing the dcformations are so small that they are unlikely to be of any consequence. Even with the soft facing the deformations arc small and would only be a cause of concern in special situations. The reinforced walls currently popular in New Zealand are being built using segmental block facings. The most common type of blocks are known as “Keystone” blocks. Facings of this type will have high vertical stiffness but low bending stiffness, so that the deformations to be expected will be between those shown in Figs 3 and 4. A number of R.E. walls using clay fill and geogrid reinforcement have been built in New Zealand in recent years. The clay has been of residual origin,
4 DOWNDRAG ON FACING ELEMENTS
including both volcanic and sedimentary (siltstone/sandstone) parent rock. Wall heights have been from 4m to about 12m. As far as the author is aware, the walls are performing satisfactorily, and deformations have been within expected, acceptable limits.
The author became interested in this issue because of observations made of a R.E. wall recently constructed in New Zealand using clay fill and "Keystone" segmental facing. Near the base of the wall there are cracks in a number of the Keystone blocks. These are not affecting the performance of the wall, but they are clearly undesirable, if only for aesthetic reasons. The wall is about 12m high, and built of the clay from which Sample 1 (see Table 1) was taken. The different stiffness of the clay fill and the concrete facing means that the facing will attract higher stresses than its gravity weight and the soil in the immediate vicinity of the wall will have less stress than its gravity weight. This effect has also been investigated as part of the finite element study. A wall facing 20cm thick with an E value typical of concrete (25,000 MPa) was adopted. Soil modulus values of 10 and 100 MPa, were selected, representative of a fairly plastic clay fill, and a very high quality, dense, granular fill. The results of the analysis are summarised in Figures 5 and 6. Figure 5 shows the results as plots of vertical stress over the height of the wall at three locations, namely in the wall itself, lOcm and 10m from the wall. This illustrates very clearly the way in which the wall attracts stresses much higher than its gravity weight. The "gravity" stress at the base of the wall should be about 150 kPa; instead it is over 800 kPa. Immediately adjacent to the wall the stress is almost zero. and 10m away it is the gravity value (6m of soil at 20 kN/m3 = 120 kPa). The soil immediately adjacent to the wall is thus largely supported by the geogrids and its weight is transferred to the concrete facing. Figure 6 shows the vertical stresses on horizontal planes; curves are given for the base and mid-height of the wall. This again illustrates the way in which the wall attracts high stress and the soil immediately adjacent to the wall has a very low stress. The stress has virtually reached the gravity weight value at a distance of 4m. This was the length of the geogrids assumed for the analysis, although the results in Figures 5 and 6 are probably not greatly influenced by the length of the reinforcement. It should be noted also that the results are not much influenced by soil modulus values from 10 to 1OOMPa; it is only when the modulus approaches that of concrete that the picture presented in Figures 5 and 6 will start to change. Figure 5 shows a maximum vertical stress in the facing of about 800 kPa. This is for a wall of height 6m. For a wall of the height in which the cracking was observed (12m) the stress would be expected to be about double this figure. Although this is many times greater than the stress due to the gravity weight of the blocks, it is only a small fraction of the expected crushing strength of the Keystone blocks,
Figure 4. Influence of soil modulus on deflection,for a stiff (concrete) facing.
498
is a danger of significant pore pressures developing during construction. A research programme is currently underway to investigate this situation and hopefully to provide guidelines to local engineers on how to avoid such an occurrence. Because of the variability of local residual soils, earth works are normally controlled by means of undrained shear strength (S,) and air voids (av), (Pickens, 1980). A lower limit on Su, usually about 15OkPa, prevents the soil being placed too wet, and an upper limit on a,, usually about 6%, prevents the soil being placed too dry. The concept is illustrated in Figure 7. It is generally believed that within these limits, the likelihood of significant pore pressures developing during construction is small, but no systematic studies have been undertaken to verify this. Current research is aimed at rectifying this situation. Samples are being prepared at a range of water contents; they are then compacted at varying densities to give samples with differing air voids and shear strength. Pore pressure response is then measured in a triaxial cell. Figure 8 shows typical results for one particular water content.
Figure 6. Vertical stresses at the base and mid-height of the wall.
which is in the vicinity of 20 MPa. Hence the additional stress in itself is not sufficient to account for the cracking observed in the blocks. It is probable that differential settlement due to non-uniform foundation conditions has induced bending stresses along the wall, and that a combination of these stresses and the additional vertical load has caused the cracking. 5 POTENTIAL FOR PORE PRESSURE DEVELOPMENT DURING CONSTRUCTION Most clays in New Zealand exist in their natural state at water contents considerably higher than their optimum water content. This means that unless they can be dried significantly prior to compaction there
Figure 8 Influence ofalr vods on pore pressure
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6 CONCLUSION None of the factors investigated and discussed in this paper present any real obstacles to the use of clay fill in geogrid reinforced earth walls. Only the need to keep deformations to very tight tolerances, or inability to dry the clay to an appropriate water content, could rule out the use of clay. REFERENCES
Figure 9. Pore pressure parameter B related to undrained shear strength and air voids.
The expected relationship between the pore pressure parameter B and the air voids and shear strength is shown conceptually in Figure 9. At the time of writing this paper insufficient data is available to establish the precise form the relationship should take, and whether it is independent of soil type.
500
Leshchinsky, D. 2000. Design dilemma: use of peak or residual strength of soil. Geotextiles and Geomembranes, April, 2000. Netlon Limited 1996. The design of reinforced soil structures using Tensar Geogrids. Netlon publication dated March, 1996. Pickens, G.A. I980 Alternative compaction specifications for non-uniform fill materials. 3Id Australia - New Zealand Conf: on Geomechanics, Perth, Vol. 1 23 1-2 15. Zornberg, J.G., NSitar & J.K.Mitchell 1989 Performance of geosynthetic reinforced slopes at failure. Journal of Geotechnical and Geoenvironmental Eng., ASCE I24(8), 670683. Zornberg, J.G., N.Sitar & J.K.Mitchell2000 Closure of discussion on the above paper. Journal of Geotechnical and GeoenvironnientalEng., ASCE 126(3), 285-286. Wesley, L.D. & C.Davidson 2000. Selection of soil strength parameters for geogrid reinforced walls. Proc. of the 2''d Asian Geosynihetics Conf: Kuala Lumpur, 2000, VoI.2 1-6.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lSBN 90 2651 863 3
Evaluation of confining effect in geogrid-reinforced retaining wall based on two-dimensional model test N. Yasufuku, H. Ochiai, Y. Ninomiya, K. Omine & M. Nakashima Department of Civil Engineering, Kyushu University, Fukuoka, Japan
T. Kawamura Department of Civil Engineering, Shinshu University, Fukuoka, Japan
ABSTRACT: Two-dimensional small model tests for simulating the geogrid-reinforced retaining wall have been performed to clarify the mobilized confining effects, comparing with the tensile ones in the model retaining wall. The reinforced effects are estimated by the changes of the active earth pressure in the retaining wall without and with geogrid-reinforcement. The characteristics of the resulting confining effect are investigated paying attention to the space of geogrid and the unit length of geogrid, which is prepared by cutting the reinforced material where the whole length is always same. 1 INTRODUCTION In current practical problems, the reinforcing effects, which are applied to the stability analysis of a geogrid-reinforced structure, are generally evaluated based on the tensile force of geogrid. Fukuda et al. ( 1 986) reported, based on the in-situ measurements, that the tensile force of a geogrid, which should be mobilized for the stability of a structure, was not mobilized in soil, although the structure maintained sufficient stability. Tatsuoka et al. (1996) reported that a strong earthquake inflicted little damage on a reinforced structure. These studies suggest the existence of an additional reinforcing effect other than the tensile effect due to tensile force of a geogrid. In a previous study, the reinforcing effects of geogridreinforced soil were experimentally examined. As an important result, the existence of an additional reinforcing effect other than the tensile effect in laboratory tests were confirmed, and the additional effect were defined as the confining effect (Ochiai et al., 1996, 1998; Kawarnura et al., 2000). The reinforced effects of geogrid-reinforced soil are generally evaluated by the tensile effect alone due to the tensile force of a geogrid. However, it has been found from a series of laboratory testing that the confining effect mobilized in the reinforced soil mass exists, which is independent on the tensile force of a geogrid. The quantitative evaluation is requested to apply the confining effects to the general geogrid-reinforced soil structure. In this study, two-dimensional small model tests for simulating the geogrid-reinforced retaining wall have been performed to clarify the mobilized confining effects, comparing with the tensile ones in the model retaining wall. The reinforced effects are esti-
mated by the changes of the active earth pressure in the retaining wall without and with the geogridreinforcement. The changes in resulting confining effect related to the active earth pressure are investigated paying attention to the space between two reinforced materials and the unit length of a piece of reinforced material. Based on the experimental data, the importance to introduce the confining effect into the design procedure in geogrid-reinforced retaining wall is discussed. 2 CONFINING EFFECT IN GEOGRlD REINFORCED SOIL The reinforced effccts of geogrid-reinforced soil are often evaluated by only the tensile effect due to the tensile force of a geogrid. Ochiai et al., (1996, 1998) carried out a series of the special shear tests to clarify the reinforced effects on the sliding plane of the geogrid-reinforced soil mass. It was found that the confining effect that is independent on the tensile effect exists, and as schematically shown in Figure I , when simply introducing the confining effect into a design method, the effect should be evaluated as the increment of the internal friction angle or the apparent increment of the normal stress as follows: PI? (sin&,,$+ AS
s = c +-
cos 6)+(1 + &,tan$
(1)
where s is the shere strength of reinforced soil, PR is the mobilized tensile force of the reinforcement, As is the area of the sliding plane, 4 is the internal friction angle of soil, 0 is the angle between the reinforcement and the sliding plane, on is the normal stress on the sliding plane and p is introduced to evaluate the degree of the confining effect. The con-
50 1
Figure 1. Relationship between s and tensile and confining effects.
G,,considering
both of
fining effect is considered to be the effect of the restriction of the soil by the geogrid, and, as a result, the confining stresses around the geogrid apparently increase. In the following, the characteristics of the confining effects are experimentally investigated through the behaviour of the active earth pressure in the geogrid-reinforced retaining wall, paying attention to the vertical space in each layer of geogrid and the type of geogrid, that is, the difference between sheet type and grid type.
horizontal deformation and force acting on the wall during the rotation are measured by the dial gage and load cell, respectively, which are mounted to the loading rod. In this study, two types of plastic materials shown in Figure 3 are used as the modeled reinforced materials, in which one is sheet type plastic film which is quite smooth named OHP film and the other is grid type plastic material. The length of each material horizontally installed in the specimen was chosen as 25cm in length. Two-dimensional small model tests for simulating the geogrid-reinforced retaining wall have been performed to clarify the mobilized confining effects, comparing with the tensile ones in the model retaining wall. The reinforced effects are qualified by the changes of the earth pressure in the retaining wall without and with reinforced materials, which are easily calculated by the measured horizontal force. The changes in resulting confining effect are investigated, paying attention to the vertical space of each layer of reinforced materials and the unit length of the reinforced material, which is prepared by cutting. However, even though the reinforced material with a fixed length is separated to two pieces, three pieces, five pieces and ten pieces by cutting, the total length is always kept as 25cm. It is noted that all the reinforced materials are horizontally installed between the top of the specimen and
3 TEST APPARATUS AND TEST PROCEDURE Schematic view of the two-dimensional small model test apparatus used in this study is shown in Figure 2, for simulating the geogrid-reinforced retaining wall. The length and effective height of the model ground are 50cni and 3Ocm. respectively. The vertical wall, which is supported by a loading rod with a frictionless roller and a load cell, can be rotated to the left hand-side in relation to the hinge joint. The motor through the loading rod controls the speed of the movement of the wall. The specimen is made of two types of aluminum rods, in which the diameters are 1.6mm and 3.0mm with the length of 50mm, where both rods are mixed and the three-fifth of them per weight is in the 1.61nm aluminuin rod. The
Figure 2. Schematic view of test apparatus.
Figure 3 Two types of reinforced materials used, Upper sheet type reinforced material (OHP film), Lower grid type reinforced material
Figure 4. Schematic view of splitted reinforced material
5 02
the position of the hinge joint, in which the vertical space of each layer of reinforced materials was always kept constant. Figure 4 shows the schematic view of the reinforced material splitted by cutting The sheet type geogrids without and with cutting are shown in Figs.4(a) to (c), respectively, where AL is the unit length of a piece of reinforced material. Each piece of reinforced material is put one upon another with lubricant grease as shown in Fig.Lt(c) to keep the total length of 25cm.
4 EFFECT OF SPLITS OF GEOGRIDS ON THE REINFORCED EFFECT Figure 5 shows the typical test results of the horizontal earth pressure against the angle of retaining wall, 8, in Fig.2 in which the wall is counterclockwise rotated by the movement of the loading rod. The results of horizontally laying the grid type reinforced material in five layers with same vertical space in each layer are shown in this figure, together with the result without reinforced material. Note that the difference in the results with reinforced material is in the length of a piece of the splitted one. It is clear that 1) the horizontal earth pressure, PM, for all the cases gradually decreases with the decreasing wall gradient, 8, and then converge a certain value in each case, and 2) the magnitude of the converged PM increases as AL becomes smaller, that is, the number of the splitted reinforced material becomes greater. This means that the reinforced effect decreases with the decreasing AL. One of the main reasons is considered to be due to the reduction of the tensile force mobilized in reinforced material. Figures 6(a) and (b) show the results of the sheet type geogrid laid in five layers and the results of the model ground with three layers of grid type reinforced material, respectively, which have similar tendency with that in Fig.5, in spite of the different number of the layers laid in the model ground and the different type of reinforced material.
Figure 5. Effect of splitting of grid type reinforced material in 5 layers on horizontal earth pressure against wall gradient.
503
Figure 6. Effect of splitting of reinforced material on horizontal earth pressure against wall gradient, a) sheet type material in 5 layers, b) grid type materials in 3 layers.
5 PARAMETERS TO EVALUATE THE CONFINING EFFECTS Figure 7(a) shows the schematic diagram in Figs.5 and 6 to explain the definition of the parameters PMO and APM.PMOdefines the mobilized horizontal earth pressure in active condition in the case of nonreinforced soil acting on the wall, which can be seen that PMO is achieved at quite small movement of the wall as shown in Figs.5 and 6. APMis also defined as the difference between PMOand the horizontal earth pressure in the case of reinforced soil mobilized in the active condition. When the reinforced effect becomes greater, APM tends to become larger. In this study, the ratio APM/PMois used as a parameter to express the degree of the reinforced effect. Figure 7(b) shows the definition of the parameters, H, Ah and L. The height of the model wall is given by parameter H, and H is 30cm in the model used. Ah expresses the vertical space in each layer of reinforced material, in which Ah in each layer is set to keep constant as shown in Fig.7(b). The parameter being normalized space, Ah/H, is used to express the laying vertical space in each layer. For example, when laying the reinforced material in five layers,
25cm is splitted into infinite number of pieces. It can be seen from this figure that AP~/PMo, reflecting the reinforced effect, gradually decreases with the decreasing AL/L, and then tends to converge to a certain value when AL/L reaches about 0.2, in other words, when splitting the reinforced material into around 10 pieces. In this case, it is clear that the conis about 0.2. The decrease of verge value of A PM/PMO this reinforced effect is considered to be caused by the decrease of the mobilized tensile force in reinforced material by splitting. When AL/L becomes less than about 0.2, the tensile force acting on the reinforced material seems to be totally small and thus in this situation, the reinforced effect caused by the tensile force approach to zero. Nevertheless, as shown in this figure, the reinforced effect still exists, which is represented by the converge values of APM/PMo.In this study, such reinforced effect calls the confining one as the reinforced effect excluding the effect mobilized by the tensile force. Based on the result in Figure 8, the confining effect is expressed as the value of AP&’Mo at AL/L=O, which is approximately determined as 0.2 by the extrapolation from the result in the APM/PMo-AL/Lrelationship. It is important to point out that the similar tendency can be obtained in the cases of the different layers of reinforced material and the different type of reinforced material as shown in Figures 9(a) and (b), although the converge values of APM/PM~ reflecting the degree of the confining effect differs from each other.
Figure 7. Definition of parameters to evaluate the confining effect.
Ah/H equals to 0.16 in this model-reinforced wall. Further, the parameter L is given by the length of a side in the sliding wedge derived when assuming the mechanism of Rankin active earth pressure in the non-reinforced ground. The defined length L is easily calculated by the height of the wall, H, and the averaged friction angle of the ground, I$‘,such that: U
7 CONFINING EFFECTS RELATED TO THE SPACE OF EACH LAYER OF REINFORCED MATERIALS
In this study, normalized length AL/L is used to express the degree of the splitting of the reinforced material, where AL is defined as a length of a piece of the reinforced material after splitting as mentioned above (see Fig.4).
Based on the results in Figures 8 and 9, the relationship between APM/PM~ at AL/L=O as the confining effect and the normalized space, Ah/H can be depicted. Figure 10 shows the results in both types of
6 CHARACTERISTICS OF CONFINING EFFECT Based on the results in Figures 6 and 7, the relationship between the reinforced effect, APMIPMo,and the normalized length of a piece of reinforced material, AL /L, is shown in Figure 8, where this is the case which horizontally lays the grid type reinforced material in five layers with same vertical space in each layer, that is, Ah/H equals to 0.16. Note that AL/L in horizontal axis is defined as the ratio of the unit length of reinforced material AL to the length I, given by Eq. (2). For instance, AL/L=O in this figure means that the reinforced material with the length of
Figure 8. Relationship between reinforced effect and normalized unit length of reinforcement material in the case of 5 layers.
5 04
Figure 11. Ratio of confining effect to total reinforced one against normalized space in each layer.
are roughly 19 and 1 1 degrees, respectively, and thus the greater the friction angle becomes, the relatively lager the confining effect becomes. The difference in the confining effect caused by the sort of reinforced materials, however, becomes gradually smaller with the decreasing the space of each layer. There seems to be a reason that the effect restraining the movement of thc aluminum rod mass is surpassed, irrespective of the types Of reinforced materials, when the distance among the reinforced materials becomes relatively small. Figure 1 1 shows the ratio of the confining effect to the total reinforced one against the normalized vertical space in each layer of reinforced materials, Ah/H. The total reinforced effect including the tensile force one is given by APM/PMoin the case without splitting the reinforced materials. It is clear from this figure that the ratio of the confining effect to the total reinforced one is in the range from 20% to 35% depending on the types of the reinforced materials. In order to make clear whether such tendency is realistic or not, the more practical experiments and the considerations will be still needed.
Figure 9 Relationship between reinforced effect and normalized unit length of reinforcement materials, a) case of sheet type material in 5 layers, b) case of grid type material in 3 layers
8 CONCLUSIONS
Figure 10. Relationship between confining effect and normalized space in each layer of sheet type and grid type reinforced materials.
Two-dimensional small model tests for simulating the geogrid-reinforced retaining wall using the aluminum rods were performed to clarify the mobilized reinforced effect in relation to the active earth pressure. The following main conclusions are obtained as follows:
reinforced materials. It is found that when the vertical space of a layer decreases from about 0.5 to 0.08, the corresponding confining effect increases from about 0.05 to 0.25. The confining effect in both type of reinforced materials differ from each other, especially at the region in relatively larger vertical space. Main reason of such phenomenon is considered to be in the difference of the frictional angle between aluminum rod mass and the reinforced materials. It may be said in this connection that the frictional angles in grid type and sheet type reinforced materials
1 ) The horizontal earth pressure for all the tests cases gradually decreases with the decreasing wall gradient and then converge a certain value depending on the type of reinforced materials, the number of the layers and the unit length of a piece of reinforced material laid in the ground. 505
2) The total reinforced effect, which is expressed as the difference between the horizontal active earth pressure in non-reinforced case and that in reinforced case, decreases with the decreasing unit length of reinforced material by splitting even though the whole length keeps constant. 3 ) The reinforced effect still exists even though each reinforced material with a certain length is splitted into infinite number of pieces. This reinforced effect is considered to be the confining effect due to reinforcement. 4)The confining effect tends to increase with the decreasing space of reinforced material laid in the model ground and the effect is expressed as a function of the normalized space of reinforced material, irrespective of a sort of reinforced materials when the normalized space becomes smaller than a certain value. 5) In this model retaining wall, the ratio of the confining effect related to the active earth pressure to the total reinforced effect is in the range from 20% to 35% depending on the types of the rein-
forced materials and the vertical space in each layer of geogrid.
REFFERENCES Fukuda, N , Yamanouchi, T & Miura, N 1986 Comparative studies of design and construction of a steep reinforced embankment, Geotextiles and Geomembranes, Vol 4 296284 Kawamura, T , Ochiai, H , Yasufuku, N & Hirai, T 2000 Confining effect of geogrid-reinforced soil Introduction into design method, Proc 2"d European Geosynthetics Conference 185-190 Ochiai, H , Yasufuku, N , Yamaji, T , Guang-Li Xu & Hirai, T 1996 Experimental evaluation of reinforcement in geogridsoil structure, Proc of Int Symp on Earth Reinforcement (IS Kyushu 96), Vol 1 249-254 Ochiai, H , Yasufuku, N Kawamura, T ,Hirai, T & Yamaji, T 1998 Effect of end-restraint in geogrid-soil structures, Proc of 6"' Int Conf on Geosynthetics 545-550 Tatsuoka, F , Koseki, J & Tateyama, M 1996 Performance of reinforced soil structures during the 1995 Hyogo-ken Nanbu Earthquake, Proc of Int Symp on Earth Reinforcement (IS Kyushu 96), Vol 1 973-1008
5 06
Landmarks in Earth Reinforcement, Ochiai et al. (eds), @ 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Applicability of an elasto-plastic model for reinforced soil structures T. Yoshida & K. Mori The Kansai Electric Power Co. Ltd, Japan
A. Iizuka Kobe University, Japan
E Maegawa & S. Amano Newjec Inc., Japan
ABSTRACT: The reinforcement technique using flexible materials such as geosynthetics has been widely employed in the current engineering practice. However, the reinforcement mechanism has not revealed sufficiently, i.e. the mechanical interaction between the reinforcements and soils, the confining effect induced by reinforcements and so forth, and then it has not yet led into a proposal of rational design procedure. This paper reports in-situ loading tests carried out on the compacted soil structure reinforced by flexible beltlike materials. Elastoplastic finite element simulations were also performed considering dilatancy characteristics of compacted soils during shearing. Throughout comparison with the field records monitored in the site and the results obtained from the conventional elastic computation, discussed are the applicability of the elasto-plastic analysis and the role of dilatancy characteristics of compacted soils in the reinforcement mechanism.
1 INTRODUCTION
2 ANALYSIS TECHNIQUE
The reinforcement technique for the soil structures has been widely employed. However, the analysis technique has not established yet, and it is expected to establish the rational design method. Stability of the reinforced soil structures is thought to be improved through complex mechanical interaction between soils and reinforcement materials. Dilatancy characteristics would play an important role in the interaction. Therefore, in order to simulate the behavior of the reinforced soil structure, an elastoplastic model proposed by Sekiguchi-Ohta (Sekiguchi and Ohta, 1977) was employed, in which dilatancy characteristics is taken into consideration. The model was originally proposed for saturated soil materials. But, in this study, an attempt was made to apply the model to compacted soils by introducing an idea of equivalent over-consolidation ratio (Hirata, M et al. 1999). And the behavior of a reinforced soil structure backfilled with compacted soil was simulated. The idea introduced is based on the similarity of dilatancy characteristics during shearing between compacted soils and heavily overconsolidated clays. Throughout comparison with the monitored records when a giant trailer passed on the reinforced soil structure, applicability of the model was discussed in this paper. Moreover, conventional elastic simulations were also carried out to examine the reinforcement effect. The effect is expected to be brought by confining the dilatancy characteristics of soils by reinforcement materials.
2.1
Elasto-plastic model
Numerical analysis was carried out by the “DACSAR’, in which an elasto-plastic model proposed by Sekiguchi-Ohta is incorporated. Basically the model was constructed to analyze saturated normally consolidated clayey materials and based on the following principles.
1) Volumetric strain of normally consolidated clay is given by linearly summing up volumetric changes due to consolidation and dilatancy. 2) Volumetric strain of normal-consolidated clay is determined by initial and present effective stress states, and independent of stress path. The yield function f is expressed as follows, A-U pf In - Dy* 1 +eo PO where D is dilatancy coefficient proposed by Shibata (1963). q* is normalized stress ratio expressed as,
f =---
+
where sij is deviatoric stress tensor (sij = crij -p f & j ) . By introducing normalized stress ratio y*, it is possible to consider anisotropic shear properties induced
507
Figure 1. Process to evaluate equivalent OCR. Figure 2. Cross section of the retaining structure.
by rotation of principal stresses. Then, the stress and strain relationship is expressed by employing the associated flow rule. 2.2 Application for the compacted soil While Sekiguchi and Ohta’s model has been developed for naturally consolidated-saturated clays, materials that are dealt with here are artificially compacted and unsaturated soils. However, since the unsaturated compacted soils, in general, show positive dilatancy (dilatation) and strain softening behavior like overconsolidated clay under shearing, we may consider the over-consolidated clay equivalent to the compacted soil and apply Sekiguchi and Ohta’s model to it if the mechanical property of the compacted soil can be characterized from viewpoint of over-consolidated saturated clays (see, Ohta and Hata, 1977, Ohta et al., 1978). Characteristics of strain softening, strength and stiffness depend on the value of OCR. Then, it is assumed that the degree of compaction can be evaluated by the equivalent overconsolidation rate OCR. Figure 1 shows the process to determine the equivalent OCR. When void ratio and vertical stress of insitu soil is evaluated e f and o f , respectively. The soil is assumed to be in the present stress state after overconsolidated to predecessor stress 00 Then equivalent OCR is evaluated as ao/os. 3 TESTSITE The site, where a series of in-situ monitoring were carried out, was the reinforced retaining wall structure as shown in Figure 2. It was reinforced by Terre Armee method, flexible beltlike materials (steel strip) and concrete plates (skin plate) were used as facing of the wall structure. It is a fairly flexible structure if comparing with the conventional gravitytype retaining wall structure. The merits of Terre Armee method would be summarized as, 1) not require larger work volume, 2) not need highly skilled technique, 3) not need longer executive period and 4) more applicable to a narrower site. The retaining retaining structure was 8.75 m high, and 11 steel strips were installed in 508
each cross section. It was backfilled with compacted soils that were compound made up of clay, sand and gravel. After completion of construction, the mechanical performance of reinforced soil structure was investigated by passing a giant trailer on the structure (Nishikata U. et al. 1999 & 1998). In order to monitor the tension stress (strain) acting to strips when the giant trailer was passed through on the structure, strain gauges were installed on the strips at a certain interval as indicated in Figure 2. Vertical load of 67kN/m2 was applied to the structure and the maximum force acting to the strips reached 25 kN when a giant trailer was passed on the structure.
4
INPUT PARAMETERS
The soils were sampled from the site and a series of constant volume shear box tests (CV-SBT test) were carried. First, very loose disturbed soil specimens, which were prepared at the in-situ water content (8%) were consolidated in the shear box by the vertical pressure of 78.4, 156.8, 313 and 616 kPa and then sheared under the constant volume after completion of the consolidation. Thus obtained results of consolidation process were potted on void ratio e and vertical stress log 0’ diagram and then the virgin compression line was drawn as shown in Figure 1. The compression index h was determined from the gradient of virgin compression line. And the effective stress paths during shear were obtained from the shear process of the test and plotted in Figure 3. Next, the undisturbed soil sample of which natural water contents was around 8% was subjected to the consolidation (oL0 = 156.8 and 616 kPa) and shear tests in the same manner. The effective stress path obtained from shear process of the test was also plotted in Figure 4. Input parameters needed in the analysis (Table 1) were determined as follows. The increase ratio of undrained shear strength S,, /oLo was determined from shear test results (Figure 3). On the other hand, Su/oL0 has been theoretically derived from the elasto-plastic constitutive
Table 2. Input parameters. h K
D
M A U’ 0
I
0
3
U
x
,
3
0
3
4
effective normal stress,
0 U
[
)
~
6
0
0
7
~
KO
(Fa)
e0
OCR
Figure 3. Effective stress paths (loose & disturbed samples).
0.159 0.014 0.058 1.978 0.912 0.309 0.448 0.253 47(mean value)
shear were calculated by the elasto-plastic constitutive model and compared with experimental values. Calculated (theoretical) stress paths are drawn and indicated by solid lines in Figures 3 and 4.
5 103
0
2co
303
4co
effective normal stress,
503 U
600
(@a)
Figure 4. Effective stress paths (undisturbed samples).
model by Sekiguchi and Ohta as (see, Ohta, Nishihara and Morita 1985), S,,
- --
0:o
1
+ 2Ko A4 exp(-A)
(3)
343
where S, and oA0. are undrained shear strength and vertical preconsolidation stress, respectively. Moreover, when the following empirical relations are introduced,
M = 1.75A (Karube) KO= 0.44
+
(4)
PI 0.42 . - (Massarsch) 100
sin$’ = 0.81 - 0.233 . log P I
(Kenny)
(5) (6)
three unknown parameters M , A and KOcan be specified. Thus determined input parameters are summarized in Table 2 together with other input parameters, V’ = Ko/l KO. In order to confirm validity of input parameters, then the effective stress paths under constant volume
Elasto-plastic finite element simulation was carried out. Finite element model is shown in Figure 5. In the simulation, strips, skin plates and rock mass were assumed to be elastic materials. Uniformly distributed load of 67kN/m2 by the giant trailer was applied as the surcharge load on the surface as shown in Figure 5. Figure 6 shows distributions of tension force working to the strips Sl-S5, when the giant trailer just passed on the monitoring cross section (when load of 67kN/m2 was just applied). In the figure, the distributions of tension force obtained not only from the elasto-plastic simulation but also from the linearly elastic simulation are compared with monitored values. It seems that the elasto-plastic simulation relatively explains monitored values, except for strip S2 where monitored stresses were quite small comparing with those of other strips. The tension force calculated from conventional linearly elastic simulation is fairly larger than those obtained from the field monitoring.
+
Table 1. Input parameters. h K
D M A KO e0
V’ OCR
FINITE ELEMENT SIMULATION
700
= 0.434Cc Cc; compression index = 0.434Cs Cs; swelling index = hA/M(l eo); dilatancy coefficient = 6 sin @ / ( 3 - sin 4’); critical state parameter = I - K / A ; irreversibility ratio coefficient of static earth pressure at rest at the completion of virgin consolidation initial void ratio effective Poisson’s ratio over-consolidationratio
+
509
soils to explain the reinforcement effect of an actual retaining wall structure backfilled with compacted soils. Throughout comparison with the field records, applicability of the model was examined. The following conclusions would be pointed out. The way that an elasto-plastic constitutive model for saturated clays is applied to unsaturated compacted soils is shown. Shear behaviors of compacted soils can be modeled into heavily over-consolidated clays by introducing “equivalent OCR”. Dilatancy characteristics play an important role in reinforcement mechanism of the soil structures. REFERENCES
6 DISCUSSION Tension force working to the strips is examined. The elasto-plastic simulation in which dilatancy characteristics is taken into account predicts fairly smaller value of tension force than conventional linearly elastic simulation and results in being consistent with monitored values. This suggests that it is important to consider dilatancy characteristics of compacted soils and the reinforcement effect is brought by preventing the dilation of soils due to shearing. For establishment of rational design procedure of reinforcement of soil structures, dilatancy characteristics of compacted soils is not negligible.
7 CONCLUSION An attempt was made to apply an elasto-plastic model considering dilatancy characteristics of compacted
510
Hirata, M. et al. (1999): The numerical simulatioiz of geosyizthetics reinforced soil structures using elasto-plastic dilatancy models. Journal of Geotechnical Engineering of JSCE, 11148. 179-192 (in Japanese). Karube, D. ( 1975): Unstandardized trinxial testing procedures and related subjects for inquiry, Proc. 20th Symposium on Geotechnical Engineering, 45-60 (in Japanese). Kenny, T. C. (1959): Discussion on Proc. Paper 1732 (Wu, 19-58),Proc. ASCE, Vol. 85, SM3, pp. 67-79. Massarsch, K. R. (1979): Lateral earth pressure in normally consolidated clay, Design Parameters in Geotechnical Engineering, Proc. 7th European Conference of Soil Mechanics and Foundation Engineering, Vol. 2, pp. 245-249. Nishikata U. et al. ( 1 997): A result of field measurements at special loud test by Terre Armee method. Proceedings of the 52th Annual Conference of JSCE, 235-236 (in Japanese). Nishikata. U et al. (1998): Field measurements ofsoil structure reinforced by Terre Armee method during special load test. Proceeding of the 33th Japan National Conference on Geotechnical Engineering, JGS. 578-579 (in Japanese). Ohta, H. and Hata, S. (1977): Strength of dynamically compacted soils, Proc. 9th ICSMFE, Tokyo, Vol. 1 , 239-242. Ohta, H., Hiura, Y., and Kuniyasu, I. (1978): Strength mobilized in compacted earth structures at failure, Proc. Symposium on Soil Reinforcing and Stabilizing Techniques in Engineering Practice, Sydney, 385-401. Ohta, H., Nishihara, A. and Morita, Y. (1985): Undrained stabilit): of KO-consolidated clays, Proc. 1 lth ICSMFE, San Francisco, Vo 1. 2 , 6 13-6 16. Sekiguchi, H. and Ohta, H. ( 1 977): induced anisotropy and time dependency in clays, Proc. 9th ICSMFE, Specialty Session 9, Tokyo, 229-237. Shibata, T. (1963): On the volume changes of normally consolidated clays, Annuals, Disaster Prevention Research Institute, Kyoto University, No. 6, 128-134, (in Japanese).
4 Foundations
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Landmarks in Earth Reinforcement, Ochiai ef al. (eds), 0 2001 Swefs & Zeiflinger, lSBN 90 2651 863 3
The settlements of a continuous foundation footing resting on the geogrid-reinforced sand layer J. Adamczyk P.B. P. “GEOSTAFF”, Warsaw, Poland
T. Adamczyk Warsaw University of Technology, Environmental Engineering Faculty, Poland
ABSTRACT: The paper presents the values of settlements for building founded on the geogrid-reinforced sand layer. The settlements of the continuous foundation were also determined at the similar load for the following conditions direct foundation on the silty loam subsoil, foundation resting on the non reinforced soil layer and foundation with several layers of geonet. The foundation resting on the soil layer shows the settlements several time lower when compared with the foundation resting directly on the silty loam subsoil. 1 INTRODUCTION
Table 1. Material parameters
1. I Foundation conditions
The subsoil of foundation Silty loam Light compacted sand Mediumcompacted sand Layer reinforced
During modernization of the industrial plant the necessity has arisen to apply the continuous foundation footing directly into the silty loam subsoil. Existing machinery, equipment and installations made another method of walls foundation of the modernized object impossible. The width of the foundation footing was limited to 120 cm. 1.2 Types of foundationfooting taken into consideration
v
Y
02 026
[kN/m’] 20 175
[kPa] 17 0
14 28
80
026
18
0
28
150
025
142
[“I
3 Q A -32
(Dluzewski, 1997). The method of finite elements was used in relation to deflections description. The elasto-plastic model of the subsoil was adapted based on Culomb-Mohr plasticity conditions. The independent plastic flow law were applied assuming that materials within a plastic range of stress are un-compressible. In all numerical analyses sixnoded triangular isoparametric elements were used of second stage shape functions. The settlement calculations were carried out for the foundation footing of the width 120 cm founded as follows:
The following methods of continuous foundation footing were taken under consideration:
-
c 4 )
E [MPaI 15 45
directly on silty loam subsoil,
- on the sand cushion, - on the sand cushion reinforced with 4 layers of geonet. The dimensions of the foundation footing crosssection: 120 cm x 40 cm. The dimensions of the reinforcing geonet layers: width 360 cm, spacing 30 cm, thickness 0.26 cm. The loads applied to the footing: p = 0.1 - 0.5 MPa. For material parameters, see Table 1.
a) b) c) d)
directly on silty loam, on sand cushion of modulus E = 45 MPa on sand cushion of modulus E = 80 MPa on sand cushion of modulus E = 45 MPa reinforced with 3 layers of geonet e) on sand cushion of modulus E = 80 MPa reinforced with 3 layers of geonet f) on sand cushion of modulus E = 45 MPa reinforced with 4 layers of geonet g) on sand cushion of modulus E = 80 MPa reinforced with 4 layers of geonet.
2 SETTLEMENT OF FOUNDATION FOOTING 2.1 Numerical model The calculations were peformed assuming the planar deflection concept using HYDRO-GEO software 513
Figure 5. Settlement of a foundation footing resting on unreinforced sand cushion.
The settlements were calculated for the foundation resting on sand cushion of 150 cm in thickness and 1 190 cm in width. To variable densities of the cushion were taken into consideration. 2.4 Foundation resting on reinforced sand cushion
514
The results of settlements for different types of footing foundations are given in tables.
of geonet vertically spaced every 30 cm. The sand incorporated in the cushion will possess the modulus of elasticity above E 2 80 MPa. At this method of foundation the footing may transmit the load p = 0.1 MPa. Notwithstanding the application of sand cushion, the weak layer of silty loam underlying the cushion possess significant influence on a settlement values. The computations completed show, that maximum effect of application the sand cushion exists when loads on foundation footing are p > 0.30 MPa. At these loads the difference between settlements calculated for the reinforced and un-reinforced cushion reaches few centimetres. At lower loads, those differences in settlements are ranging from 2 to 3 cm. A significant effect of the cushion application has the compressibility of the underlying soil layer. The silty loam existing beneath the designed footing made impossible to aplly the loads above 0.1 MPa. The calculations completed show, that insufficient compaction of sand in a cushion causes its additional settlement of about few millimetres, at the loads below p < 0.3 MPa and of few centimetres at loads above 0.3 MPa. The compaction of sand in the cushion has higher influence on settlements values of the un-reinforced cushion when compared to one reinforced with geonet.
Table 2. Settlements of foundation resting directly on the subsoil. Load IMPaI 01 02 03 04 05
Settlement s [cml -2 2 -8 5 -26 1 -84 9 -130 7
Table 3. Settlements of foundation resting on sand cushion. Load [MPaI 01 02 03
04 05
Sand cushion modulus E = 45 MPa Settlement S [cm] -1 5 -4 2 -8 8 -IS 8 -29 6
Sand cushion modulus E = 80 MPa Settlement S [cm] -1 2 -3 4 -6 9 -12 3 -20 3
Table 4. Settlements of foundation resting on sand cushion reinforced with 4 layers of geonet. Load [Mpal
Sand cushion modulus E = 45 MPa Settlement s [cm] -1 3 -3 4 -6 1 -10 0 -15 2
I
0.1 0.2 0.3 0.4 0.5
Sand cushion modulus E = 80 MPa Settlement S_[Cn)I -1 1 -3 0 -5,7 -9 5 -14 6
3 CONCLUSIONS
- The application of sand cushion reinforced
”
-
Designed 120 cm wide continuous foundation footing is resting on silty loam subsoil. Usually for cohesive soils the maximum settlements are limited to 3% of the footing width. At the direct foundations the settlements are to be limited to slim = 3.6 cm (Brzqkata and Hung Son, 2000). For the industrial plants foundations, due to the existing machinery and equipment, such high values of settlements can not be accepted. Therefore, the maximum allowable settlement of 1% the footing width has been accepted slim = 1.2 cm. For such settlement value the foundation footing under consideration will be founded on the cushion of 150 cm in thickness, reinforced with 4 layers
with 4 layers of geonet allowed the foundation of footing on soft silty loam subsoil. The compressibility of underlying subsoil exerts significant influence of a sand cushion application. Insufficient compaction of sand in a cushion exerts higher influence on the settlements values of an un-reinforced cushion than one reinforced with a geonet.
REFERENCES Brzqkala B & Hung Son N ,2000 0 poduszkach, poszewkach i materacach XI1 National Conference of Soil Mechanics and Foundation Techniques, Szczecin-Miqdzyzdroje, Poland, May 2000 Dluzewski M J , 1997 Hydrogeo-program of finite elements for geotechnics and environmental engineering Wydawnictwa Politechniki Warszawskiej, Poland Warsaw, 1997
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Field investigations on a soft ground of Bangladesh reinforced by granular piles M. Alamgir Department of Civil Engineering, Bangladesh Institute of Technology (BIT),Khulna 9203, Bangladesh
S.M. Zaher Assistant Engineer, Local Government Engineering Department (LGED), Paikgacha, Khulna, Bangladesh
ABSTRACT: The effectiveness of granular piles in improving a typical soft ground of Bangladesh, was investigated in the field and thus presented in this paper. Sand piles and stone columns of 200mm diameter and 6m long were installed by dry-displacement method in single and group pattern with triangular arrangement at 7 5 0 m spacing. The effectiveness was measured by the plate load test on reinforced ground. The field measurement shows that the bearing capacity of soft ground is increased significantly due to the installation of granular piles irrespective of the types of granular materials. The result reveals that the stone columns can carry 2.4 times higher load than that of the sand piles. The investigation further shows that the spacing ratio of granular piles is to be less than 2.5 to get the group effect. Standard penetration test results show that the consistency of sub-soil was increased by 2 to 3 times due to the installation of granular piles.
1 INTRODUCTION
1999a,b, Alamgir et al. 2001). The applicability of such ground improvement technique to improve the soft cohesive soil deposits containing organic is needed to investigate critically. The effectiveness of granular piles in improving soft ground exists in the south-western region of Bangladesh, was investigated in the field and thus presented in this paper. Sand piles and stone columns of 200mm diameter and 6m long were installed by dry-displacement method in single and group pattern with triangular arrangement at 750mm spacing in a typical soft ground site exists in BIT campus, BIT, Khulna, Bangladesh. Sand prepared by mixing of two-thirds of local sand (available in the south-western region of Bangladesh) and onethirds of Sylhet sand (a widely used construction sand of Bangladesh) was used in sand piles and crushed stone of 19mm down well graded was used in stone columns. Single sand pile, seven sand piles in group, single stone column and three stone columns in group were installed. Effectiveness was measured by the plate load test on reinforced ground. The field measurement shows that the load carrying capacity of soft ground is increased significantly due to the installation of granular piles irrespective of the types of granular materials. The result reveals that the stone columns can C ~ T Yaround 2.4 times higher load than that of the sand piles. The investigation shows that the spacing ratio of granular piles is to be less than 2.5 to get the group effect. The change of soil consistency along the depth was also examined by conducting Standard penetration test (SPT) after ten months of ground improvement. SPT results show that the penetration resistance in-
The development of modern foundation practices, namely, earth reinforcement, to overcome the limitations of the conventional foundation system, has been proved to be viable both technically and economically for the construction works in marginal sites. Earth reinforcement of soft ground by granular piles is considered as one of the versatile and cost effective method. Installation of granular piles transforms the soils into a stiffer composite mass with intervening native soil providing lower overall compressibility and higher shear strength. In the recent years, this technique has also been adopted in Bangladesh in various projects to improve the marginal sites. However, the effectiveness of this method has not been well recorded or monitored due to the lack of required field investigations and the monitoring system. At the present time, more granular piles projects in the USA have been constructed in silty sands rather than cohesive soils. World wide the reverse is true. Reinforcement of soft cohesive soils for construction purposes by granular piles have been established for the last few decades (Engelhardt & Golding 1975, Mitchell & Huber 1985, Okiawa et al. 1992, Bergado & Miura 1994, Alamgir 1996). The installation technique has the big influence on the performance of this ground improvement method (Datye 1978, Barksdale & Bachus 1983, Aboshi & Suematsu 1985). There is a record of successful application of the technique to improve the finegrained soft ground site in a water front structure at Fakirhat, Khulna, Bangladesh (Alamgir & Zaher 517
creases along the depth by 2 to 3 times, due to the installation of granular piles.
2 CONSTRUCTION OF GRANULAR PILES The sub-soil condition of the site and the installation of granular piles are discussed here. The soft ground is reinforced by two types of granular piles, namely, sand piles and stone columns. 2.1 Location of project site This field investigation is performed within the campus of BIT, Khulna, Bangladesh, which can be treated as an ideal place for such field experiment since the sub-soil consists of soft soil to a great depth. The campus is located in the south-western part of the country, as shown in Figure 1.
2.2 Sub-soil condition of the site In the upper horizons, the sub-soil of vast areas of Bangladesh is composed of very soft fine grained soil deposits of Recent origin. In the south-western coastal districts, sub-soils are consist of fine grained soil deposits predominantly peat and muck, due to presence of World’s biggest mangrove forest, the Sundarbans, which was extended in these regions in the past. The field and laboratory investigations reveal that the sub-soil of BIT campus consists of finegrained soil of very soft to soft consistency till the depth of 65ft. A layer of filling sand of very loose to medium relative density exists at the top loft, after that a layer of organic clay of black and dark gray exists at a depth of 15 to 25 ft. The physical and mechanical properties of sub-soil up to 65ft. depth are shown in Table 1.
Figure I . Project site shown i n Bangladesh map.
of several factors specific to the problem at hand (Mitchell 1981). Schlosser and Juran (1979) made an excellent comment that when dealing with techniques of soil improvement, experience has almost always preceded theory. However, in this field study, granular piles, which has a proven record of success as a versatile and cost effective ground improvement technique and have been started to use successfully in Bangladesh in several marginal sites, is considered. Here, granular piles cylindrical in shape were installed in single and group patter. Granular piles are categorized as sand piles and Stone columns based on the types of granular materials.
2.3 Selection of granular piles Selection of the most suitable ground improvement technique in any case only be made after evaluation Table 1. Geotechnical engineering properties of sub-soil of the site. Physical properties Depth
soil strata
(ft.1 0-5 5-10 10-15 15-20 20-25 25-30 30-35 35-40 40-45 45-50 50-55 55-60 60-65
Fine sand Organic clay Clay
Silty clay
Water content, Liquid Limit, Plastic limit (w%, wl%, w,%) 33 ___ ---39 ___ --45, 59, 31 58, 77, 39 223, 112,55 36, 51, 39 36,47, 31 46,42, 32 47,49, 33 24, 37, 36 47, 50, 35 39,48,34 45, 50, 36 3
,
9
,
Unit weight y (Fim’)), Specific gravity, G, -______ , 2.75 -______ , 2.73 25.56, 2.73 17.50, 2.57 7.46, 2.50 14.93, 2.50 18.58, 2.71 13.96, 2.67 14.25, 2.88 13.45, 2.64 14.47, 2.62 13.80, 2.65 14.50, 2.62
Compressibility Properties
Strength properties
Initial void ratio, eo
Undrained shear N-Value strength, S, (kPa) ____ 6 ____ 2 12.0 2 26.0 4 30.0 4 43.0 7 44.0 7 25.0 4 40.0 7 37.0 6 46.0 8 55.0 11 48.0 8
Compression index, C‘
-----
-__ -__
1.706 2.170 7.962 1.207 1.404 1.501 1.464 1.568 1.474 1.502 1.480
0.257 0.391 1.308 0.249 2.176 0.137 0.154 0.169 0.156 0.166 0.154
518
Coefficient of consolidation C, (m’hec) -----
3.83xlO-/ 5.00~10-~ 3.66~10.~ 7.20~10.~ 12.2~10.~ 8.83~10‘~ 9.96~10.~ 7.81~10-~ 14.8~10” 6.60~10-~ 1 3 . 6 1~0-7
2.4 Materials of granular piles Two types of granular materials, namely, sand and stone chips, were considered for the construction of granular piles. Sylhet sand, a yellowish-brown river sand of Sylhet, Bangladesh having FM=2.507, Dl0=0.23mm, D30=0.43mm7 D60=0.73mm, C,=3.17 and C,=1.1025 and a Local sand, available in the south-western region of Bangladesh having FM= 0.69, Dlo=0.O9mm7 D30=O. 12mm, D60=0.2mm, C,=2.20 and C,=0.80, were used in sand piles. Here, FM=Fineness modulus, Dlo=Effective diameter of particle size of which 10% sample is smaller, D30= Effective diameter of particle size of which 30% sample is smaller, D ~ o =Effective diameter of particle size of which 60% sample is smaller, C,=Coefficient of uniformity and C,=Co-efficient of curvature. To increase the cost effectiveness, a combined prepared by one-thirds of Sylhet sand and two-thirds of Local sand were used. Such combination provided an effective and cheap material to improve marginal sites (Alamgir & Zaher 1999a, b). 19mm down well graded stone chips of Sylhet, Bangladesh was used in stone columns. It is originated from crushed bolder and color is whitish and light gray. The grain size distribution of granular materials is shown in Figure 2. 2.5 Installation of granular piles Granular piles with a diameter of 200mm and a length of 6m were installed in a group and single patter. The layout of the constructed sand piles and stone columns is shown in Figure 3. The granular piles were constructed by dry-displacement method. The installation equipment mainly consists of a 1500rpm traditional rig machine, a two end open casing pipe of 200mm diameter, 61n long and 8mm thick. The granular materials were compacted by a hammer of weight 450kg, 175mm diameter and 3m long. The installation procedures as shown in Figure 4 in a schematic diagram, are briefly described as:
1. The casing pipe was placed vertically at the designated point and inserted into the ground about 450mm depth manually. A plug of coarse sand about 750mm is then made at the bottom of casing pipe to seal the pipe. 2. The soil beneath the plug of casing pipe was then displaced by dropping the hammer inside the casing and the casing was then driven to the designated depth by its own weight. 3. At this stage the sand plug was broken by giving excess energy and additional water, the hammer was then withdrawn from the casing. 4. Casing was then lifted by lm from its original position and the designated granular materials were then poured to have a l m thickness and compacted by the hammer to obtain designated compaction.
5 . Casing was then withdrawn further and hole was poured by the same granular materials to have same thickness and compacted. In general, free fallheight of hammer was 0.75 to I.Om and the thickness each layer was about l m and 0.65 to 0.751~1 before and after compaction, respectively. 6. Step 5 was then continued till the granular piles reached the ground surface to have a completed granular pile. 3 FIELD INVESTIGATION To establish the effectiveness of granular piles in reinforcing soft fine-grained soil, some field investigation were carried out. Plate load tests, with a square plate of 300x300mm7were conducted on natural and reinforced ground at a depth of 1.2m from the existing ground surface. On reinforced ground, plate was placed on the top of granular piles for single patter, on the top of middle sand piles and on the top of soil encircled by stone columns for group pattern. In natural ground the load intensity was increased from 10.9 to 98.lkPa at an interval of lO.9kPa. The load intensity was increased from 32.7 to 228.9kPa at an interval of 32.7kPa and 21.8 to 218kPa at an interval 5 19
of 2 1.8Wa for single and group sand piles, respectively. In case of stone columns, the load intensity was increased from 32.7 to 523.2 at an interval of around 5OkPa and 21.8 to 130.8kPa at an interval of 21.8kPa for single and group, respectively. In each load increment, settlements were measured till the rate of deformation less than 0.25mmlhr. Load increment was continued till the failure or up to 25mm settlement. The resul are represented in Figure 5. To depict the change of consistency of sub-soil, standard penetration test was performed at the two locations of the reinforced ground till the depth of 30ft. after ten months of granular piles installation. 520
columns plate was placed on soil in plate load test. Since capacity for natural ground and group stone columns are almost same, this result again indicates that group effect can not obtained for a spacing ratio of 2.5. This finding is agreed with other researchers.
The bore hole one is located in between the peripheral and central sand piles while bore hole two is encircled by the three stone columns. The results indicate the remarkable improvement of consistency of sub-soil along the depth.
4.3 Comparisons of settlement response
4 RESULTS AND DISCUSSIONS
A comparison of load carrying capacity due to the installation of sand piles over stone columns is described here. For the same settlement, the load carrying capacity of single stone column is 2.4 times higher than that of single sand pile. In case of group, it can be seen that the sand piles carried more loads than that of stone columns. The reason is that the test conditions for group stone columns is different from the group sand piles and single stone column. In case of group stone columns, plate load test was done on the top of the excavated ground encircled by the three stone columns. The test results obtained from both the group sand piles and stone columns reinforced ground, it is revealed that group effect can not expected for a spacing ratio 22.5.
The results obtained from field investigations by plate load tests and sub-soil investigations are presented and hence discussed in this chapter. Results obtained from reinforced ground are compared with those of natural ground. 4. I Response of ground reinforced by sand piles The load-settlement response of ground reinforced by sand piles is shown in Figure 6. This figure shows that load intensity taken by single and group sand piles are 230 and 220 kPa, respectively, corresponding to 25mm settlement, while it is 98Wa for natural ground. The improvement for the installation of single and group sand piles is almost same, which is 2.3 times greater than that of natural ground, which reveals that the group can not be obtained for the spacing ratio larger than 2.5.
4.4 Improvement of sub-soil Conditions The penetration resistance obtained on natural and reinforced ground is shown in Figure 8. This figure reveals that the penetration resistance of the soft ground increased significantly due to the installation of granular piles. N-value ranges from 2 to 7 for natural ground but in reinforced ground it increases from 5 to 11 and 5 to 12 for bore hole one and two, respectively. Result also shows that the increment of N-value does not depend on granular materials.
4.2 Response qf ground reinforced by stone coluinizs Figure 7 shows the load-settlement response of stone column-reinforced soft ground. This figure shows that corresponding to 22mm settlement, the load intensity taken by single and group stone columns are 525 and 110 kPa, respectively, against of 95kPa for natural ground. The improvement for the installation of single stone column is 5.5 times greater than that of natural ground, while it is 1.15 times greater in case group stone columns. In case of group stone
5 CONCLUDING REMARKS Based on this study the following conclusions can be made: 1. Field observation reveals that the dry-displacement method is a suitable installation technique for the construction of granular piles in a soft fine-grained soil. 2. Plate load test result shows that the granular piles improved substantially the bearing capacity of natural ground. 3. The stone columns constructed by 19mm down well graded stone chips provide better improvement. The load carrying capacity is around 2.4 times higher than that of sand piles. 4. The result indicates that the spacing ratio of granular piles should be less than 2.5 to obtain group effects. 5. Sub-soil investigation shows that the shear strength of soil was increased along the depth due to the installation of granular piles.
52 1
6 ACKNOWLEDGEMENT The support of Ministry of Science and Technology, Government of Bangladesh, for this study by providing a grant to the first author in a Research and Development Project in the year of 1998 and 1999 is grateful1y acknowledged. REFERENCES Aboshi, H. & N. Suematsu 1985. Sand compaction pile method: State-of-the-art paper. Proc. 3rd Iiit. Geotechnical Senzinnr on Soil Improvenient Methods, Nanyang Tech. Institute, Singapore. Alamgir, M. 1996. Analysis of soft ground reillforced by columnar inclusions. Ph.D. Thesis, Department of Civil Engineering, Saga university, Saga, Japan. Alamgir, M. & S.M. Zaher 1999a. Field performance of granular columns installed to improve fine grained soil deposits. Proc. Civil and Environmental Engineering CoilferenceNew Frontiers and Challenges, 8-12 Nov. 1999, 2(IV):5766, Bangkok, Thailand. Alamgir, M. & S.M. Zaher 1999b. Field performance of sand piles in improving soft ground. Paper presented in the 43rd Annual Convention of IEB, 5-8 Murch 1999, Dhaka, Bangladesh.
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Alamgir, M., S.M. Zaher, M.A. Haque, M.M. Sobhan & M.A. Razzaque 2001. Study on the performance of recently adopted some foundation systems in the soft ground of Khulna region. Souvenir 45th Annual Convention of IEB, 17-19 Feb. 2001,76-8 1, Khulna, Bangladesh. Barksdale, R.D. & R.C. Bachus 1983. Design and construction of stone coluinns, Vol.I . Report No.FHWA/RD-83/026, NTIS, Virginia, USA. Bergado, D.T. & N. Miura 1994. Improvement techniques for soft and subsiding ground. In N. Miura, M.R. Madhav & K. Koga (ed.), Lowlands Development and Muiiugenzent: 103134. Rotterdam: Balkema. Datye, K.R. 1978. Special construction techniques. Proc. IGS Con$ on Geotechnical Engineering, New Delhi, India: 30-34. Engelhardt, K. & H.C. Golding 1975. Field testing to evaluate stone column performance in a seismic area. Geotechrzique. 25(1):61-69. Mitchell, J.K. 1981. Soil improvement. State of the Art Report. Proc. loth ICSMFE, Stockholm, Sweden, 4509-567. Mitchell, J.K. & T.R. Huber 1985. Performance of stone column foundation. J. Geotechnical Engineering, ASCE. 1 11(2):205-223. Schlosser, F. and I. Juran 1979. Parametres de calcul des sols artificiellement ameliores: Rapport general Seance 8. Proc. SmEuropean Conf. on SMFE, 8: 1-29.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets 6: Zeitlinger, ISBN 90 2651 863 3
Bearing capacity considering stiffness of reinforcement material K. Arai & M. Kamon Fukui University, Fukui, Japan
S. Nomura & Y. Yokota Maeda Kosen Co. Ltd., Fukui, Japan
ABSTRACT: This paper proposes a practical procedure for estimating the bearing capacity of strip footing on reinforced ground. The procedure aims to fill a gap existing between classical FEM and conventional limit equilibrium analysis, by creating a definite collapse mechanism analogous to a slip surface. Based on MohrCoulomb yield criterion, a simple non-associated flow rule, a smeared shear band approach, and on an improved initial stress method, the procedure provides an explicit collapse mechanism represented by stress yield condition. Since the collapse mechanism is supported by a displacement field and a stress field, the procedure enables to perform stability analysis taking the stiffness and deformation into consideration.
1 INTRODUCTION Conventional stability analysis based on limit equilibrium method tends to become uncertain, when a soil stratum consists of multiple layers, or when it includes other materials having quite different stiffness like earth reinforcement materials. This is because limit equilibrium method evaluates the material properties only by its final strength. The method represents kinematical conditions only by using the mechanically reasonable shape of a slip surface, and does not explicitly allow to consider the stiffness and deformation of materials, which seem to play an important role for evaluating earth reinforcement methods, and which may affect the global collapse mode. To compensate for these defects, a lot of trials have been made by applying other analytical methods, for instance, FEM and limit analysis. In order to utilize these analytical methods in the practical design, the methods are expected to provide an explicit collapse mode as well as a slip surface used in stability analysis, because the results by these methods should be related closely to the conventional solutions. Classical FEM does not necessarily provide a reasonable collapse mechanism. Subjected to MohrCoulomb material, the limit analysis method has not completely overcome the difficulty that the limit theorems cannot be proven without the normality rule in plasticity, and that the normality rule may not hold for the material. In spite of many researches in recent years, the accurate description of localization phenomenon in soils is still open to question. For instance, the bifurcation analysis which tries to simulate actual localized deformation, seems to give a promising view, while the analysis may not give rea-
523
sonable solutions for complicated boundary value problems like bearing capacity. Still now the stability analysis considering both reasonable collapse mode and material stiffness, remains a thorny subject for practical design work. Using a modification of the smeared shear band approach (Pietruszczak et al., 1981) which is based on estimating average mechanical properties of elastic solid and shear band, and using a new calculation scheme for nonlinear FE analysis, this paper aims to develop a practical procedure for estimating the bearing capacity, which enables to create a reasonable collapse mode supported by a displacement field. This paper treats only a centrally and vertically loaded strip footing on flat subsoil under the plane strain condition.
2 CONSTITUTIVE RELATIONSHIP 2.1 Yield criterion To relate the proposed procedure to conventional stability analysis, Mohr-Coulomb and Coulomb yield criteria are employed respectively to plane strain soil mass and friction interface between structure and soil. For the friction interface we employ the thin layer finite element as shown in Fig. 1 (Desai et al., 1984). Mohr-Coulomb:
{(ox+oj)sin @+2ccos @}=O Coulomb:
F c I ~ s t l -C-Q tan$=0
where ox, o,, and zxy: stress components, or and w: normal and shear stresses in friction interface (see Fig. l), and c and 9: cohesion and friction angle. Both quadrilateral plane strain and thin layer finite elements are built up from four constant strain triangles, and a set of stresses is regarded constant within each element.
2.2 Coulomb inte$ace Both for Mohr-Coulomb and Coulomb materials, a linear elastic response is assumed before yielding. Fig. 2 schematically illustrates the relationship between stress vector { o } and strain vector { E } . When applying further footing pressure after a stress state has reached the yield surface, the stress state will move along the yield surface as seen in Fig. 2. This is because normal stress or becomes larger with the increase in footing pressure, and because yielding shear stress increases with normal stress for frictional material. Point B in Fig. 2 corresponds to a plastic equilibrium state at an arbitrary position within a yield region. At the elasto-plastic state from point A to B shown in Fig. 2, we employ the simplest non-associated flow rule or plastic potential Qc defined by Fig. 3 (Mroz, 1980) QC =I zst I - g - ot tanv (3) where v: dilatancy angle (see Fig. 3) and g : a hypobethetical parameter which is not cited cause Qc is used only by its differential form. For the thin layer element shown, Eqs. (2) and (3) give the elasto-plastic stress-strain relationship as (Zienkiewicz et al. 1969)
{ 60}=[D,sre~] {6 ~ e P }
Figure 1 . Interface element.
Figure 2, Stress-strain relationship (Coulomb material),
(4)
{ 60}={ 60s, &or,6Tsr)T {6 & ~ } = ~ E{. F P ,6&rcP, GysreP}T [ D s r c ~ =[D]-[D] ] { aFc/o'{ O } } x { a Q c / 3 { o } ) T [D]/ {dFc/o'{o}}T x[D]{ aQc/a{o }}
where { 60) and { ~ E C P } : stress and elasto-plastic strain increments ( ~ i 21, ~ . [ D , ~: ~e~asto-p~astic ~I stress-strain matrix in local coordinate s-t in Fig. 1 , [ D ]: elastic matrix, E: Young's modulus, p: Poisson's ratio, and G: rigidity modulus.
Figure 3. Non-associated flow rule (Coulomb material)
2.3 Mohr-Coulomb material When shearing a finite size of soil element, it is well known that we often observe a shear band or slip surface as shown in Fig. 4 (a). Despite many theoretical and experimental studies concerning the mechanism of shear band formation, we have not reached a final agreement with regard to inclination angle of shear band a defined in Fig. 4 (a) (e.g. Vardoulakis et al., 1980). Since our main concern is to Figure 4. Shear band formation.
5 24
get a practical design procedure, we employ the most fundamental expression as
a= n/4+$/2 (51 Without introducing a separate interface element corresponding to a shear band, Pietruszczak et al. (1981) proposed the smeared shear band approach which evaluated the average stress-strain response of solid and shear band. This approach assumes elastic response of solid and purely plastic response of shear band. Herein we assume elastic response of solid and elasto-plastic response of shear band, because this postulate yields a convenient constitutive behavior as described later. According to the procedure by Pietruszczak et al. (1981), the average stress-strain matrix is given as follows. Assume that a plane strain solid element reaches a yield state, and that a shear band has been created as shown in Fig. 4 (a). Regarding the shear band as a thin layer element shown in Fig. 1, strains in shear band are given in local coordinate s-t as (Eq. 4)
Figure 5. A set of two slip surfaces.
{ 6&eP}=[D,sreP]-I{ 60) (6) Assuming elastic response except in shear band, strains in solid region are { 6&e}=[D]-1{ 60) (7) Superposing these two strains by Pietruszczak et al. (198 I), average stress-strain matrix of the whole element [D,Pv] and that in global coordinate x-y [Dxya~j] are [D,pv]=( [D,fP]-llId& XCOSP) 1 - 1
Figure 6. Isolation of slip surface.
cOsp+[D]-yl-t/&
(8)
where t: thickness of shear band, and [ U :coordinate transformation matrix. Both [ D s p ] and [D,p>] include no current stress components. The average matrix calculated above becomes close to the stressstrain matrix for the stratified or cross anisotropic material which is built up from the Coulomb material represented by a thin layer element as illustrated in Fig. 4 (b).
Figure 7. Direction of slip surface in a element.
2.4 Direction of shear band
wedge below strip footing base, which is represented by a series of interface elements in Fig. 9 shown later. Thus we employ the right-hand side shear band B-B' defined in Fig. 5 within the active wedge in Fig. 9, and assume the left-hand side shear band AA' in Fig. 5 outside of the active wedge as seen in Fig. 6. Referring to Fig. 7, the direction of A-A' or B-B' line in Fig. 6 is given as
Generally a set of two shear bands or slip surfaces A-A' and B-B' is possible for a finite soil element according to the principal stress state as shown in Fig. 5. In practical problems we must select one of these two shear bands. Since our object is to get a practical method for stability analysis, we make efficient use of the conventional collapse mode as illustrated in Fig. 6. Many experimental researches have observed actually the active wedge developing in model test. A lot of bifurcation analyses also have obtained the active wedge, while the analyses have not got the complete collapse mode as illustrated in Fig. 6. Based on these results, we assume the active
p= - a - 0
: A-A'line
= a - 0 : B-B'line (10) where p: inclination angle of shear band, and 0: angle of the major principal stress from vertical axis
525
(Fig. 7). Note that compressive stress is positive here and that shear stress ~ . is~ negative r along A-A' line in Fig. 6 and positive along B-B' line.
{ r }I = {
GO}1-
{ GE}~+{ GB}
[a-J( { slf)
= { GO)[- ( [ D ] r - [ D x ~ r)[Bli{ l
+cj[BLT{0 0 } j A j )} r=O 3 NUMERICAL PROCEDURE
(1 1)
where { r J r :residual, [B]c matrix for calculating : strain components from nodal displacements, global stiffness matrix, { Sf}: load increment vector, AL: area of the element, and suffixes i a n d j denote element number. The iteration procedure by CGM is as follows. 1) Set the trial values of {GO}.2) Calculate gradient {g}flgiven by Eq. (12), where n designates an iterag}/l T{ g } ) ~ tion number. 3) { d}ii=-{g)"+{d}ii-J-{ l{g}f2-JT{g}lr-l. 4) { 0 0 } ~ ~ + ~ = { 0 0 } f ~ +5 h) Repeat ~~{d}~~. 1) to 4) until {r}r becomes sufficiently small. { g } n and hi2 are calculated analytically. Conclusively the numerical steps during a typical load increment are summarized as follows. 1 ) Performing an elastic analysis by using actual load increment { SJ}, calculate (GE} and { E ) in Fig. 8. 2) Find the yield finite elements in which { GE} violates the yield criterion. 3) For the yield elements, calculate yield stress { oA} both from { G ~ )and the preceding stress state. 4) Concerning { (TA}, calculate direction of the major principal stress 8, and find shear band inclination angle p by Eq. (10) (Fig. 8). 5 ) Calculate [ D X p ]by Eq. (9). 6) Determine {GO} by CGM. 7) Again, find the yield finite elements by performing an elastic analysis by use of both { S j } and (GO} determined at 6). When finding new yield elements, determine {GO} by CGM subjected to the total yield elements including the new yield elements. Repeat this procedure until no new yield elements are found. 8) Based on the final results at 7), calculate necessary state variables ( O B } , settlements, and so on.
[a
3.1 Fundamental aspect Fig. 8 defines actual stress of initial state { GI}, yield stress { GA},actual stress of plastic equilibrium state { GB},elastic stress { GE}, virtual initial stress {(TO}, total strain { E } , elastic strain ( E " } , and elasto-plastic strain { E V } . 1) For the convenience of solving bearing capacity problems, we apply footing pressure by many loading stages subdivided. In the application of initial stress method, we use the same stiffness matrix throughout all the loading stages, because we assume linear response of subsoil both before and after yielding. 2) Yield stress { O A } is isolated by Zienkiewicz et al. (1969) 3) To determine the direction of shear band as shown in Fig. 7, it is necessary to find direction of the major principal stress 0. We decide 0 by using yield stress { o A } , and use it throughout the succeeding loading stages.
3.2 Iniproved initial stress method The original initial stress method often provides unstable results for our problem. The incremental procedure which treats the nonlinearity as piecewise linear, does not create the collapse mode as illustrated in Fig. 6. These difficulties are avoided by introducing an iteration scheme based on the conjugate gradient method (CGM, Flecher et al. 1964) instead of the original initial stress method. CGM is often used for solving simultaneous linear equations with quite many unknowns, for the purpose of reducing computational effort. The constitutive model employed here, which is a linear equation also at elastoplastic state, enables to apply effectively CGM. Since { G O } = {GE) - {GB}, the basic equation is given as
4 CASE STUDIES We discuss the effect of earth reinforcement, concerning the bearing capacity of a rigid strip footing on a homogeneous and ponderable c-$ soil stratum. FE meshing and material parameters are given in Fig. 9, in which y: unit weight, T: thickness of footing and A: area of cross section of truss material. As described before, the active wedge below footing base is assumed by a series of interface elements, so as to isolate the direction of shear band (Fig. 5). Angle CI in Fig. 9 is given by Eq. ( 5 ) , regarding the vertical footing pressure as the major principal stress. Actual initial stress vector { O I } is given as: vertical stress q)/=overburden pressure, horizontal stress o,I=Ko(T~,I,and z,,r=O, where KO=1-sin$. For convenience, we introduce 'footing pressure ratio R' defined as
R = 4 /4u Figure 8. Initial stress method.
526
(12)
footing: I3980 MPa, /I M.16, T-1 m
K7
footing pressure ratio R
0
interface: (3-3.68 MPa, Ee9.8 MPa, /I =0.33 c=9.8 kPa. 6 =M'.v M
U
h
0.5
1
1.5
2
I
I I I tI
I I I
I I I
I
I
612=1.5 m
Figure 10. R-settlement curves.
18.5 m
Figure 9. FE meshing.
where q: current footing pressure applied, and qu: ultimate bearing capacity given by Terzaghi. R is increased step by step at each loading stage subdivided. Fig. 10 shows the relationship between R and footing settlement calculated by the proposed procedure. When (I exceeds 2 5 O , the iteration by CGM does not converge at a certain loading step. This suggests that we cannot find a plastic equilibrium state B shown in Fig. 2 for larger R, which is considered as the bearing capacity. When CGM converges at every loading stage, we regard the bearing capacity as the footing pressure at which settlement increases remarkably, or as the pressure at which Rsettlement curve merges into the approximate straight line observed after R has exceeded the critical pressure, as shown in Fig. 10. This is because the straight line means the plastic flow of foundation subsoil. The bearing capacity in terms of R is called 'critical footing pressure ratio Rc;. As shown in Fig. 10, this earth reinforcement method increases RCr from I (no reinforcement) to 1.3. Fig. 11 compares the yield region respectively for reinforced ground with natural ground, in which the bold solid line in each finite element represents the direction of shear band as shown in Fig. 4 (a), and that the element has yielded. Fig. 11 (a) seems to produce a collapse mode analogous to Prandtl mechanism given for imponderable subsoil. More complete collapse mode is observed for larger R. The collapse mode in Fig. 11 is supported by the displacement field shown in Fig. 12 and stress field shown in Fig. 13. That is, the collapse mode is created by considering the weight of subsoil, stiffness of footing and subsoil, friction between footing and subsoil, and stress concentration at the edge of rigid footing, most of which are ignored in Prandtl and Terzaghi approaches. The yield region in Fig. 11 tends to distribute deeply below footing. Despite the lateral plastic flow as illustrated in Fig. 11, the vertical pressure must reach lower subsoil due to the vertical equilibrium condition, and the pressure makes lower subsoil yield. Although conventional stability analysis provides little infor-
527
(a) Natural ground (R=1.5)
(b) Reinforced ground (R= 1.5) Figure 1 1. Yield region.
mation about yield state except on the location of slip surface, there is the possibility that the region lower than the slip surface may yield. In Fig. 12 we observe little deformation of lower subsoil. The proposed procedure represents a collapse mode by yield condition of stresses, which is the same as conventional stability analysis and different from most application of conventional FEM using strain distribution. The conventional design scheme based on stability analysis, uses the collapse mechanism assumed without reinforcement, and evaluates the reinforcement only by its final strength. The scheme neglects the stiffness of reinforcement material which may restrict the deformation of subsoil and which may largely contribute to the improvement of bearing capacity. This example proves the possibility of applying the proposed procedure to the stabil-
tional stability analysis and classical FEM. The procedure employs Mohr-Coulomb and Coulomb yield criteria respectively for soil mass and friction interface between soil and structure. By assuming a linear elastic response before yielding and a simple non-associated flow rule after yielding, and by employing a modified smeared shear band approach and an improved initial stress method, the procedure provides an explicit collapse mechanism like a slip surface. At the collapse mode, a stress yield criterion is satisfied as well as along a slip surface supposed in conventional stability analysis. Case studies show that the proposed procedure provides a solution close to the conventional solution, and show the possibility that the procedure gives a reasonable estimate of bearing capacity on reinforced ground. REFERENCES Desai, C. S., Zaman, M. M., Lightner, J. G. & Siriwardane, H. J. 1984. Thin-layer element for interfaces and joints. 1121. J. Numer. Anal. Methods Geomech. 8: 19-43. Flecher, R. & Reeves, C. M. 1964. Function minimization by conjugate gradient. Cornputer J . 7: 149-154 Mroz, Z. 1980. Deformation and flow of granular materials. Mechanics of Solids (the Rodney Hill 60th Anniversary Volume) Pergamon Press, Oxford. 119-132. Ortiz, M., Leroy, Y. & Needleman, A. 1987. A finite element method for localized failure analysis. Computer Methods Mech. Eizg. 61: - 214. Pietruszczak, S. & Mroz, Z. 1981. Finite element analysis of deformation of strain-softening materials. Irzt. J. Nutn. Methods Eng. 17:327-334 Vardoulakis, I. 1980. Shear band inclination and shear modulus of sand in biaxial tests. Int. J. Num. Ana. Methods Geonzech. 4:103-1 19. Zienkiewicz, 0. C., Valliappan, S. & King, I. P. 1969. Elastoplastic solutions of engineering problems 'initial stress', finite element approach. Int. J. Nilmer. Met}zod.y Eng. I : 75-100.
ity analysis of earth reinforcement, which takes the stiffness and displacement of material into consideration.
5 CONCLUSIONS This paper proposed a numerical procedure for arialyzing the bearing capacity Of strip footing*The procedure aims to fill a gap existing between conven-
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Reinforcement of soft sub-grade for high-speed railroads using geocell S.D. Cho, J.M. Kim, M. Chung & S.H. Yoon Korea Institute of Construction Technology, Koyang, Korea
Y.Y. Kim E&S Engineering, Seoul, Korea
ABSTRACT: This paper presents the results of field plate loading tests and laboratory dynamic loading tests carried out to evaluate the performance of geocell where it is used to reinforce soft sub-grade for high-speed railroads. The factors selected in the tests include the type of infill material, the number of geocell mattress layers, the thickness of cover soil, the stiffness of the original foundation soil, and the presence of geotextile separating the geocell mattress and the neighboring foundation soil layer. The results of the tests confirmed the effect of geocell mattresses in improving load-bearing capacity of soft soil. The settlement rates of geocell reinforced railroad structure subjected repeated loadings were revealed from the dynamic loading tests. This paper deals with field plate loading tests and laboratory dynamic loading tests designed to investigate the performance of geocell when it is applied to soft sub-grade for high-speed railroad. The relative efficacy of geocell was compared with the reduced thickness of reinforced roadbed of sub-ballast and aggregate layer.
1 INSTRUCTION Geocell or geocell mattress is referred to as a threedimensional soil confinement system consisting of a series of interlocking cells. Geocell may provide a practical means to reinforce soft foundation soils for the construction of high-speed railroads, in cases where the conventional soft ground improvement techniques are not favored due to time and cost constraints. Reinforcing soft ground with geocell mattresses is easy and fast. Geocell mattresses, first, are positioned on a surface in their stretched form. Then, each of cells is filled with coarse grains or concrete that is selected in such a way to better serve the purposes of its application. When geocell mattresses are used in load support applications, cover soil is usually applied on the top of the geocell mattresses. Improved load bearing capacity of a soft ground with an aide of geocell is a combined result of the hoop strength of the cell walls, the passive resistance of adjacent cells, and the frictional interaction between the infill and the cell walls (Presto and Intersol Engineering Inc. 1997). The performance of geocell reinforcement, therefore, is influenced by factors such as: raw material, shape, and size of a geocell; type of infill and compacting effort for infill; and applied load. Some literatures report on the use of geocell: geocell made of woven and non-woven geotextile (Rajagopal et al. 1999); large scale model tests on polymeric geocell mattress (Barthusrst & Jarrett 1988); and reinforcement of embankment using geocell mattresses (Bush et al. 1990; Cowland & Wong 1993). However, more researches are yet to be needed to better understand the efficiency of geocell with many different influencing factors.
2 EXPERIMENTAL PROGRAM
2.1 Field plate loading tests The site for plate loading tests was situated near the Inchon international airport where its construction was ongoing. For all the tests, the shallow surface soil was excavated to a depth where the underlying silty clay layer was exposed. A deposit of silty clay is typical in this region and exists with the thickness of 3 to 10 m. The shear strength of this silty clay layer measured at its near-surface depth from vane shear tests was in the range of 5.9 and 12.7 kPa. A series of plate loading tests were performed with the combination of influencing factors. The factors selected were the type of infill material, the number of geocell mattress layers, the thickness of cover soil, the stiffness of the original foundation soil, and the presence of geotextile separating the geocell mattress and the neighboring foundation soil layer. Geocell mattresses used were made of textured high-density polyethylene and provided by its local agent. The dimensions of a geocell mattress in its expanded position were approximately 2.4 m in width, 6.1 m in length, and 0.2 m in thickness. The dimensions of an individual cell were 406 mm in 529
len th and 488 mm in width and the area was 991 cm . The infill materials used were crushed stone and sandy soil. For each of the plate loading tests, the same cover material as the one for infill was employed. The crushed stone was classified as GW by the unified classification system with the maximum size of 60 mm. The maximum dry density was 2.06 Mg/cm3. The sandy soil was classified as SW and its maximum dry density was 2.06 Mg/cm3. The infill and the cover material were compacted at its relative density of 90% and 95% respectively.
F
2.2 Laboratory dynamic loading tests Dynamic loading tests were carried out using a test setup simulating the real field conditions. A schematic of the test setup is shown in Figure 1. In a 1.0m wide, 1.4m long, and 2.0m deep chamber, there were, from the bottom, a 0.5 m thick silty clay layer, a 0.2 m thick geocell mattress layer and 0.05 m thick cover soil, an aggregate layer of crushed stone, a 0.2 m thick sub-ballast, and a 0.35 m thick ballast. The thickness of the aggregate layer varied during the tests: 0.25, 0.35, 0.45, and 0.55 m. The combined layer of sub-ballast and aggregate layers constitutes a reinforced roadbed. A sheet of nonwoven geotextile with a tensile strength of 29.4 kN/m was placed between the geocell mattress and the underlying clay for separation. The silty clay was obtained from the site where the plate loading tests were performed. Crushed stone with the maximum size of 125 mm was used as the infill for the geocell mattress and the overlying components. The maximum sinusoidal load of 117.7 kN was applied at the frequency of 3.5 Hz as many as 80,000 times on the ballast through a 0.27mx0.8m rectangular steel plate. The loading conditions were
determined from the preliminary tests reflecting the local design criteria for high-speed railroads and the limitations of the test setup. The local desi n s ecifications require K30 no less than 68.6 MN/m . Otherwise, the original foundation soil should be reinforced using proper means. If K-30 is between 68.6 and 107.9 MN/m3, the required thickness for sub-ballast and aggregate layer is 0.2 m and 0.6 m, respectively. The second series of dynamic loading tests were performed to investigate the effect of geocell reinforcement in reducing the thickness of reinforced roadbed. The loading conditions and test chamber were the same as those for the previous dynamic loading tests. Since these tests were designed to simulate the foundation soil condition of K.70 no less than 68.6 MN/m3, sandy soil instead of clay was placed in the bottom of the chamber and compacted to have a K30 value in the vicinity of 68.6 MN/m3. Geocell mattress was not employed. An aggregate layer was placed immediately above the sandy soil foundation with the thickness of either 0.4 or 0.6 m. The thickness of sub-ballast and ballast was also 0.2 m and 0.35 m respectively.
B P
3 RESULTS AND DISCUSSION 3.1 Field plate loading tests Sub-grade reaction modulus K ~ (MN/m3) o and strain modulus E,, (MN/m2) were calculated from the load versus settlement curves of plate loading tests. K30 is obtained from the normal stress 0 corresponding to a settlement of 1.25 mm. The subscript 30 means the diameter of the loading plate. E is the gradient of the secant modulus between the points 0.3xqrlO1and 0 . 7 ~ qusing , ~ ~the ~ following equation (German standard, DIN I8 134):
s = a, +a,x 0, + a2x 0;
(2)
where r = radius of the loading plate (mm), o;,,, = the maximum average normal stress of the loading cycle (MN/m2), 00 = the average normal stress below the plate (MN/m2), s = settlement (mm), and ao, a ] , a2 = factors. Figure 2 shows the results of the plate loading tests performed with respect to a number of geocell layer (1 or 2 layers) and the type of infill with a constant cover thickness of 50 mm. The legends in Figure 2 are explained in Table 1. Without the presence of geocell mattress, the soil exhibited an ultimate load bearing capacity of 58.8 W a at about 30 mm settlement. In cases where the geocell mattresses were placed on the same soil, the effect of geocell
Figure 1 . A schematic of test setup for laboratory dynamic loading tests.
530
some extent, outperformed crushed stone at both Sites A and B, which is against a general tendency of the effect of angularity of grain in load supporting mechanism. It may be because crushed stone than sandy soil was relatively easier to penetrate into the soft foundation soil during load application. In the tests with one layer of geocell and cover soil of varying thickness (Case I , and Cases 5-8), the load intensity and Settlement curves were almost linear as observed in Figure 2. The curves exhibited increasing K30 and E,, as the thickness of cover soil increased. The combined effect of the densification of infill during the compaction of cover soil and the load distribution effect of cover soil itself are believed to be responsible for increased values of K-70 and E,,. When the same material for infill and cover was used, one layer of geocell mattress with thicker than 200 rnrn cover soil led to better load supporting capacity in terms of the indices than two layers of geocell mattresses with 50 rnrn cover soil. This indicates that, in a practical sense, a cost efficient design of geocell reinforcement can be achieved by determining an appropriate thickness of cover that would meet the design requirements. Figure 3 shows the results of the plate loading tests undertaken to investigate the improvement of load bearing capacity by means of the filling of sandy soil on soft foundation soil. The thickness of fill varied between 0.25 m and 1.00 m. No geocell reinforcement was employed in this case. The value of K30 was: 47.8 MN/m3 for a 0.25 m thick filling; 63.6 MN/m3 for 0.50 m; 73.1 MN/m3 for 0.75 m; and 91.4 MN/m3 for 1.OO m. Proportional relation between load intensity and settlement as observed in the cases of geocell reinforcement was found only when the fill thickness was equal to or more than 0.75 m. The comparison of these K.70 values with
Figure 2. The results of plate loading tests as a function of the number of geocell layer and the type of infill: the thickness of cover was fixed at 50 mm. Table 1 . Description of test conditions. Denotation Test Description Case 1 1 geocell layer + 5 cm cover : sandy soil" Case 2 1 geocell layer + 5 cm cover : crushed stone Case 3 2 geocell layers + 5 cm cover : sandy soil Case 4 2 geocell layers + 5 cm cover : crushed stone Case 5 1 geocell layer + 10 cm cover : sandy soil Case 6 1 geocell layer + 20 cm cover : sandy soil Case 7 1 geocell layer + 30 cm cover : sandy soil Case 8 1 geocell layer + 40 cm cover : sandy soil Case 9 without geocell reinforcement Note: $: indicates the type of cover soil Table 2. Calculated values of Kjo and E,, from plate loading tests. Test Conditions Case 1 Case 2 Case 3 Case 4 Case 5 Case 6 Case 7 Case 8 Case 9
Site A Kqn E,> (MPa/m) (MPa) 62.8 7.0 57.9 7.8 76.5 7.9 69.6 8.1 73.6 9.2 74.5 9.7 81.4 13.9 107.9 16.5 1.0 20.6
Site B K?O E,. (MPdm) (MPa) 56.9 4.8 4.0 55.9 72.6 6.9 58.8 7.8 68.6 7.0 74.5 8.1 75.5 10.5 79.4 13.6 16.7 2.3
mattresses in increasing load-bearing capacity was obvious as expected. The gradients of the load intensity and settlement curves did not show distinct inflection points and were almost linear in the range of the applied load intensity. The parameters K.30 and E, were calculated as indices of how the geocell reinforced soil behaved under the given test conditions and are listed in Table 2. The effect of infill material can be seen from the comparison of K30 and E,, of Case 1 with those ofCase 3: and Case 2 versus Case 4. Sandy soil, to
Figure 3. The results of plate loading tests with varying thickness of fill overlying soft clay foundation.
53 1
those for Site A shown in Table 2 indicates that one layer of geocell mattress with a 0.1 m cover (Case 5 in Table 1) is equivalent to a 0.75 m thick fill without geocell reinforcement. It can be said that reduction of fill thickness can result from the use of geocell mattresses, which in turn, leads to reduction of the weight of foundation structures for high-speed railroads. This will be of help in increasing the stability of the structures and in saving construction time. When geocell mattresses placed on a soft foundation soil are subjected to dynamic loadings, infiltration of clay particles into the geocell mattresses, or mud pumping, might occur without a separator at the interface of these two layers. Geotextile is commonly used in this case as a separator. Plate loading tests under the condition of Case 5 in Table 1 were carried out at Site B with and without geotextile. The results are shown in Figure 4. Without geotextile, E,, was 7.0 MPa as shown in Table 2. In the presence of geotextile, load-bearing capacity of geocell reinforced soil significantly decreased. Strain modulus E,, was calculated to be 1.0 and 0.9 MPa. The reduction of load bearing capacity with geotextile is probably because geotextile might have prevented the infill from penetrating into the underlying soft soil during compaction. This penetration of the infill formed a replaced layer immediately beneath the bottom of the geocell mattress, producing a more resistance to the external forces during plate loading tests. The plate loading tests did not provide an information about whether this adverse effect of geotextile would be only valid for the given test conditions or not. Since the geotextile used had, however, a marginal strength, it is not unreasonable to predict that reduction of load bearing capacity is instantaneous during construction and may disappear under real loading conditions.
532
3.2 Dynamic loading tests During the dynamic loading tests, settlement was measured at the three different points as shown in Figure 1: on the top of and beneath the ballast layer; and on the top of the clay layer. The settlement rates with respect to the different thickness of the aggregate layer are plotted in Figures 5 to 8. Examination of the figures revealed a consistent pattern in the settlement rates, but a wide range of variations in the magnitude of settlement. The total settlements appear to increase at relatively rapid rates until the number of load ap-
Figure 9. Total settlement versus number of load application with respect to varying thickness of aggregate layer.
ing stage are thought to be due to the reorientation and cracking of the grains in the geocell mattress. Anomaly in the magnitude of settlement of each constituting layer appears to be caused by inconsistency involved in the test preparation processes. The results of the second set of dynamic loading tests in terms of total settlement and number of load application are depicted in Figure 9. The data points in open symbols were extracted from the con-esponding tests results in Figures 5 to 8. Closed symbols are for the tests with 0.4 and 0.6 m thick aggregate layers overlying the sandy soil foundation. Up to the load applications of 80,000 times, the magnitude and rate of total settlement for the case of a 0.35 m thick aggregate layer over a geocell mattress (Figure 6) were similar to those for a 0.6 m thick aggregate layer over the sandy soil foundation. With an aid of geocell reinforcement, replacement of soft clay with sandy soil could be spared. In other aspects, the combined layer of geocell mattress and soft clay soil in this case is equivalent to the replaced sandy soil foundation with a 0.6 m thick aggregate layer. Therefore, it can be said that geocell reinforcement resulted in reduction of the thickness of aggregate layer by 0.25 m or so. Even in the cases where clay foundation soil was replaced by sandy soil, the total settlement rates did not reach plateau values as the number of load application exceeded 80,000 times and approached 200,000 times. There is no reason to doubt the effect of geocell reinforcement in increasing load-bearing capacity of soft soil. However, more studies with different conditions are needed to better understand the reduction of roadbed thickness under dynamic loadings.
plication reached approximately 20,000 times. Then, they continued to increase monotonically with relatively slow rates for the rest of the load application. Settlement in the ballast layer constituted the vast majority of total settlement. The case of Figure 6 was exceptional, but the ballast settlement took up about a half of the total settlement. For both the subballast and the clay layer, the magnitude of settlement approached a relatively small plateau value and the subsequent increments were marginal. The sharp settlement rates of the ballast layer in the early load-
533
4 CONCLUSIONS The effect of geocell reinforcement in improving load-bearing capacity of soft foundation soil was confirmed by plate loading tests. The values of subgrade reaction modulus and strain modulus significantly increased and no inflection points were found in the load intensity versus settlement curves. One layer of geocell mattress with a 0.1 m thick cover was equivalent to a 0.75 m thick fill without geocell. The use of geotextile between the geocell mattress and the underlying soft soil foundation led to reduction in load-bearing capacity. However this adverse effect of geotextile is thought to be instantaneous during construction and overcome under real loading situations. The results of the test performed as a function of the number of geocell mattress and varying thickness of cover indicated that a cost efficient design of geocell reinforcement could be achieved by determining an appropriate thickness of cover. Dynamic loading tests were carried out using a test setup simulating the real loading conditions of high-speed railroads. The settlement rates under dynamic loadings exhibited a consistent pattern. The
534
total settlement increased at sharp rates up to the loading number of 20,000 times and, then, continued to increase monotonically with relatively slow rates for the rest of the load application. Settlement in the uppermost ballast layer constitutes the vast majority of total settlement.
REFERENCES Barthurst, R.J. & Jarrett, P.M. 1988. Large-scale model tests of geocomposite mattresses over peat subgrades. Trarisportation research Record I 188: 28-36. Bush, D.I., Jenner, C.G., & Bassett, R.H. 1990. The design and construction of geocell foundation mattresses supporting embankments over soft ground. Geotextiles and Geomembrcine.s 9: 83-98. Cowland, J.W. & Wong, S.C.K. 1993. Performance of a road embankment on soft clay supported on a geocell mattress foundation Geotextiles ciiicl Geomenibranes 12: 687-705. Presto Products Geosystems & Intersol Engineering Inc. 1977. Manual for the Geoweb Cellular Confinement System: General Reference Material. Rajagopal, K., Krishnaswamy, N.R. & Latha, G.M. 1999. Behavior of sand confined with single and multiple geocells. Geotextiles mid Geonienzkranes 17: 171- i 84.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, lS5N 90 2651 863 3
Behaviour of footings on reinforced sloped fills A.K. Choudhary & B.P. Verrna Depurtment of Civil Engineering, Regional Institute of Technology, Junzshedpur, India
ABSTRACT: A series of bearing capacity tests with footings placed on unreinforced and reinforced slope were performed to investigate the performance of footing located at the crest of a sloped fill influenced by the presence of a reinforcing layer within the body of the fill. Fly ash; a waste product coming out of the thermal power plants was used as fill material. The results of the investigation indicate that fly ash can be satisfactorily used for such geotechnical applications. An analysis has also been proposed to obtain the ultimate bearing capacity of the footing on reinforced sloped fill. Comparison of the theoretical analysis with test results shows good agreement. 1 INTRODUCTION
steeper slope resulting in minimum of material requirement and land acquisition. The beneficial effects of incorporating tensile reinforcement in soil fills has been described by several researchers since the pioneering work by Vidal. As of now; reinforced earth technique has become a powerful construction technique for civil engineers that has demonstrated its value in wide variety of practical applications such as steep reinforcement slopes, retaining walls, shallow slips in clay embankment, embankments on soft soils, working platform and unpaved roads etc. The advent of improved polymer materials has further added a new dimension to the efficiency of design and construction of such reinforced earth structures. The reinforcing techniques concentrate on the use of geogrids in improving the performance (i.e. the load-carrying capacity and settlement) of the abutment fill. Geogrids are essentially synthetic reinforcing mesh structures which have been used to enhance the bearing capacity of soils. Examples such applications are given by Binquet and Lee (1975 a, b), Basset and Last (1978), Akinmusuru and Akinbolade (198 l), Fragaszy and Lawton (1984) and Verma and Char (1986). Although reinforced soil has become popular in recent years, the problem of the behaviour of a footing loaded in the vicinity of the crest of reinforced slope has received only limited attention (Gnanendran & Selvadurai 1989, Huang & Tatsuoka 1994, Manjunath et al. 1994). The scope of the investigation in all these studies was particular to the specific problem in hand and purely experimental in nature. Therefore; in the present investigation, after conducting many model tests, theoretical analysis of the problem has been done and based on theoretical
In some engineering practice, heavy structures such as bridge abutments or other traffic facilities have to be constructed close to the crest of a slope. The estimation of the ultimate load capacity of such foundation is important for safe and efficient design of such foundations (Meyerhof 1963; Winterkorn and Fang 1975, Shields et al. 1977). The positioning of the footing in relation to the edge of the embankment fill is another important aspect; which has implications not only on the safety but also on the economy and efficiency of the overall design of such structures. In view of the escalating cost of construction with the use of conventional materials, utilization of industrial wastes for such major geotechnical application has been highly emphasised during last few decades to bring economy in construction. To overcome difficulties in getting conventional material; fly ash has been successfully used as a structural fill or embankment material in a number of construction projects throughout the world. Difficulty in using fly ash is that it has low bearing capacity. In recent years, reinforced earth technique has been looked upon as a cost-effective solutions to many such soil and foundation engineering problems. Reinforced fly ash embankments shall, therefore; have promising potential in the days to come, where the fly ash will provide the bulk of the mass and the reinforcement will provide the necessary strength to the mass of the geotechnical system. Reinforcement of the embankment in such situation will not only offer better performance in terms of their load supporting capacity but also can be made stable under an externally applied load at relatively 535
analysis, equations have been suggested for determination of ultimate bearin capacity of the reinforced slopes.
2 EXPERIMENTAL PROGRAMME The model tests were conducted in an open ended masonry tank having dimensions of 1.80 m x 1.20 m x 0.30 m with its frontage made of 12 mm thick perspex sheet, which facilitated the observation of failure surface at the end of each experiment. The rigid footing was modelled by a seasoned teak wood section measuring 300 mm x 100 mm (L x B) in plan area. A schematic view of the test facility alongwith the equivalent loading diagram for calculating the load on the footing at failure is shown in Figure 1. The fly ash used for the model tests were procured directly from the electrostatic precipitators of Tata Iron & Steel Company Limited, Jamshedpur. The standard Proctor's density and the corresponding O.M.C. was found to be 9.34 kN/m3 and 48% respectively. The value of apparent cohesion(c) and the angle of internal friction (@)were about 20 kPa and 14' respectively. Commercially available polypropylene model geogrids, 0.27 mm thick and 300 rmn wide having tensile strength equal to 4.0 kN/m and tie-soil friction angle (@p)equal to 35' were used as reinforcing elements. The models were prepared by compacting the fly ash at O.M.C. in layers of 150 mm thickness. The placement dry density was about 90% of the Standard Proctor's density. The compacted fly ash was first built upto the desired height and the reinforcement was placed at the specified location and the filling was then continued upto a height of 750 mm. The longitudinal sides of the test tank were covered with thin and transparent sheets coated with white grease in order to reduce friction between the soil and the sides of the test tank and to induce a state near plane strain in the tested soil mass. The fly ash was compacted to the required level with reinforcing elements buried at specified locations. Extra soil was cut out from one side of the fill with the help of a sharp edged trowel to form the slope of 45'. The model footing was then placed in position at the surface of the compacted fill and then loaded gradually until failure.The corresponding settlements were recorded by using two dial guages placed diagonally opposite. The following parameters were chosen in the present study : 1. The ratio of the distance of the edge of the footing from the crest of the slope (De) to the width of the footing (B) i.e. DJl3. 2. The ratio of embedment depth of the reinforcing layer (Z) to the width of the footing (B) i.e. Z/B. The ratio De/B was varied from 1.0 to 3.0 while the ratio Z/B was varied in the range of 0.25 to 3.0. 536
The length of reinforcing element was kept constant as L = 10B throughout the test programme. 3 THEORETICAL ANALYSIS 3.1 Bearing capacity of unreinforced slopes Of the various theories available for determination of bearing capacity of slopes, the theory presented by Meyerhof (1 957) with certain modifications in context of compacted fly ash slopes seems to be in good agreement with experimental results. The bearing capacity is generally computed for a footing either on top of a slope or on face of slope as
Theoretical bearing capacity calculated with the above equation gave results higher than those obtained experimentally. Therefore, the above equation needs to be modified keeping in view the compressibility of fly ash. In order to take the effect of compressibility of fly ash into acocunt, it is proposed to modify equation ( 1 ) by incorporating the soil compressibility factor in equation (1) as suggested by Vesic (1 973) from the anology of expansion of
cavities. The Meyerhofs equation in its modified form for bearing capacity of footings on slopes may therefore, be written as 0 '
= cNcqFcc+ $4 YBNyqFyc
(2)
0.3
where: N,,, N,, are Meyerhof s bearing capacity factors and are functions of DJB for different inclinations of slope. The variation of N,, and N,, are presented in the form of curves (Meyerhof 1957). y is the unit weight of the soil. Fcc,F,, are soil compressibility factors as proposed by Vesic (1973) and are functions of soil friction angle (@)and rigidity index (Ir). 0 ' is the ultimate bearing capacity for unreinforced fly ash slope.
2 4 0.2
4 4 0.1
3.2 Bearing capacity of reinforced slopes
0 ,
Since the primary objective is to evaluate the efficiency of the reinforcement, it is convenient to present the results for the reinforced soil system with respect to an identical unreinforced system. The most comprehensive work relating to the bearing capacity of reinforced subgrades is presented by Binquet and Lee(1975b) and the same has been extended for the case of footings resting on the top of a slope by modifying some of the assumptions to make it more realistic. The prominent among these assumptions are : 1. At the plane separating the downward and lateral movements of soil mass, the geogrids are assumed to undergo two right angle bend around two frictionless roller. 2. The geogrid-soil friction co-efficient is assumed to vary with respect to depth as per the following relation : fc
= mf
'
3
I
I
51
(3)
0
f = Geogrid-soil friction coeffcient = tan @M f c = Mobilized geogrid-soil friction co-efficient m = Mobilization factor given by;
m = (1.O - Z/B) 0.15 + 0.85 (4.a) for Z/B 5 1.0
(44
(3.0 - Z/B) 0.25 + 0.20 (4.b) for Z/B > 1 .O
(4b)
I
4
I
1
1
2
3
4
z/B
5
Figure 2b. Variation of k,/B & Xo/B with Z/B (after Binquet & Lee, 1975 b).
The corresponding value of tie force (Trj) may be obtained by substituting the value of ultimate bearing capacity for unreinforced slope ('0) into the expression for tie force i.e.
Incorporating the above mobilization factor into the expression for tie pull out frictional resistance of any tie placed at a depth 2 (Binquet & Lee 1975b), the mobilized tie pull out frictional resistance (FB) is given by
FB= 2 f e ( ~ ~ ~ ) [ ~ l ~ ~ q o ( q , / q O ) + y L(L0 - X0)ZI
2 t /B
Figure 2a. Variation of A , , A2 & A3 with Z/B.
where :
=LI
f
f
L
1
Trj=
'o[('R/'Q)
-11 (AiB - AzZ)
(6)
Considering the limiting equilibrium condition and equating the developed tie force (Eq.6) to the mobilized tie pull out frictional resistance (Eq.5) and on rearranging, we get:
(5) qR
= ' 0 + [ ~ ~ ~ ( L D RA~B'R >L{
+y (Lo - X0)ZH / (AlB 537
(7)
where : A I , A2 and A3 are dimensionless forces and Lo and XOdenotes the locii of the points of maximum shear stress and the point where the vertical stress intensity is only I % of the applied bearing pressure. All these quantities are functions of depth ratio (Z/B) and their variation with depth ratio are presented in the form of curves as shown in Figure 2a. and Figure 2b. The term (LDR) refers to linear density ratio and represents the total width of reinforcement ties per unit length of the footing. On substitution of the various values, Equation (7) can be solved for the desired bearing pressure q ~ The . theoretical bearing capacity values, qtll for the various cases are tabulated in Table 1.
crease in the load carrying capacity is primarily due to improved redistribution of load provided by the geogrid. While at greater depth of embedment ((Z/B)>2.0)), the presence of the geogrid reinforcement does not lead to a significant improvement in the load carrying capacity as major portion of failure surface passes through unreinforced portion and consequently, frictional resistance offered by the geogrid is less. The detailed outline of the results are presented in Table 1. From Table 1 it can also be seen that for a given depth ratio, there is a marked increase in the bearing capacity as the edge distance is increased from B to 3B. However, for a given depth ratio there is marginal decrease in the BCR values as the edge distance is increased.
Table 1. Outline of the results.
S1.No 1 2 3 4 5 6 7 8 9
ZIB
U.R 0.25 0.50 0.75 1 .O 1S O 2.0 2.50 3.0
Experimental B.C. q,, (kP,) D, = B D,=2B D,=3B 65 76 88 94 99 107 122 128 140 128 140 154 131 137 149 I14 126 139 104 115 128 91 105 115 82 94 106
D, = B 72.75 87.18 93.63 103.12 113.80 109.73 106.77 99.52 94.30
Theoretical B.C. q,], (W,) D, = 2B D, = 3B 77.71 81.16 97.23 93.1 1 99.93 104.30 109.88 114.57 121.01 126.03 121.08 116.43 113.18 117.64 105.46 109.60 99.97 103.92
Note : U.R = Unreinforced, B.C = Bearing capacity
4 RESULTS AND DISCUSSIONS It is convenient to present results for the reinforced slope with respect to the corresponding results obtained for its unreinforced counterpart. A term bearing capacity ratio(BCR) has, therefore, been introduced to analyse the test data and is defined as follows :
in which; q~ and ‘0 are bearing capacity for the reinforced and unreinforced cases respectively. With the help of proposed analysis, the bearing capacities of footings on reinforced/unreinforced slopes were calculated for varying edge distances and depth ratios. The results obtained analytically are presented in terms of BCR. Finally, the results were compared with the experimentally obtained BCR values for the various cases. The variation of BCR with depth ratio for varying edge distance, are presented in Figure 3. From Figure 3, it can be seen that for any given edge distance, there is a considerable increase in the ultimate bearing capacity of the footing resting on reinforced slope as compared to that of unreinforced slope. However, the increase in load carrying capacity is significant upto a depth ratio of 1.0 and the increase becomes marginal beyong depth ratio of 2.0. At lesser depth of embedment ((Z/B)
538
Figure 3. Variation of BCR with ZIB.
5 CONCLUSIONS 1 . Fly ash can be successfully used even in steep faced embankments and the load carrying capacity of such embankments can be enhanced by 100% by incorporating a geogrid reinforcement 2. The edge distance affects the load carrying capacity of unreinforced as well as reinforced slopes
and load carrying capacity increases with increase in edge distance for both the cases. 3. The optimum location of the geogrid reinforcement is at a depth between 0.5 and 1.0 times the width of the footing. 4. The location of the geogrid layer at a depth greater than twice the width of the footing does not lead to significant improvement in the load carrying capacity. 5. Comparison of the analytically obtained results with model test data shows a good agreement and equation (7) can be used to evaluate bearing capacity of footings resting on reinforced slopes. The results of the present investigation are sufficiently encouraging and may be extended to prototype.
6 ACKNOWLEDGEMENT Authors are thankful to the Principal, R.I.T., Jamshedpur for providing facilities to work in Geotechnical Engineering Laboratory of the Institute, Department of Atomic Energy, Government of India for providing financial assistance and authorities of The Tata Iron & Steel Company Limited, Jamshedpur for providing fly ash from their electrostatic precipitators. REFERENCES Akinmusuru, J.O. & Akinbolade, J.A. 1981 .Stability of loaded footings on reinforced soils. J.Geotechnica1 Engg. Divission, ASCE, vol. 107, No.GT6: 8 I9 - 827. Basset, R.H & Last, N.C. 1978. Reinforcing earth below footings and emabarkments. Proc symp. earth reinforcement, ASCE, Pittsburg: 202-23 1.
Binquet, J. & Lee, K.L. 197%. Bearing capacity tests on reinforced earth slabs. J. Geotechnical Engg. Division,ASCE, Vol. 101 ,No.GTI 2: 1241- 1255. Binquet, J. & Lee K.L.1975b. Bearing capacity analysis of reinforced earth slabs. J. Geotechnical Engg. Division, ASCE, Vol. 101 No.GT12:1257-1276. Choudhary, A.K. & Verma, B.P. 1999. Stability of loaded footings on reinforced fly ash slopes, Proc. Indian Geotechnical ConferenceJGC-99: 145-147. Choudhary, A.K. & Verma, B.P. 2000. Footings on reinforced sloped fills. Proc. Indian Geotechnical Society, IGC-2000: 331 -332. Fragaszy, R.3 & Lowton,E. Bearing capacity of reinforced sand subgrates . J. Geotechnical Engg. Division, ASCE, vol. 110, NO 1 0 : 1500-1507 Huang, C.C., Tatsuoka, F. & Yasuhiko,S. 1994. Failure mechanism of reinforced sand slopes loaded with a footing. Soils and foundations, vo1.34, No.2:27-40. Meyerhof, G.G.. 1957. The ultimate bearing capacity offoundation on slopes. Proc. 4'h ICSMFE, Vol. 1 : 384-386. Meyerhof, G.G. 1963. Some recent research on the bearing capacity of foundation on slopes. Soils & Foundations, 1 :30-37. Selvadurai, A.P.S. & Gnanendran, C.T. 1989. An experimental Study of footings located on a sloped fill, influence of a soil reinforcement layer, Proc. Canadian Geotechnical J., Vol. 261467-473. Shields, D.S., Scott, I.D., Bauer, G.E., Deschenes, J.H. & Barsvary, A.K. 1977. Bearing capacity of foundation near slopes. Proc. 9'h International Conference on Soil Mechanics and Foundation Engineering, Tokyo, Vo1.2:7 15-720. Verma , B.P & Char, A.N.R. 1986. Bearing capacity tests on reinforced sand subgrades. J. Geotechnical Engg. Division ASCE, vol 1 12: 701 -706. Verma, B.P. & Char, A.N.R. 1988. Modelling for Bearing Capacity Analysis of reinforced sand subgrades. Proc. International Symp. on theory and practice of earth reinforcement, Japan: 245-250. Verma, B.P. & Jha, J.N. 1992. Three dimensional model footing tests for improving subgrades below existing footings. Proc. International Symp. on Earth reinforcement practice. Fukuoka, Japan : 707-71 1 Vesic, A.S. 1973. Analysis of ultimate loads of shallow foundations, J. Soil mechanics & foundation Engg., ASCE, V01.99, NO. SMI: 45-73.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
The capacity of reinforced subsoil loaded by uplifted foundation E. Dembicki Ceotechnical Department, Technical University Gdansk, Poland
R. Duszynski Hydraulic Structures and Water Resources Management Department, Technical University Gdansk, Poland
ABSTRACT: The uplift behaviour of mushroom foundations embedded in cohesionless soil media, with and without geosynthetics reinforcement, is investigated in small-scale model tests. Two configurations of geomaterial reinforcement and four types of geomaterials are compared. The most effective is geocomposite located directly at the base of a foundation. Model tests made by authors are presented in details. The model test procedure and results are discussed. bility, and very good interaction between soil and reinforcement.
1 INTRODUCTION 1.1 Foundations There are many engineering structures, which foundations are loaded by high uplifting forces. Typical examples are electric lines towers, radio and TV transmitter towers, high chimneys. It is obvious that uplift capacity of the footing foundation is much lower (approximately equal sum of weights of foundation and soil lying in the region of displacement) than bearing capacity. In that case it is useful because of an economical reason, to strengthen subsoil by designing the construction with some tensile elements. These elements placed at the proper depth around the foundation allow transferring much higher uplifting loads with the appropriate factor of safety. Many parameters have an effect on decision what types of foundations use. Mainly under the influence of soil type and construction characteristic foundation is chosen. The most popular in Poland is mushroom foundation. This kind of foundation has many advantages: - could be prefabricated in a few sizes, - could be mounted directly on site, - is much cheaper then other types of foundations, - has comparatively high capacity.
2 MODELTESTS
2.1 Test facility Over 80 model tests with 2 types of foundations and 4 types of geosynthetics reinforcement were conducted. Tests were made in a wooden-steel box with dimensions 185 cm x 230 cm and 120 cm depth. Pullout loads was effected through the steel rode articulated to the top of foundation (Fig. I). The foundation was pulled out with displacement rate of 0.43 mdsec. A load cell of capacity 6000 N and accuracy ~f:0.1 % was mounted between the model of foundation and pullout steel rod. Displace-
1.2 Geosynthethics reinforcement Application of geosynthetics inclusions is a very well known method of soil reinforcement that increases the resistance of soil due to interaction of soil and tensile materials. Both foundations and anchors can be loaded by higher pullout forces due to use of geosynthetics reinforcement, which has got high mechanical and chemical resistance, high dura-
Figure 1. Test facility.
54 1
ment was recorded by two gauges with 100 mm measurement range and accuracy & 0.1 %. All data was recorded in the intervals of 1 sec with PCcomputer. 2.2 Models of rizushroornfoundutions Two types of aluminium foundation models were used. The base part of foundation has diameter of 20 cm, the shaft has diameter of 8 cm. Both parts were jointed with four screws. One model has 40 cm height the other one has 60 cm. Model geometry is shown in Figure 2. All dimensions are presented in Table 1. Table 1. Dimensions of foundations model. Model
A B
H cm 40 60
Dimensions r d cm cm 3.5 4 3.5 4
Weight R cm 10 10
N 74 98
Figure 3. Types of reinforcement used. Table 2. Properties of geomaterials used as reinforcement. Material
Aperture Thickness Weight
mm 20x20 Geogrid Geotextile ----Geonet 6x6 Geocomposite 6x6
mm 1.37 0.47 2.50 5.47
g/m' 250 200 450 1100
Max strength along across kN/m 35.0 20.1 34.82 22.1 3.28 2.26 34.82 22.1
Two schemes of reinforcement were adopted (Fig. 4). In the first scheme the reinforcement was placed directly at the base of a mushroom foundation. In the second scheme, the reinforcement was placed at the height H/5 over the base at the layer of soil. load cell
4
foundation.
Figure 2. The foundation model.
geosynthetic
2.3 Geosynthetics reinforcement The sheet of geomaterial used as reinforcement has circle shape with outer diameter of 50 cm. There was a circle hole with diameter of 8 cm in the centre of material that allows putting the reinforcement around the foundation shaft (Fig. 3). The diameter of reinforcement was chosen after analysis of soil geotextile interaction, which provides the proper anchorage of geomaterial in the soil. Four kinds of geomaterials were used as reinforcement. First three were fabricated geomaterials, the fourth one was geocomposite made of two geonets and one layer of geotextile between them (Table 2).
542
Scheme I
Scheme II
Figure 4. The manners of reinforcement use.
2.4 Test mediurn Uniformly graded Baltic Sea sand "Lubiatowo" was chosen as the test medium. Sand samples were laboratory tested. The properties of chosen medium are presented in Table 3. To fill the box the technique of sand raining through the travelling funnel was employed in order to achieve homogeneous sand beds around foundation models.
Table 3. Properties of sand "Lubiatowo". Parameter Maximum dry unit weight Minimum dry unit weight Effective grain size Uniformity coefficient Natural water content
Symbol ydlnnh (kN/m-') ydmlIl(kN/m3) DlO (mm) C,,
The comparison of the various types of the reinforcement show that the higher uplift resistance is
Value obtained with the geocomposite. The use of the 17.5 geonet or the geogrid gave the smallest increase of 14.3 0.152 uplift resistance. This relation is because of the interI .43 action between the soil and the geomaterial. The Wll (%I 0.1 1 geocomposite composed of two sheets of the geonet 7 (de.1 27
3 TEST PROCEDURE
A model of a foundation was placed in the box on the 25 cm thick layer of sand. The sand bed around the foundation was prepared by sand raining technique. Next the load cell, displacement gauges and loading system were connected to the top of foundation. The pull out displacement was controlled by the computer and has value 0.43 &sec. The same procedure was adopted for tests with the reinforcement. First the model was positioned in the box, then the reinforcement was placed around the base of the foundation. Next the sand bed was prepared. All recorded parameters like pull out force and total displacements were measured and recorded by PC-Computer every one second.
and one sheet of the geotextile between them has the best parameters such a friction between the soil and the material surface. Tests results for both height of model foundations and for the reinforcement placed on the base of foundation (diagram 11) are presented on Figures 5 and 6. The values of uplift resistance for both model height and for each reinforcement placed (diagram I1 and 111) are presented in Table 4. Table 4. Results of tests on the mushroom foundation with and without reinforcement. Model
Geomaterial reinforcement Grid/Net Textile Composite A 11;k 832 1280 1460 A III:** 748 740 750 B I1 1616 1885 2474 2734 B 111 1689 1678 1655 * Foundation model H=40 cm, reinforcement on the base of foundation; *'!: Foundation model H=60 cm, reinforcement on the base of foundation
4 RESULTS AND DISCUSSION As a result of model tests, values of pullout loads and displacements of the foundation were obtained. Recorded data were used to plot load-displacement curves for each considered model. The diagrams are described below: Diagram I: No reinforcement around the foundation. Diagram 11: Layer of the geomaterial reinforcement placed directly at the base of the foundation. Diagram 111: Layer of the geomaterial reinforcement placed at depth H/5 over the base of the foundation. An analysis of results shows, that for both height of foundation models (40 cm and 60 cm), the behaviour of the model embedded in the nonreinforced sand was similar to that which is embedded in sand reinforced with geotextile layer placed at depth H/5 over the base of the foundation (diagram 111). There was no noticeable increase of uplifted foundation capacity due to such reinforcement. The load-displacement curves have the same character. The load-displacement curves of the models with any type of the geomaterial reinforcement (diagram 11) differed significantly from those in unreinforced or reinforced by diagram I11 conditions. For each type of reinforcement the uplift resistance of foundation increase more rapidly than for the unreinforced case. The peak value of resistance is reached at the higher value of displacement. 543
Unreinforced Q (N) 71 1
5 CONCLUSIONS
The observations and conclusions drawn in this study are valid only for the type of foundations and reinforcement proposed by the authors.
Based on the laboratory investigations carried out on model mushroom foundations with two different height embedded in sand without and with four types of reinforcement placed in two different layers, the following main conclusions are drawn: the load displacement relationship in the unreinforced case is similar to that observed in the reinforced one, the reinforcement placed directly at the base of foundation give noticeable increase of uplift resistance, the geocomposite reinforcement was the most effective in enhancing uplift resistance of mushroom foundations embedded in reinforced soil, the maximum increase in capacity was over 200% of that in unreinforced conditions for the geocomposite placed on the base of the foundation with 40 cm height.
REFERENCES Bolt A. 1998. The modelling of foundations for supporting structures. XLVI Scientific Paper of Technical University of Gdansk. Gdansk: TUG. Dembicki E. & Duszynski R & Sokolowski P. 1998. The influence of soil reinforcement on the uplift capacity of mushroom foundations (in Polish). Structure subsoil collaboration; Proc. intern. sytnp., Wigiy, 18-20 June 1998. Biaiystok. Dembicki E. & Duszynski R. 1998. Laboratory test on e f i ciency of geomaterial reinforcement of soil loaded by an uplifted mushroom foundation. 1998. (in Polish). The Maritime Engineering and Geotechniques, September 1998. Parashar S.P. & Krishnaswamy N.R. 1994. Plate Anchors with Geosynthetics. Fifth International Conference on Geotextiles, Geomembranes and Related Products; Proc. intern. symp Singapore, 5-9 September 1994.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Practical aspects of the design of deep geotextile coated sand columns for the foundation of a dike on very soft soils M. Geduhn & M. Raithel
Engineering ofice Kempfert + Partner Geotechnik, Germany
H.-G. Kempfert Institute of Geotechnique, University of Kassel, Germany
ABSTRACT: This paper presents an investigation of a fundamental suitability of a new foundation system “geotextile coated sand columns“ (GCC/GSM) to the foundation of a dike on very soft soils (sludge). This bearing system is a further development of the already known coluinn foundations such as vibro displacement piles and granular piles. In the future, the foundation system GCC/GSM is intended to be used for water engineering projects in very soft soils. 1 INTRODUCTION
Table 1. Decisives model laws (index m=model, n=prototype)
The foundation system “geotextile coated sand columns” (GCC/GSM) is a further development of the already known column-foundations such as vibro displacement piles and granular piles. In contrary to conventional column foundations, coated columns can be used as a ground improvement in very soft soils, for example peat or sludge (undrained shear strength cu < 15 kN/m2). On the basis of the known procedure for calculation and dimensioning of gravel and sand columns an analytical calculation model has been developed taking the geotextile coating into account (Raithel & Kempfert 1999). The new foundation system has proved worth as regards many traffic-road and railway projects in Germany and the Netherlands. In the future, this bearing system is intended to be used for water engineering projects in very soft soils. For this reason, the fundamental suitability of the above mentioned GCCSystem to a dike foundation on a very soft soil (sludge) as a basic foundation system has been investigated.
Term Settlement
unit cm
model law s,= l/h .s,l
Geotextil stiffness
kN/m
J,, = Ilk2 . J,
Force
kN
P,,=
Load
kN/ni2
qn= l/h
11x3
. P, ‘
(3,
2.1.2 Soil mechanical parameters and geotextil The test soils were a very soft soil called sludge and a medium grained sand. The soil parameters are shown in table 2. The coating in the small scale model test was a synthetic material of poly-ester with a stiffness of J = 20 kN/m. The true scale stiffness of the geotextil coating is J = 2000 kN/m. Table 2. Soil mechanical parameters
r‘lr
w
cp‘
C’ Eoed [kN/m2] [kN/mZ]
[kN/m3]
[%I]
[“I
6 / 16
65
34
5
10/18
0.7
38
2
Soil type Sludge
2 SMALL SCALE MODEL TESTS (SCALE 1:10)
Sand
1800” 28500
2.1 Model test boundary conditions 2.1.1 Model theory Because of the principle of similarity, mechanical behavor of soil can be investigated in a small scale. To obtain the scaling law relationships de-cisive parameters of the system have to be derived with the dimensional analysis to dimensionless pro-ducts, which must have the same value in both systems (model scale to true scale) (Gortler 1975). Hence these model laws are decisive:
2.2 Rotational symmetric nzodel tests On basis of the present knowledge of the small scale model tests for the GCC-foundation system (Raithel 1999), 8 rotational symmetric model tests at a scale of 1:lO had been carried out under static loading. The concept of the rotational symmetric model tests was based on the “unit cell concept”, which consid-
545
ers a single column in a virtual infinite column grid. The influence area of a single column in a triangular grid is a hexagonal element, which can be transformed into a circular element with an equivalent area. Therefore an axially symmetric model is received. According to the assumption that there is a geotextile coated columns grid of 15 % and a columns diameter of d = 0.80 m in practice, it follows in small scale model test that the ratio of the influence area AE to the column area AS to As/AE= 50.5 cm* / 336.6 cm2 = 0.15 = 15%, and the model column diameter was d = 0.08m. Because there exist almost a linear relation between settlement and column length, the length is insignificant in the model test. To analyse the bearing and deformation behavior of the foundation system a column length of 0.5 m was taken, assuming a 5 m column length and sludge thickness in nature.
2.2.1 Results of the rotational symmetric model tests Figure 1 shows the settlements with and without GCC-system. Due to the interaction between cover filling, columns, geotextile and soft layer, the foundation system shows flexible and self regulating bearing behavior. The results of the static loading tests proved the self regulating bearing behavior of the system which is due to the various interactions. = Settlement without GCC / SettleThe factorp ent with GCC) of ground improvement in very soft soil amounts t o p = 2.6. Comparing the results of analytical approach with the model investigations (Figure 2), it can be shown that the analytical method (Raithel / Kempfert 2000) can provide a good approximation to measured values of the settlement. The differences are due to the small scale effect caused by transferring the cohesive behavior of the soft soil. The radial support is guaranteed 0
5
10
Load [kN/m*] 15
20
25
30
0
50
100
Load 150
[kN/11i2]
200
250
300
350
0
Figure 2. Comparison with the settlement in model test relating to true scale and with the analytical model calculation.
through the composite between the coating and the surrounding soft soil, because the geotextile is under ring tension force. In addition, the shear strength of the surrounding sludge were measured. Table 3 shows the results of shear strenght measurements before and after loading the GCC-system. Table 3 shows an increase of the shear strength of the sludge after loading by an average factor of 2. Therefore, the developed foundation system GCC/ GSM offers a possibility of an enormous settlement reduction, accelerates the settlement rate and improves the shear strength of the surrounding very soft soil. 2.3 Model tests for columns installation method There are two construction methods in practice have: the boring method and the vibro displacement method. The vibro displacement method is more economical and ecological friendly. But the compaction of the soft soil under vibration may lead to soil
35
0.0 0.5 I .0
E
I
2.0
--
2.5
-+
6
5
c i)
2
Table 3. Results of shear strength [kN/m2] measurements of the sludge before and after loading
1.5 --
Test number
3.0 -3.5
--
4.0
--
Figure 1. Settlements with and without GCC in rotational model tests scale 1:lO under static loading.
546
mean value
before loading
after loading
C"
47
C"
2.4 4.9 4.1 7.9 7.8 7.4 4.0 4.0
7.0 7.0 16.0 15.0 14.0 25.0 3.0 4.0
8.0 10.0 11.0 16.5 13.0 10.0 12.0 5.5
9, 26.0 19.0 21.0 26.0 26.0 26.0 15.0 7.0
5.3
11.0
11.0
21.0
deformations. Moreover, the effect of local liquefaction of the surrounding soft soil is not investigated at present. Therefore, 4 model tests for columns installation method in sludge, scale l :10, under vibration had been carried out in a geotextile coated columns grid (As / AE = 15%). Using the dimensional analysis, the measurement results of the small scale model tests and the in situ situation are directly transferable. In Table 4 characteristic values are shown for the vibro displacement method in practice and in model test. The vibration will be produced from two eccentric masses which work in reverse rotation under an angular velocity. The dimensions of the test container were: length 2.0 m, width 0.8 m and height 1.3 m. The sludge thickness was 0.5 m. Figure 3 shows the top view of the principle model boundary conditions. The deformations at the surface were measured after the columns had been installed in rows. Addition-ally, the shear strength at different depths were measured before and after 6 rows of columns had been installed.
3. An obvious liquefaction of the soft soil by the compaction energy was not observed. 4. A skew adjustment of the columns in sludge as a result of the column installation under vibrations had not been manifested. To present the first 2 above mentioned effects clearly, a typical deformation is shown at the surface is in Figure 4. The heaving of the surface of the sludge as shown in Figure 4 can be expressed as percentage of the column depth (C = 0.50 m). Table 5 shows the results of all 4 model tests as a mean matrix value. From Figure 4 and Table 5 , it can be shown that heaving of the soft soil between the columns rows due to the compaction energy is expected to be up to 4% of the columns depth. The heave can be increase up to 8% of the columns depth by further installation of additional rows of columns. Table 6 shows the results of shear strength measurements at different depths. The shear strenght in sludge before the installation of the columns in comparison with after 6 rows of columns had been installed shows a increase by a factor of 2.5 to 4.5. To summerise the results of the model test: the effect of liquefaction of the soft soil by the compaction energy was not observed. Further, an increase in the shear strenght of the surrounding soft soil by an average factor of 3.5 was measured, which shows the stabilization effect of the installation method.
2.3.1 Results of the model tests for columns installation method During the installation of the 15% geotextile coated column grid, the following effects were observed:
1. The compaction by vibration lead to the heaving of the soft soil between the column grid and before installing the grid. 2. The heaving of the soft soil produced wave like deformations at the surface of the grid. Table 4. Characteristic values for the vibro displacement method in practice (true scale) and in model test magnitude
true scale
model scale
transmission
[min.']
2400.0
static moment
[Nm]
340.0
0.34
2 140.0
2.14
centrifugal force [ kN] weight
ltl
frequency range [Hz]
2400.0
6.0
0.006
30 - 40
30 - 40
Figure 4. Heaving of the surface of the very soft soil after 3 rows of columns had been built.
Table 5 . Heave [cm] at the surface of the sludge as a mean matrix value.
Rowofcolumn 1
2
3
4
5
6
build in 1. row buildin2. row build in 3. row build in4. row build in 5. row build in 6. row
1.7 1.9 1.9 1.9 1.9 1.9
1.1 2.4 2.6 2.6 2.7 2.7
0.6 1.4 2.6 3.3 3.5 3.5
0.2 0.6 1.8 3.4 4.0 4.3
0.2 0.3 1.0 3.2 3.7 3.7
0.1 0.1 0.4 1.0 2.4 3.7
increase [%]
3.8
5.4
7,O
8.6
7.4
7.4
*)
"reference to the column depth ( I = 0.50 m)
Figure 3. The top view of the principle model boundaries of the test container with finished column grid As 1 AE= 15%.
547
Table 6. Results of shear strength c, [kN/m2] measurements in the sludge at different depths before and after the installation of sand columns depth [cm] c, before GCC c, after GCC
10
25
40
1.7 5.9
I .7 4.5
1.6 7.2
3 TRUE SCALE TESTS At present the foundation of a dike on very soft soils by using the foundation system “geotextile coated sand columns” (GCC/GSM) will be realized at the river “Elbe” in Germany (Hamburg). Hence, an investigation by a true scale test in a german harbour was conducted by the company “Mobius BauGesellschaft GmbH & Co.” (Mobius, unpupl.). 30 geotextile coated sand columns with a diameter of d = 0.80 m in a columns grid of 20% were installed. The columns depth was C = 5.0 m on average. The vibro displacement method was chosen from a offshore pontoon. The soil parameters of the very soft soil are comparable with the model test soil shown in table 2. During the manufacturing of the 20% geotextile coated column grid, the following effects were observed: E
H
E
The compaction by vibration leads to the heaving of the soft soil between the columns grid. It was estimated up to 2-4% of the columns depth. A skew adjustment of the columns in sludge as a result of the column installation under vibrations had not been manifested. An obvious liquefaction of the soft soil by the compaction energy could not be disco-vered. The average factor of the increase in the shear strenght is 1.5.
The investigation in the true scale tests confirms the results of the small scale model tests. Based on the true scale tests and the model tests (scale 1:10) results the foundation system GCC/GSM is practicable for water engineering projects in connection with very soft soils, e.g. sludge.
4 SUMMARY On basis of the present knowledge of small scale model tests for the foundation system “geotextile
548
coated sand columns” (GCC/GSM) two kinds of model tests at a scale 1:lO had been carried out in very soft soil. The 8 rotational symmetric model tests under static loading proven the possibility of an enormous settlement reduction, acceleration of the settlement rate and increase in the shear strength of the surrounding very soft soil. It can be shown that by using the analytical model a good approximation of the settlement measurements can be obtained. The radial support is guarantee through the comfined effect of the coating and the surrounding soft soil, because the geotextile is under ring tension force. On the basis of interaction between cover filling, columns, geotextile and soft soil layer, the foundation system shows flexible and self regulating bearing behavior. The 4 columns installation method model tests under vibrations show that the compaction by vibration leads to the heaving of the soft soil between the columns grid. An effect of liquefaction of the soft soil by the compacting energy was not be observed. Rather an increase in the shear strength of the surrounding soft soil with an average factor of 3.5 was measured, which shows the stabilisation effect of the installation method. A skew adjustment of the columns in sludge as a result of vibrations has not been manifested. Moreover, an investigation in the true scale test confirms the above mentioned results of the small scale tests. From the results of the true and small model test, it can be concluded that the new foundation system “GCC/GSM” can be practicably applied for water engineering projects on very soft soils. REFERENCES Gortler, H. 1975. Dinzen.sionsana1y.se. Ingenieurwissenschaftliche Bibliothek. Berlin. Springer Verlag. Mobius, 2000. Geokunststoffummantelte Sandsaulen System Mobius (GSM) im Wasserbau. Unpublished. Raithel, M. & Kempfert, H.-G. 1999. Beinessung votz geokunststoffumnwntelterz Saizdsiiuleiz. Bautechnik 76. Heft 1 1: 983991. Raithel, M. 1999. Zum Trag- und Verforiizuizg.svel.l?alten von geokuizststc~ffummantelteizSandsauleiz. Schriftenrei he Geotechnik. Universitat Gh Kassel. Heft 6. Juli 1999. Raithel, M. & Kempfert, H.-G. 2000. Calculation Models f o r Darn Foundations with Geotextile Coated Sand Columns. An International Conference on Geotechnical & Geological Engineering GeoEng 2000. Melbourne.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Bearing characteristics of clay reinforced by a sandwiched geogrid-sand system H. Ghiassian & M. Jahannia Civil Engineering Department, Iran University of Science and Technology, Iran
ABSTRACT: A study of the bearing capacity and compressibility characteristics of a cohesive soil, reinforced by a geogrid layer and supporting square footing loads, has been conducted. The lack of adequate frictional resistance between the clay and the reinforcing element was compensated by using a thin sand layer encapsulating the geogrid. In this way, the tensile force induced in the geogrid was transferred to the bulk clay medium through the sand particles and the soil reinforcement was accomplished. Experiments were conduced on specimens with 0.15 x 0.15 x 0.15 m dimension, the footing size of 3.7 x 3.7 cm, which was loaded under a strain controlled condition. All specimens were saturated and presumably loaded under an undrained condition. The results confirmed the effectiveness of the sand layer in improving the bearing capacity and settlement characteristics of the footing. A comparison was also made between the theoretical and experimental bearing capacity values for unreinforced cases, indicating that the experimental values are smaller than the theoretical predictions, in which Terzaghi's equation showed a closer prediction than that by the general bearing capacity equation. Depth ratio of the first reinforcement layer (u/B) Vertical spacing between reinforcement layers (h/B) Number of reinforcement layers (N) Footing dimension (B, L) Geogrid layer dimension ratio (b/B, 1/B) Geogrid modulus Relative density of sand.
1 INTRODUCTION More advances in soil reinforcement techniques and applications, iterested researchers in examining the reinforced soil under footings and foundations. Binquet and Lee (1975) reported an early study on the bearing capacity of shallow foundations placed on a reinforced ground. They presented design charts for strip footings on underlying sandy layers reinforced by metal strips. Although this technique appeared economical compared to other improvement methods, the corrosion of metal reinforcing elements convinced researchers to use other materials such as geosynthetics which perform more satisfactorily. The use of geosynthetics for the reinforced soil under footings is not as extensive as for other applications such as retaining walls, embankments, etc. More research have been conducted on the bearing sand as by Guido et al. (1985, 1986, 1987), Khing, et al. (1992, 1993), Takemura et al. (1992), Omar et al. (1993 a, b), Yetimoglo et al. (1994), and Adams and Collin (1997). Few studies on the bearing clay are also reported in the literature by Ingold and Miller (1982), Milligan and Love (1984), Sakti and Das (1987), Dawson and Lee (1988), Das (1989), Sah (1990), Mandal and Sah (1992), and Shin et al. (1993). The use of geogrids was mainly favored in these studies due to their relatively high modulus compared to geotextiles. According to these studies, the bearing capacity depends on the following parameters:
The results of the above studies show that the bearing capacity of a footing for the case of reinforced soil is 1.8 to 4 times that of the footing for the unreinforced case. Two studies for square and strip footings on clay, which was reinforced by geogrids, are reported by Mandal and Sah (1992), and Shin et al. (1993) respectively. These results show that the improvement of the bearing capacity for two reinforced and unreinforced cases is a ratio about 1.5, much smaller value than that obtained for sand. According to Shin's explanation for this difference, the bearing capacity increase is basically coming from two sources; the increase of soil modulus and pullout resistance of reinforcement. The pullout resistance is composed of two phenomena; first, the frictional resistance between the soil and the geogrid, and second, the passive resistance developed between the transversal strips of the geogrid and the soil. These mobilized resistances, particularly the passive resistance, depend on the friction angle of soil. Therefore, the pullout resistance of geogrid embedded in satu549
Table 1. Summary of tests (B = 3.7 cm, u/B= 0.4, b/B=4).
rated clay at the undrained condition (@=O condition) is normally expected to be smaller than that in sand. It appears that if the lack of sufficient pullout resistance in clay could somehow be resolved, more improvement in the bearing capacity from the reinforcement should be accomplished. A study by Sridharan et al. (1991) on the pullout resistance of metal bars, which were confined by a thin cylindrical layer of sand placed in a cohesive medium, showed that a 15 mm diameter sand layer was sufficient to increase the pullout resistance to the amount as if the whole medium were composed of sand. Similarly, it would be expected if a reinforcing element is sandwiched between a thin sand layer and used in a clay soil beneath a footing, the bearing capacity of the footing should increase due to the improvement of the pullout resistance of the element. The study presented here examines this hypothesis.
B-I B-2 B-3 B -4 B-5
2 TEST MATERIALS The type of soil used in this study was clay with 70% passing the No. 200 sieve. The Unified Classification of the soil was CL, and other properties were determined to be LL=26%, PI=8%, G,=2.67. A clean sand was used to encapsulate the reinforcing element that was a layer of geogrid made by Huesker Synthetic Company and labeled Fortrac Geogrid 55/30-20. Its tensile strength was 55 W / m at 12.5% strain, and with an aperture size of 20 x 20 mm.
I
0
I
0.5 0 0.5 0
U U
R R R
All specimens were reinforced by one layer of the geogrid with the dimension of b=4B (B=3.7 cm) and the depth of u=0.4B measured from the bottom of the footing. These numbers were selected according to the study by Shin and et al. (1993). Table 1 shows the information about the performed tests. Each test was proceeded until a clear drop in the axial load occurred or the total settlement of the plate reached to 1.5 cm. In test B-5, instead of using the sand lens, the geogrid was nailed to the clay (see Fig. 2) in order to increase its pullout capacity and evaluate its effects on the bearing capacity of the footing. It should be noted that in all tests, the geogrid reached to its pullout capacity, far less than its tensile strength. At the end of each test, three soil samples were taken by a 1.5 inch sampler from the section bellow the geogrid for determination of the unconfined compression strength, and the degree of saturation of the specimen.
5 DISCUSSION The load-deformation characteristic and the type of instability were investigated in each test. Also, the type of failure in the geogrid layer was examined after removing the overlying soil layer; that appeared to be a pullout failure in all tests. As shown in Table 1, five tests were performed. Two dimensionless parameters; BCR, (bearing capacity ratio at ultimate load), and PRS (percentage reduction in settlement), as introduced in previous studies, are used here for analyzing the results:
3 TEST APPARATUS The model footing was a steel square plate with 3.7 cm width and 1 cm thickness loaded under the strain controlled condition in an unconfined compression test equipment. The load was applied at the rate of 1.5 mm/min on the plate. The soil specimen was prepared in a cubic metal box with the dimension of 15 cm. This box was made sufficiently rigid to prevent any important lateral deformation on its vertical sides under the footing loads. The total thickness (height) of the clay layer was 12 cm, a little more than three times the model footing width, therefore minimizing the boundary condition effects. All tests were performed on saturated clay specimens.
4,l
4 1
qu = ultimate bearing capacity of footing on unreinforced soil qu(R) = ultimate bearing capacity of footing on reinforced soil S, = settlement of footing on unreinforced soil at pressure of ql, S, = settlement of footing on reinforced soil at pressure of qu.
4 SPECIMEN PREPARATION
Figure 1 presents the load-settlement plots of the tests. The ultimate bearing capacity value for each test was obtained based on the method proposed by Brand et al. (1972), which defines the ultimate load at a point where the plot of bearing load against the settlement becomes practically linear. According to these curves, the type of failure is expected to be the local shear or punching because there is no distinct
The dry clay was first mixed with 22% water (a water content between the plastic and liquid limits). The mix was then allowed to cure in a plastic bag for a week so that water could thoroughly be distributed in the clay, and a homogenous soil sample be made. The soil was then compacted in 2cm layers to a predetermined dry unit weight of 16.48 kN/m3.
550
Figure 2. The use of nails for increasing the pullout resistance of geogrid.
Figure 1. Load-settlement results of the footing on reinforced saturated clay. Table 2. Summary of test results.
S,/B=27
B-3 ~B-4
B-5
0.5 0
76 70
1.49 1.37
2.5 24
14.5 16
46.3 40.7
break on the curves. Observations also confirmed this failure mode as there was very slight soil swell around the footing, and the plate just penetrated into the underlying soil. Table 2 the results obtained from Fig. 1. As can be realized from the comparison between the results of tests B-2 and B-1 , the 0.5 cm sand lens without the geogrid has a negligible effect on the bearing capacity (BCR,=1.06). Therefore, the main contribution of the sand lens on the bearing capacity can be attributed to the improvement of the pullout resistance of the geogrid, as can be seen from the cornparison between the results of tests B-3 and B-4. In test B-3 in which only the geogrid is used, the bearing capacity has increased 25% and the settlement has decreased 33% (Le., BCR,=1.25, PRS=33). But, in test B-4, with a thin layer of sand placed around the geogrid, the results have improved to 49% increase in the bearing capacity and 46% decrease in the settlement (BCR,=l.49, PRS=46.3). Consequently, the presence of the sand layer around the reinforcing element appears to have important effects on both the bearing capacity and settlement characteristics of the reinforced saturated clay. In order to evaluate the effect of the frictional resistance of the geogrid and its anchorage in the clay, test B-5 was conducted. Clearly, if the ultimate tensile strength of the reinforcing element is mobilized, the maximum benefit will result. Otherwise, the amount of the increased bearing capacity will depend on the amount of the frictional strength mobilized at the soil-element interface. These forces, in turn, depend on the effective length and dimension of the element, as well as the interface friction angle 55 1
and type of the element. In this test, an idea was examined to increase the pullout resistance of the element. Instead of using the sand layer around the geogrid, 20 nails with 8 cm length and 4 mm diameter were uniformly and vertically inserted through the openings at the perimeter of the geogrid, and embedded inside the soil specimen as shown in Fig. 2. The objective was to create some passive resistances against the geogrid displacement, and consequently, to increase the frictional resistance of the element. The inspection of the soil specimen and the geogrid at the end of the test revealed that the displacement of the geogrid did occur despite the nail resistance, and the bearing capacity increased 37% and the settlement decreased about 40%, indicating that the sand layer had performed more effectively than the nails. The results in Fig. 1 also show that the ratios of the ultimate settlement (S/B) in the reinforced and unreinforced cases are almost identical varying between 24% and 27%. This result is similar to those reported by Shin et al. (1 993). It should be noted that since the size of the square footing (3.7 cm) is fairly small compared to the grid aperture of the geogrid (2 cm), it would be expected that significant scale effects influence the results. Therefore, it is probably not possible to obtain quantitative information from these tests. However, the influence of the sand lens on the improved bearing capacity and settlement characteristics of the footing can be realized in the results. The Mechanism of failure in all tests was the punching failure. This behavior was observed during the tests, and also reflected on the trend of the plots in Fig. 1. No distinctive swell was observed around the footing in any test, and the soil underneath the footing settled vertically downward along with the overlying plate. 6 BEARING CAPACITY PREDICTIONS A comparison was made between the experimental results of the bearing capacity of the footing placed on an unreinforced soil with the theoretical predictions
Terzaphi's Equation 11
2 3
= 1.3<:-+
c,,* N,. ,
N , =5.7
(b=O
General Bearing Capacity Equation (Das, 1995)
q,, = 1.2* C,, * N c , qu (Experiment, Test B-1) qu (Terzaghi's equation) q,(General equation)
N,. =5.14
(b=O
=51 kPa = 62 kPa = 74 Wa.
The theoretical expressions predict more bearing capacities than the experimental value with the Terzaghi's equation being in a better agreement.
7 CONCLUSIONS In order to increase the pullout resistance of geogrid layer embedded in a reinforced bearing saturated clay medium, an idea of placing a sandy soil around the reinforcing element (geogrid) was experimentally examined. The following conclusions can be made from this study: The existence of the sand layer around the geogrid has pronounced effects on the increase of the bearing capacity and decrease of the settlement at the failure. The presence of the sand layer itself (no geogrid) has a negligible effect on the bearing capacity. The presence of the sand layer does not change the settlement at the failure considerably. The failure mode is the punching failure with little if any surface heave. The theoretical predictions of the bearing capacity overestimate the experimental values with Terzaghi's equation giving a closer answer. REFERENCES Adams, M. T., and Collin, J. G. (1997). "Large model spread footing load tests on geosynthetic reinforced soil foundations.", Jour. of Geotech. and Geoeiivir. Engrg., ASCE, Vol. 123, NO. 1, PP. 66-72. Binquet, J., and Lee, K. L. (197%). "Bearing capacity tests on reinforced earth mass.", Jour. of Geotech. Eiigrg., ASCE, Vol. 101, NO. GT12, PP. 1241-1255. Binquet, J., and Lee, K. L. (1975a). "Bearing capacity tests on reinforced earth mass.", Jour. of Geotech. Engrg., ASCE, Vol. 101, NO. GT12, PP. 1257-12 Brand, E. W., Muctabhant, C., and Taechathummarak, A. (1972). "Load test on small foundation in soft clay.", Proc. Specialty Conference on Performarice of Earth-Supported Structures, ASCE, Vol. 1 , Part. 2, PP. 903-928. Das, B. M., (1989). "Foundation on sand underlain by soft clay with geotextile at sand-clay interface.", Geosynthetic'89 Conf., Vol. 1, San Diego, PP. 203-2 13.
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Das, B. M., ( 1995). "Principles of foundation engineering.", 3rd Editions, PWS Publishing Company, Boston. Dawson, A., and lee, R. (1988). "Full scale foundation trials on grid reinforced clay." Geotech. Special Publication, No. 18, ASCE, New York, PP. 127-147. Guido, V. A., Biesiadecki, G. L., and Sullivan, M. J. (1985). "Bearing capacity of a geotextile reinforced foundation.", Proc. 11th Int. Con$ Soil Mech. and Found. Engrg., San Francisco, Vol. 3, PP. 1777-1780. Guido, V. A., Dong, K. G., and Sweeny, A. (1986). "Comparison of geogrid and geotextile reinforced earth slabs", Can. JOUKof Geotech.,Vol. 23, No. 1, PP. 435-440. Guido, V. A., Knueppel, J. d., and Sweeny, M. A. (1987). "Plate loading tests on geogrid-reinforced earth slabs.", Proc. Geo,iyntlzetics'87Coizf., New Orleans, PP. 2 16-225. Ingold, T. S., and Miles, K. S. (1982). "Analytical and laboratory investigations of reinforced clay.", Proc. ofthe 2nd Int. Con$ on Geotextiles, Vol. 3, Las Vegas, PP. 587-592. Khing, K. H., Das, B. M., Puri, V. K., Cook, E. E., and Yen, S. C. ( I 992). "Bearing capacity of two closely-spaced strip foundations on geogrid-reinforced sand.", Proc. Earth Reinforced Pract., Balkenia, Kyushu, Japan, PP. 6 19-624. Khing, K. H., Das, B. M., Puri, V. K., Cook, E. E., and Yen, S. C. (1993). "The bearing capacity of a strip foundation on geogrid-reinforced sand.", Jour. of Geotextiles and Geomenibranes, Vol. 12, No. 4, PP. 351-361. Mandal, J. N., and Sah, H. S. (1992). "Bearing capacity tests on geogrid-reinforced clay.", Jour. qf Geotextiles and Geomernbraiies, Vol. l l , PP. 327-333. Milligan, G. W. E., and Love, J. P. (1984). "Model testing of geogrids under an aggregate layer on soft ground.", Proc. Synz. on Polymer Grid Reinforcement in Civil Eng., Netlon, London, No. 4.2. Omar, M. T., Das, B. M., Puri, V. K., Cook, E. E., and Yen, S. C. (19934. "Ultimate bearing capacity of rectangular foundations on geogrid-reinforced sand.", Geotech. Testing Journal, ASTM, Vol. 16, No. 2, PP. 246-252. Omar, M. T., Da5, B. M., Puri V. K., and Yen, S. C. (1993b). "Ultimate bearing capacity of shallow foundations on sand with geogrid reinforcement.", Can. Jour. of Geotech. Engrg., Vol. 30, No. 3, PP. 545-549. Sah, S. H. (1990)."Experimental studies on bearing capacity of geosynthetics reinforced saturated clay.", M. Technology Dissertation, UT, Bombay. Sakti, J., and Das, B. M. (1987). "Model tests for strip foundation on clay reinforced with geotextile layers.", Transportation Research Record No. 1153, National Academy of Sciences, Washington, D. C., PP. 40-45. Shin E. C, Das, B. M., Puri, V. K. Cook, E. E., and Yen, S. C. (1993). "Bearing capacity of a strip foundation on geogridreinforced clay.", Geotech. Testing Jour., ASTM, Vol. 16, NO. 4, PP. 534-541. Sridharan, A., Srinivasa Murthy, B. B., Bindumadhava, and Revanasiddappa, K. (1991). "Technique for using finegrained soil in reinforced earth.", Jour. of Geotech. Engrg., ASCE, Vol. 117, No. 8, PP. 1147-1 190. Takemura, J., Okamura, M., Suesmasa, N., and Kimura, T. ( 1 992). "Bearing capacity and deformations of sand reinforced with geogrids.", Proc. Earth Reinforced Pract., Balkema, Fukuoka, Kyushu, Japan, PP. 695-700. Yetimoglo, T., Wu, J. T. H., and Saglamar, A. (1994). "Bearing capacity of rectangular footings on geogrid-reinforced sand.", Jour. of Geotech. Engrg., ASCE, Vol. 120, No. 12, PP. 2083-2099.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Multi-layered reinforced granular soil resting on soft soil - tension membrane effect C. Ghosh & K. Yasuhara Department of Urban and Civil Engineering, Ibaraki University, Japan
M.R. Madhav Department of Civil Engineering, IIT-Kanpur, India
ABSTRACT: The effect of applied surface loading on reinforced soil structures have been conveniently obtained in the form of increase in load bearing capacity/stability enhancement which are mainly expressed in terms of degree of improvement compared to the unreinforced ones. In this paper with the help of Pastern a k s model, the granular fill have been modified to include the effect of no. of layers of reinforcement in plain strain loading situation. Nonlinear hyperbolic responses of the granular and soft soil are introduced into the formulation. Parametric results reveal that reinforcement action is significant at lower stiffness of the granular fill and tensile forces are maximized at the center of the foundation. Reinforcement layer of 3 times the footing width placed within significant depth below footing is sufficient for effective improvement. tensible, e.g. planar geosynthetics or inextensible, eg. iron or aluminium metal strips) on the internal system modification could be accounted for. Some way of looking into such behaviour could be in the form of interaction shear transfer, soil-reinforcement interlock, and lateral bearing effect in case of grid reinforcement or adhesion effect in case of reinforced clay. Pasternak type model has been found suitable to describe the mechanical response of such foundation system. This model has been modified to include the nonlinear response of the soft soil as well as granular fill (Ghosh and Madhav, 1999). An extension of the same for single layer reinforced granular fill has been presented by Ghosh and Madhav (1994a, I994b). This paper presents “tension membrane” effect of multi-layer reinforced system for plain strain loading condition.
1 INTRODUCTION In order to build structures on soft soil granular fill are often spread out so that underlying soft soil experience less pressure. Else due to low load bearing capacity soft soil often requires replacement or modification of its nature by mechanical or chemical method. Use of planar reinforcement (e.g. woven and non-woven geosynthetics, geogrids, etc.) in granular soil helps the load spreading efficiency and thus ensures betterment in the load settlement behaviour. Usually 2 to 3 layers of such materials extending 2 to 3 times the footing width are placed within depth of 1 to 1.5 times the footing width. Quite extensive experimental investigations are reported in the literature but analytical treatise to this kind of reinforced soil system is very few. Moreover realistic analytical explanations are yet to be far from the behaviour traditionally obtained in the small-scale model tests. Efforts have been made by many to evaluate the nature of tensile forces both experimentally and numerically. Considerably umpteen numbers of analytical and mathematical models are available in the literature to recast above problems from the viewpoint of realistic quantification and prediction of the actual behaviour. The basic aim of all such approaches have been primarily to predict the amount of tensile force developed within the reinforcement so that a choice could be made for the design geosynthetic requirement. Almost same has been the approach for the stability requirement of the reinforced earth retaining wall system, reinforced slope or embankment structures. In none of these investigations, the effect of tensile reinforcement (ex-
2 SOIL-REINFORCEMENT INTERACTIONS Schematic of soil-reinforcement interactions in granular fill-soft soil foundation system are shown in Fig. 1. Interaction between granular soil and planar reinforcement layer has been represented in terms of “tension” and “confinement” effect, both separately and together, which is here in called “combined” effect. Ideally a reinforcement layer becomes effective in tension as well as confinement of the surrounding soil. Confinement of the soil mainly derives from the inward shear forces at the soil-reinforcement interface. In this paper “tension” effect of reinforcement has been formulated for plain strain loading.
553
Granular soil Soft soil
Soft soil
QR-c
QR-T
Soft soil
Tension effect
Confinement effect Combined effect
Soft soil
Figure 1. Shematic of soil-reinforcement interaction mechanisms in foundation beds.
3 TENSION EFFECT
monly adopted. Herein 3 layers are being used for the analysis. Bottommost layer is placed at the interface of granular soil and soft soil. Position of each layer and granular soil are defined with shear modulus (G), thickness of soil (h) and interface friction 0.) at both faces of the reinforcement. As per Pasternaks concept granular soil is assumed incompressible in the vertical direction and hence all the reinforcement layers will remain parallel to each other, both before and after the load application. The reinforced soil element is discretised in Fig. 3b. The unknowns are, the surface deformation (w) and tensile forces in respective layers. The governing equations for plain strain loading conditions are as follows:
In Fig. 2, schematic of granular bed and reinforcement are shown. Due to applied load (in case of uniformly loaded footing)/ displacement (in case of rigid footing), the foundation deforms. The granular soil is assumed as Pasternak shear layer in 1Dimension. As a consequence same amount of deformation will be transmitted to the soft soil, which is conveniently idealized by Winkler spring. With increasing deformation, the reinforcement will develop tensile forces, which is due to frictional bond between granular soil and reinforcement. Below a formulation of the same is being presented for multilayer reinforced foundation bed.
4 FORMULATIONS 1-pI2tan6 91 =q2 1+plItan6
Positions of reinforcement are shown in Fig.3a. In most of the design practice, 2 to 3 layers are com-
554
T,cos8
d'w
l + p , , tan6 dx'
(Ib)
Loamisp.
*I Granular soil., G
Surface deformation, w
4
I
............................
,,,,,,
/
,,,.
-
,
,
,
<
<
1
-
I_r.....‘
..............
-
................
............ k
....................
t
............................. -...-.
4
.”...................
t
Reinforcement inTension.
.
Figure 2. Tensile membrane mechanism - two layers reinforced foundation bed.
q3 = q4
1 -p2’ tan0 1 + p21tan0
T, cos0 d’w 1+ p 2 1 tan0 dx’
(14
q5 = q6
1-p3‘ tan0 1+pU3, tan0
Tl cos0 d’w 1 + p 3 ,tan0 dx’
(10
dT,= - ( p , , c o s ~ - s i n ~ ) -q (p,2cosB-sin0)q2 , dx
(1h) dT’
-=- (p2, cos 0 - sin 0)q3 -
dx
(,u2’cos 0 - sin 0)q4
3 =- (p3,cos 0 - sin e>q, - (p32 cos 6 dx
-
Figure 3. Definition sketch - (a) Reinforced granular fill overlying soft soil, (b) Reinforced soil element.
sin 0)q6 (1.9
The above equations are interdependent. They can be conveniently expressed into equations with four unknowns as w, T1, T2 and T3. Combining Eqns. 1, the final expression for applied load, q is,
555
footing, displacement is taken as uniform within the footing zone. For the reinforcement, tensile forces at the ends are taken as zero or it can be specified if the effect of end anchorage is required.
4-
:
T’cosB 1-p12tan8 1+p2,tan6 I + p l 1 tan8
i
1-p,’ tan8 l + p u ,tan8 ,
1:
1+
T3cos8 l + p u , tan8 ,
1-p” tan6
5 RESULTS AND DISCUSSIONS (2)
I+p2,tan8 1 l
Now substituting Eqns. 1 a and Ib into Eqn. lh, we get for T1 as, dT,
(pI2cos8+sin8)cos6
dx
1-pI2tan8
-+T1[
,
- {(p, cos 8 - sin 8 ) + ( p 1 cos 2 8 + sin 8 )
(3) From Eqn. Ij, we get for T2 as,
5 + T2 (p2’cos 8 + sin 8 ) dx {(p2,cos 8 - sin 8 ) +
d’w. _ cos8 tan8
dx’
cos 6 + sin 8 )
(4) Similarly from Eqn. I j, clT,
~-
dx
i
T,cos8 d’w = - { ( p 3cos8-sin8) , l + p u ,tan8 , dx’
1-pi2tan8 1 + p 3 ]tan 8
I
+ (p3?cos 8 + sin 8 ) } k,w
(5)
Load-settlement responses of the reinforced foundation are shown in Fig. 4. In this case only ‘tension membrane‘ effect are depicted in qualitative terms for rigid strip footing. With increasing number of reinforcement layers, load carried by the footing also increased. This model cannot identify optimum no. of reinforcement layers required for the design. The nonlinear parameter (B, = k,B/p, ) for soft soil is taken as 20 and for granular fill B, =O. Interface friction coefficients at both top and bottom faces are taken as 0.3. Lengths of all reinforcement layers, L are taken 3 times the footing width, B. Normalised load q’ (=q/k,B) is plotted against settlement of the induced displacement, WO(=w/B). Effect of granular soil stiffness is presented in*Fig. 4. For ten fold increase in shear stiffness (G (=GWk,B’) = 0.05 to 0.5) load carried by footing also increased significantly. With stiffer granular soil effect of tensile reinforcement is small. However, reinforcement layers contribute significantly in the confinement enhancement of the granular soil (Ghosh and Madhav, 1994b). Settlement-distance profiles of the rigid strip and uniformly loaded strip footing are shown in Fig. 5 and Fig. 6. Load carried by the rigid strip footing for given settlement are calculated from the integration of the surface deformation profile and they are shown in Fig. 5. With more reinforcing layers, the footing carries more load. This is being indicated as larger spreading of surface deformation profile outside the footing base. Load spread is more with higher WO which means membrane action is effective at larger displacement of the footing. Relative reduction in the settlement of uniformly loaded footing at the center is more at lower load intensity (Fig. 6). However, surface deformation is more or less limited within 2.5 times the footing width.
6 CONCLUSIONS
Now introducing hyperbolic nonlinearity for the soft soil (b, ) as well as for the granular soil (b\), all the four equations are expressed in non dimensional finite difference form. Gauss-Seidal iteration technique has been used to solve the equations. Minimum step size of 0.05 was found sufficient. Displacement boundary conditions were taken as that slope at the centre as well as at the far extent of the granular fill are zero. For uniformly loaded strip footing, applied load within the footing width is taken as uniform and zero at elsewhere. For rigid
Various parametric studies reveal that with suitable selection of B, and B, model tests results can be compared with the present analysis. Model parameters, like k, is usually evaluated as initial slope of the load-settlement plot of soft soil. Shear stiffness of the granular soil (G) are obtained from standard laboratory or field tests. Interface friction coefficients are obtained from the pullout or modified direct shear tests.
556
-
---
- Reinforced U Footing settlement, W,, Figure 4. Load vs. settlement response - effect of number of reinforcing layers (N) and shear stiffness of granular soil (G).
Distance from center of footing, X
On r e inf o t c ed WO
Layers
q
3
0.0372
3
0-0475
Figure 5. Surface deformation profile for rigid strip footing - effect of number of reinforcing layers (N).
557
Distance from center of footing, X
0.5
1.5
1.0
2.0
2.5
3.0
n reinfo r ced
Uniform strip &=&=0.3 G'=0.05 L=3 B,=lO, B,=O
Figure 6. Surface deformation profile for uniformly loaded strip footing - effect of number of reinforcing layers (N).
REFERENCES
soft soil system - confinement effect. Ceotex-tilesurzd GeonzeiizD~-ane.s.13(1 1): 741-759. Ghosh, c. and Madhav, M, R. 1 9 9 4 ~~. ~ i ~ fill f ~ Ghosh, ~ c. ~ and~Madhav, d M. R. 1999. Nonlinear subgrade modeling of Granular fill-soft soil foundation system. b?dian soft soil syste,n - Inembrane effect. Geote.vtiles alld GeCeotechnical Jourizal. 29(4): 339-361. mzenzhr~iiies.1 3, (1 1 ): 727-74 1. Ghosh, C. and Madhav, M. R. 1994b. Reinforced granular fill
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Prediction of the behaviour of a geogrid reinforced sloped fill under footing load C.T. Gnanendran School of Civil Engineering, Uiziversity College ADFA, University of New South Wales, Northcott Drive, Cunberra, ACT 2600, Australia
ABSTRACT: This paper summarizes the results of a numerical study on the performance of a model footing located near the crest of a sloped fill reinforced with a layer of geogrid reinforcement. Predicting the loadsettlement behaviour of the footing and the stabilising force contributed by the reinforcement proved difficult due to the effects of compaction during construction. The potential of modelling the compaction effects by varying the stiffness parameter K of the fill using both Janbu’s and Duncan and Chang’s models as well as using higher KO value for setting the initial horizontal stresses are examined. The predicted behaviour of the reinforced sloped fill for these different scenarios are discussed in comparison to the measured responses. This study indicates that the load-settlement behaviour of the footing could be predicted well by adopting Janbu’s equation for the stress dependent stiffness with high K values in the order of 4000 and KO= 1.0 for setting the initial horizontal stresses. However, the tensile force in the reinforcement could not be predicted satisfactorily. 1 INTRODUCTION Foundations located on sloped fills are used quite extensively as supports for bridge abutments and geosynthetic reinforcement are often used within the body of the fill to enhance its load carrying capacity and improve the load-settlement behaviour. Knowledge of the load carrying capacity of the reinforced sloped fill (or the load-settlement behaviour of foundations supported on it) and tensile forces/strains developed in the geosynthetic reinforcement are essential requirements for efficient design of these structures, however, predicting these responses had often been difficult. The load settlement behaviour of a footing 10cated near the crest of a sloped fill reinforced with a layer of geogrid reinforcement and the progressive development of stabilising force in the geogrid reinforcement with increasing footing load were studied until failure through a laboratory model (Fig. 1). The behaviour of this reinforced sloped fill under footing load was back analysed using an elasto-plastic, nonlinear finite element model with the objective of predicting the load-settlement behaviour of the footing and the stabilising tensile force developed in the geogrid reinforcement. However, difficulties were encountered in predicting the load-settlement behaviour of the footing and the reinforcement strain/force development simultaneously from a single analysis apparently due to the effects of compaction. Different strategies were adopted for accounting the cornpaction effects by adopting different soil models and 559
varying different model parameters. Details of the finite element numerical model and the predicted behaviour of the reinforced sloped fill under footing loading for different scenarios of model parameters to account the compaction effects and their relative merits are discussed in comparison to the measured performance of the footing and geogrid reinforcement. 2 BRIEF DESCRIPTION OF MODEL TESTING A displacement controlled footing load test on a sloped fill reinforced with a layer of geogrid reinforcement was examined with an experimental setup shown in Fig. 1 (see Selvadurai and Gnanendran, 1989; and Gnanendran and Selvadurai, 2001 for further details). The testing was carried out in a reinforced concrete tank measuring 1500 mm long, 880 mm wide and 1200 mm depth with the model strip foundation made of a steel box section measuring 104 mm wide and 870 mm long in plan. The sides of the test tank was fitted with polished stainless steel sheets to minimize end effects and the bottom of the model strip footing was made rough by applying timber bonding glue and spraying Ottawa sand. Mortar sand (SP - poorly graded sand with effective diameter, Dlo = 0.27 mm; coefficient of uniformity, C, = 3.0; and the curvature coefficient, C, = 0.95) was used as the fill material for model testing. The moisture content of the mortar sand was maintained between 4% and 5%. The bulk density of the mortar
P
-.
- Geogrtd Reinforcement
'
/
f
-Reinforced concrete
E
Compacted sand
~
,
,Laboratory
I All dimensions are in mm
Figure 1. Configuration of the geogrid reinforced sloped fill (modified after Gnanendran and Selvadurai, 200 1).
sand in its compacted state was kept constant at 17.6 kN/m3 in all the experiments. Shear strength parameters of the soil were determined to be approximately c = 5 kPa and Cp = 40". An extruded polypropylene biaxial Tensar BXI 200 (SS2) geogrid, measuring 870 mm wide by 740 mrn long, instrumented with a number of strain gauges was used as reinforcement of the sloped fill. The depth of embedment of the geogrid reinforcement was selected to be d = 0.7B based on the studies reported by Selvadurai and Gnanendran ( 1989) which was expected to give the maximum improvement in the ultimate bearing capacity of the footing. 3 NUMERICAL MODELLING AND SELECTION OF PARAMETERS To obtain reasonable stresses and strains in the sloped fill, it is necessary to consider both stress dependent stiffness characteristics of granular materials as well as plastic failure. Janbu's equation was used to account for the stress-dependent stiffness of the fill.
ses but rn was assumed to be 0.5 for all the analyses reported in this paper. The geotextile reinforcement was modelled as a series of linear elastic bar elements, whose axial stiffness (4 is a representative value per unit width of sloped fill. Based on wide-strip tensile tests, J = 450 kN/m for 0 - 2% strain, J = 600 kN/m for 0 I % strain, J = 700 kN/m for 0 - 0.5% strain and initial tangent stiffness = 1080 kN/m; and q =32 kN/m (Mylleville, 1991). The reinforcement-soil interface was modelled using nodal compatibility joint elements, assumed to be rigid plastic and non-dilatant (i.e., y~= 0). The geogrid-fill interface friction angle was taken to be 40". Provision was made for slip between the reinforcement and the soil by incorporating interface slip elements above and below the reinforcement. Thus, slip could occur independently above and/or below the reinforcement. A modified version of the program AFENA (Carter and Balaam, 1995) was used for these analyses.
4 DETAILS OF ANALYSES For the analyses of geosynthetic reinforced embankments on soft soils and retaining walls, modelling the granular fills using Janbu's stiffness parameter K ranging between about 500 and 1000 has been found to give reasonably good predictions (e.g., Rowe and Soderman, 1987; Rowe and Mylleville, 1989; Gnanendran and Rowe, 1995; and Ho and Rowe, 1996). Consequently, two different analyses using K = 500 and 1000, hereafter referred as Runs 1 and 2 respectively, of the reinforced sloped fill under footing load were performed to study the load - settlement behaviour of the footing and the gradual develop350
300
,-.250
. E
where E is the Young's modulus of the soil, P, is the atmospheric pressure, c ~ 3 is the minor principal stress, and K and YIZ are the empirical parameters. However, maximum and minimum E values of 25 and 2 MPa were assumed for the fill in all the analyses reported in this paper. Plastic failure of the fill was modelled using a Mohr-Coulomb failure criterion and a non-associated flow rule with a cohesion intercept c', friction angle Cp' and dilatancy angle w. The properties of the fill material were obtained based on laboratory tests and the values used in the analyses were: y = 17.6 kN/m3, c' = 5, @' = 40", w = 9" and v = 0.35. K was varied for different analy560
5
200
1
U 0 -
m 150 c
-
U 0
100
A
Experimental (aner Gnanendran and Selvadurat, 2001) Analysis - Run 1 (KO= 0.36, Janbu - K = 500, J = 1080 kN/m) Analysis - Run 2
(KO= 0.36, Janbu - K =1000. J = 1080 kN/m)
Analysis - Run 3 (KO= 0.36, Janbu - K = 1000, J = 600 kN/m)
50
0 0
10
20
30
40
50
60
70
80
Footing settlement (rnm)
Figure 2. Variation of footing loading versus settlement for Runs 1 to 6 - comparison with experimental data.
ment of stabilising force in the reinforcement. Although, both these analyses predicted the ultimate load capacity of the reinforced sloped fill satisfactorily; the predicted load-settlement behaviour of the footing was found to be quite different from that measured in the laboratory (Fig. 2). Despite using the higher initial tangent stiffness of J = 1080 kNhn in both these analysis, the measured load-settlement response was much stiffer compared to that indicated by the analyses. However, to study the effect of using a more appropriate stiffness for the reinforcement, another analysis was performed with J = 600 kN/nz, i.e., stiffness corresponding to 0-1% strain, referred as Run 3.
400
4.2 Eflects of changiizg KO In all the analyses discussed above, the coefficient of earth pressure at rest, KO,was considered to be ac-
- Run 3 (KO= 0.36, Janbu -
.. . c
300 -
5 5
250
20 -
'
I
p,: ,<
-
U
'
200 -
I
CI)
I -
'.
..- .
i
NOTE. J = 600 kN/m for all ihese analyses Analysis - Run 7 ( & = 1.0, Janbu . K = 5000) Analysis - Run 8 ( & = 1.O,Janbu - K = 4000) Analysis - Run 9 (K, = 0.36, Duncan and Chang - K = 5000) Analysis - Run 10 (KO= 0.36, Duncan and Chang - K = 10000) . Anaiysis - Run 11 (KOii 1.0, Duncan and Chang - K = 10000) . j Analysis - Run I2 = 1.0, Duncan and Chang .K = 15000)
__ d/f
-
''
,
The predicted load-settlement behaviour of the footing by all three analyses (i.e., Runs 1, 2 and 3) were different from the measured response and this difference is attributed primarily to the effects of compaction which was not considered in the numerical model. To verify whether the compaction effect could be modelled satisfactorily by changing the stiffness of the compacted fill, a series of analyses were carried out for increasing K values. Three analyses, hereafter referred as Runs 4, 5 and 6 corresponding to K = 2000, 4000 and 5000 respectively, were carried out in this series to model the effects of compaction on the stiffness of the soil. As evident from Fig. 2, the agreement between the predicted and measured footing load versus settlement behaviour improved with increasing K value. However, as will be discussed later, the comparison between the predicted and measured reinforcement force/strain worsened with increasing K value (see Fig. 3).
Analysis
...a
1
4.1 Effects of changing Janbu 's K parameter
..
L ... . .... ...--. '.* .. .. ...
350 -
,
...
(7
Footing settlement (mm)
Figure 4. Variation of footing loading versus settlement for Runs 7 to 12 - comparison with experimental data.
cording to Jaky's relationship of KO= I-sin @< However, compaction could induce higher initial horizontal stresses in the fill, prior to the application of footing load. To assess whether the influence of compaction could be accounted for approximately with higher initial horizontal stresses using a higher K , value, two additional analyses were carried out with K , = 1 for the cases of Janbu's stiffness parameter K = 4000 and K = 5000, referred as Runs 7 and 8 respectively (Fig. 4). 4.3 Influence of using Duncan and Cliang Hyperbolic model Additional analyses were cai-ried out to verify whether any other well-known soil model could yield better predictions for the behaviour of the reinforced sloped fill under footing loading and the Duncan and Chang (1970) hyperbolic model was chosen for this purpose. The tangential Young's modulus of the fill given by the following equation was used:
K =1000. J = 600 kN/rn)
Analysis - R u n 6 (KO= 0.36, Janbu - K = 5000, J = 600 kN/rn)
where, S=
-4
-3
-2
-1
0
1
2
3
Relative displacement from centreline of footing ( x B )
Figure 3. Variation of tensile force across the geogrid reinforcement for Runs 1 to 6 - comparison with experimental data.
rf ( l - s i n @ ) ( q -41) 2ccos@+203sin@
(3)
K, iz and cf are non-dimensional constants. However, plastic failure of the soil was modelled using the Mohr-Coulomb criterion similar to the previous analyses. Here again, the influence of increasing stiffness of the fill on the behaviour of the reinforced sloped fill under footing loading was investigated with K
56 1
values of 5000 and 10000 in Runs 9 and 10 respectively. However, n and rf were assumed to be 0.5 and 0.8 for all the cases considered in this series. To examine the influence of using higher initial horizontal stresses with KO = I , two additional analyses, referred as Runs 11 and 12, were also performed for the cases of K = 10000 and 15000 respectively.
5 RESULTS AND DISCUSSION The variation of footing load versus settlement predicted from Runs 1 to 6 are compared with the corresponding experimental data in Fig. 2. The predicted tensile force distribution across the geosynthetic reinforcement at ultimate footing load for each of these six analysis cases is compared with the measured response in Fig. 3. It can be observed from Figs. 2 and 3 that Run 2 with Janbu’s K = 500 for the fill and reinforcement J = 1080 kN/m predicted the ultimate footing load satisfactorily (difference of only 5.6%) but under estimated the maximum reinforcement force at ultimate footing load by about 30%. When Janbu’s K was increased to 1000 (i.e., Run 2), the analysis gave better predictions of both the ultimate footing load and maximum geogrid force at ultimate condition, i.e., under prediction of ultimate load by about 4.3% and over prediction of maximum geogrid force by about 11%. Although the reinforced soil system indicated a stiffer response in Run 2 compared to that of Run 1, the predicted footing load-settlement response was still significantly less stiff compared to that obtained in the experiments. This indicates that the predicted settlement at a given footing load would be much higher than the measured value (e.g., the predicted settlements at 250 kPa footing loading would be approximately 4 1 mm and 33 mm from Runs 1 and 2 whereas the measured settlement was only 7 mm). For the analyses and design of reinforced soil structures it is common practice to use secant modulus rather than the initial tangent stiffness for the geosynthetic. Consequently, the secant stiffness of 600 kN/m corresponding to 0 to 1% strain was used in all the analyses reported in this paper except Runs 1 and 2. It is worth noting that the measured geogrid strains were less than 1 % in the experiments (Gnanendran and Selvadurai, 2001) and the stiffness corresponding to 0 to 1% strain range is therefore appropriate for use in the analyses. With Janbu’s K = 1000 for the fill, decrease of the geogrid stiffness from J = 1080 in Run 2 to 600 kN/m in Run 3 resulted in a decrease in the overall stiffness of the reinforced soil system as expected. Run 3 over predicted the ultimate footing load by about 13.6% but the maximum reinforcement force at ultimate footing load was reasonably well predicted. However, the maximum footing load predicted in Run 3 was at a significantly larger settle-
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ment compared to those of both Runs 1 and 2 and at much higher settlement than that observed in the experiment. Despite the fact that the value of Janbu’s K commonly used for analysing reinforced soil structures range between 500 and 1000, all the three analyses (i.e., Runs 1 to 3) failed to predict the measured load-settlement and geogrid strain responses satisfactorily. This is attributed to the effects of compaction on the stress-strain characteristics of the reinforced soil system. To model the compaction effects approximately, higher Janbu’s K values of 2000, 4000 and 5000 were attempted for modelling the fill material behaviour in Runs 4, 5 and 6. The use of higher K values resulted in improved prediction of the load-settlement behaviour of the footing (Fig. 2), the use of K = 5000 in Run 6 giving the best load settlement response followed by that of Run 5 and subsequently that of Run 4. However, the geogrid strain distribution at ultimate footing load predicted by the analyses deviated further from the measured response, the maximum reinforcement force was under predicted by as high as 48% in Run 6. It is therefore apparent that the compaction effects could not be modelled satisfactorily by adopting higher values for Janbu’s stiffness parameter K for the fill material. Compaction could induce higher initial horizontal stresses than that usually estimated assuming geostatic K, ( 2 I-sin 4’)condition. This is more likely to be the situation in laboratory model tests where rigid boundaries are used in the experimental set up compared to field conditions. To verify whether the compaction effects could be modelled satisfactorily with the combination of higher stiffness for the fill (i.e., higher values for the Janbu’s stiffness constant K> and higher KOvalue for setting up the initial horizontal stresses, two additional analyses were performed. In both these analyses, i.e., Runs 7 and 8, KO was assumed to be 1.0 but Janbu’s stiffness parameter was assumed to be K = 4000 and 5000 respectively for the two analyses. The footing load versus settlement behaviour predicted from the two analyses had closer agreement with the experimental results than the analyses discussed previously (i.e., Runs 1 to 6 - see Figs. 4 and 2). However, the maximum reinforcement force at ultimate footing load was under predicted by 74% and 46% in Runs 7 and 8. This indicates that the predicted reinforcement forces from Runs 7 and 8 are lower than those obtained for the corresponding analyses Runs 5 and 6 where the usual geostatic K, ( 2 1-sin 4 ’)value was assumed (Figs. 5 and 3) for setting the initial horizontal stresses. Of all the analyses cases considered so far, Run 8 with K, = 1.0 and Janbu’s K = 4000 appears to give the best prediction of the load settlement response and tensile force distribution across the reinforcement. This analysis gave correct prediction of the ul-
timate footing load, indicated a load-settlement response closer to that obtained from the experiment but under predicted the maximum reinforcement force by about 46%. The tensile force distribution across the reinforcement predicted by different analyses indicate a more localised reinforcing effect close to the footing even for cases where the maximum tensile force was well predicted (e.g., Run 3) whereas the experiment indicated a fairly spread distribution (Figs. 3 and 5). This observation suggests that an additional factor such as the occurrence of reinforcement pre-tensioning during compaction and/or the reinforcement stiffness increasing with vertical (or confining) stress need to be considered for predicting all the responses accurately from a single analysis. However, these have to be verified experimentally prior to use in the analysis and are beyond the scope of this paper. Analyses using the Duncan and Chang hyperbolic model for the fill material with K = SO00 and 10000, but with the usual geostatic KO(= I-sin 4’) for setting the initial horizontal stresses, also predicted the ultimate footing load accurately (the difference being less than 2% compared to the experimental data for both Runs 9 and 10). The predicted load-settlement responses were quite different from the experimental data, the ultimate (maximum) loads occurring at significantly larger settlements but the prediction improving with increasing K value from 5000 in Run 9 to 10000 in Run 10. Here again the predicted tensile force across the reinforcement at ultimate footing load was much lower than the measured value; the difference of the maximum force being 56 and 5 1% respectively for Runs 9 and 10 respectively compared to the experimental data. When K, was increased to I .0 in Run 11 to account for higher initial horizontal stresses induced by compaction, the ultimate footing load was over predicted by about 17% and the predicted loadsettlement behaviour of the footing showed a stiffer response but still significantly different from the experimental result (Fig. 4). The ultimate footing load was over predicted by about 20% when the Duncan and Chang stiffness parameter K was increased further to 15000 in Run 12. However, the load - settlement response predicted from Run 12 agreed better with the experimental data for smaller settlements compared to that of Run 11 and it indicated a relatively plastic type response afterwards whereas the experimental data indicated a strain softening type of behaviour. The predicted settlements at a typical footing loading of, say, 250 kPa were approximately 9, 12 and 15.5 mm from Runs 12, 11 and 10 while the measured settlement was 7 mm. However, the maximum footing loads occurred at a significantly higher settlement of about 30 mm in Runs 11 and 12 and hence the predicted tensile force distributions 563
.. .
Analysis - Run 9 (KO= 0.36, Duncan and Chang - K = 5000; J = 600 kN/m) Analysis - Run 10 (K,= 0.36, Duncan and Chang - K =10000; J = 600 kN/m) Analysis - Run 11 (KO= 1.0, Duncan and Chang -
K = 10000; J = 600 kN/m)-
Analysis - Run 12 (K,= 1.O, Duncan and Chang - K = 15000; J = 600 kN/m)i
-4
-3
.2
-1
0
1
2
3
Relative displacement from centreline of footing ( x B )
Figure 5. Variation of tensile force across the geogrid reinforcement for Runs 7 to 12 - comparison with experimental data.
across the reinforcement at ultimate footing load were relatively better than Runs 7 to 10 analyses (see Fig. 5). Runs 11 and 12 under predicted the maximum tensile force in the reinforcement at ultimate footing load by 21 and 17% respectively compared to 46% in Run 8. For the load-settlement behaviour of the footing, Run 8 gave the best prediction and it accurately predicted the ultimate footing load. But it under estimated the reinforcement force significantly. In summary, this study indicates that the behaviour of the geogrid reinforced sloped fill subjected to footing loading could be predicted well by adopting Janbu’ s equation for the stress dependent stiffness with high K values in the order of 4000 and KO= 1.0 for setting the initial horizontal stresses. However, the tensile force in the reinforcement could not be predicted satisfactorily and it is believed that other factors such as the development of pretensioning force during compaction and/or stress dependent stiffness for the reinforcement need to be considered to improve the prediction. Duncan and Chang hyperbolic model with higher stiffness constant in the order of 10000 to 15000 and KO= 1 .O gave relatively better prediction than the other cases for the reinforcement force but still couldn’t predict the load settlement behaviour of the footing at higher footing loading close to ultimate and the reinforcement force satisfactorily from a single analysis.
6 SUMMARY AND CONCLUSION The load-settlement behaviour of a footing located near the crest of a sloped fill reinforced with a layer of geogrid reinforcement and the stabilising force contributed progressively by the geogrid reinforcement were studied until failure of the footing through a laboratory model. An elasto-plastic nonlinear finite element model was used to analyse this problem with the objective of predicting the Ioad-
settlement behaviour of the footing and the stabilising tensile force developed in the geogrid reinforcement. The stress dependent stiffness characteristic of the fill was modeled using Janbu's equation. Details of the model and the parameters selected for analyses are discussed in the paper. Difficulties were encountered for predicting the load-settlement behaviour of the footing and the reinforcement straidforce simultaneously from a single analysis apparently due to the effects of compaction. Different strategies such as varying Janbu's stiffness parameter K of the fill, using a higher K, value for setting the initial horizontal stresses in the fill and adopting Duncan and Chang hyperbolic model for the fill were attempted for considering the compaction effects. The predicted behaviour of the reinforced sloped fill under footing loading for these different scenarios of model parameters are discussed in comparison to the measured loadsettlement behaviour of the footing and the tensile force developed in the geogrid reinforcement. This numerical investigation indicates that the load-settlement behaviour of the geogrid reinforced sloped fill subjected to footing loading could be predicted well by adopting Janbu's equation for the stress dependent stiffness with high K values of the order of 4000 and KO = 1.0 for setting the initial horizontal stresses. However, the tensile force in the reinforcement could not be predicted satisfactorily and it is believed that other factors such as the development of pretensioning force during compaction also need to be incorporated. Analysis using Duncan and Chang hyperbolic model with higher stiffness constant K in the order of 10000 to 15000 coupled with KO = 1.0 predicted the load-settlement behaviour of the footing satisfactorily under typical working stress conditions. However, the ultimate footing
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load, load-settlement behaviour at higher loading condition closer to the ultimate and the reinforcement force couldn't be predicted satisfactorily using this approach.
REFERENCES Carter, J.P. and Balaam, N.P. (1995). AFENA Version 5.0 - A general finite element algorithm - User's Manual, School of Civil and Mining Engineering, University of Sydney, N.S.W. 2006, Australia. Duncan, J.M. and Chang, C.Y. (1970). Nonlinear analysis of stress and strain in soils. Journal of Soil Mechanics and Foundation Engineering Division, ASCE, 96(5), pp. 12691653. Gnanendran, C.T. and Rowe, R.K. (1 995). Predicted undrained behaviour of the Sackville test embankment. Proceedings of the GEOSYNTHETICS '95 Conference, Nashville, Tennessee, USA, Vol. 1, pp. 53-66. Publisher - Industrial Fabrics Association International, St. Paul, MN, USA. Gnanendran, C.T. and Selvadurai, A.P.S. (2001). Strain measurement and interpretation of stabilising force in geogrid reinforcement. Technical Note - Geotextiles and Geomembranes Journal, 19: pp. 177-194. Ho, S.K. and Rowe, R.K. (1996). Effect of wall geometry on the behaviour of reinforced soil walls. Geotextiles and Geomembranes, 14: pp. 521-541. Mylleville, B.L.J. (1 991). "Behaviour of heavily reinforced embankments on soft foundations." Ph.D. Thesis, University of Western Ontario, London, Ontario, Canada. Rowe, R.K. and Mylleville, B.L.J. (1989). "Consideration of strain in the design of reinforced embankments." Proceedings of Geosynthetic '89 Conference, San Diego, U.S.A., pp. 124-135. Rowe, R.K. and Soderman, K.L. (1984). Comparison of predicted and observed behaviour of two test embankments. Geotextiles and Geomembranes, 1, pp. 143-160. Selvadurai, A.P.S. and Gnanendran, C.T. (1989). An experimental study of a footing located on a sloped fill: influence of soil reinforcement layer. Canadian Geotechnical Journal, 26: 467-473.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Performance of geotextile-reinforced shallow foundations used in Bangladesh M.A. Haque Public Works Department, Khulna, Bangladesh & Former Graduate Student, BIT, Khulna, Bangladesh
M. Alamgir & M. Salim Department of Civil Engineering, Bangladesh Institute of Technology (BIT), Khulna 9203, Bangladesh
M.H. Kabir Department of Civil Engineering, Bangladesh University of Engineering & Technology, Dhaka, Bangladesh ABSTRACT: This paper describes the field performance of geotextile-reinforced shallow foundation system constructed in a highly compressible soil deposits in the south-western region of Bangladesh to bear the buildings. The settlement in all the twenty eight constructed buildings were measured for the last three years starting from the completion of the foundation system. The measured settlements of the constructed buildings are more than that of the predicted values except the three buildings. The largest settlement, 600mm, was measured at a 75 square meters staff quarter while the lowest value was measured as 50mm at the Ladies hostel. The field measurement shows that uniform settlement occurred in all the buildings except two residential buildings. The results reveal that the geotextile-reinforced foundation system works perfectly in preventing the differential and total settlements of the buildings resting in a highly compressible soil provided the design is done properly based on the actual soil condition of the site and the material properties. Khulna medical college campus as an alternative of conventional foundation. Twenty eight buildings of different categories were constructed and thus monitored for a long period to depict the performance of geotextile-reinforced foundation system (Haque 2000). This paper describes the field performance of geotextile-reinforced shallow foundation systeni used to bear the buildings constructed in a highly compressible soil deposits in the south-western region of Bangladesh. Here the focus is mainly given on the measurement of total and differential settlements of the buildings occur under full design load. In the constructed twenty eight buildings, the settlement measured for 3 years, are more than that of the predicted values except the three buildings, namely, Academic building, Boys hostel and Ladies hostel. The largest settlement, 6OOmm, was measured at a 75 square meters staff quarter, which is four times higher than that of prediction. The lowest settlement was measured as 50mm at the ladies hostel. Almost uniform settlement occurred in all the buildings except two residential buildings. The sub-soil investigation reveals that the soil conditions around the residential buildings, where maximum settlement occurred, are different than that of general soil condition of the site. which was considered in the design. The results reveal that the geotextile-reinforced foundation system works perfectly in preventing the differential settlement and as well as the total settlement of the buildings resting in a highly compressible soil provided the design is done properly.
1 INTRODUCTION Soil reinforcement has become a major pal? of geotechnical engineering practice over the last thirty five years and its use is growing rapidly as world wide development poses an increasing demand for land reclamation and the utilization of soft foundation soils. The modern form of soil reinforcement was introduced by Henry Vidal, a French architect and an engineer in the 1960s. Vidal’s concept (Vidal 1969) was for a composite material formed flat reinforcing strips laid horizontally in a frictional soil, the interaction between the soil and the reinforcing members being solely by friction generated by gravity. The use of geosynthetic is a rapidly expanding area of geotechnical technology with numerous new products and applications being produced with any one year (Raymond & Giroud 1993). The geosynthetic-reinforced granular fill soft soil systems are now being used very frequently for shallow foundations (Shukla 1994). Experimental study on the use of geosynthetics in foundation bed was started as early as 1970 by Yamanouchi (Yamanouchi 1970). Such reinforced soil systems provide improved bearing capacity and reduced settlements by distributing the imposed loads over a wide area of weak sub-soil (Binquet & Lee 1975, Sakti & Das 1987, Ghosh & Madhav 1994, Otani et al. 1994). Soil reinforced provides numerous other indirect benefits such as speedy construction time, ease in construction etc. In Bangladesh, geotexlile have been used in a shallow soil-reinforced foundation system to bear the buildings loads in a highly compressible organic soil at
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2 STATEMENT OF THE PROBLEM The public works department (PWD) of Bangladesh undertaken the construction of Khulna medical college complex buildings at Sonadanga, Khulna in 1992. Sub-surface investigation was carried out at the pro-ject site. Based on the site condition, a shallow foundation system using geotextiles was employed in all constructed buildings at the campus. 2.1 Location and soil condition ofproject site Khulna city is located about 332 km south-west of capital Dhaka as shown in Figure 1. The Khulna medical college complex is situated at Sonadanga of Khulna Metropolitan City and within the old 250 Bed Hospital Complex. The sub-soil profile of Khulna city is characterized by brownish gray to gray. Soft to medium soft clay with silt and very few gray loose fine sand with organic contents which reveals that the sediments are sub-recent to recent in age. It also shows the regular sequence of lithological composition and consistency of the formation of same nature at least up to the 30m depth in all the explode bore holes. Sub-soil investigation (FCL 1984, CRTS 1995) shows, geologically the project site lies on a thick highly compressible fine-grained clay and organic soil deposits. The ground water table varies with the season. During the rainy season the water table lies at ground surface to 150mm depth while during dry season the water table goes down to 1 .0 to I .5m below the ground surface. The sub-surface soil is composed of pure clay and some where it is black organic clay. This layer extends to 2.50m depth and then starts organic clay, organic silt and clayey silt, which extends up to 15m.
2.2 Foundation system Sub-soil investigation reveals that the soils arc too weak to support a traditional economic spread foundation at shallow depth. Considering the soil type and its initial properties, materials availability, equipment, technical advantages, economy and finally with a hope and confidence, the authority came to a decision to introduce a new foundation system, geotextile-reinforced shallow foundation. By introducing geotextiles reinforcement, same traditional spread foundation was provided at shallow depth with improving sub-soil by sand mixed with khoa (brick aggregates) and sand compaction with geotextiles. The spread foundation rests 011 the compacted composite bed. The foundation system of two representative buildings are described below. In the foundation system of Academic building having plinth area 3900 square meter, the total area covered by geotextiles is 7800 square meter. As per design provision, the geotextile is extended 4.5111 bcyond the plinth area of the buildings in all sides
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Figure 1 . Project site shown in Bangladesh map
and the gaps between the front and mid block, mid block and gallery and galleries and blocks. were covered by geotextiles. After lajiing the geotextile on natural ground at a depth 3m from the existing ground level, a 600mm layer of densified brick aggregates and sands mixed were placed. Then I . l m sand (FM2I .O) and 0.8m sand (FM22.5) layers were placed. Both the layers were compacted by vibroroller to obtain the designated degree of compaction. RCC footing having a cement concrete at the base was placed in the foundation. The plinth level raised to a vertical level at 1.20m above existing ground lcvel (EGL). Typical cross section of the foundation in the Academic Building is shown in Figure 2. In the foundation system of 75 square meter of staff quarter, the area covered by geotextiles is 520 square meter by extending geotcxtile 4.5m beyond the plinth area ofthe buildings in all sides. Alter laying the geotextile on natural ground at a depth 2.225m from the existing ground level, a 300nim layer of densified brick aggregates and sands mixed were placed. Then I . I m sand (FM2I .O) and 400mm sand (FM22.5) layers were placed. Both the layers were compacted by vibro-roller to obtain the designated degree of compaction. RCC continuous footing having was then placed in the foundation. The plinth level raised to a vertical lcvel 1.5m above existing ground level (EGL). A typical cross section of the foundation system constructed in the 75 square meter staff quarter is shown in Figure 3.
Figure 3. Typical cross of the foundation system used at 75 square meter staff quarter.
2.3 Constructed buildings Total twenty nine buildings for different types are considered for the construction in Khulna medical college campus. As suggested from the sub-soil conditions, similar type foundation system were employed. Table 1 shows the list of already constructed 28 different types buildings. Out of these 28 buildings, the largest one is the Academic Building of 3900 square meter plinth area, constructed with reinforced cement concrete (RCC) frame. In addition of academic building, other 27 buildings, plinth areas vary from 1530 to 46 square meter were constructed.
performance is mainly evaluated based on the observed settlement response of the constructed buildings. To observe the settlement from the beginning of the construction work, permanent Bench Mark pillars were established in front of each building. The settlement of all the buildings are measured at about every 6 months interval from July, 1997 to August, 2000, as shown in Table 2. The settlement measurements have been taken for total 19 constructed buildings. In general, from the table it can be seen that the settlement increases with time. The large amount of settlement occurs within the first one year. After that, although the settlement continues, the rate of settlement decreases.
3 FIELD INVESTIGATION
Field investigation was carried out to depict the insitu performance of geotextile-reinforced foundation system adopted in the constructed buildings. The
Table 1 List of the buildings constructed with geotextile-reinforced foundation system
S1 No 1 2 3 4 5 6 7 8
Name of the Building
Academic Building Boys & Girls Hostel Prof &Assoc Prof Quarter Assistant Professor Quarter Lecturer & Doctor’s Quarter Second Class Staffs Quarter Third Class Staffs Quarter Fourth Class Staffs Quarter
No of Building
Plinth Area (square meter)
No of Stories
Unit in each Building
1 2 6 3
3900 1530 140 116 93 75 56 46
4 4 4 4 4 4 4 4
_------
5 1 3 7
567
4 4 8 8 8
8
Table 2. Measured settlement of the constructed buildings. S1.
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18
19
Settlement (mm) Measured on the Following Seven Dates
Name of the Building
No.
15.9.97
15.3.98
15.9.98
26.11.98
30.6.99
30.12.99
30.8.2000
0 0 30
0 0 30 216 235 223 213 244 329 302 298 366 408 326 213 34 1 372 396
5 5 37 244 259 262 253 277 363 323 354 408 524 393 283 445 445 482
10 10 37 262 27 1 298 296 305 381 344 378 436 600 413 329 485 52 1 524
357
415
427
15 20 40 275 282 298 305 305 387 354 398 442 600 41 6 357 493 526 536 442
25 30 75 285 29 1 298 316 305 392 364 426 449 600 420 390 502 532 546 466
60 50 126 300 300 300 327 306 400 372 440 453 600 426 414 510 537 558 483
Academic Building Ladies Hostel Boys Hostel Prof 's Quarter No 1 Prof 's Quarter No 2 Prof 's Quarter No 3 Prof 's Quarter No 4 Asstt Prof 's Qr No 1 Asstt Prof 's Qr No 2 Lecturer's Qr No 1 Lecturer's Qr No 2 Lecturer's Qr No 3 Second Class Staff Qr 3rdClass Staff Qr No 1 3rdClass Staff Qr No 2 4"' Class Staff Qr No 1 4"' Class Staff Qr No 2 4'" Class Staff Qr No 3 4'h Class Staff Qr. No.4
155 180 165 155 186 265 27 1 219 332 296 256 158 293 268 314 302
4.2 Settlement response of residential buildings
4 RESULTS AND DISCUSSIONS
After 3 years of construction, the settlement of residential buildings were found as much higher than that of predicted. The time-settlement profiles of four types residential buildings are shown in Figures 4 to 7 . Very soft dark gray and grayish brown clay exists up to 3m depth from existing ground surface. Next 6m is very dark gray organic clay and the further 9m is gray/dark gray silty clay with organic traces. The geotextile was laid just on the very dark gray organic clay, which is most compressible layer. Due to the self weight of the building and improved foundation layer on geotextile, the bottom layer compressed deadly. But the structure surrounding with geotextile (4.5m around the building) settles uniformly. As a result, no differential settlement occurred in any place and no cracks observed in any portion of the building.
The performance of geotextile reinforced foundation system are studied based on the measured settlements. The foundation system shows the negligible settlement case and also four times larger settlement than that of predicted. The causes of settlement of these two extreme cases are identified here and hence discussed in the following sections. 4.1 Settlement response of academic building The foundation of Academic building was constructed for five-story at front and north block, threestory for both galleries and cafeteria. In first phase construction was completed three stories at front block and partly four stories at north block. Full load has already been loaded at the gallery. But cafeteria stands up with one story load. From the loading condition described above, it is clear that the two blocks of the original academic building was not yet been loaded fully as per the design consideration. Only two galleries was loaded as per the design. Even though the full load has not been imposed yet, 60% of the settlement would be expected to occur at this loading stage. But only 6Omm settlement was recorded till August 30. 2000, which is much lower than that of predicted value. Sub-soil report shows that a relatively good soil condition exists in this particular site. Also the load intensity of all blocks are distributed over a larger area than the building area due to the extension of geotextiles by 4.5m in all the sides from the boundary of the building.
4.3 Causes of variation between the predicted and nieasured settlements If one endeavor to inquire about the cause of excessive settlement, then it is revealed that, in Academic building the existing imposed load is now below the assessed vertical load. In this building, the geotextile area is large, as a result the bearing area is also augmented, and the loading intensity is comparatively low. on the foundation. Since the improved soil bed on geotextiles acts as a mat. moreover the soil condition around the Academic building area is comparatively more eminent than the residential areas. Besides th almost all the buildings in the residential 568
area settles excessively. The sub-soil condition shows that there exists a organic clayey silt layer in between the depth of 2.4 and 6.4111 and again in between 13 to 18m. Due to these two thick compressible layers, the structures fall into serious settlement problem. There is a compressible layer just beneath the foundation, which will be obviously compressed when greater load imposed on it. Again due to the smaller nature of the residential buildings, the load intensity spreads over a smaller area. In residential buildings the plinth area of individual buildings is small and also the buildings were constructed individually, In spite of that just now the load intensity is more in the residential area than the Academic building. Due to this over and above load and weak compressible soil foundation, the quarters settled more quickly but the settlement occurs at uniformly in every place of the buildings and cracks in the buildings do not visible.
4.4 Performance of geotextiles The settlement behaviour depicts clearly that the geotextiles performed perfectly in the foundation system. If geotextile has not worked accordingly with pull up or expansion then differential settlement could be occurred. As a result cracking symptoms inevitably is to be shown. Any differential settlement except two residential quarters were observed. The differential settlement in those two buildings occurred for the different consistency in horizontal soil layer due to the presence of deep ditch, which was not encountered in the sub-soil report. However, no cracks were observed in any one of the constructed buildings. 5 CONCLUDING REMARKS Based on the present study the following conclusion can be made:
569
Consultancy, Research and Testing Services (CRTS) 1995 Sub-soil investigation report at the site of Khulna medical college, Sonadaga, Khulna, Department of Civil Engineering, Bangladesh Institute of Technology (BIT), Khulna Foundation Consultant Limited (FCL) 1984 Soil investigation report on the proposed site of Khulna Medical college complex, Sonadanga, Khulna Haque, M A 2000 Monitoring the performance of bidding foundation improved by geotextiles in Khulna medical college M Engg Thesis, Department of Civil Engineering, Bangladesh Institute of Technology (BIT), Khulna Ghosh, C & M R Madhav 1994 Settlement response of a reinforced shallow earth bed Geotextiles and Geomenibranes, I 3 643-556 McGown, A 1993 Geosynthetics case histories In G P Raymond and J P Giroud (ed ), International Society for Soil Mechanics and Foundation Engineering, Technical Committee Report (TCS), Geotextiles and Geosynthetics Otani, J , 1-1 Ochiai & Y Miyata 1994 Bearing capacity of geogrid reinforced grounds Proc 5th Int Corg Geotextiles, Geoniembranes and Related Products, Singapore, 1 117-120 Sakti, J P & B M Das 1987 Model tests for strip foundation on clay reinforced with geotextile layers Transpor Research Record, 1 153 40-45 Shukla, S K 1994 Fotindation model for reinforced gtzlnirlar fill-soft soil systeni and its settlement response Ph D Thesis, Department of Civil Engineering, Indian Institute of Technology (IIT), Kanpur, India Vidal, I4 1969 The principal of reinforced earth Highway Research Record No 282 Yamanouchi, T 1970 Experimental study on the improvement of the bearing capacity of soft clay ground by laying a resinous net Proc Symp Foirndations Interbedded Sands, Perth 102-108
1. Field investigation shows that the measured settlements are 2 to 4 times higher than that of predicted values in all the constructed buildings except three buildings. 2. Despite the large settlements, no differential settlements and cracks were observed, that reveals the good performance of geotextiles in preventing differential settlement even in very soft ground. 3 . The field investigation depicted the fact that the exact sub-soil profile in every building site is to be determined individually to use in design. 4. In Bangladesh, geotextiles are almost new materials in shallow foundation. So in the contract provision to monitor the performance of the foundation should be included.
6 ACKNOWLEDGEMENT The support of Ministry of Science and Technology, Government of Bangladesh, by providing National Science and Technology (NST) Fellowship in the year 1998-99-2000 to the second author is gratefully acknowledged. REFERENCES Binquet, A M & K L Lee 1975 Bearing capacity tests on reinforced earth slabs J. Geotechnical Engineering, ASCE, 101 1241-1255
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Landmarks in Earth Reinforcement, Ochiai et a1 (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Effects of tensile and bending rigidities of reinforcement in reinforcing soil structures and ground N. Kotake Toyo Construction Co. Ltd., Japan
F. Tatsuoka Department of Civil Engineering, University of Tokyo, Japan
T. Tanaka Department of Biological & Environmental Engineering, University of Tokyo, Japan
M.S.A. Siddiquee Department of Civil Engineering, Bangladesh University of Engineering and Technology,Bangladesh
C.C. Huang National Cheng Kung University, Taiwan ABSTRACT: A nonlinear elasto-plastic FEM analysis was conducted to simulate a series of plane strain laboratory model tests on the bearing capacity of reinforced soil. The model ground was made of dense Toyoura sand and reinforced with linear, tensile reinforcing members placed horizontally beneath a strip footing. The FEM model for sand considered pressure-dependency and anisotropy in the internal friction angle, strain softening, dilatancy and shear banding. The FEM simulation agreed reasonably with the physical experiment with respect to the load-settlement relations and shear banding patterns. Effects of tensile and bending rigidities of reinforcement on the reinforcing mechanism for different reinforcing patterns were evaluated by examining the tensile force and bending moment mobilized in reinforcement and the strain fields of soil obtained from the FEM analysis. 1 INTRODUCTION Experimental research works have been extensively conducted over the last two decades for the purpose to evaluate the reinforcing mechanism of reinforced soil foundation so as to be able to predict the ultimate bearing capacity and to optimize the arrangement of reinforcement beneath the footing (e.g., Huang and Tatsuoka 1988, 1990). Huang and Hong (2000) analyzed the experimental results available so far, and suggested that the increase in the bearing capacity by reinforcing may be predicted by evaluating the “deep-footing mechanism” and the “wideslab mechanism.” On the other hand, it has been difficult to obtain realistic solutions for the deformation and failure of a reinforced soil foundation from numerical analysis. One of the reasons may be that the bearing capacity of a footing on sand has been obtained only for an idealized material model, while realistic solutions have not been obtained for most of the classical boundary value problems in geotechnical engineering due to computational difficulties in applying numerical analysis for a realistic non-linear soil model. The recent development in the numerical approach revealed that the bearing capacity characteristics of a footing on dense sand observed in physical model tests could be reasonably simulated by FEM only when based on appropriate modelling of sand properties (Tatsuoka et al., 1991; Siddiquee et
al., 1999). In their work, a number of factors that affect the strength and deformation of sand were taken into account, including; a) confining pressure; b) anisotropy; c) non-linear strain-hardening and strainsoftening; d) dilatancy; and e) strain localization into a shear band(s). For numerical simulation of reinforced soil structures, the interaction between the reinforcement and the surrounding soil should be considered in addition to the soil properties described above. By adequately modelling the reinforcement in the FEM procedure, the behaviour of unreinforced, single-layer reinforced and multi-layer reinforced sand specimens in plane strain compression tests was reasonably simulated with regard to the global stress-strain relations and shear banding (Kotake et al., 1999; Peng et al., 2000). This paper presents results from a numerical simulation of well-controlled laboratory model tests on reinforced soil foundation. The FEM results are compared with the physical experimental ones with regard to the load-settlement relations, improvement of the bearing capacity by reinforcing, shear band development in the reinforced zone for different reinforcing patterns. Then, the effects of tensile and bending rigidities of reinforcement in reinforcing soil structures and ground examined with regard to the tensile force and bending moment mobilized in reinforcement and the strain fields of soil are presented.
57 1
2 LABORATORY MODEL TESTS In a series of laboratory model tests on the bearing capacity of reinforced soil foundation conducted by Huang and Tatsuoka (1988, 1990), different reinforcing patterns were examined to evaluate the effects of the length, the number of layers, the horizontal spacing and the rigidity and rupture strength of reinforcement. A couple of reference tests were conducted on unreinforced sand ground. Figure 1 shows the setup of the laboratory model tests under plane strain conditions. Each model ground was produced by pluviating air-dried Toyoura sand to have very similar relative density values in a range of Dr= 80 - 86%. The model ground was 183 cm in width, 40 cm in length, and 74 cm in depth. A rigid footing, 10 cm wide and 39.8 cm long, with a rough base, guided against tilting and translation, was placed on the model ground surface. The sidewalls of the sand box consisted of 3 cm thick transparent acrylic plates with steel stiffeners to ensure the plane strain condition. The inside surface of the acryl plate was well lubricated by means of a silicone grease layer placed between the plate and a latex rubber membrane sheet in contact with sand. Displacement fields on the intermediate principal strain (€2) plane of the model ground were obtained from displacements at the nodal points of the 1 cmsquare meshes drawn on the outer surface of membrane. Reinforcing strips, having a thickness of 0.5 mm and a width of 3 mm, were made of phosphor ronze that were highly linear elastic. Their surfaces were made rough by gluing the particles of the model sand. The precise tensile forces were measured by means of strain gauges attached to the strips. Among the laboratory model tests, one test on unreinforced sand ground and two groups of tests onreinforced sand ground with short and long rein-
forcement layers (Figure 2) were selected for simulation. Group-a was conducted to study the effectiveness of short reinforcement layers having a length L equal to the footing width B (L/B=l), and also the effects of the number of reinforcement layers (n = I , 2, 3 and 5 ) and the depth of reinforced zone (DR/B =0.3, 0.6, 0.9 and 1.5) on the bearing capacity characteristics; Group-b was to study into the effects of the length of reinforcement layers, which extended laterally to the outside of the footing width (L/B=l, 2, 3.5 and 6) with a number of layers of n=3. In the case of L/B=3.5, L/B=2 only for 1st layer. 3 FEMMODEL A nonlinear elasto-plastic FEM analysis was conducted to simulate the laboratory model tests. The FEM model for sand considered pressure-dependency and anisotropy in the internal friction angle, strain-softening, dilatancy and shear banding. The details of the sand model have been reported in several papers by the authors (Siddiquee et al., 1999; Kotake et al., 1999; Peng et al., 2000). The FEM mesh is shown in Figure 3. With the symmetric nature of the problem, only a half domain
Figure 3. E M mesh. Figure 1. Setup of the model tests.
572
of the model ground was analyzed. Lateral and bottom boundaries of the analysis domain were assumed to be perfectly smooth. The total number of the plane elements and the nodal points are 540 and 58 1, respectively. &-stress condition is assumed for the initial stress state of the homogeneous level ground. The value of K, was obtained based on the empirical equation K,=0.52 . e, (Okochi and Tatsuoka, 1984). The unit weight of sand yd of 15.58 kN/m3 and the initial void ratio e, of 0.66 were used as representative values. Loading on the footing was simulated under displacement control. Uniform vertical displacements were prescribed to the nodal points along the footing base while lateral movement of those nodal points were fixed to simulate the fully rough base. The displacement increment was 0.00 1 cdstep, i.e., O.Ol%/step of the footing width B, that has been found to be small enough to keep accuracy and numerical stability. Layers of reinforcing strips were modeled by beam elements for fundamental cases. It was assumed that the tensile and bending rigidities of the reinforcement layer in the simulation were equal to Figure 4. Normalized load-settlement relations for unreinforced and reinforced ground. the respective average value of the strips placed in each layer in the physical experiment. The area A inforced zone. It is seen that the FEM analysis can and the moment of inertia I of the beam elements per reasonably simulate the physical experiment in this unit length to the €2 direction in each layer were thus respect. calculated based on the number of strips placed in each layer. For each layer, they were A = 9 . 0 ~ 1 0 - ~ For the ground reinforced with long reinforcement layers (Group-b), the results of the physical cm2 and 1 = 1 . 8 7 5 ~ 1 0cm4 - ~ per unit length to the €2 experiment show that in order to increase the beardirection, respectively. The behaviour of the reining capacity as well as the initial rigidity, increasing forcing strips was simply assumed to be linearly the reinforcement length L from B to 6B was found elastic based on the measurements on stress-strain to be not so efficient as increasing the number of relations of the strip used in the physical experishort reinforcement layers of L=B as was examined ments. The Young’s modulus of the reinforcement E in Group-a. In the physical experiment, the effect of was taken equal to 1 . 2 2 ~ 1 0kN/m2, ~ which is the using reinforcement longer than the footing width on same with that of the reinforcing material used in the the total increase in the bearing capacity becomes physical experiment. less significant as L becomes longer than B, and it looks that the upper bound is reached as L exceeds 2B. It may be seen that the FEM analysis can simu4 LOAD-SETTLEMENT RELATIONS late this behaviour rather accurately. However, the FEM analysis largely overestimated the values of N, Figure 4 compares normalized load-settlement relafor the cases of L/B>2. tions in term of N=2q/(yd B) and S/B for the unreinIt is the general tendency that the FEM results forced and the reinforced ground obtained from the exhibit the larger pre-peak stiffness and the smaller physical experiment and the FEM analysis. In the settlement at peak state compared to those from the FEM analysis, the footing pressure q was obtained physical experimental tests. A number of factors by simply averaging the vertical stresses at the gauss must be related to the discrepancy seen in the loadpoints in the soil elements, which were immediately settlement relations. beneath the footing base. For the settlement S, the vertical displacements of the footing base were used. For the ground reinforced with short reinforce5 STRAIN FIELDS ment layers (Group-a), the results of the physical experiment clearly show that the initial stiffness, the Strain fields in the €2 plane obtained from the dispeak footing load, and the settlement at peak load of placements measured in the physical experiments reinforced ground increase with the increase in the with a sufficient accuracy were compared with those number of reinforcement layers, i.e., the depth of re-
573
obtained from the corresponding FEM analyses. Figure 5. shows the contours of maximum shear strains z71ax( = E ] - E ~ ) of a set of unreinforced ground. and reinforced ground with different length (L/B=l, 2, 3.5 and 6) and n=3 of reinforcement, respectively. Here, all of the strain values are accumulated ones from the initial condition. In the test with L/B=l, tensile reinforcement layers were placed immediately beneath the footing to restrain potential large strains forming a wedge beneath the footing as in the unreinforced ground. By reinforcing the ground as shown in Figure 5, only small strains were induced inside the zone reinforced to a depth of DR immediately beneath the footing. Instead, intensely sheared bands were formed along the lateral faces of the reinforced zone. By placing reinforcement layers longer than B, the shear bands were spread into wider areas than in the case with L/B=l. This attributes to the rigidity of reinforcement, which spread the footing load more widely, and to the tensile forces that were mobilized in the reinforcement along with deformation of sand, which restrained the shear band development as in the unreinforced ground. On the other hand, the strain fields did not change so largely by increasing L/B to larger than 2. It may be seen that the shear bands diffused due to the interaction between the shear bands and the reinforcement layers. Despite the above, a wedge was formed beneath the reinforced zone that is similar to the one in Group-a (L/B= 1). It is seen from these results that the shear banding characteristics observed in the physical experiment on unreinforced and reinforced ground were well simulated by the present FEM analyses.
6 REINFORCEMENT FORCES Figure 6 shows tensile forces in the reinforcing strip of each layer at steps with an increment of S/B=O.Ol measured in the physical experiment with L/B=3.5 and n=3. Figure 7 shows tensile forces and bending moment obtained from the corresponding FEM analysis. It is seen that the FEM analysis can capture the major characteristics of the mobilized tensile forces seen in the physical experiment. The distribution of the tensile forces indicates the following points : a) It is seen that the tensile force in the reinforcing layers is largest at or around the centerline, and it becomes much smaller outside the footing width. The slope of the distribution curve of tensile force becomes steeper as the loading proceeds. This indicates that the interaction force between soil and reinforcement, that is the shear force mobilized at the interface between the two, increases with the increase in mobilized tensile force in the reinforcement.
574
b) The largest tensile force is induced in the bottom layer because the layer is deformed to the largest extent together with the adjacent soil zone beneath the reinforced zone that supported the bearing pressure without reinforcement. c) The tensile force in the central part becomes the maximum at or around the peak footing load and then exhibits unloading. This behaviour may be associated with unloading in the shear stress induced in the soil in contact to the reinforcement. d) As seen in the strain fields, the effect of restraining the potential tensile strains in soil is largest immediately beneath the footing. On the other hand, the extended portions of reinforcement beyond the footing width can contribute only in a secondary manner to the increase in the bearing capacity. This can be explained that the interaction force between the reinforcement and the surrounding soil can be mobilized effectively under sufficiently large normal stresses on the reinforcement, which is not the case with the part of reinforcement located outside the footing width. e) The bending moment is mobilized gradually in the pre-peak regime, while it increases rapidly after the peak footing load. It is particularly the case at the location where the intense shear band crosses the reinforcement layer. The increase in
Figure 6. Tensile forces in reinforcing strips obtained from physical experiment, LIB=3.5 and n=3.
Figure 7. Tensile force and bending moment obtained from FEM analysis, WB=3.5 and n=3.
575
the bending moment that is due to the shear deformation in the shear band is largest immediately beneath the footing edge.
7 FEM ANALYSIS ON BENDING RIGIDITY Contribution of the bending rigidity E1 of reinforcement can be evaluated by conducting FEM analysis without bending rigidity of reinforcement. Figure 8 shows normalized load-settlement relations for the reinforcing patterns of L/B=1 and 2 with n=3 obtained from the FEM analyses performed to compare the cases between with and without bending rigidity, obtained by using either beam or truss element for the reinforcement model. In the case with short reinforcement layers (L/B=l), the bending moment was scarcely induced in the pre-peak regime in the physical experiment. Correspondingly, the load-settlement relations obtained by using either beam or truss elements were very similar not only in the pre-peak regime but also in the post-peak regime. This result indicates that for this type of reinforcing pattern, the tensile rigidity is the controlling property for reinforcing members. On the other hand, in the case with longer reinforcement layers (L/B=2), the bending moment was to some extent induced in the physical experiment. Correspondingly, in the numerical analysis the difference in the peak footing load by the effect of the bending rigidity of the reinforcement was considerably large. It was found that the load-settlement relation without bending rigidity of the reinforcement was similar only up to S/B=0.05, by which shear bands had developed intensely beneath the footing edges and had crossed the reinforcement layers. Consequently, it can be concluded that the bending rigidity of reinforcement can contribute effectively to the increase in the ultimate bearing capacity of reinforced soil by restraining the shear band development only when shear bands have developed and have crossed the reinforcement layers. So, it is relevant not to expect the effects of reinforcement bending rigidity when dealing with the behaviour of reinforced soil before the shear band development.
8 CONCLUSIONS From the results of analysis in the present study, the following conclusions can be derived:
a) The numerical results from the present FEM of unreinforced as well as reinforced soil foundation agreed reasonably well with the physical experiment with respect to the reinforcing effects for
576
Figure 8. Normalized load-settlement relations when using either beam or truss element.
various reinforcing patterns using different numbers of layers and lengths of reinforcement. b) The tensile rigidity of reinforcement is the very important property for full range of loading. The bending rigidity of reinforcement becomes important after intense shear bands develop crossing the reinforcement. REFERENCES Huang, C-C. & Tatsuoka, F. 1988. Prediction of bearing capacity in level sandy ground reinforced with strip reinforcement. Proc. Int. Symp. Theory and Practice of Earth Reinforcement: 191-196. Huang, C-C. & Tatsuoka, F. 1990. Bearing capacity of reinforced horizontal sandy ground. Geotextiles and Geomembranes. 9(5 1-82): 236-267. Huang, C-C. & Hong, L-L. 2000. Ultimate bearing capacity and settlement of footings on reinforced sandy ground. Soils and Foundations 40(5): 65-73. Kotake, N., Tatsuoka, F., Tanaka, T., Siddiquee, M.S.A. & Yamauchi, H. 1999. An insight into the failure of reinforced sand in plane strain compression by FEM simulation. Soils and Foundations 39(5): 103-130. Okochi, Y. & Tatsuoka, F. 1984. Some factors affecting K<)values of sand measured in triaxial cell. Soils and Foundations 24(3): 52-68. Peng, F., Kotake, N., Tatsuoka, F., Hirakawa, D. & Tanaka, T. 2000. Plane strain compression behavior of geogridreinforced sand and its numerical analysis. Soils and Foundations. 40(3): 55-74. Siddiquee, M.S.A., Tanaka, T., Tatsuoka. F., Tani, K. & Morinioto, T. 1999. Numerical simulation of bearing capacity characteristics of strip footing on sand. Soils and Foundations 39(4): 93- 109. Tatsuoka, F., Okahara, M.. Tanaka, T., Tani, K., Morimoto, T. & Siddiquee, M.S.A. 1991. Progressive failure and particle size effect in bearing capacity of a footing on sand. McLean et al. (ed.), Geotec. Engrg. Cong.; ASCE Geotech. Special Publication 27-2: 788-802,
Landmarks in Earth Reinforcernenf,Ochiai et al. (eds), Q2001 Swets & Zeithger, ISBN 90 2651 863 3
Collapse loads on reinforced foundation soils Radoslaw L. Michalowski Uniyersity of M d z i g a n , Anrz Arbor. lLSX
Xuemei Xin Umversitj ofMirhigarz, Aniz Arbor, USA
ABSTRACT: Unpaved roads and reinforced fills are among conirnon applications of geosynthctics in thcir reinforcement function. Calculations of limit loads on rcinforccd foundation soils are typically based on very approximate assumptions, often based on small-scale experiments. The kinematic approach of limit analysis is used here to indicatc what a rational solution to thc limit load niight look like. Preliminary results for a foundation soil reinforced with one layer of reinforcement are presented.
1 INTRODUCTION Geosynthetic reinforcement is often used to improve performance oT paved and unpaved roads, and, more generally, to reduce settlement and increase bearing capacity of engineered fills. Limit loads on foundation soils have been a subject of research since the 19SO’s, and a substantial body of literature is available in regard to the strip footings on uniform and isotropic soils. An application of traditional mcthods to stability analysis of reinforced soils requires some modifications. Kinematic method of limit analysis offers a common framework for calculations of limit loads for homogeneous and inhoinogeneous soils, isotropic and anisotropic, and natural and reinforced soils. Two approaches can be clearly distinguished in stability calculations of reinforced soils: continuum analysis and the structural approach. In the first one the reinforced soil is first homogenized to form an anisotropic c~ntinuum,and the calculations of bearing capacity are performed using the finite element or the slip line method. The second approach involves limit analysis where both the soil and reinforcement arc considered as two separate structural components. The latter approach will be focused on in this paper.
2 STABILITY ANALYSIS OF REINFORCED FOUNDATION SOIL
2. I Kinematic approach Limit analysis is a common framework within which solutions to many stability problems can be found. Although kinematic approach guarantees the upper bound (thus unsafe) to the true limit load. it is more 577
commonly applied than the lower bound, since the kinematically admissible collapse mechanism are easier to construct and optimize than statically admissible strcss fields. Calculations of the limit load are based on the upper-bound theorem, which states that the work dissipation rate is not less than the work rate of external forces in any kjneniatically admissible mechanism
The left-hand side of eq. (1) represents the rate of work dissipation in the mechanism, while the terms on the right-hand side show the work rate of the limit load on boundary S; and the work rate of soil weight. Hence, an upper estimate of the limit load can be calculated from eq. (1) once the dissipation rate and the work of the soil weight are known. Equation (1) will be used later to arrive at a reasonable formula for limit loads over reinforced foundation soils.
2.2 Collapse pattern Reinforcement is limited here to one layer of geosynthetics placed in a granular fill. In an earlier paper by Huaiig & Tatsuoka (1990) two effects were distinguished depending on the length of the reinforcement: “a deep footing effect” associated primarily with short reinforccmcnt, and a “wide slab effect.” Only long reinforcement is considered in this paper. The mechanism postulated for the interaction of soil and reinforcement is similar in shape to the mechanism without rcinforcement, as indicated by earlier experiments (Michalowski 1998). This mechanism is shown schematically in Fig. l(a), and the velocities of the blocks are indicated in the hodograph in Fig. l(b).
Figure 2. The effect of reinforcement depth d/B on the bearing capacity of strip footings (one layer of reinforcement).
2.3 Preliminary results
Figure 1. Collapse of reinforced soil: (a) failure mechanism, and (b) hodograph.
Collapse of reinforcement can occur in two ways: rupture or pull out. Calculations of bearing capacity in case of reinforcement rupture are somewhat easier to address using a homogenization approach (Michalowski & Zhao 1995). Here the focus is on the pull out mechanism. Calculations of work dissipation rate during pull out require that the traction on reinforcement be known. However, the stress distribution is unknown, and, to make the calculations possible, a realistic yet approximate distribution of traction on reinforcement had to be assumed. This assumption relaxes the rigorous character of limit analysis, though the result is still expected to be a reasonable solution. Following the distribution of the mean stress in a plasticity-based solution for a bearing capacity problem, we assume that the vertical stress distribution on reinforcement is constant across the block immediately below the footing (Fig. I(a)), and constant across the block adjacent to the boundary (the last block in the mechanism, the first one being the one immediately below the footing). This traction, while assumed constant across these two blocks, is increasing with depth. The normal stress on reinforcement is then assumed to change in a linear fashion in the “transition” zone (between the first and last blocks). Having made this assumption, calculations of the work dissipation rate during reinforcement pull out become possible, and the framework of the kinematic approach of lirnit analysis can be used.
578
Having the normal stress distribution on the reinforcement, the work dissipation rate was calculated, and eq. (1) was used to indicate whether the model yields reasonable results. Calculations were performed for different reinforcement depth d (internal friction Cp = 35’, geosynthetic/soil interaction friction angle C p = 26’, b/B = 6). The function plotted in Fig. 2 indicates the percent increase in bearing capacity as function of the depth of reinforcement below the footing. As the calculations were performed for a surface footing, there is no increase of limit load if the geosynthetic is placed at d = 0. The benefit from the reinforcement then increases gradually up to the depth of about 0.8 of the footing width, and then drops off very rapidly after that. The maxiinum benefit from one-layer reinforcement appears to be about 40% in terms of the bearing capacity increase. Similar calculations for cohesive soils (Cp = 0, c > 0) indicated about 30% increases in bearing capacity and optimum reinforcement depth of about 0.45B. These results seem to be very reasonable when compared with experimental results available. The optimum depth of reinforcement was reported ranging from about 0.4 d/B for cohesive soils (Sakti and Das 1987) to about 1 for granular soils (Yang et al. 1994). 3 BEARING CAPACITY While attempts have been made in the past to derive a formula for bearing capacity increase associated with the rupture of reinforcement (Giroud & Noiray 1981), no pull out mechanism has been included in such analyses. In the following, we are suggesting that a general formula can be derived from the structure of the energy rate balance equation. Considerations are limited here to granular fill (no cohesion) and one layer of reinforcement, but they can be easily extended to a more general case.
The left-hand side of eq. (1) contains only the work dissipation rate due to reinforcement pull out, since soil has no cohesion. The terms on the righthand side include the work rate of the distributed limit load p , and the soil weight y. Since the traction on the reinforcement is dependent on p and y,then the work dissipation rate will have two terms. Equation (1) can be rewritten here in a simple form
D,,
d, 2 q,+
4-
(2)
Assuming that the footing moves downward with velocity v,, the work rate of the limit load is
while the work rate of the soil weighty is
w, =CS,y., 2
LLll
(4)
,=I
where p i s the bearing pressure, v, is the vertical component of the velocity of the first block, v,''''~'is the vertical component of velocity of block i, y is the unit weight of the soil, Si is the area of block i, and n is the number of blocks in the mechanism. Substituting equations (3) and (4) into (2),and solving for bearing pressure p, onc can present the result in the following form
wherefis the reinforcement roughness coefficient " f tan
Q,,
tan 4 where qW is the soil-reinforcement interface friction angle. Coefficient N, is adoptcd herc after a recent limit analysis-based proposal (Michalowski 1997) N = e(0.65+5 l i I . r n @ ) tan # Y (6) The remaining coefficients in equation (5) have heen evaluated through a series of numerical calculations for variety of internal friction angles.
4 NUMERICAL CALCULATIONS Several series of calculations were perfornicd in or-der to evaluate coefficients M,, and M , Each series was performed for a variety of internal friction angles and different depth of die reinforcement (and f = 0.75).
Because the geometry of the collapse mechanism (Fig. l(a)) is not known a pi-ioi-i, an optimization technique was used to arrive at the minimum bearing capacity for each set of internal friction angle and the reinforcement depth. The geornetry of the mechanism was variable in the optimization process.
Figure 3. Cocfficicnt M , as iiinctioii of tan@
The form of eq, (5) is the direct implication of the work balance equation. The numerical values of the terms in ( 5 ) were extracted from the optimization computations, and the coefficients M , and were plotted as functions of the internal friction angle and the reinforcement depth. Figure 3 shows coefficient M , as a function of tan$. As expected, M , i s an increasing function of internal friction angle. An interesting plot is that showing (d/BjM, as function of relative reinforcement depth CUB(Fig. 4). For the fill up to about 30" fd/B)M, increases rather slowly with an increase in CUB?and it drops to zero at depth where thc mechanism is not influenced by the reinforcement. For larger internal friction angles (d/B)A4, increases more rapidly with an increase in reinforcement depth, and it drops to 7ero at much larger depths. For practical reasons coefficient M , was approximated with an exponential function (using the least squares technique). The resulting formula can he written here as M = el l i i 3 9 8 c a n g tan @ (7) It should be noted that eq. ( 7 ) can be used only when the reinforccmcnt is placed at a depth where it clearly contributes to bearing capacity. This depth depends on the internal friction angle of the soil and it varies from zero to 0.65 for $ = 30*, 1.0 for Q, = 40",etc. These depths can be incurred from Fig. 4. Coefficient M p was also found based on the optimization calculations of p. It was also found to be a nonlinear function of Q,, but it appears to be approximately linear in tan$ M , = 0.22 tan #
(8)
Coefficients in ( 7 ) and (8) now can be used directly in calculations of bearing capacity of surface footings using the formula in ( 5 ) .
579
The analysis has been restricted here to surface footings on a granular fill, but the concept can easily be extended to footings on cohesive-frictional soils reinforced with geosynthetics. The work is underway to generalize this approach to include surcharge load, soil cohesion, and multiple layers of reinforcement. REFERENCES Giroud, J.P. & Noiray, L. 1981. Geotextile-reinforced unpaved road design. Jrnl. Geotechnical Engineering Division, 107: I 233- 1254. Huang, C.C. & Tatsuoka, F. 1990. Bearing capacity of reinforced horizontal sandy ground. Geotextiles and Geonzeinhrarzes, 9: 5 1-82. Michalowski, R. L. 1997. An estimate of the influence of soil weight on bearing capacity using limit analysis. Soils & Foundations, 37(4): 57-64 Michalowski, R. L. 1998. Limit analysis in stability calculations of reinforced soil structures. Geotextiles and Ceomenzbranes, 16: 3 1 1-331 Michalowski, R. L. & Zhao, A. 1995. Continuum versus structural approach to stability of reinforced soil. J. Ceotech. Engrg., ASCE, 121: 152-162. Sakti, J. P. & Das, B. M. 1987. Model tests for strip foundation on clay reinforced with geotextile layers. Transportation Research Board, 1153: 40-45 Yang, J., Ochiai, H. & Hayashi, S. 1994. Experimental study on bearing capacity of geogrid reinforced foundation ground. Proceedings of Jqxinese Society of Civil Engineers, 499: 3-28.
Figure 4. Product (d/B)M,as function of relative depth of reinforcement d/B for different internal friction angles (I$).
5 CONCLUSIONS A consistent use of limit analysis was presented to calculate the bearing capacity of reinforced fills.
580
Landmarks in Earth Reinforcement,Ochiai et al. (eds), @ 200 1 Swets & Zeitlinger, ISBN 90 265 1 863 3
Model loading tests on the footing reinforced with prestressed micropiles Kinya Miura Associate Professor, Asian Institute qf Technology, Bangkok, Thailand
Yukihiro Tsukada Chief Manager, Toizoku Construction Branch, Ministry of Construction, Japan
Yoshinori Otani Research Engineer, Nit-me, Cu. Ltd., Osaku, Japan
Mizuho Ishito Graduate Studenr, Hokkaidv University, Sapporo, Japan
Guan-Lin You Graduate Student, Asian Institute of Technology, Bangkok, Thailand
ABSTRACT: To investigate the mechanism of the bearing capacity of footing reinforced with a group of micropiles, a series of loading tests were carried out on the model footirigs with and without micropiles. To mobilize the network effect appropriately and improve the bearing capacity, a tensile stress was induced in the micropiles at the beginning of the loading process in this series of model loading tests. As a result, the bearing capacity was improved more than 100% with the prestress of 7 0 8 of the pullout resistance, not only in dense and dilative ground, but also in the loose and contractive ground. The network effect of a group of nlicropiles on the bearing capacity is mobilized positively with the appropriate confinement of the grourid material beneath the footing. In this process the micropiles not only provide load bearing capacity directory through thcir skin friction, but also raise the base pressure on the footing with the confinement.
1 INTRODUCTION Micropiles are now widely used both as structural supports in foundations as well as for in-situ earth reinforcement. Pioneered by Lizzi (197 1, 1978) in Italy, micropiles now enjoy worldwide recognition (US Development of Transportation, 1997. 8r Tsukada, 1997). Micropiles are claimed to cause minimum disturbance to structures, subsoil and the environment. Furthermore, they can be cast-in-place replacement piles with small diameters and can be easily installed in pre-drilled boreholes containing steel rods as reinforcement and grouted under pressure. It is thus not surprising that micropiles are considered as promising foundation elements in improving the bearing capacity of existing foundations which are deteriorating for one reason or another. The Hyogoken-Nambu earthquake in 1995 in Japan caused extensive damages to the foundations of bridges and this in turn has triggered research and development on the use of micropiles in strengthening foundations. Additionally, the design concept of micropiles offers a wide range of flexibilities by which they can withstand axial and/or lateral loads. ‘They can be considered either as a single component in a composite soilipile mass or a small diameter substitute in a conventional pile. Micropiles can sustain sufficient load by friction ils there are grouted piles installed under controlled pressures. Because of their flexibitity and their installation in small diameters, they can be used conveniently as a group in reinforcing
spread footings. The interaction between the footing and the micropile group makes it susceptible to large loads and displacements under earthquake type of destruction. In the reinforcement of existing foundations, the micropiles and footing are considered to be a piled-raft foundation in their performance. Although the applications of micropiles are increasing in various situations, their mechanism of developing the bearing capacity is not yet fully understood. Thus the aim of the scrics of study on model micropile is to investigate some important aspects which classify and quantify the development of bearing capacity in micropile foundations and to develop new method to improve the performance of micropile foundations. In previous study, the method o i the model loading tests has been reported by Y. Tsukada et al. (1999), and the test results on three series of model tests (or. footing test, micropile test; and micropile foundation test) have been reported by Miura et al. (2000). Three types of micropiles with different bending stiffness and surface roughness but without prestress were used in previous study, which hereafter are called non-preswessed micropiles to distinguish them with the prestressed micropiles in this study. In this study, the model micropiles were vertically installed beneath the footing in the artificiai sand ground. And the behavior is observed under some series tests of micropile foundation in dense and medium-dense ground under vertical and inclined loads. To examine the mechanism of the interaction and improve bearing capacity effectively,
58 1
ginning of the model loading tests. The soil material enclosed by the micropile group and the footing was confined due to the prestress, as a result the stiffness of the soil material and base pressure of the footing were increased. The overall bearing capacity was improved in both vertical load and inclined load. 2 METHOD FOR MODEL LOADING TESTS In this study, a newly developed apparatus for the model micropile foundation test has been introduced to apply prestress onto the micropiles as shown in Figure 1, which also made it possible to measure the load carried by the micropile group and the load by the footing separately. So the mechanism of interaction between the footing and micropiles was examined. The other main characteristics of the tests are summarized below, for the details of the testing method, refer to Y. Tsukada et al. (1999). The model footing is made of stainless steel with a diameter of 40 mm, which is reinforced with a group of micropiles; five types of model footings are available to fit the need of variation of the pile number and pile inclination angle in the micropile group. Three types of model micropiles were used in previous non-prestressed study. Two types, S-Stype and S-R-Type, are made of stainless steel with high bending stiffness of EZ = 1.28~10-INm’ ( E = 2 . 1 ~ 1 0MPa); ~ and the other type, P-R-Type, is made of lastic with low bending stiffness of EZ = 3pNm2 ( E = 3 . 1 ~ 1 0MPa). 2.50~10~ Two rough surface types, S-R-Type and P-R-Type, were coated with thin sand layer so as to mobilize sufficient skin
Figure 1. Newly developed device for the inducement of prestress on micropile group.
Value ~ - - __ ~___ (a) Sand Particles Grain density, p, 2.717 g/cm3 1.610 g/cm3 Max dry density, ptl,,I,l, Min dry density, P,,,,~~,~ 1.255 gkm’ Mean grain size, Ds0 0.18 mm Uniformity coefficient; U, 1.82 (b) Grounds Type of ground Dense Medium Loose Relative density, Dr 9 5 2 2 % 6522% 5022% Angle of friction, gd 38.5 deg 36.2deg 34.8 deg Parameter-
~
~
friction with ground; and sand particles were glued to the micropile surface. In current study, S-R type micropile was used for prestressed micropile foundation test. The model micropile foundation was set up on the surface of the model sand ground formed in a mold. Oven dried silica sand was deposited through air with a nozzle, and tapped with a rubber hummer so as to obtain prescribed three different relative densities: dense, medium, and loose grounds. The physical and mechanical properties of the model sands are shown in Table 1. In this study on model micropile foundation, eight S-R type micropiles were used to reinforce the model footing, which were vertically installed. Dense and medium-dense model grounds were used to compare the effect of grounds. Both vertical load and horizontal load were applied in this study. The load ratio k is defined as horizontal load over vertical load. For vertical loading test, k = 0. The purpose of applying horizontal load is to model the effect of footing reinforced with micropiles under seismic loading condition. Shown in Figure 2 is a photo of the new device after a loading test.
Figure 2. Photo of the test device.
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3 TEST RESULTS AND ANALYSIS To assess the degree of improvement of the bearing capacity with micropiles quantitatively, the network effect index Z,lc was newly introduced (Fig. 3). The I,,, of unit means that the bearing capacity of the footing reinforced with micropiles is equal to the summation of the surface footing and the micropile group. If the confining effect by the interaction between the subsoil, footing and the micropile group is positive, the interaction improves the bearing capacity, and the value of I,,, becomes larger than unit. In the non-prestress micropile model tests, significant effect of the relative density on the bearing capacity was recognized, in the tests of surface footing, a group of micropiles and foundation reinforced with micropiles. In dense ground, due to the dilatant behavior of ground material, bearing capacity was remarkably high compared with those in medium and loose grounds (Fig. 4). Because of the increase in confining pressure on the surface of micropiles due to the dilatant behavior of dense ground material, the skin friction of micropiles is remarkably increased. It was found that the confinement of the ground material with the interaction between the footing and the micropile group plays an important role in the improvement of the bearing capacity. The confinement increases the confining stress in the ground material beneath the footing and then the base pressure on the footing as shown in Figure 6. The confining effect is a function of the geometry
and stiffness of the micropile group and the mechanical properties of ground materiaI. However, the improvement of bearing capacity, which is positive effect of the confining effect, was not mobilized in loose and medium grounds and also in the early stage of loading with small displacement in dense ground (Figs. 4&5). To induce the confining effect more effectively even under small displacement, it is necessary to raise the confining stress on the ground material beneath the footing at initial stage. Then, in the series of model loading test on the micropile foundation, some amount of
Figure 5. Effect of relative settlement on network effect index
I,,,.
Figure 6. Improvement of bearing capacity due to the confining effect on ground material beneath footing; (a) in dense ground, (b) in loose ground.
Figure 3. Definition of network effect index for bearing capacity o fiiiicropile foundation.
Figure 7. Concept of prestress on micropile group and induce base pressure and confining pressure.
Figure 4. Effect of relative density on network effect index I,,,,.
583
prestress was induced to the micropile group. Shown in Figure 7 is the illustration of how to apply prestress; as a reaction of the prestress, the initial base pressure and confining stress was simultaneously increased. The prestress test was conducted on the footing reinforced with eight S-R type micropiles on the dense and medium-dense sand. The load on the micropile group could be measured with the load cell inside the model micropile foundation; the base pressure was evaluated from the difference between the total load and the load carried by micropile group. The degree of prestress was indicated with of the prestress on the mithe ratio -Q,,~~~~/Q,,*/,,,,~,~ cropile foundation to its ultimate value in micropile test. The inclination of the load is indicated by k. Under vertical loading condition ( k = O), the observed load-displacement behavior is shown in Figures 8 & 9 for dense and medium-dense ground, respectively; where the averaged base pressure ql,, the equivalent base pressure Q,n,,/A~for the load carried by micropiles and the base pressure evaluated as qj (= ql, - Q,,l/,/Af)are plotted against settlement S13.In the case of non-prestressed micropile group (Q,,l~~~/Q,,,,,,,,cll = 0) the base pressure is rather low and does not increase with On both Figure 8, Effect of prestress on the behavior ofniicro pile faundense and medium grounds. Due to the k~duced dations on dense ground; (a) total load, (b) load on micropile prestress on micropile the load on micropile group group, (c) base pressure. becomes negative and base pressure increases at initial condition. And the base pressure increases notably and as a result the total load increases at a certain displacement. The load carried by micropile group Q,,,/, decreases with the prestress and the micropile group gains the margin until the ultimate condition. Shown in Figures 10 & 11 is the improvement of bearing capacity with the prestress in dense and medium-dense sand grounds, respectively. With the increase of prestress, the load on micropile group was reduced monotonically, on the other hand the base pressure on the footing increased; this trend is recognized at all the levels of displacement until S,/D = 20%. As a result the base pressure increases with the prestress even under initial stage of loading ( S J D = 5%). The bearing capacity was increased almost 100% with the prestress equivalent to a half of ultimate bearing load of micropile group. Under both vertical and horizontal loads, the observed load-displacement behavior in dense sand is shown in Figures 12 & 13 for k = 0.3, 0.6, respectively. The observed variations of load-displacement are similar to those observed in vertical loading test, which demonstrates the effect of prestress on the improvement of bearing capacity under seismic loading condition. Shown in Figures 14 & 15 is the influence of k on the load bearing capacity in dense and medium-dense sand, respectively. It is found Figure 9. Effect of prestress on the behavior of micro pile founthat wheIl the k increases the of imdations on medium ground; (a) total load. (b) load on micropile provement induced by prestress decreases. group, (c) base pressure.
584
Figure 10. Improvement of bearing capacity with inducement of prestress on micropile group in dense ground.
Figure 12. Effect of prestress on the behavior of micro pile foundations on k = 0.3; (a) total load, (b) load on micropile group, (c) base pressure.
Figure 11. Improvement of bearing capacity with inducement of prestress on micropile group in medium dense ground.
Figure 13. Effect of prestress on the behavior of micro pile foundations on k = 0.6; (a) total load, (b) load on micropile group, (c) base pressure.
585
4 CONCLUSIONS To investigate the effect of prestress on the improvement of the bearing capacity of footing reinforced with a group of micropiles, a series of model micropile foundation tests has been carried out. The circular footings were reinforced with a group of SR type micropiles, and were subjected to different prestress in both dense and medium-dense sand. Based on the observed load-displacement behaviors from the comparative examinations under different loading conditions, the following concluding remarks were drawn: Prestress increases the interaction between the footing and the micropile group, and this interaction was significantly effective on the confinement of ground material and on the improvement of the bearing capacity of footing in dense sand. It was found that the prestress on the micropile group is effective on the improvement of bearing capacity of micropile foundation in both dense and medium sand. It was effective even in initial loading stage with small displacement on both dense and medium sand ground. Under the test condition, the prestress equivalent to a half of the ultimate bearing load on the micropile group increased the total bearing capacity of micropile foundation by 100%. Prestress increases the overall bearing capacity of micropile foundation, but the load carried by micropile decreases with the prestress. It indicates that the footing is bearing more load with improved contact between footing and soil under high prestress. Prestress is also effective in increasing bearing capacity of footing under horizontal force.
Figure 14. Influence of k on the load bearing capacity in dense sand.
REFERENCES
~i~~~~ 15. ~ dium-dense sand.
Lizzi, F. 1971. Special Patented Systems of Underpinning and more Generally, Subsoil Strengthening by Means Of Pali Radice (Root Piles) with Special Reference to Problems Arising from the Construction of Subways in Built-up Area, Special Lecture given at university of Illiiiois at UrbunaClzainpaign, etc. Lizzi, F. 1978. “Reticulated Root Piles to Correct Land Slides,” Proceedings cf ASCE Coi$ereiice, Chicago, Illinois, October 16-20. 1-25. Miura K., Tsukada Y., You G. L., Ishito M., Otani Y. & Tsubokawa Y. 2000. Model Investigation On The Bearing Mechanism Of Footing Regarding The Interaction Between The Footing And A Group Of Micropiles. Proceedings qf’ tl?e 3‘“ in te rtz CItiorz a I coizfe rerice on g ro iiiz cl impro veineizt techniques, September 2000, Singapore. 255-262 Tsukada, Y. 1997. State-of-the-Art: Application of Micropiles in Japan, Proceeditzgs qf 1st International Workshop on Micropiles, Deep Foundations Institute, 265-279. Tsukada, Y, Miura, K. & Tsubokawa, Y. 1999. Model Loading Tests on Micropile Foundation on Sand Ground, Japanese Geotechnical Society, T.wclzi-to-kiso, 47( I), 35-38 US. Department Of Transportation, 1997. Federal Highway Administration: ~ of k on fthe load lbearing ~capacity ~in me- ~ ~ ~ Drilled and Grouted Micropiles: State-oj Practice Review, Vols. 1-4.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Mechanical properties of soilbags and their applications to earth reinforcement Hajime Matsuoka & Sihong Liu Department of Civil Engineering, Nagoya Institute of Technology, Nagoya, Japan
Keizaburou Y amaguchi Toslzo Engineering Co., Ltd., Yokohama, Japan
ABSTRACT: In this paper, we first present the mechanical properties of soilbags and their verifications by unconfmed compression tests on both model and real soilbags. Then, we report the applications of soilbags in earth reinforcement, such as the reinforcement of soft foundations where the N-values are as low as 1-2 and the construction of retaining walls. When reinforcing soft foundation, soilbags have also an effect of reducing traffic-induced vibration in addition to the increase of bearing capacity. Through this study, we believe that the reinforcing method using soilbags will become a promising method for earth reinforcement.
1 INTRODUCTION For a long time, people have used soilbags to prevent a flow of soils from floodwater and build temporary structures in case of emergency. But, it seems no applications of soilbags in building a permanent structure. As a result of the studies on soilbags by Matsuoka et al. ( 1 999, ZOOO), many advantages of soilbags, such as improving bearing capacity of soft ground, being friendly to our environment, reducing traffic-induced vibration and so on, have been elucidated and soilbags have been applied to various aspects of earth reinforcement. In this paper, we first present the mechanical properties of soilbags and their verifications by unconfined compression tests on both model and real soilbags. Then, we report several application cases in earth reinforcement. 2 STRENGTH PROPERTY OF SOILBAGS AND ITS EXPERIMENTAL VERIFICATIONS Figure 1. Stresses acting on two-dimensional model soilbag and On particles inside the soilbag.
Let us consider the stresses acting on a 2D soilbag. Figure l(a) shows a 2D soilbag subjected to the principal stresses, and 03. Under the application of 01, and 03, the total perimeter of the bag usually extends and a tension T takes place in the bag. This tension produces an additional stress that acts on the soil particles inside the soilbag, whose components are expressed as B,),= 2 T I B I
o O =32 T l H I
soil particles inside the soilbag are the combined result Of the applied Stresses and the ad& tionallY Produced Stresses by T, as shown in Figure At the following equation holds (Matsuoka et 2000a): 2T l+sin@ 0, += K,, o3+%),where K,, = - (2) B I-sin@
(
II\
\'I
By comparing Eq.(2) with the strength expression of 01f = a31 -+ 2C for a cohesive-frictional ma-
where B and H are the width and height of the soilbag, respectively. Thus, the stresses acting on the
K ~ j
6
terial, we can obtain the expression of the apparent cohesion c of soilbags due to their tension T.
587
This suggests an interesting fact that an apparent cohesion produces in the soilbag due to the tension of bag although the original materials inside the soilbag are frictional materials such as sands and crushed stones. In order to verify the above strength property of soilbags, a number of biaxial compression tests were performed on 2D paper-wrapped model soilbags. The model soilbag has dimensions of 15cm wide by 3.75cm high. The materials inside it are two kinds of aluminum rods with diameters of 1.6mm and 3mm and a length of SOmm that are mixed in a ratio of 3:2 by weight. The angle of internal friction, @ , of this aluminum rod mass is about 25". Owing to the limitation of the loading capacity in our biaxial compression test apparatus, a very weak paper with a tensile strength of about 0.36kN/m is used in the model soilbag. To minimize the boundary influence of the apparatus, three model soilbags are piled up to construct a specimen, as shown in Figure 2. The results of these tests are expressed with the solid Mohr's stress circles at failure in Figure 3. It can be seen from these solid Mohr's stress circles that the soilbags have an apparent cohesion. From Eq. (3), it is calculated that the apparent cohesion, c, of the model soilbags is 14kPa. The solid straight line in Figure 3 is drawn by using c=14kPa and @ =25" . It is approximately tangent to the solid Mohr's stress circles, indicating that Equations (2) and (3) can represent reasonably the strength characteristic of soilbags. Furthermore, the broken Mohr's stress diagrams in Figure 3 represent the stresses of the material inside the soilbags at failure after combining the additional stresses that are produced by the tension, Tf = 0.36kN/m, of the paper-made bag. The envelop of these broken Mohr's stress circles is a straight line passing through the origin with an inclined angle of 25", which is equal to the angle of internal friction of the material inside the soilbags. This indicates that the materials inside the soilbags are kept to be the frictional material although it is wrapped up. Figure 4 shows the unconfined compression test on real soilbags with initial dimensions of 40cmx 40cm x lOcm and Table 1 summarizes the results of this test on soilbags comprising of various different bags and materials inside the bags. The load, F, of soilbag at failure in Table 1 can be predicted from Eq. (2) by taking into consideration 03'0 at the unconfined compression test.
Figure 2. Schematic illustration of biaxial compression test on 2D model soilbags.
Figure 3. Results of biaxial compression tests on 2D model soilbags.
Figure 4. Unconfined compression test on real soilbags to confirm their extremely high strength to resist external forces.
where L is the length of soilbag. For example, in the case that the bag is made of PE and the materials in-
588
side the bag is crushed stones, B=L=0.4m, H=O.lm, T = 12kN/m and K, = 5.55 ( @ =44"), thus F=204kN. This predicted value agrees nearly with the measured value, 230-29OkN. Some discrepancy is originated from the three-dimensional effect of real soilbags and the difference in soilbag dimensions at
Table 1. Summary of the results of unconfined compression t e m on various soilhags. Bag material
I
Materials inside soilbag
1
gad
1
I _ -
dl
fdllure (k!i')
PE
230-290
I
i
PP Crushed
1
550-650
.~
PP
PP Dry coal ash
shirash
StOneS
I
680
1
266
PP Wet coal ash
1
510
PE (UVR) Volcanic
PP Volcanic
I
__
rich
850
ach
1
l"..
5
K
Note: PE - Whitc bag made of polyethylene with tension strength of about 12kN/ni; PP - Green bag made of polypropylene with tension strength of ahour 20kN/m; PE (UVR) - White bag made ofpolyethyiene having a property of resisting ultraviolet ray. One soilbag has an initial size of40cm x 40cm x I Ocm.
failure from its initial ones. It is found from this prediction that the failure load of this soilbag is about 42 timcs larger than the tensile strength of the PE bag itself that i s equal to 4.8kN (=12kN/m x 0.4m). Thcrcfore, it is understood that the soilbag has an extremely higher strength than that originating from the tension of the bag. 3 APPLICATIONS OF SOIL,BAGS TO EARTH REINFORCEMENT 3.1 Reiiqorceunent of sofi.,foundation
u p to now, we have applied soilbags to reinforce moi-e than 30 soft grounds with the N-values as low as 1-2, where one- or two-floor buildings are constructed. Figures 5 and 6 show a typical arrangement of soilbags under the building foundation and a typical real constiwtion field, respectively. To minimize the external forces on soft grounds, the lower two layers of soilbags are connected by both scwiiig the bottoms and tying the openings of two bags. Undcr the whole raft foundation, two layers of soilbags are placed but not connectcd. One soilbag is about 40cm long by 40cm wide by lOcm high. The materials inside the soilbag inay be clxished stones, excavated soils. asphalt wastes and so on. The bag is usually made of polyethylene that can sustain for a long time if it i s protected from sunlight. The gaps between soilbags must be filled with small soilbags or soils. And, every layer of soilbags must be compacted using a vibrator to eitsure the tension of bags cffect quickly when the soilbags arc subjected to external forces from the buildings. To illustrate the effectiveness of this reinforcing method, wc estimate the bearing capacity of ground by considering that the sliding failure would not take place within the assembly of soilbags but would take place within the soft ground below the soilbags. This assumption is based on the fact that soilbags have very high strengths as shown in Table 1 and are not casy to slide between them. The load of the upper building is assumed to act on the 40cm wide footing that is then transmitted to the soft ground through the assembly of soilbags with an spreading angle of about 45". For a soft ground of c,,=I 8.6kPa. =0 and -/ =16kN/n,3, the bearillg capacity is 38,9kNim bcforc the while it is increased to be
589
Figure 5. Typical arrangcmei~t of soilbags to reinforce soft foundation with N-values as low as 1-2.
Figure 6. Typical construction field of using soilhags tu reinforce Soft f@Undationwith N-values as IOW 1-2: (id) soilbags under footing; (b) soilbags under riift foundation.
305kN/m after the reinforcement by soilbags. The effectiveness of this reinforcing method is thus illustrated. In the application of this method, we have encountered an extremely weak foundation where the ground was waterlogged and a work man settled into it as deep as 30cm, as shown in Figure 7. In this case, the arrangement of soilbags as shown in Figure 8 was implemented, i.e. 3 layers of 4-connected soilbags were first placed into the ground below water level, then 5 layers of such soilbags were further placed above the ground water. The materials inside the soilbag were crushed stones. Consequently, the reinforced ground can even withstand a heavy construction machine like backhoe. Clearly, if the crushed stones were directly put in the waterlogged ground, then they would sink into the ground.
Figure 7. Waterloggedi weak ground where soilbags are to be used.
3.2 Construction of retaining walls Two retaining walls have been constructed using soilbags: one was constructed such on a very weak foundation that it is impossible to build a reinforced concrete retaining wall. It is about 2m high, 50m long and inclined to horizontal with an angle of 80"; the other was constructed on a loam foundation with a height of about 4.5m, a total length of about 21m and an inclined angle of 75". Herein, we present the latter one in detail. Figure 9 (a) shows the arrangement of soilbags of it. In this case, 6 soilbags are connected behind the wall in the lower part and 10
Figure 8. Arrangement of soilbags for waterlogged weak foundationofFigure7.
(b) Completion of piling up soilbag viewing in vertical direction
(a) Arrangement of connected soilbags
(c) Piling up soilbags from upper view
(d) Completion of castellated retaining wall
Figure 9. Construction of a retaining wall with 30,000 soilbags.
590
Figure 11. Measured accelerations in Z-direction: (a) no reinforcement, (b) with reinforcement.
soilbags are connected in the upper part. One soilbag has dime nsions of 40cm in length by 40cm in width by lOcm in height. The materials inside the bag are construction wastes with tiles that are in free charge. The bag is made of polyethylene. The soilbags are piled up and compacted with a vibrator layer by layer with an overlapped arrangement in vertical direction, as seen in Figure 9 (b). Figure 9(c) shows the top surface of this retaining wall before lining with asphalt. It is seen from Figure 9(c) that there are some construction wastes spread over the surface of soilbags, which serve as filling the gaps among soilbags and preventing the bags to be broken during compaction. As seen in Figure 9(c), a row of soilbags on the top surface is removed to make a trench to drain ground surface water. Since the polyethylene-made bag is sensitive to sunlight, a thin layer of steel-reinforced mortar is cast on the outside surface of the wall, as shown in Figure 9(d). In this project, about 30,000 soilbags were used.
Figure 12. Relationship between ratio of acceleration spectrum and frequency in x, y and z directions measured at point P2.
4 EFFECT OF REDUCING TRAFFIC-INDUCED VIBRATION In addition to the improvement of bearing capacity, the reinforcement by soilbags has an effect of reducing traffic-induced vibration as well. Herein, we present an example of measurement of the trafficinduced vibration in a field where the soft ground has been reinforced by soilbags (Matsuoka et al. 2000b). Figure 10 illustrates the locations of two measuring points. The measuring point P2 is located on the 1st floor of a building where the foundation is reinforced with soilbags, while the measuring point PI is located on the ground outside the building that is not reinforced with soilbags. The directions along and perpendicular to the traffic road are denoted as x- and y-directions, respectively. The vertical direc591
tion is denoted as z-direction. Figure 11 presents the measured results plotted as acceleration in z direction versus time. The similar results are obtained in both x- and y-directions. It is clearly seen from Figure 11 that the accelerations induced by traffic vibration decrease greatly when the foundation is reinforced by soilbags. Furthermore, Figure 12 gives the relationship between the ratio of acceleration spectrum at point P?-to that at point PI and vibration frequency in z-, x- and y-directions. It is seen from Figure 12 that, within the vibration frequency ranging from 1 to IOHz, the acceleration spectrum at point P2 is decreased by half less than that at point PI (the ratio of them is about 0.4) in any of three directions. It is said that the vibration with a frequency ranging from 5 to 8Hz is the most sensitive to human's feeling and the resonance frequency of a wood-made living house is less than 10Hz. These facts suggest that soilbags are very suitable to the reinforcement for foundations of our houses.
5 CONCLUDING REMARKS The main results of this paper may be summarized as follows: 1) Due to the tensile force of the soilbag, an apparent cohesion, c, produces in the soilbag although the original materials inside the soilbag are frictional materials such as sands, crushed stones and so on. This can explain such a surprising test result that a single soilbag of 40cm x 40cm x lOcm, consisting of crushed stones and polyethylene bag, can sustain a vertical load of 230 - 290kN. 2) The soilbags can be used to effectively reinforce soft foundation and build retaining walls. This earth reinforcing method is cost saving, friendly to our environment. 3) When reinforcing soft foundation by using soilbags, there is also an effect of reducing trafficinduced vibration in addition to the increase of bearing capacity.
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Through this study, we believe deeply that the reinforcing method using soilbags will become a very promising method for the earth reinforcement.
6 ACKNOWLEDGEMENTS The authors would like to acknowledge the cooperation in the experimental work provided by all the related former and current students of Nagoya Institute of Technology, in particular, Mr. H. Kodama, Mr. Y. Iizuka, Mr. J. Nakamura, Mr. T. Hasebe and Mr. L. Shimao. The authors also wish to express their sincere gratitude to Prof. Y. Chen of Suzhou Institute of Urban Construction and Environmen Protection, China, who has been a visiting researcher of Nagoya Institute of Technology. REFERENCES Matsuoka, H. & Liu, S.H. (1999): Bearing capacity improvement by wrapping a part of foundation, Japan Society of Civil Engineers (JSCE), No.6 17/III-46, 235249 (in Japanese). Matsuoka, H. & Liu, S.H. (1 999): A method of bearing capacity improvement utilizing soilbags, Proc. of X I Panamerican Con5 on Soil Mechanics and Geotechnical Eng., Foz do Iguassu, Brazil, 1-8. Matsuoka, H., Liu, S.H., Kodama, H. & Oka, T. (1999): Strength properties of assembly of model soilbags and applications of soilbags in real engineering practice, Proc. of the 34'" Japan National Con5 on Geotechnical Engcneering, K-14, 898, 1797-1798 (in Japanese). Matsuoka, H., Chen, Y., Kodama, H., Yamaji, Y. & Tanaka, R. (2000a): Mechanical properties of soilbags and unconfined compression tests on model and real soilbags, Proc. oj'the 35"' Japan National Coiif: on Geotechnical Engineering, 544, 1075-1076 (in Japanese). Matsuoka, H., Yamaguchi, K., Maeda, K. & Kodama, H. (2000b): Building vibration reduction by improving foundation using soilbags, Proc. ( . t h e 35"' Japan Nutioiial Cot$ on Geotechiiicnl Engineering, 546, 10791080 (in Japanese).
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets 6: Zeitlinger, ISBN 90 2651 863 3
Behavior of reinforced foundation under uplift and push-in loadings model tests and analyses T. Nakai, T. Teranishi, M. Hinokio & K. Adachi Nagoyn Institute of Technology, Nagoya, Japan
ABSTRACT: A pile foundation with reinforced bars has been proposed by Matsuo and Ueno and is put practical use for increasing uplift bearing capacity of transmission tower and others. For investigating the mechanism of such type of reinforced foundations under not only uplift loading but also push-in loading, two- dimensional model tests and the corresponding numerical analyses were performed. Model tests and numerical simulations are done with different stiffness of reinforcements and different insertion direction of reinforcements. It is shown through the experimental and numerical study on uplift bearing capacity that though the flexible reinforcements works as the tensile reinforcements, and the stiff reinforcements as the bending reinforcements, the reinforcements protruded diagonally downward is the most effective for both cases. On the other hand, the experimental and analytical results show that the foundation with flexible reinforcements is not so effective against push-in loading, though the stiff reinforcements protruded diagonally downward work most efficiently.
1 INTRODUCTION Foundations with reinforcements protruded diagonally downward were developed and put into practice in order to increase the uplift bearing capacity of electric transmission tower and others (Matsuo and Ueno, 1989; Tokyo Electric Power Company and Dai Nippon Construction, 1990). Numerical simulation was also carried out to investigate the mechanism of reinforcement, and its results were presented at the previous symposium in 1996 (Nakai and Ueno, 1996). The numerical results shows that the reinforcements protruded diagonally downward are the most effective regardless of the stiffness of the reinforcements. In the symposium, there were active comments and questions from the floor about the most effective direction of the stiff reinforcements They were that the most effective direction of the stiff reinforcements should be diagonally upward, being different from the numerical results. After then we carried out small scale model tests as well as numerical analysis and showed experimentally that the reinforcements protruded diagonally downward are the most effective against uplift load regardless of the stiffness in the same way as the numerical analyses (Nakai et al., 1999). In the present study, we perform the model tests and the elastoplastic finite element analyses not only under uplift loading but also under push-in loading to investigate the influence of the stiffness of reinforcements and the insertion direction of reinforcements on the effect of reinforcements. In addition to 593
the behavior under push-in loading, the following two points are newly considered in the present study: one is to employ an elastoplastic constitutive model for geomaterials which takes into account the influence of the density and/or the confining pressure as well as the soil dilatancy and others more elaborately, and the other is to measure the axial forces and the bending moments of the reinforcements with the strain gages in the model tests. 2 DESCRIPTION OF MEDEL TESTS
As shown in Figure 1 , the foundation with the length of 23cm and the wide of 6cm is set up in the 2 dimensional ground. The penetration depth of the foundation is 18cm. The ground is made of a mass of aluminum rods in which two kinds of rods having diameter of 1.6 and 3.0mm are mixed in the weight 3:2. Such model ground exhibits the behavior like dense and/or medium dense sand with negative and positive dilatancy. The dots in Figure 2 show the observed stress-strain-dilatancy relations of biaxial tests on the aluminum rods mass under constant minor principal stress ( ol=19.6kPa) and constant major principal stress ( o2=19.6kPa). The reinforcements with the length of Scm are protruded to three different directions from the lower part of foundation at three levels (see Fig. 1(b)). Two kinds of thickness (3mm and 0.2mm) of aluminum plate are used as the reinforcements. Aluminum plates on which aluminum rods of 1.6mm in diameter are
glued at equal spacing of 1cm are employed. The friction angle between the reinforcement and the model ground is about 20". Upward or downward displacement is imposed continuously to the foundation. Displacement and uplift or push-in load of the foundation are measured by a displacement transducer and load cell. Axial force and bending moment of the reinforcement can be measured by the strain gages that are glued at the both sides of the reinforcement (The positions of the strain gages are lcm, 2.5cm and 4cm from the side of the foundation). The measured data are recorded in a personal computer though a data logger. The movements of ground as a whole can be known by taking photo with a digital camera. 3 METHODS OF ANALYSIS Plane strain finite element analyses under drained condition are carried out in the same scale as the model tests. Finite element meshes for the three cases are shown in Figure 3. Elastoplastic constitutive model for sand named subloading t;- model (Nakai et al., 2001) is used. This model can describe properly the following typical characteristics of sand in the same way as the previous model named t;-sand model (Nakai, 1989), regardless of small numbers of parameters: (i) Influence of intermediate principal stress on the deformation and strength of sand. (ii) Influence of stress path on the direction of plastic flow. (iii) Negative and positive dilatancy.
(b) Arrangement of reinforcements Figure 1 . Outline of model tests.
In addition to these points, the new model can take into consideration (iv) Influence of density and/or confining pressure. The values of soil parameters of the new model for the aluminum rods mass are listed in Table 1.
The "lid
Figure 2. Stress ratio and volumetric strain vs. deviatric strain in bi-axial tests on aluminurn rods mass.
Figure 3. Finite element meshes for three cases.
594
in Figure
are the
re-
Table 1. Values of soil parameters for aluminum rods mass.
h K = 98kPa) Rcs (comp.)
c ~ N C( p
E
0.008 0.004 0.3 I .8 I .2
sults corresponding to the observed ones, and the dotted curves are the calculated results in which the initial confining pressure is assumed to be two orders smaller in magnitude. This is because the initial confining pressure in model tests is much smaller than that in the bi-axial tests. We can see that the constitutive model describes strain softening behavior as well as the influence of the confining pressure. The initial state of the model ground is created by simulating the one-dimensional self-consolidation. The foundation is assumed to be an elastic material with enough stiffness. The reinforcements are simulated by beam elements: Axial stiffness and bending stiffness a;e assumed as EA=8.44 x IO'kN and EI=2.8 1 x IO'kPa for flexible reinforcements, and EA=1.27 x 104kN and EI=9.50 x 10'kPa for stiff reinforcements, respectively. In order to evaluate the friction between the foundation and the ground and between the reinforcements and the ground, an elastoplastic joint element is inserted between them (Nakai, 1985). The elastoplastic joint element can describe the slip behavior on the interface between structure and soil, which is described schematically in Figure 4. Here, (p\ and p,,j are the shear and normal stresses on the interface, and (w, and w,,) the shear and normal relative displacements on the interface. The friction angle 6 used in the analysis between the foundation and the ground is determined to be 14", and those between the reinforcements and the ground 20" from the slip tests of the foundation and the reinforcement on the aluminum rods mass. Upward or downward displacement is increasingly applied on the top of the foundation in every case.
4 RESULTS AND DISCUSSIONS Figures 5 (a)-(c) show the observed results of uplift test of the foundation with stiff reinforcements. Here, diagram (a) is the relationships between uplift load P and upward displacement d of the foundation, diagram (b) is the distributions of axial force in the reinforcement and diagram (cj the distribution of bending moment in the reinforcement. Figures 6 (a)(c) are the corresponding computed results. Further, for the cases with flexible reinforcements, respectively. The self-weight of the model foundation is ncluded in the observed uplift load in Figures 5 and 7. The dotted horizontal lines in the observed uplift load - displacement relation (diagram (a)) in these
595
Figure 4. Schematic slip behavior at inteil'ace.
figures indicate the self-weight of the model foundation. Not only computed results but also experimental results show that the foundation with reinforcements protruded diagonally downward is the most effective against uplift load in every case. In diagram (b) and (c) in Figures 7 and 8 with flexible reinforcements, though the bending moment is almost zero in every case, tensile axial force of diagonally downward reinforcements near the foundation is the largest. We can then see experimentally and numerically that the flexible reinforcements protruded diagonally downward work most effectively as tensile reinforcements. From diagrams (bj and (c) in Figures 5 and 6 with stiff reinforcements, we can see that the reinforcements work as bending reinforcements as well as tensile reinforcements. Although the mechanism of reinforcing of stiff reinforcement is different from that of flexible reinforcement, reinforcements protruded diagonally downward are the most effective against uplift load, in the same way as the flexible reinforcements. Figures 9 and 10 show the observed and computed results of push-in tests of the foundation with stiff reinforcements, respectively. Figures 1 1 and 12 are the results of push-in tests in case of flexible reinforcements. The resistance against push-in loading becomes larger when the reinforcements are protruded horizontally or diagonally downward in the
597
the stiff reinforcements protruded diagonally downward is pulled up from O.Ocm to l.Ocm. Figure 14 shows the observed and computed movements when the same foundation is pushed down from O.Ocm to 2.5cm. The observed movements are shown as photos that are taken by means of multiple exposures. The computed movements show the distributions of the magnitude of total displacements. We can see that there are good qualitative agreements between the observed and the computed movements.
5 CONCLUSIONS Experimental and numerical study on uplift and push-in bearing capacity of reinforced foundation has been done. The numerical results in which mechanical behavior of the soil and the reinforcement and frictional behavior between the soil and the reinforcement are taken into account properly describe well the experimental results. The flexible reinforcements work as tensile reinforcements, and the stiff reinforcements as tensile and bending reinforcements. The reinforcements protruded diagonally downward are the most effective against uplift loading regardless of the stiffness of reinforcements. On the other hand, the stiff reinforcements protruded horizontally or diagonally downward are effective against push-in loading, but the flexible reinforcements are not so effective.
6 ACKNOWLEDGEMENTS The assistance of K. Iwasawa of Nagoya Institute of technology in performing numerical analysis and model tests is greatly acknowledged. REFERENCES Matsuo, M. and Ueno, M. 1989. Development of ground reinforcing type foundation. Plnc. 12"' ICSMFE 2: 1205-1208. Nakai, T., 1985. Finite Element Computations for active and passive earth pressure of retaining wall. Soils rrrici Fouridations 25(3): 98-1 12. Nakai, T. 1989. An isotropic hardening elastoplastic model for sand considering the stress path dependency i n threedimensional stresses. Soi1.s arid Foirndatio17.s29( 1): 1 19-137. Nakai, T., Hinokio, M., Hoshikawa, T., Yoshida, H. and Chowdhury, E. Q. 2001. Shear behavior of sand under inonotonic and cyclic loadings and its elastoplastic niodeling. Proc. 10th IACMAC. 1 : 367-372. Nakai, T., Isobe, Y., Suzuta, and Teranishi, T. 1999. Numerical simulation and model tests of reinforced foundation under uplift loading. Proc. NUMOC 111 1 :45 1-456. Nakai, T. and Ueno, M. 1996. Numerical study on uplift bearing capacity of caisson type pile. Proc. Irit. S ~ n i or1 . Ear-tli Reiilforcernerit. 1 : 629-634. Tokyo Electric Power Company and Dai Nippon Construction 1990. Report on bearing capacity of criissoii type foundutinii Mritli reii$irciizg hcirs (4) (in Japanese).
Figure 14. Movements of ground under push-in loading.
analyses and the model tests. On the other hand, we can see from the experimental and computed results that the flexible reinforcements are not so effective against push-in loading regardless of their insertion direction, even though the tensile force acts on the reinforcements. i.e., The tensile force of reinforcements to increase bearing capacity of foundation works more effectively against uplift loading than against push-in loading. The bending moment is efficient push-in loading as well as uplift loading. Figure 13 shows the observed and computed movements of the ground, when the foundation with
598
Landmarks in Earth Reinforcement, Ochiai et al. (eds), Q 2001 Swets E: Z~itlinger,fSBN 90 2651 863 3
Characteristics of geogrid reinforced cohesive soil and its analytical method E. Ogisako ~nstituteof Technology, ~
~
i~orporation, ~ i z Tokyo, ~ Japan
K. Ryokai ~ h ~ b a t Co. o ~ Ltd., h ~ ~ h ~Japan ~ a ,
ABSTRACT: The characteristics of interaction between cohesive soil and geogrid is investigated on the basis of the pull-out test results of geogrid in soil. The pull-out resistance acting between cohesive soil and geogrid can be divided into two parts; the one is dependent on a displacement of geogrid while the other is independent of it. The analytical method introducing its characteristics of interaction between cohesive soil and geogrid into a finite element method is proposed and its validity is verified. From the comparison between the analytical values and the experimental values, it is proved that the behavior of reinforced cohesive soil by geogrid can be analysed well by means of the proposed method.
GBlO) and p o ~ y ~ e r g r i(SR80) d whose property is shown in Table 1. And two kind of soil whose property is shown in Table 2 are used in tests.
1 INTRODU~TIO~ It is important to grasp the characteristics of the pullout resistance as an interaction between soil and geogrid in the design and analysis of the geogrid reinforced soil structures. Particularly taking that a soil in the field is used as a soil reinforced by geogrid into consideration, it is supposed to be very important to estimate the pull-out resistance of geogrid in a cohesive soil. In this paper, first, the characteristics of interaction between cohesive soil and geogrid is investigated on the basis of the pull-out test results of geogrid in soil. Next, the analytical method introducing its characteristics of interaction between cohesive soil and geogrid into a finite element method is proposed and its validity is verified.
Figure 1. Apparatus of pull-out test Table 1 . Property of geogrld.
2 CHARACTERISTICS OF INTERACTION BETWEEN SOIL AND GEOGRID
Kind Mesh (mm) Tensile strength (kN/m)
2.1 Method ofpull-out test The apparatus of pull-out test is shown in Figure 1. The dimension of soil box is 600mm in length with 3OOmm in width and 160mm in height. A geogrid is laid in the center of the soil box and a constant overburden pressure is applied and then the geogrid is pulled out. The rate of pull-out is lmdmin. The overburden pressures are varied 4 to 5 steps. The displacements of geogrid nodes and the pull-out force of top of geogrid are measured during tests. The geogrids used in tests are FRP geogrid (GB5,
100x30
GBlO 100x30
49
98
GB5
SR80 166x82 83
Table 2. Property of soil.
599
Kind
Silt 1
Sift 2
Gravel Grain size (%) Sand Silt & Clay Wet density (g/cm’> Cohesion (kPa) Angle of internal friction (degree)
31 37.5 59.3 1.55 9.6 28.9
64 25.2 68.4 1.45 21.9 11.6
2.2 Relationship between pull-out resistance and displacement
2.3 Relationship between pull-out resistance and overburden pressure
The relationship between pull-out resistance stress and average displacement defined by following equations is shown in Figure 2. r=--F 2A
The ultimate value of pull-out resistance in the relationship between TI and U shown in Figure 3 is defined as Tlult. And so, the ultimate pull-out resistance q , l t is presented as Tult=To+Tlult. The relationship between overburden pressure onand pull-out resistance TO, Tlult and TuIt respectively in each case are shown in Figure 4 to Figure 7. It is proved that the relationship between overburden pressure on and pull-out resistance TO, Tlult and Tult respectively is almost linear in spite of a kind of reinforcing materials and a kind of soils. That is, a following friction law is concluded between the ultimate pull-out resistance and overburden pressure.
ui - i=l
n Where, F : Pull-out force A : Area of reinforcing material in soil Ui : displacement of nodes of geogrid n : number of nodes of geogrid in soil The pull-out resistance stress T increases to some value with almost zero displacement in initial stage. After then T increases as development of displacement but comes near a constant value. This tendency is also same in different overburden pressures. Then the pull-out resistance mobilized in the stage with zero displacement is defined as TO and the pullout resistance mobilized after then dependent on a displacement is defined as T I . The relationship between T I and U is shown in Figure 3. This relationship is closely resembled by a hyperbola shown as solid lines in Figure 3.
Where, c* is a cohesive component of pull-out resistance T ~and I ~ 6 is a frictional component of it.
Figure 4. Relationship between pull-out resistance strength and overburden pressure (GBIO : Silt 1).
Figure 2. Relationship between pull-out resistance stress and displacement (GBIO : Silt 1).
Figure 3 Relationship between pull-out resistance stress z1 and displacement (GB 1 0 Silt 1 )
Figure 5 Relationship between pull-out resistance strength and overburden pressure (GB5 Silt 1)
600
Figure 8. Modeling of interaction between soil and geogrid
tance acting between soil and geogrid increases to TO with zero displacement and after then the relationship between ~1 and displacement U can be closely resembled by a hyperbola as shown Figure 3. (1) in the case O ~ T S T O
K, =xG10'1(kN/m2/m)
(5)
(2) in the case O ~ T > T O /
\2
Where, K,, is an initial shear stiffness. And a following friction law is concluded between the pull-out resistance TO and ~ l and ~ overl ~ burden prcssure.
ro = co * +o)? tan So
3 ANALYTICAL METHOD OF GEOGRID REINFORCED SOIL
The relationship between the pull-out resistance acting on a geogrid and the displacement of geogrid can be closely resembled by a hyperbola as mentioned above. Thc inethod introducing this relationship into the analytical method by FEM is investigated. In the analysis by FEM a geogrid is represented by a plane truss element which transmits only an axial force, and joint elements are set between soil and geogrid in order to express the pull-out resistance acting between both as shown Figure 8 (Ogisako et al. 1998). The relationship between the pull-out resistance and the displacement abovementioned is introduced into the joint element. The following relationship is concluded between a shear stress T acting in joint element and a displacement U . z = K,u
(4)
The shear stiffness K,s in above equation is defined as following because the pull-out shear resis-
(7)
From a procedure above, the characteristics of pull-out shear resistance between soil and geogrid obtained from pull-out tests can be introduced into FEM analysis. 4 ANALYSES OF GEOGRID REINFORCED COHESIVE SOIL GROUND FEM analyses for the model experiments of geogrid reinforced cohesive soil ground (Ogisako & Ryokai 1998) are performed by means of the proposed analytical method above. From the comparison between the analytical values and the experimental values, a validity of the analytical method is investigated. 4.1 Outline of model experiment The apparatus of model experiment is shown in Figure 9. The dimension of soil box is 1500mm in length with 3OOmm in width and 800mm in height. A geogrid with L in length is laid in the depth with D of soil ground and a vertical load with lOOnim width is applied. The soil used in experiments is silt
60 1
Figure 10. FEM mesh of model experiment (in the case of L/B=5 and D/B=0.3).
Figure 9. Apparatus of model experiment
1 whose property is shown in Table 2. And the used geogrid is FRP geogrid whose tensile strength is 12 kN/m and tensile stiffness is 392 kN/m. A settlement of load plate, a vertical displacement at a surface of ground and a strain of reinforcing material are measured during experiment.
Wherc,
4.2 Condition of analysis
The input parameters are determined from triaxial compression tests (UU tests) results as shown in Table 3 . And the parameters for the interaction between soil and geogrid are determined from pull-out tests results as shown in Table 4.
(1) Analytical modcls and cases The analyses are performed for the cases that a length of geogrid stays in L/B=5 and a laying depth varies D/B=0.3 to 1.0. And both the cases used the analytical method considered the interaction between soil and geogrid abovementioned (hereafter called the proposed method) and the cases used a method in which the interaction is not considered that is, soil is represented by a solid element and a geogrid is represented by a plane truss element only (hcreafter called an ordinary method) are performed. An example of the analytical models is shown in Figure 10. In analyses two dimensional plane strain condition is assumed and as the boundary condition a horizontal direction is fixed and a vertical direction is free in sides, both direction is fixed in bottom.
: Major principal stress : Minor principal stress c : Cohesion 4 : Angle of internal friction P, : Atmospheric pressure Rf : Failure ratio K, n : Constants determined experimentally.
01
(53
4.3 Analytical results
( I ) Strain of geogrid The relationship between strain of geogrid and settlement of load plate is shown in Figure 11 and Figure 12. In these figures the strains at the position of a center, a middle (10 cm apart from a center) and a edge (20 cm apart from a center) of geogrid are Table 3. Input parameters of soil. K
(2) Input parameters As a stress-strain relationship Duncan-Chang’s nonlinear elastic model (Duncan & Chang 1970) is used. In this model an elastic modulus is indicated as a following equation.
560
n 044
R
t
090
c (kPa) 96
Poisson’s ratio 0.333
d,
(degree) 28.9
Density (kN/m3) 15.2
Table 4 Input parameters of interaction between soil and geogrid Initial shear stiffness (kN/m’/m) 679 1
602
6
Vertical stiffness (kN/m’/m)
CO* (kPa)
(degree)
10”
0.0
22.0
cI* 61 (kPa) (degree)
1.0
24.8
The distributions of the strain of geogrid in each settlement of load plate respectively are shown in Figure 13 and Figure 14. The analytical values by the proposed method in shown Figure 13 is larger at the center and is smaller at the edge, so its tendency is same as the experiments. The analytical values also agree well with the experimental values quantitatively. On the other hand, the analytical values by an ordinary method are overestimated at the center and underestimated at the edge. The tendency abovementioned is also same in other cases. Accordingly it is supposed that the proposed analytical method is able to estimate the strain of geogrid in the experiment both qualitatively and quantitatively than an ordinary method.
shown respectively. Figure 11 indicates the comparison between the analytical values by the proposed method and the experimental values, while Figure 12 indicates the comparison between the analytical values by an ordinary method and the experimental values. The analytical values by the proposed method increase as a settlement increases. Its tendency is same as the experiments and the analytical values agree well with the experimental values quantitatively. On the other hand, the analytical values by an ordinary inethod differ from the experimental values that is, the strain at the edge decreases as a settlement increases and the strain at the center is overestimated and the strain at the edge is underestimated.
(2) Deformation The comparison between the analytical values by the proposed method and the experimental values in a distribution of a displacement at a ground surface is shown in Figure 15 and Figure 16. The ground settles near by the load plate and heaves up apart from
Figure 11. Relationship between strain of geogrid and settleinent of load plate (comparison between analytical values by proposed method and experimental values in the case of L/B=5 and D/B=0.5). Figure 13 Distribution of strain of geogrid (comparison between analytical values by proposed method and experimental values in the case of L/B=5 and D/B=O 5 )
Figure 12. Relationship between strain of geogrid and settlement of load plate (comparison between analytical values by ordinary method and experimental values in the case o ~ L / B = ~ and D/B=0.5).
603
Figure 14. Distribution of strain of geogrid (comparison between analytical values by ordinary method and experimental values in the case of L/B=5 and D/B=0.5).
5 CONCLUSIONS
Figure 15 Distribution of displacement at a ground surface (comparison betwcen analytical values by proposed method and experimental values in the case of L/B=5 and D/B=O 5)
Figure 16 Distribution of displacement at a ground surface (comparison between analytical values by proposed method and experimental values in the case of L/B=5 and D/B=O 8)
the load plate as a settlement of the load plate increases. Particularly in the case that a laying depth of geogrid is deep, a magnitude of heaving becomes large at the side of the load plate as a settlement increases. Such tendency agrees well with the experiment. Further the analytical values also agree well with experimental values quantitatively.
604
The interaction characteristics between cohesive soil and geogrid is investigated on the basis of the geogrid pull-out test results. Further the analytical method introducing its interaction characteristics into finite element method is proposed and its validity is verified. The following is concluded. 1 . The pull-out resistance acting between cohesive soil and geogrid can be divided into two parts; the one is dependent on a displacement of geogrid while the other is independent of it. And the relationship between the former and the displacement can be closely resembled by a hyperbola. Further the linear relationship is concluded between the ultimate strength Of pull-out resistance and the overburden pressure. 2. As the method that can be taken the interaction between cohesive soil and geogrid into consideration, the analytical method introducing the characteristics of pull-out resistance obtained from the pullout tests into finite element method is proposed and is applied to the model experiments of the geogrid reinforced cohesive soil ground. From the comparison between the analytical results by the proposed method and those by an ordinary method, it is proved that the analytical values of the geogrid strain and the displacement at a ground surface by the proposed method agree with the experimental values better than an ordinary method. Accordingly the validity of the proposed analytical method is verified. REFERENCES Duncan, J M & C Y Chang 1970 Nonlinear analysis of stress and strain in soils Jozrrnal of Soil Mechanics and Fozmdutions Division,ASCE, SM5 1629-1653 Ogisako, E , K Ryokai & Y Sakai 1998 Analysis of geogrid reinforced embankment of cohesive soil Proceedings of the 531-d annual con$erence of the Japan society of civil engineers, 3 720-72 1 (in Japanese) Ogisako, E & K Ryokai 1998 Bearing capacity ofgeogrid reinforced ground of cohesive soil Proceedings ofrhe 33rd Japan national conference on geotechnrcal engineering 2383-2384 (in Japanese)
Landmarks in Earth Reinforcement, Ochiai et a/. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Field loading test on the footing reinforced with prestressed micropiles Yoshinori Otani Research Engineer, Hirose, Co. Ltd., Osaka, Japan
Kinya Miura Associate Professor, Asian Institute of Technology, Bangkok, Thailand
Mizuho Ishito Graduate student, Hokkaido University, Sapporo, Japan
Guan-Lin You Graduate student, Asian Institute of Technology, Bangkok, Thailand
Yukihiro Tsukada Chief Manager, Tohoku Construction Branch, Ministvy of Construction ABSTRACT: To reveal the mechanism of bearing capacity of the footing reinforced with prestressed micropiles, large-scaled field tests were conducted with the experience in the previous series of laboratory model study. A series of loading tests were carried out on uniform medium stiff loam ground on three types of square surface footings ( 1 m x 1m); surface footing, surface footing with eight non-prestressed micropiles of 3m in length, and surface footing with eight prestressed micropiles. In the preliminary loading tests, the bearing capacity and pullout resistance of micropile, and its applicability and durability of the prestress were examined on single micropile. On the surface footings reinforced with micropiles, both horizontal and vertical loading tests were conducted. The performance of footing was improved with the reinforcement of micropiles, and prestress was notably effective on the stabilization of footing. The network effect was recognized in both non-prestressed and prestressed micropile foundation. The prestress raised the base pressure at early stage, and the load bearing capacity in both horizontal and vertical directions. is mobilized positively with the appropriate confinement of the ground material beneath the footing. The micropiles not only provide load bearing capacity directory through their skin friction, but also raise the base pressure on the footing with the confinement by the interaction between the footing and a group of micropiles. Also it was demonstrated that the bearing capacity is improved more efficiently with the prestress, which induces the confinement on the ground material beneath the footing at early stage of the loading process. In this study, the large-scaled field tests on the footings reinforced with micropiles were conducted on natural uniform loam ground. The purpose is to examine the findings obtained in the series of laboratory model loading tests, and to clarify the mechanism of network effect and the effect of prestress in micropiling.
1 INTRODUCTION Micropiles, which was pioneered by Lizzi ( 1971, 1978) in Italy, are now widely used both as structural supports in foundations as well as for in-situ earth reinforcement. Micropiles are considered as promising foundation elements in improving the bearing capacity of existing foundations which are deteriorating for one reason or another with minimum disturbance to structures, subsoil and the environment. Triggered by the destructive HyogokenNambu earthquake in 1995 in Japan, the research and development on the use of micropiles in strengthening foundations have been focused (Tsukada. 1997). To clarify the mechanism of the bearing capacity and develop new rational method to improve the performance of micropile foundations. a study on model micropile has been conducted by authors continuously. In previous study, the method of model loading tests has been reported by Tsukada et al. ( 1 999). the test results on three series of model tests (footing test, micropile test, and micropile foundation test) have been reported by Miura et al. (2000), and the effect of prestress on the improvement of bearing capacity has also been investigated in model micropile foundation study by Miura et a1 (2001). From these studies, it is found that the network effect of a group of micropiles on the bearing capacity
2 IN-SITU SOIL PROPERIES The large-scale field tests were conducted inside the factory of Hirose Co. Ltd, N339, Jyoba, Shirai-chyo, Inba-gun Chiba, Japan; the plan layout of the investigation is shown in Figure 1 . The soil profile and the SPT N-value of the upper 10 meters are shown in Figure 2. The subsoils consist of fill, loam, cemented clay, sandy clay, and fine sand, in order 605
from ground surface to the 10 m depth. The fill, loam and clay are soft; the N values obtained are less than 5. The soil properties are tabulated in Table 1.
3 TEST METHOD 3.1 Test apparatus The micropile (MP in short) used for Single MP (SMP) test and Group MP test is same in the specification; the diameter is 100 mm, length 3.0 m and reinforced by 0 3 2 steel bar, as shown in Figure 3 . The MPs were instrumented with strain gauges arranged for the bending in two directions at three sections of -0.2"', -1 .4"', and -2.4"'. The footing (FT in short) used is made of steel and 1000 x 1000 mm square in plane shape (Fig. 3). The loading apparatus was set up according to the requirements of each of loading tests. Two I600x190 beams were used as reaction beams in the loading and pullout tests of S-MP. In the prestress test of single MP (FT-S-PSMP), a 50t jack was placed on the footing to apply prestress on the MP. Shown in Figure 4 is the loading apparatus of S-MP test. The loading apparatus for horizontal test of MP groups is shown in Figure 5. After prestressing the MPs group in the FT-PSMP (footing plus eight prestressed MPs), H-300x300 was used as the load transfer bar. A 50t jack was used to apply horizontal load; a load cell of 200kN in capacity was used to monitor the load.
Figure 2. Geotechnical boring log of the test site Table 1 . Soil properties of the investigation site
I Loam
Soils Depth (m) Wet density
(g/cm') Dry density Pcl (g/cm') Water content
w,, (%)
1 Cemented 1 FineSand
2.0-2.8m
Clay 3.7-4.5m
8.0-8.8m
I .63 1
1.721
1.905
1,ooj
1,144
1.442
62.7
50.6
32.2
Total stress (b ("\ I 17 6 1
\.'
,
I437
I 0.04 I
I
I
I 33.7
I 37.0
I
I
I
Effecrtive . stress
I 28.4
B value
I 0.92-0.96 I 0.94-0.96 I 0.92-0.96
8 (9 r \ I
Figure 3. Micropile, footing and their connection
606
kN and settlement S,,, = 18 1.43 mm. Yielding load is identified as PJ)= 90.0 kN, and the ultimate load was estimated as PI,= 313.8 kN by Weibull curve. Shown in Figure 7 is the comprehensive test result in the S-MP loading test with maximum applied load P,,, = 30 kN and settlement S,,, = 28.69 mm. Yielding load is identified as PJ,= 16.0 kN, and the ultimate load is estimated as PI, = 30.3. And shown in Figure 8 is the axial force distribution. At section 1, friction cut technology was applied for the need of tensile prestress (refer to Fig. 3).
Figure 6. Comprehensive test results in FT test
Figure 5. Loading apparatus of horizontal test
3.2 Measuring system Load Measuring: The variation of load was controlled by a hydraulic jack; Monitoring the load was controlled by the highly accurate load cell. Displacement Measuring: Displacement was measured by a 100 mm DT-IOOE displacement gauge, whose minimum reading is 1/100 mm. The DT-IOOE displacement gauge was connected to an ASW-5OC automatic exchanger, and recorded by TDS-60 1 multi-functional digital meter. Strain Measuring: To measure the stain of MP, the end of the strain gauge was connected to the ASW-5OC automatic exchanger, and recorded by TDS-60 1 multi-functional digital meter. The average value of the two strain gauge reading at the same section was used under each of loads. Time Recording: Time and the elapsed time were recorded from TDS-601 to a personal computer through the entire tests.
Figure 7. Comprehensive test results in S-MP loading test
4 TEST RESULTS 4.1 Preliminary loading tests FT Test, S-MP loading and pullout tests were conducted in a preliminary loading test programs. Shown in Figure 6 is the comprehensive test result in the FT test with maximum applied load P,,, = 2 10.0
Figure 8 Axial force distribution along MP in S-MP loading test
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4.3
MP group prestress test
In the FT- PSMP group, there are two phase prestress, as in the FT-S-PSMP prestress test. In the 1'' prestress test, the maximum prestress load was T,,, = 7.5 kN, and T,,,, = 9.6 kN in the 2"d prestress test; the loads were applied to all the eight micropile begun with pile 1 and 5 simultaneously. The straintime relationship at section 1 is shown in Figure 11. The top figure shows the MPs when they were subjected to the applied prestress. 4.4 Horizontal loading test on the footings reinforced with MP group After the FT-PSMP prestress test, the horizontal loading test between the FT- PSMP group and FTMP group (footing plus eight non-prestressed MPs) was conducted. The loading procedure was stopped when the horizontal movement of the FT- MP group reached 27.45 mm to minimized the damage to the footings. Shown in Figure 12 is the load-movement variation for FT-PSMP and FT-MP group, respectively.
Figure 9. Comprehensive test results in S-MP pullout test
Shown in Figure 9 is the comprehensive test result in the S-MP pullout test with maximum applied load T,,,, = 14.3 kN and settlement S ,, = 19.52 mm. Yielding load is identified as T,, = 10.0 kN, and the ultimate load is estimated as T,= 14.3 kN. 4.2 Single MP prestress test (FT-S-PSMP Prestress Test) FT-S-PSMP test includes two-phase prestress. In the 1" prestress test, the maximum prestress load was l;,7ux = 7.5 kN. When T,,,, = 7.5 kN, E/ = 49 p; after the load was released, €1 = 22 p ( T = 3.4 kN). The maximum displacement of footing was 0.48 mm, and the extension of the MP was 0.67 mm. In the FT-S-PSMP 2"d prestress test, the maximum prestress load was T,,,Or= 9.6 kN, with €1 = 57 p; after releasing the load, E/ = 25 p ( T = 4.2 kN). Then, the pullout test was conducted; the pullout load was increased by 2 kN a step, up to the maxiinum load T,,,, = 30 kN, the maximum displacement is S,,,, = 27.06 mm. The variation of load and displacement during the full procedure of FT-S-PSMP prestress test is shown in Figure 10.
0
10
20
30
40
Figure 1 1 . Strain variation at section 1 during the entire prestress in FT-PSMP 2"" prestress test Horizontal load (kN) 0
5
10
15
20
25
10
50
Elapsed Time t (IIOUT) Figure 12. Movement of FT-PSMP and FT-MP groups under horizontal loading
Figure 10. Full procedure of FT-S-PSMP prestress test
608
Figure 13. Load-settlement relations of FT, FT+8MPs, FT-MP, and FT-PSMP
Figure 14. Comparison ofthe pullout load with the skin resistance in S-MP loading test
4.5 Vertical loading tests the footings reinforced with MP group
stage when the settlement is small; the curve of FTMP group is even lower than that of FT+8MPs, which may be due to loose contact between FT and the ground surface at the early stage. But with the increase of the load, the load-settlement curve of FT+8MPs becomes lower, since the confinement of the FT and MPs in the MP group was induced with the increased stress in the MPs and subsoils. In this figure, the curve of the FT-PSMP group is always higher that of the FT-MP group, which strongly demonstrates the effect of prestress is positive on the improvement of veering capacity and reducing settlement.
After the horizontal loading test, the vertical loading tests were conducted. The first applied load was 60 kN, the final applied load was up to 600 kN and 560 kN with the maximum settlement of 227.235 mm and 243.25 mm for FT-PSMP and FT-MP group, respectively (Fig. 13). Yielding load is judged as Py = 240 kN, the ultimate load is estimated as P,, = 600 kN . 5 ANALYSIS AND DISCUSSION
5.1.3 Effect of prestress on bearing capacity Effect of prestress on single MP: The maximum pullout load in S-MP pullout test was 14 kN, it was 30 kN as low as in FT-S-PSMP test. So the skin resistance of single PSMP can be more than twice the skin resistance of ordinary single MP. The difference in skin resistance between S-MP and FT-SPSMP indicates the significant effect of prestress on increasing the confining stress of the subsoil around the MP and on increasing the frictional force between MP and the subsoils. After prestress, the stress in MP remains in tension side, which induced the upward stress onto the soil; however the base plate subsides, which applied downward stress onto the subsoil. So the subsoil beneath the FT was in confinement, and the stress hence increased. Meanwhile, due to the densification of the soil. the frictional angle may have increased. Therefore, the frictional resistance of MP was notably increased as observed in the test on FT-S-MP. Effect of prestress on MP group: In the vertical loading tests, the maximum applied load in the FTPSMP group is about 7% higher than that in the FTMP test (Fig. 13), and both the load carried by FT and the load by MPs are generally higher in the FTPSMP due to the increased confining pressure in the FT-PSMP group. But the improvement of bearing capacity by the prestress in MP group tests is lower
5.1 Mechanism of inicropile in bearing load 5.1.1 Bearing mechanism of single MP From the Loading test on S-MP, the maximum applied load is 30 kN, among which point bearing capacity and the skin resistance are estimated 16.61 kN and 13.39 kN, respectively. From the S-MP pullout test, the ultimate bearing capacity is 14.3 kN, this result is well agreed with the skin resistance of 13.39 kN in the loading test, which testifies the accuracy and reliability of the strain gauges in the in-situ tests. Figure 14 shows the comparison of the pullout load in S-MP pullout test with the skin resistance in SM P loading test. That the skin resistance is slightly higher in the S-MP pullout test than in the S-MP loading test may be due to the increased fitness between the MP and subsoils after the loading test. 5.1.2 Bearing mechanism of M P group The maximum applied load of a single MP is 30 kN as stated in previous section, and the summation of the load carried by eight MPs can be estimated as 8 x 30 = 240 kN. To investigate the MP group effect, the load carried by FT and the load taken by eight MPs are summated, 210 + 240 = 450 kN. The Ioadsettlement relationships of FT+8MPs, FT-MP and FT-PSMP are shown in Figure 13. From this figure, M P group effect is not clear at the early loading 609
6 CONCLUSIONS
than that in single MP tests, since there was a better confining effect in the center MP in the FT-S-PSMP test than in the peripheral MPs in the FT-PSMP test. The other reason lies in that the soft subsoils tend to move laterally under relatively large load so as to reduce the confining effect. 5.2 Mechanism of micropile in movement control 5.2.1 Micropile in vertical settlement control One of the major features of MP is its excellent performance in settlement control. As shown in Figure 13, under the maximum applied load of 210 kN in FT test, the settlement of 250 mm in FT test is reduced to only 12 mm and 4 mm in FT-MP test and FT-PSMP test, respectively, under the same load. 5.2.2 Eflect ofprestress on movement control Vertical Settlement: From Figure 13, the maximum settlement in FT-MP test is 243.25 mm with a maximum load of 560 kN, while it is only 227.23 mm in FT-PSMP test with a maximum load of 600 kN. In other word, the coefficient of subgrade reaction was 1.86x104kN/m3and 3.97x104kN/m’ in the tests on FT-PSMP and FT-MP, respectively. This comparison means the settlement was suppressed at early stage offloading and become almost half due to the effect of prestress on MPs in the FT-PSMP group. Horizontal Movement: The effect of prestress on horizontal movement control is remarkable compared with on vertical loading. The movement in the FT-MP group seems linearly increased with load as shown in Figure 12. The coefficient of subgrade reaction was 1.01x103kN/m3 and 17.1x103kN/m3 in the tests on FT-PSMP and FT-MP, respectively. This clearly showed the remarkable effect of prestress also in horizontal loading. As shown in Figure 12, the failure pattern of FTMP under horizontal load looks like punch-in, while the failure pattern of FT-PSMP is local failure. The difference in failure mode demonstrates that the subsoil composite was increased from soft soil of low strength to medium soil of medium strength due to the application of prestress on MPs.
610
The bearing capacity of the footing is greatly increased when it was reinforced with a group of micropiles, no matter prestressed or non-prestressed. The effect of prestress on the improvement of bearing capacity was significant; the pullout resistance of S-PSMP was more than twice that of S-MP, and 7% vertical bearing capacity is increased in prestressed MP group (FT-PSMP) than in the nonprestressed MP group (FT-MP). The effect of prestress in movement control was also remarkable in micropile group. The horizontal movement of prestressed micropile group (FTPSMP) was less than tenth those of the non-prestress micropile group (FT-MP), compared in the coefficient of subgrade reaction. In vertical loading, the coefficient of subgrade reaction was twice in the prestressed micropile (FT-PSMP) compared with the non-prestress micropile group (FT-MP). REFERENCES Lizzi, F 1971 Special Patented Systems of Underpinning and more Generally, Subsoil Strengthening by Means Of Pall Radice (Root Piles) with Special Reference to Problems Arising from the Construction of Subways in Built-up Area, Special Lecture given at itniversity of Illinois at UrbanaChanzpaign, etc Lizzi, F 1978 “Reticulated Root Piles to Correct Land Slides,” Proceedings of ASCE Conference, Chicago, Illinois, October 16-20 1-25 Miura K , Tsukada Y , You G L , Ishito M , Otani Y & Tsubokawa Y 2000 Model Investigation On The Bearing Mechanism Of Footing Regarding The Interaction Between The Footing And A Group Of Micropiles, Proceedings of the 3Id international conference on ground improvement techniqites, September 2000, Singapore 255-262 Miura K , Tsukada Y , Otani Y , lshito M & You G L 2001 Model loading tests on the footing reinforced with prestressed micropiles, International Symposiiini of Earth Reinforcement, Fukuoka, Japan Tsukada, Y 1997 State-of-the-Art Application of Micropiles in Japan, Proceedings of 1st International Workshop on Micropiles, Deep Foundations Institute, 265-279 Tsukada, Y, Miura, K & Tsubokawa, Y 1999 Model Loading Tests on Micropile Foundation on Sand Ground, Japanese Geotechnical Society, Tsuchi-to-kiso, 47( I), 35-38
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, lSBN 90 2651 863 3
Jet grouting application for quay restoration in Tunisia A. Sfar Office de Ia Marine Marchande et des Ports, 2060 La Goulette, Tunisia
M. Bouassida Ecole Nationale d’hgknieurs de Tunis, ENIT BP 3 7 Le Belvkdire 1002, Tunis, Tunisia
ABSTRACT: Few years ago, the quay of Sfax harbor was made-up because of the increase of the traffic ships. After this amenity, unstopped settlements occurred in the revetment layer of the roadbed. From the soil investigation conducted near to the quay, the absence of a filter, to prevent the migration of fine particles from the material fill, was observed. Due to the presence of cavities in the soil, the measured permeability was roughly 1 cm/s. To remedy at this situation, the stabilization of the revetment wall, using the jet-grouting technique, was decided. A micro-concrete and a cement grout were jetted in boreholes, with 105 mm diameter, at depths ranging from 1 I m to 13 m. A remarkable decrease of the permeability was observed after water tests carried in the stabilized quay. 1 INTRODUCTION
2 DESCRIPTION OF THE TRADE QUAY
The trade harbor of Sfax is located at the Tunisian East coast, at 34’43’ north latitude and 10’46’ East longitude, where the seafloors have weak slopes. The basin of the harbor covers a total area of 65 hectares with 2600 meters of quay. The length of specialized quays is 1460 meters. Roadbeds of this harbor have a surface of 24 hectares, and include 22,700 m’ of covered surfaces. Sfax, as second town of Tunisia, offers the harbor ensuring the most important traffic (by 30% of the general import/export traffic) composed of 63% of solid jumble, 23% of liquid jumble and 14% of various wares. The harbor was opened to the trade in 1891. To follow the evolution of the traffic, this harbor knew several extensions and amenities among which, particularly, the elongation of the trade northwest quay and the extension of the roadbeds behind. The principal basin of the quay is limited by two quay walls. The first one, with total length of 580 m, was built by prefabricated caissons filled by reinforced concrete whose foundation rests at depth 1Im hydro, and leveled at 2.5 m hydro. In 1950, the construction of the quay was done 20 m in front of the old one with a foundation at -6 m hydro. A protection of the angle of the quay, of 40 cm in width, was done by masonry and cutting stones. Surges that reach the entrance of the basin come &om the southeasterly sector. The intensity of currents tide reaches 50 c d s at marine depth of 1.5 to 4 m.
After the recent extending of the trade quay of Sfax harbor, unstopped settlements occurred in the revetment layer of the roadbed. These settlements were pronounced and localized in the revetment layer. Despite several reparations, continuous settlements took place progressively until 15 cin on average. From several borings carried out in the roadbed, it was confirmed the absence of a filter between the revetment wall of the quay and the material fill with an important percentage of fine particles. The rule of the filter, to protect the loose of fine particles from the fill, was not ensured. This process, that would be amplified by the tide phenomena, created a new arrangement of gravel and blocks from which settlements occurred. From the diagnostic of the quay, a deposit of fine particles occurred on the joints. That confirms the assumed loose of fine particles due to the absence of a filter behind the quay. The observed degradations in the coronation beam come from two different facts. The first one is relevant to the constitutive material of the beam. The second fact is related to the marine water effect on mortar joints. As executed in masonry quarry stones, the beam can not resist to the transmitted solicitations. As a matter of fact, if submitted to a direct or a frontal shock of ships the deterioration of cemented mortar joints can not be avoided in time. This phenomena will be aggravated because of the absence of defenses.
61 1
In addition, the mortar cement joints being thin, their desegregation becomes easy under the aggressive action of the marine water. Even if under moderate shocks, the two mentioned facts contributed largely to the progressive deterioration of the coronation beam.
3 ANALYSIS OF THE DISORDERS In order to clarify the origin of disorders, an investigation program was adopted. The latter included a topographic survey on a band width of 25 m behind the quay, a bathymetric survey on a band width of 20 m in front of the quay, and four boreholes executed until 15 m depth. Also, an investigation by divers was done with a video film production. The subsidence area, where localized settlements occurred, is located at 20 in in width from the quay border. Because of that, the origin of the observed settlements could not be attributed to the compressibility of the roadbed material mainly constituted by quarry wastes. Therefore, settlements are rather due to insufficient stability conditions of the quay wall as hole, or to the loose of fine particles through voids existing between stones of the revetment wall. From the soil investigation carried behind the quay, it appeared a material fill constituted by gravel and calcareous blocks. The study of stability of the recent quay wall was carried regarding sliding on its base and overturning. As shown in figure 1, stability conditions require a fi-iction angle of the material fill equal to 37". This condition is normally satisfied from the description of the filled material behind the quay whose friction angle is by 42". It is then concluded that the stability of the quay wall was not compromised, Studi (1 996).
Figure 1. Computed safety factors from the stability study of the quay wall.
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4 REMEDIATION From the diagnostic and the conducted surveys above described, a solution of the quay restoration, including roadbeds and the beam coronation, should be advanced. The solution of a new quay construction is inadequate, and it should not be analyzed because settlements will not stop in the roadbeds of the existent quay whose stability was verified. By the construction of a new quay the port exploitation will be more complicated. To remedy at this situation, with minimized disturbance of the harbor activity, repairing was the preferred solution. For the latter, two alternatives were envisaged. The first one needs the insertion of a filter, with selected materials, between the revetment wall and the roadbed. The second alternative is the consolidation of the roadbed mass using the jet grouting technique. 4.1 Thefirst alternative: insertion of a.filter A geotextile film or blocks of rock (of 0 to100 kg) is the filter material to insert between the revetment wall (behind the quay) and the fill roadbed. This alternative needs to evacuate the existing fill, to execute the filter, and to replace both the excavated fill and the existing circuits. The total cost of this alternative is by 2800 thousands Tunisian dinars. 4.2 The second alternative: grouting in the revetrnent wall By grouting a cementing agent in the roadbed behind the quay, voids will be significantly reduced. Assuming 40% of porosity in the rock blocks, the volume of grouted materials to inject is by 30 cubic meters per linear meter of the quay. Therefore, a monolithic mass will be obtained. Depending on the coefficient of permeability of the soil roadbed, the use of a cement mortar will contribute to a less restoration cost. For this reason, and to make a better estimation of this operation, the execution both of permeability and grouting (of a cement mortar) tests was recommended. The total cost of this alternative is by 2900 thousands Tunisian dinars. A comparison between the two restoration alternatives is presented in table1 . The costs of the two restoration alternatives are similar. The main technical advantage of the first alternative is a new construction of the quay. But, it needs a complete stop of the harbor exploitation (quay and roadbeds) in the area to restore. While, by the second alternative, the consolidation of roadbeds is carried without having recourse to opened excavations.
Table I . Comparison between the two restoration alternatives. Criteria First alternative Cost (per linear 2800 thousands meter of quay) Tunisian dinars * The execution is relatively easy for local contractors * a complete substitution : better perAdvantages formances * The port exploitation is significantly compromised * Demolition and Disadvantages replacement of the existent circuits : longer time of executi on.
Second alternative 2900 thousands Tunisian dinars * Without disturbance of the port exploitation * Without risk for stability of the existing foundations * Difficult estimation of quantities * Risk in loose of the grouted materials * Adequate execution by few contractors * Permanent control of the execution
Figure 3. Adopted methodology for jet grouting treatment.
Because of the advantage in time execution, the second alternative was finally adopted.
at 12 m depth. The grouted material was a cement of bentonite. 5.3 Equipments and description of the process
5 JET GROUTING APPLICATION The jet grouting treatment, planned in two phases, should be executed in vertical boreholes, with 105 mm diameter, along the elevation of the revetment wall of the quay. The jet grouting parameters, per meter in depth, were to reach a maximum volume of six cubic meters of the grouted material, or a maximum grout pressure of 0.2 MPa that should be maintained during one minute. The grouted material is a mixing of micro-concrete and mortar. As illustrated in figure 2 (dimensions are given in meter) drilled boreholes, in regular square mesh 4mx4m, will be executed respectively until 11 m depth for the external line and 13 m depth for the internal line. 5. I
Thefirst phase of soil treatment
It was conducted successively along the internal and the external lines. Primary boreholes were executing by jetting a micro-concrete grout. Then, secondary boreholes were executed between the primary ones by jetting a cement grout. As illustrated in figure 3, primary and secondary boreholes are distanced 8 in. At the end of the first phase, Lefranc water tests, Cassan 1993, and laboratory permeability tests were performed to check the efficiency of the jet grouting treatment.
From the roadbed surface to the predicted depth, boreholes were drilled by a device SR6, using a temporary casing (wire line) with 105 mm in diameter. After, the driller device was removed, and the jet grouting string was installed within the temporary casing. From the borehole base, the temporary casing was remote on 1 m depth along which the jetting was executed at lower pressure controlled by a manometer. The jet grouting strings, made in PVC material, had a nominal diameter larger than 26 mm. The grouting device referenced “Moyno 2000-Miro 1” pump materials such as mortar, micro-concrete, and grouts. Such pumps are equipped with devices controlling the injected pressure and the debit (or volume) of the grouted material. The jet grouting criteria (maximum volume or maximum pressure of grout above mentioned) were adopted both during the execution of primary and secondary boreholes. 5.4 Materials For the grout mixes, potable water was used fi-om the distribution circuit of Sfax town. For microconcrete the used sand comes fi-oin a quarry with parameters: calcareous content is 30%, sand equivalent is greater than 80% and fineness modulus ranges froin 2 to 3. The cement comes fi-om the “Artificial Tunisian Cements” factory whose characteristics comply with the French AFNOR standard. Based on the observed decrease in volume of the grouted materials, from the internal line to the external one, an important volume of voids was filled.
5.2 The secondphase of soil treatment In the second phase, the grouting was achieved along the intermediate line located between those executed in the first phase, through drilled boreholes
613
Figure 2. Illustration of the jet grouting alternative.
614
Finally, the recourse to the jet grouting technique was a successfd operation to restore the quay of Sfax harbor without stopping the trade activity.
6 CONCLUSIONS From the quay restoration experienced in the trade quay of Sfax harbor, two main remarks should be hold. 1) Based on recorded results, a remarkable decrease of the injected volume of micro-concrete or cement grout was obtained following the treatment sequences. 2) From laboratory and in-situ permeability tests, conducted in the treated soil, the measured coefficient of permeability was by 10" cm/s. That was enough lower regarding the required value by the technical specifications for the quay restoration.
REFERENCES M. Cassan 1993. Aide-MCmoire d'hydraulique souterraine. 2"lle edition, Presses des Ponts & Chausses. Paris. W. Sondermann, P.S. Toth 2000. State ofthe art of jet grouting shown on different applications. 4"'Int . Conf GIGS. Helsinki, June 7-9. 181-1 94. Finnish Geotechnical Society. Studi 1996. Etude de remise en etat des quais de commerce et de phosphate du port de commerce de Sfax. Rapport (( Diagnostic & Solutions B. Tunisic.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Use of technologies based on soil reinforcement when trunk pipeline repair Yu.1. Spector, N.M. Rakhmatullin, V.V. Novoselov & N.F. Shchepin Ufa State Petroleum Technological University, Russia
ABSTRACT: Mehods of bottom bed sections protection in places of river crossing with underwater pipelines and other buried communications are given. The methods described are based on structures use which allow both to stop the erosion process and promote the restoration of the initial bottom topography. It is well known that the process of change of river bed relief is continuous. In this connection, an emergency situation occurs as a result of soil washing and evacuation at some sections of underwater pipelines under operation, which can provoke pipeline failure and ecologic disaster. That’s why timely detection of sections where processes of bottom erosion are progressing in a zone of piping and the following restoration of bed topography is extremely actual.
1 USE OF ANTIEROSION COVER (AEC) IN A RIVER BED One of the methods of bed washout protection as well as restoration of the washed out soil is a method based on use of antierosion rubber-cloth cover. The above cover was successfully used in bank slopes strengthening. AEC is a rhomblike ribbed construction (Fig. 1) with a rib height 10-12 cm. Rhomb diagonals are chosen in dependence on water stream velocity. Material for AEC manufacturing is a strip of conveyor belt with almost indifferent floatability. Thanks to increase of hydraulic friction in place of protective cover laying, loss of water flow energy
occurs which in its turn causes a process of deposition the alluviations, being before in motion, in cells of AEC. Protective cover (Fig. 2) consists of a set of closely laid separate elements (1) arranged along a pipeline axis (2) in such a way that they overlap the whole section (3) being subjected to washout. To fix the cover in a bed in a given position, the construction is equipped with anchor beams (4) which are not fixed stiffly to a bottom soil and can vertically deepen as a result of possible underminning.
/ Figure 1. AEC construction.
Figure 2. Protective cover.
617
AEC construction shown in Figure 2 which provides protection of a bottom section subjected to bed erosion, has three technological ways of functioning: 1. Preliminary filling of a washed out bottom section up to a full restoration of the initial bed topography with the following laying of antierosion construction onto the restored section. This way is used in cases of significant soil washing with water flow and demands obligatory underwater-technical jobs. 2. Laying of antierosion construction onto the washed out bottom section overlapping by its area the boarder-lines of wasout. The next step is filling an embankment ( 5 ) of loose material (sandy-gravel mixture, rock debris) hihger river afloat along the boarder-line of the construction laid in such a way that in flood period, soil from the embankment filled the cellular voids of construction. In such a way the restoration of previously washed out bottom topography occurs. This way is used at the initial stage of channel erosion and for preliminary study of velocity parameters of the given river. 3. Laying of antierosion construction with the following process of natural deposit of silt transported with flow into the cellular structure of construction. The method is effective with the significant soil volume in a moving stream. The laying of previously assembled elements of AEC is done with the use of floating means (pontoons, barges, floating cranes, etc.). Their choice and use depend on river dimensions and velocity of its flow. The maximum river depth where AEC was laid according to specially worked out technology, was 9 m with the maximum flow velocity 1 &sec.
Figure 4. AEC laying.
2 CONCLUSIONS A complex of experimental investigations carried out, as well as in situ survey of bed sections with the antierosion cover for a number of years, confirmed an efficiency of AEC. The results of above investigations are shown in a diagram (Fig. 3). REFERENCES Alperin I.E., Bykov L.C., Gurevich V.D. 1973. Strengthening of banks of navigation channels, rivers and reservoirs. Truiisport: 369 p., Moscow. Babkov V.F., Bezruk V.M. Bases of soil science and soil mechanics. 1986. Utiiversity: 239 p., Moscow. New Plastic Honeycomb Builds Better Road Base. 1986. Highway & Heavy Constructior7: v. 129, No.3, 56-57. Soil science. Edited by Sergeev E.M. 1986.Ui7iversiry: 387 p., Moscow.
Figure 3. Diagram of experimental investigation results.
618
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Analysis of two layer soil system beneath rigid footings - a global approach A. Sridharan & B.R. Srinivasa Murthy Indian Institute of Science, Bangalore, India
P. Vinod College of Engineering, Thiruvanantapui-am, India
ABSTRACT: The bearing capacity and settlement behaviour of a rigid foundation on a weak soil can be improved by providing a reinforced soil bed or a geocell mattress. Mechanisitically this will result in a two-layer soil system below a rigid footing. The upper reinforced layer will have a relatively high value of modulus of elasticity. Often the ratio of the moduli of the upper and lower layer being more than 100. There is no simple method available in literature (excluding numerical methods) to estimate the load settlement behaviour and ultimate load of such a system. The contact pressure distribution below rigid footings placed on the surface of a two layer granular soil will depend on the elastic and plastic zones formed, the thickness of the upper layer, the load level in relation to the ultimate load of the system and the ratio of elastic moduli of the soils of the two layers. Solution for this type of problem can be obtained without using the complex constitutive modelling approach. The adopted method in this investigation has been termed as ‘Global Approach’. This is more suitable for reinforced earth where modelling is complex. In this investigation, the basics of the global approach has been presented. The numerical results obtained from this approach have been presented in nondimensional form, which can be used to estimate the ultimate bearing capacity and load displacement behaviour of a two layer system below rigid footing. The results have been validated in relation to the published experimental data. involving a reinforced upper soil layer where modelling is complex.
1 INTRODUCTION
Situations are often encountered where structures are to be built on weak soils. The frequency of occurrence of such situations is assuming greater proportions day by day with increasing rate of construction activities and scarcity in the availability of ‘good’ sites at present. One method of improving the bearing capacity and load-displacement behaviour of footings placed on weak soils is to construct them over a compact reinforced soil bed or a geocell mattress. Mechanistically it then turns out to be the case of a footing resting on a two-layer soil system. The upper reinforced layer will have a high value of modulus of elasticity relative to the lower weak layer, the ratio of their modulii sometimes being even as high as 100. There is no simple method available in literature to estimate the ultimate bearing capacity and load - displacement behaviour of such two-layer systems. The finite element method can capture the complexities of the problem accurately, but it is more elaborate and has not found a wide acceptance in foundation design practice. In this paper, it is intended to present the salient features of a new, simple approach for determination of ultimate bearing capacity and settlements at various loads of a two - layer soil system beneath a footing. Such an approach is highly desirable for problems
619
2 STATEMENT OF THE PROBLEM To present the various details of the proposed approach, the example problem considered is that of a rigid circular footing of diameter D resting on the surface of a two-layered half-space with a granular material of thickness Z as the stronger upper layer. Modulus of elasticity and Poisson’s ratio of the upper and lower soils are (El, vl) and (E2, v?) respectively. The ultimate bearing capacity of a homogeneous semi-infinite mass of upper soil beneath the rigid circular footing is qul and that of lower soil is qu?. It is required to determine the ultimate bearing capacity (quL) and the settlement (86~)at a defined value of load ratio, k of the two- layer soil system, where
average pressure acting on the two-layer system k = ________________________________________--ultimate bearing capacity of the two-layer system The finite element method has been used for obtaining a simple solution to the problem.
3 CONTACT PRESSURE DISTRIBUTION BENEATH A RIGID FOOTING ON A TWOLAYER SOIL SYSTEM
4 BASICS OF A GLOBAL APPROACH
It is intended to propose a method where the most probable contact pressure distribution existing at the interface of a rigid footing and a two-layer soil system is to be used as an input parameter for the solution of the problems stated in section 2. Contact pressure distribution beneath rigid footings has been of interest to several investigators. Laboratory and field investigations (e.g., Ho and Lopes, 1969; Akai and Otsuki, 1974) on contact pressure distribution beneath rigid footings have indicated that a progressive change in contact pressure distribution, with increase in pressure beneath the centre of the footing, occurs with increase in load. Further, the pattern of distribution is a combination of that at elastic state and at plastic state at all load levels before the ultimate state, plastic state existing close to the edges of the footing and the elastic state in the interior. The most probable contact pressure distribution in such a system, due to the existence of both elastic and plastic states simultaneously, has been termed as the elastic-plastic contact pressure distribution. The elastic-plastic contact pressure distribution beneath rigid footings can be obtained by the method suggested by Schultze (1961) and Balakrishna et.al. (1992). Using the above method elastic-plastic contact pressure distribution patterns have been obtained in a non-dimensional form for various values of El/E*, Z/D and k. The typical shape of a pressure distribution pattern is shown in Fig. 1. A realistic analysis of bearing capacity or settlement of a two-layer soil system beneath a rigid footing should satisfy the appropriate, defined contact pressure distribution pattern. Specification of such a stress boundary condition forms one of the two basic features of the proposed approach which has been termed as 'Global Approach'.
As stated in Section 3, in the present study, the most probable contact pressure distribution, which is a function of EJE2, Z/D and k, is used as the loading on the circular footing [In particular for ultimate bearing capacity analysis, contact pressure distribution for the case of k = 1.0 is used.] Further, it is known that a centrally and vertically loaded rigid footing would undergo uniform vertical displacement at all load levels, the magnitude of displacement obviously increasing with increase in load ratio. This reality, namely uniform displacement of all the nodes at the soil-footing interface can be used as a constraint condition in the finite element analysis along with the specified boundary condition. This is in contrast to the conventional elasto-plastic finite element analysis in which either the stress or the displacement boundary condition is specified and the other determined, for a given material behaviour (constitutive law). On the other hand, the basis of the present approach is the simultaneous satisfaction of the known stress boundary condition (i.e., the defined contact pressure distribution) and the constraint condition of uniform displacement (though magnitude is not known apriori). For a given two layer system at a defined value of load ratio, only a unique pattern of variation of elastic modulii of soils in space would satisfy the two requirements simultaneously. This unique pattern can be obtained by performing trial finite element analysis with different patterns of variation of elastic modulii within the soil system. Such a procedure may be considered as a method of arriving at the constitutive behaviour of a given two-layer soil system at a defined value of load ratio from a global perspective since it assigns importance to the practically observed features (most probable contact pressure distribution and uniform settlement) of the system as a whole. The feasibility of this approach is examined in this paper. It is a fact that some plastic yielding occurs in soils near the edges of the footing even for small load levels. Plastic zones develop around the edges of the footing, their size increasing with increase in load, and the remaining soil is within elastic limits. At ultimate state, the plastic zones would have spread to such an extent that the remaining elastic soil does not play any significant role in sustaining the loads (Chen, 1975; Selvadurai, 1979). It is assumed in the present study that at any value of load ratio, the two-layer system is composed of distinct elastic and plastic zones, the size of the plastic zones around the edges of the footing increasing with increase in load ratio. 4.1 Size and shupe of plastic zone Jumikis (1 969) has indicated that for a centrally and vertically loaded rigid footing at failure, the rupture
Figure 1 . Elastic and plastic zones.
620
surface is symmetrical and the footing settles uniformly. Further, the rupture surface which separates elastic and plastic zones coincides remarkably well with the mathematical curve of a logarithmic spiral of the form: ri = roaee where ro = initial radius vector of the logarithmic spiral curve; ri = radius vector from the pole of the spiral curve to any point on the curve; 6 = angle of amplitude between ro and ri; and a =a dimensionless parameter. In the present study, Eqn. (1) is used for defining the cross-section of the shape of the boundary between plastic and elastic zones at all load levels (Fig. 1). Here, lroqis the distance from the edge of the footing to the point of transition between elastic and plastic zones in the plane of soil - footing interface and is a function of El&, Z/D and k; At ultimate state plastic zones would have spread to such an extent that r, = 0.5D. 4.2 Variation of elastic modulii within the plastic zone At any point P within the plastic zone, the elastic modulus (Ep)has been considered to vary as a function of the radial distance of the point P from the pole 0 in the following manner:
ultimate state, influence coefficients can be modified appropriately. 5.1 Modified influence coefficients for ultimate bearing capacity prediction It is proposed that the conventional equation for elastic settlement of a homogeneous soil beneath a rigid circular footing may be written in an analogous form to determine the settlement ( ~ u L )of a two-layer soil system beneath a rigid circular footing at ultimate state as:
(3) where quL = ultimate bearing capacity of the twolayer soil system; and IUL= modified influence coefficient of the two-layer soil system at ultimate state. ~ U in L the above expression may be considered as the settlement corresponding to the ultimate bearing capacity point in a pressure vs. settlement curve. Eqn. (3) can be written for a rigid circular footing on homogeneous semi-infinite masses of upper and lower soils respectively at ultimate state as: (4)
where the subscripts 1 and 2 refer to the upper and lower soils. IUI= Iu2 since both these refer to the modified influence coefficient of a homogeneous soil at ultimate state (which will, hereafter, be called IUH). The settlement at ultimate state of a two-layer soil system (6"~)may be expressed in terms of the settlements at ultimate state of individual soil layers under homogeneous conditions ( 6 ~ 1and 6 ~ 2 )using the weighted average method of Sridharan et al. (1990) as:
where j = I or 2 depending on whether P is within the the upper layer or the lower layer; f = a dimensionless parameter; and OP and OP' are as shown in Fig. I . The use of equivalent elastic modulii for points within plastic zone is quite reasonable since particulate soil is stable even beyond yielding. Further, the above equation satisfies the two boundary conditions, namely (i) beneath footing edges, elastic modulus is zero, which is practically true for the case of a surface loaded rigid footing on a granular soil, because of the complete flow conditions existing therein; and (ii) perfect elastic conditions prevail outside the plastic zone.
where (CF)l = cumulative influence factor for the upper layer (Fig. 2). Eliminating 6 ~ 1 6, ~ 2and 6 " ~equations (3), (4), ( 5 ) and (6) and assuming v1 = v2 the following equation for ultimate bearing capacity of a two-layer soil system (quL) is obtained:
5 ESTIMATION OF ULTIMATE BEARING CAPACITY AND SETTLEMENT USING THE GLOBAL APPROACH In elastic theory, settlements of footings are linked with applied loads and soil properties through influence coefficients. In the present investigation where a spatial variation of elastic modulii is considered for the two-layer system at all load levels up to the
62 1
(CF), 0
0.2
0.4
0.6
0.8
1.0
Figure 2. Cumulative influence factor (Sridharan et al. 1990).
In the above equation, the value of IULdepends on El/E2, and Z/D. In order to determine IULfor a given two layer system at ultimate state, the finite element method of analysis has been used with the incorporation of the features described in sections 4.1 and 4.2. It has been found that for a given two layer system, only a unique combination of 'a' and 'f [see Eqns. ( 1 ) and (2)] satisfies the two global requirements ( mentioned in the beginning of section 4) of the system simultaneously. Values of modified influence coefficients have been evaluated for vari, Z/D and are presented in ous values of E I / E ~and Table 1. It is seen from this table that modified influence coefficient (and consequently the ultimate bearing capacity) is significantly different for two-layer soil systems with different values of E I / E ~particularly for a smaller thickness of the upper layer.
= modified influence coefficient of a rigid circular footing on a two-layer soil system at load ratio = k. Values of modified influence coefficient have been evaluated using a procedure similar to that used in section 5.1. In this case, the modifjed influence coefficient ( I ~ Lnot ) only is a function of El/E2 and Z/D but also of load ratio (k) (Table 2). It is seen from this table that stratification has a pronounced effect on settlements. Further, for a given two layer soil system (with defined values of E[/E2, and Z/D), there is a significant increase in the magnitude of modified influence coefficient (and consequently settlement) with increase in load ratio. There is no simple method available in literature which accounts for the non-linearity in the pressure vs. settlement behaviour of even homogeneous soils at higher loads. In the present method, non-linear response of the soil system is taken care of in an approximate manner through the use of appropriate values of modified influence coefficient in Eqn. (8). Table 2. Modified influence coefficients (ILL)for settlement prediction
EL
z
E2 1 2
D
5
10 Tablc 1. Modified influence coefficients (1~1)for ultimate bearing capacity prediction
EL E2 1 2 5 10 20
0.25 3.056 4.966 10.732 20.540 39.100
0.50 3.053 3.903 6.149 9.265 14.310
z/D 0.75 3.056 3.503 4.572 5.839 7.650
1.00 3.056 3.328 3.886 4.511 5.412
2.00 3.056 3.111 3.215 3.359 3.650
20
5.2 Modi$ed influence coefficients .for settlernent prediction
--
0.25 0.50 0.75 1.00 2.00 0.25 0.50 0.75 1.00 2.00 0.25 0.50 0.75 1.00 2.00 0.25 0.50 0.75 1.00 2.00
0.1 0.786 1.280 1.109 0.998 0.930 0.822 2.727 1.948 1.509 1.256 0.917 4.936 3.018 2.102 1.643 1.061 8.530 4.420 2.951 2.279 1.328
0.3 0.926 1.463 1.275 1.166 1.078 0.963 3.159 2.231 1.704 1.426 1.0.59 5.826 3.439 2.333 1.839 1.201 10.73 5.145 3.214 2.468 1.451
k 0.5 1.074 1.672 1.450 1.320 1.232 1.1 12 3.638 2.578 1.937 1.598 1.208 6.797 3.967 2.621 2.040 1.552 12.89 5.943 3.505 2.675 1.600
0.7 1.282 1.995 1.711 1.545 1.446 1.322 4.359 3.024 2.237 1.863 1.418 8.237 4.653 2.989 2.315 1.564 15.70 6.958 3.936 2.971 1.798
0.9 1.678 2.571 2.167 1.974 1.853 1.718 5.574 3.738 2.739 2.299 1.817 9.952 5.632 3.600 2.797 1.963 19.92 8.661 4.775 3.419 2.198
6 COMPARISON STUDY
In order to check the validity of the numerical results obtained by the proposed approach, a comparison study has been made using the data obtained from the literature. Typical results of predicted and observed values of ultimate bearing capacity of twolayer systems are shown in Table 3. Predicted and where qkL = average pressure on the two-layer soil observed pressure vs. settlement curves for two typisystem at load ratio; k (and is equal to k.qkL) and I ~ L cal cases of layered soils are presented in Fig. 3. De-
It is proposed that settlement of a two-layer soil system (&L) beneath a rigid circular footing at any load ratio can be expressed as:
622
tailed comparison study has been presented elsewhere (Vinod, 1995). It is seen from Table 3 and Fig. 3 that the predictions are fairly accurate.
presented in non-dimensional form. The results are validated in relation to the published data. The possibility of adopting the 'Global Approach' to twolayer systems with a reinforced upper soil layer merits exploration since modelling is quite complex in this case.
7 CONCLUDING REMARKS The basics of a simple method termed 'Global Approach' to estimate the ultimate bearing capacity and settlement at various load levels of a rigid circular footing resting on a two-layer soil system with a stronger material in the upper layer is presented. The numerical results obtained form the approach are
8 ACKNOWLEDGEMENT The authors would like to thank Mr. P Raghuveer Rao for his help in preparation of this paper.
Figure 3. Comparison of predicted and observed pressure vs. settlement curves of two-layer soil systems.
623
Table 3. Predicted and observed values of ultimate bearing capacity of two-layer soil systems Source of data Milligan et al. (1986) [experiinetanl study] Bindumadhava (1990) [experimental study] Brocklehurst (1993) [large strain finite eleiiient analysis]
D(,,,) 0.30 0.15 0.25
Type of soil Granular Material overlying soft soil Sand overlying soft soil Granular material (Matusuoka yield criterion) overlying clay (von Mises vield criterion)
4u
EL
qu2 7.08
E2 15
1.90 5.07
zD
8.65
0.50 0.67 0.5
(kP4 Predicted 565.0 733.0 106.1
Observed 495.6 742.6 115.5
3.02
0.6
150.9
158.0
ClUL
REFERENCES Akai, K. & H. Otsuki 1974. Model studies on the stress distribution and the bearing capacity of soil ground. Proc. Japanese Society of Civil Engineers. 223:99-107. Balakrishna, C.K., Srinivasa Murthy, B.R. & T.S. Nagaraj 1992. Stress distribution beneath rigid foundations on sands. Int. J. for Numerical and Analytical Methods in Geomechanics. 16(1):65-72. Bindumadhava 1990. Reinforced soil technique jor sof/expansive soils. Ph.D Thesis, Indian Institute of Science, Bangalore, India. Brocklehurst, C.J. 1993. Finite element studies of reinforced and unreinforced two-layer soil systems. Ph.D Thesis, University of Oxford, UK. Chen, W.F. 1975. Limit Analysis and Soil Plasticity, Developments in Geotechnical Engineering 7, Elsevier, Amsterdam. Ho, M.M.K. & R. Lopes 1969. Contact pressure of a rigid circular foundation, Joiiriid of the Soil Mechanics and Fouridiitions Division, ASCE. 95(3):79 1-802.
624
Jumikis, A.R. 1969. Theoretical Soil Mechaizics. Van Nostrand Reinhold Company, New York. Milligan, G.W.E., Fannin, J. & D.M. Farrar 1986. Model and full-scale tests of granular layers reinforced with a geogrid. Proc. 3rd Inter-izutioizal Conference O H Geotexxtiles, IAJ8, Vienna,Austria :61-66. Schultze. E. 1961. Distribution of stress beneath a rigid foundation. Proc. 5th Intenzatiotzal Corifereiice on Soil Mechanics and Foundation Eizgiiieering. Paris. 1 :807-813. Selvadurai, A.P.S. 1979. Elastic aiialysis of soil$outzdatioiz irrteruction, Developments in Geotechnical Engineering. 17, Elsevier Scientific Publishing Company, Amsterdam. Sridharan, A., Gandhi, N.S.V.V.S.J, and S. Suresh 1990. Stiffness coefficients for layered soil systems. Jourtial of the Geotechnical Etzgiizeering Division, ASCE. 1 16(4):604-624. Vinod, P. 1995. Aizalyses of two-layer soil systems beneath rigid footiizgs. Ph.D Thesis, Indian Institute of Science, Bangalore, India.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Design on limit equilibrium of foundation reinforced with geosynthetics Z. Wang Wuhan University and Tsinghua University, People
Republic of China
X.Q. Wang Wuhan University of Technology, People
Republic of China
ABSTRACT: The effects of reinforcement on bearing capacity of foundation on reinforced earth include upward tension of vertical component and lateral restraint of horizontal component of reinforce . Based on the principle of limit equilibrium, the formula of ultimate bearing capacity is deduced. The results of calculation using the formula are coincided with that from reference given by other researchers. The length of reinforcement is given based on limit pull out analysis and puncture shear failure mode. The design tensile force of geosynthetics in the formula of ultimate bearing capacity is equal to ultimate tensile strength. A factor of safety, Fs= about 3.0, is applied to the ultimate bearing capacity and to the ultimate tensile strength of geosynthetics as well to arrive the allowable bearing capacity. The effects of numbers of reinforcements on bearing capacity and settlement of foundation are analyzed, and a conclusion is given that sand cushion reinforced with layers of geosynthetics is reasonable selection. The available design formula of foundation reinforced with geosynthetics should concern with spread of vertical stress in the cushion and surcharge modification of bearing capacity due to increase of depth of the foundation. A case history of foundation of sluice reinforced by geogrid is given.
1 INTRODUCTION
bearing capacity based on the principle of limit equilibrium and the formula of determining the length of reinforcement based on the equilibrium of anchoring segment outside shear failure (load spread) lines by analyzing two limit equilibrium modes, those are limit breaking and pullout of reinforcement. The ultimate bearing capacity calculated by the formula is very consistent with the results of model tests given by other researchers and a case of foundation of sluice reinforced with geogrid designed by the method is given.
Shallow strip foundation resting on reinforced earth is made up of single layer or several layers of reinforcements horizontally placed in the soil under the foundation. A great number of model tests show it has the advantage of improving bearing capacity, and this type of foundation has been applied in many structures. Original galvanized steel ties used as reinforcements have been replaced by geosynthetics. The design methods of this type of foundation mainly include the followings: ( I ) Analysis of stresses in reinforcement and soil by Boussinesq’s solution (Binquet and Lee, 1975)’ ( 2 ) Formula of Improving Terzaghi ultimate bearing capacity (Yamanouch and Gotoh, 1979), ( 3 ) Finite element method, (4) Analysis of reinforced foundations by the slip-line method. Many assumptions have been used in these methods. For example, method (1) supposes that reinforced earth and substratum are homogeneous medium, the tensile force of reinforcement is vertical and upward, and the length of reinforcement reaches the point where the increase of vertical stress equals to 0.1 times the load. q, method ( 2 ) supposes the deformation of reinforcement on both sides of the foundation forms a arc of circle, it’s hard to determine the radius of arc and the direction of tensile force, method ( 3 ) and method (4) have little application because of their complication. This paper deduced the formula of ultimate
2 LIMIT ANALYSIS ON BREAKING AND PULLOUT OF REINFORCEMENT When the upper layer of reinforcement below the foundation is located at a depth,zl, less than about 2/3 b. the lowest layer of reinforcement is located at a depth, z,, less than about 2b ( b = width of the foundation), the number of reinforcement layers, N, is greater than 3 and the ties is sufficiently long, in this case, the soil mass fails when the upper reinforcement break. The breaking points below the foundation are the intersection points of reinforcements and load spread lines, AC and BD. as shown in figure 1, and the spread angle (0) doesn’t increase no matter the adding of reinforcements (Love, 1987). 625
The effects of reinforcement on bearing capacity of foundation include upward tension of vertical component 2NTsina/(b+2 zn tan@) and lateral restraint of horizontal component of reinforce. And the latter can be calculated by limit equilibrium theory.
Where q. , ,? and q ,, = ultimate bearing capacity of the foundation supported on reinforced and unreinforced earth, respectively. The BCR increases with the increase of the length, L, and N of reinforcements until the mass of reinforced earth covers the whole zone of failure surface. We deduced the length L, of the whole zone of general shear failure on both sides of the foundation,
Where T= ultimate tensile strength of reinforceL,, = b( I +2 tan ( TC /4+@2j er'* tan41 (6) ment, a= angle of tensile force and horizontal surface. Let a= 45"+@2, q+ angle of friction, D,,-= In the transitional zone the failure surfaces are maximum depth of slip surface in soil. So the imcurves being arcs of a logarithmic spiral, the maxiprovement of ultimate bearing capacity by reinmum depth, D,, can be gotten by derivation of the depth, forcement, &,,
Aq,,= NT(2sin(45O+qY2)/( b+2 zn tan@)+ co~(45~+@2)tan'(45~+@2)/ D,,)
b cos4
4 42) tan4
D,,=
(2)
(7)
2cos ( n /4+@2)
Aqr = NT(2sin(45"+@2)/( b+2 zn tan@)+ cos(45 O+ @2)tan2(45"+ @2)/ DJF,
Refer to formulas (6) and (7),L, and D, are given in the table, I available from different values of 4.
(3)
Table 1 . Length and depth of slip Surface
KO)
0
10
5
15 20
25
30
35
L,, (xb) 3.00 3.50 4.14 4.97 6.06 7.53 9.58 12.53 D,,(xb) 0.71 0.79 0.89 1.01 1.16 1.35 1.59 1.90
Figure 1 . Cross section of reinforced foundation.
Where the factor of safety F, is equal to about 3.0. F, is applied to the ultimate bearing capacity and to the ultimate tensile strength of geosynthetics as well to arrive the improvement of allowable bearing capacity, Aq,.. The total improvement of allowable bearing capacity of foundation reinforced with geosynthetics, A 4,
Aq = Aq,+yz,, N q / F , + 2qz,, tan@/(b+2z,,tan@) (4) Where N,= bearing capacity factor, q= load of foundation, kPa. The third term in the formula (4) is reduction of q at level zn, because spread of load. The improvement of the ultimate bearing capacity has generally been expressed in a non-dimensional form called the bearing capacity ratio (BCR),
BCR
From Table 1 , the D,, is less than 2b, that can be reached by properly arrangement of reinforcements (2/3b
q
-
Where. T,= allowable tensile force of reinforcement, kN/m, Fsp=safety factor against pullout of re-
411
626
inforcement, F.~p=3,&= coefficient of friction between reinforcement and soil, determined by test and without test data the (Chinese Standard for application of geosynthetics in hydraulic and hydro-power engineering) SL/T225-98 takes A,= 2/3tan@for geotextile,&,= 0.8tan@for geogrid. Considering in formula (8) To F,= T, then,
Where d= embedment of foundation. in fig. I, d=O. In order to examine the limit analysis method, the ultimate bearing capacity calculated by formula (2) has been compared with the results based on model tests by Ju et al( 1996). In the model test. the uniform and dense sand has relatively density D,. = 75%, unit weight y = 1 5.5kN/m3. Width of model foundation b= O.l.m, the ultimate bearing capacity of sand foundation was q,, = 16 I .9kPa. And the friction angle e45"are determined by back analysis based on classical bearing capacity theory by Ju et al. Geonet used as reinforcement has a ultimate tensile strength T= 2 kN/m , Other parameters of the tests are length L=6b, the number of reinforcement layers is from 1 to 6, distancedz = 21 = 0.25b = 0.025m. Figure 2 gives the curves for load versus settlement. The load-settlement behavior of unreinforced sand (N=O) shows a peak in Figure 2. The ultimate bearing capacity and settlement at peak are 161.9 kPa and 8.05 mm, respectively. Table 2 gives the variation of the ultimate bearing capacity (q,,,.) and BCR at the same settlement (8.05 mm) according to the increase of N. In order to compare the test results with that from limit analysis, the improvement of ultimate bearing capacity (Aq,,) calculated by formula
Table 2. A comparison of ultimate bearing capacity of sand foundation reinforced with geonet. Calculated
N a q , q , , , kPa kPa 0 161.9
Tests
q,,r
kPa
BCR
161.9
1
43.9
205.8
1.27
196.0
1.21
2
77.3
239.2
1.48
271.6
1.68 1.70
3
105.1
267.0
1.65
274.8
4
134.8
296.7
1.83
317.4
1.96
5
151.8
313.7
1.94
319.5
1.97
6
172.6
334.5
2.07
340.8
2.11
* 0=
/1,,=0.284 in
(2), ultimate bearing capacity (q,,,., q,,,.= ql,(16I .9 kPa)+Aq,,) and BCR are also given in table 2.The results from tests and forniula(2) are very consistent with each other in the Table 2. 3 A CASE HISTORY Huangshi Hexing Sluice lies in Hubei Province of China, which was first established in 1876 and reestablished in 1956. The foundation of the sluice rest on mucky soil, the thickness of the mucky soil is 10 to 18m. below it lies thick silt layer. Crack and deformation of the sluice caused by different settlement of untreated ground and aging of thc sluice structure made it a serious hidden danger to Yangtze River dyke. So it had to be pushed over and reestablished after flood in 1998. The new foundation supported on 93 piles according to original design. then the plan of reinforced earth with geogrid have been acccpted and designed by this method. Width of the sluice foundation b=5m, load of foundation q = 280 kPa, embedment d = 1.201~1,allow bearing ccpacity of mucky soil, q o , i l OOkPa, y= 18.4kNlm' c=40kPa, b16". The number of geogrid layers is 3, Z I = 0.6m, z 3 = 1.6m , distance between each layer is 0.5m, angle of friction in sand liner@=33". The structure of sand cushion reinforced with 3 layers of geogrids is shown in Figure 3. The result of A g calculated by this method is I80kPa, which include modified surcharge offers 74.25kPa, and load spread offers 74.5kPa( e = 30') and reinforcement only offers A g,=20.25kPa. Other results are D,=5.2m, T=46.61kNlm. The length of each layer of reinforcement are 6.79m, 7.12m and 7.47m. Finally, an uniform length of 7.5m was adopted. The geogrid has ultimate tensile strength of 50kN/m. When Ag,,was calculated with formula (2) the @16" (mucky soil) was taken, because the thickness
Figure 2 Curves for load versus settlement in sand (after Ju J.W, Son S.J, Kim J.Y and Jung L.G, 1996).
627
5000 --
-+-
tensile strength should be used as tensile force at limit equilibrium analysis. 4. The best type of reinforced earth is the sand cushion reinforced with geosynthetics. In spite of considering the effects of reinforcements, the load spread and increase of surcharge must be accounted. 5 For sand cushion reinforced with multiple of geosynthetics the design methods (1) and (2) are not adopted. 6. The layout of reinforcements, 21, less than about 2/3 b, and zn, less than about 2b can be applied in design of foundation reinforced with geosynthetics. 7. The length of reinforcement is given based on limit pull out analysis and puncture shear failure mode. 8. The method of limit analysis has been verified by test result and a case history.
1 ---?I $luicechninber -
Figure 3 Sand cushion reinforced with 3 layers of geogrids in Huangshi Hexing Sluice
of sand cushion was only 1.6m more less than DI,=5.2m. The foundation of reinforced earth was accomplished in Jan. of 1999. It only took one month. The sluice has played an important role in control flood in 1999 and 2000.
REFERENCES Binquet J, and Lee K L ,1975, Bearing capacity analysis of reinforced earth slabs, J Geotech Eng Div ASCE(IOl), GT12 1257-1278 Ju J W, Son S J, Kim J Y and Jung L G , 1996, Bearing capacity of sand foundation reinforced by Geonet, Proc Int Symposium on Earth Reinforcement, Fukuoka, Japan, 603-608 Love, J P, 1987, Analytical and model studies of reinforcement of a layer of granular fill on a soft clay subgrade, Can Geotech J (24) 61 1-622 Yamanouch T, Gotoh K, 1979, A proposed practical formula of bearing capacity for earth work method on soft clay ground using a resinous mesh, Technology report of Kyushu University, 52(3) 201-207
4 CONCLUSIONS 1. The effects of reinforcement on ultimate bearing capacity include horizontal restraint and upward tension of the reinforce. 2. The horizontal restraint stress of reinfbrcement is equal to horizontal component of tensile force divided by the maximum depth of slip surface, then the improvement of ultimate bearing capacity can be deduced based on limit equilibrium theory. 3. Because of safety factor of both geosynthetics and allow bearing capacity are about 3, the ultimate
628
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
The application of ground reinforcing materials for the pile foundation M. Yamada Oriental Consultants Co., Ltd, Fukuoka, Japan
N. Iwagami Structure Engineering Center, Fukuoka, Japan
H. Ochiai Kyushu University, Fukuoka, Japan
Y. Maeda Kyushu Kyoiiritsu University, Fukuoka, Japan
Y. Igase Japan Highway Public Corporation, Tokyo, Japan ABSTRACT: This study presents results of applying ground reinforcing materials to pile foundation for the purpose of improving the horizontal bearing capacity of the pile foundation, and the compound foundation consists of ground reinforcing material and pile. The objective of this study is to verify the effectiveness of ground reinforcing material, based on results of bearing capacity experiment using a physical model of a pile in sandy ground and numerical analysis. According to the results of the experiment and numerical analysis, it was confirmed that the horizontal bearing capacity increases in proportion to the length of ground reinforcing material and the improvement depth, which is the number of layers of ground reinforcing material. Therefore, the application of ground reinforcing material is an effective way of increasing the horizontal bearing capacity of piles.
1 INTRODUCTION Pile foundation has a very small horizontal bearing capacity compared to its vertical because of its long and slender form. However, in most cases, the number and dimension of pile are determined by its horizontal bearing capacity since the new horizontal design load is about two to three times of the past seismic design load based on recent major earthquakes. Therefore, it implies the importance of evaluating the horizontal bearing capacity of pile and its methods of improvement. This study is about the application of ground reinforcing materials in improving horizontal bearing capacity of piles and the bearing capacity characteristics of hybrid foundation structure composed of ground reinforcing material and pile. This paper reveals the comparison of loading test and numerical analysis results. The loading test is performed to confirm the effectiveness of ground reinforcing material in improving horizontal bearing capacity of pile using model piles embedded in sand ground.
2 TEST SUMMARY
2.1 Pile model and ground reinforcing material A diagram of the model pile and ground reinforcing material is shown in Figure I . There are two types of One has a dipile used that is made Of mension of 50mm wide, 15mm thick and 700mm
Figure 1. Pile model and ground reinforcing material.
629
long while the other has a 50mm-wide solid section, which has larger stiffness. Pile length is set from a value where its stiffness is about two to three times the ground to a length corresponding to long length pile. Also, the pile was shaped to have a pit on the sides for installation of strainometer and measuring device cords. The ground reinforcing materials that were used are factory belts (0.6mm thick, Imm mesh) made of Teflon glass fiber. Tests were conducted according to the cases indicated in Table 1 in order to confirm the effect of length and number of layers of ground reinforcing materials used.
Range of reinforced soil
No.
LR
1 1
1
2 4
5 6 I
* 9 11
Test ground is formed using dried okagaki sand placed inside a large bucket that is 40cm wide, 236cm long, 65cm high. The sand was placed after setting the pile inside the bucket by passing three layers of sieve. During this process, the ground reinforcing material is clipped to the pile when the ground reached the installation position. Moreover, in order to approximate the actual in-situ condition, overburden soil pressure (0 v ~ )is applied on top of ground since the effect of soil weight is very small in small-scale model test. The overburden soil load is composed of ten steel weights; each has a dimension of 5cm x 40c1n x 5cm and weight equivalent to (r v~=3.8kN/m2. These are set on both sides of the ground, i.e. active and passive.
12
l3 14
1
1 z: 1 1 ::: 1 I :z 1 200
3 layers (LR=3@50=150)
300
6 layers (L~=6@50=300)
10
2.2 Test ground preparation
abbreviated designation
Lb
9 layers (L~=9@50=450)
1s
16
3-200 3-300
:-()oO
600
3-600
200 300
6-200 6-300
600 200 300
6-600 9-200 9-300
600
9-600 0-000
0
i::::
Monotonous loading process is considered where the magnitude of load is gradually increased by 0 P ~ = 4 9 Nevery 30 seconds. In between load steps, the load is held for 30 seconds in order to check if the displacement settles. Moreover, the magnitude of load is set to a certain range so that the stress of model pile is within its allowable range; thus, large displacements and ultimate state of pile are beyond the scope of this study.
2.3 Loading apparatus and method A diagram of loading apparatus is shown in Figure 2. It applies load to the pile head by moving the table plate horizontally using BF cylinder. Although this apparatus is capable of applying both horizontal and vertical loads simultaneously, only horizontal loads were applied in the test. Furthermore, specially made joints were installed at connection of rod and pile so that the pile head would have hinged connection.
3 STUDY OF TEST RESULTS 3.1 Reinforced soil depth and bearing capacity of pile Figure 3 shows the effect of number of ground reinforcing material layers in improving horizontal bearing capacity of pile using Lb=600mm ground reinforcing material. The graph implies that horizontal bearing capacity of improved grounds is larger than unimproved ground in all cases. 2s
E
y, 2 0 3:
e,
-z-
;3
15 10
c
E
g
0.5
3: 0.0
0
5
10
15
20
Displacement
25
30
35
1% (mm)
Figure 3. Effect of reinforced soil depth (Lh = 600mm).
Figure 2. Loading apparatus.
630
3.2 Ground reinforcing material length and horizontal bearing capacity of pile Figure 4 shows the effect of ground reinforcing material length Lb when reinforced soil depth is 150mm, i.e. using three layers of ground reinforcing material. Based on this, horizontal bearing capacity is greater using longer ground reinforcing materials.
3.3 Efjrect of ground reinforcing material to horizontal bearing capacity of pile Figure 5 reveals the relationship of horizontal bearing capacity ratio (i.e. ratio of improved ground with respect to unimproved ground) and ground reinforcing material length when displacement of pile head is set to 6~=30mm.Based on this result, horizontal bearing capacity ratio increases with length and number of layers of ground reinforcing materials. Also, horizontal bearing capacity is much more improved when using stiffer pile, where in this case the dimension of pile is 50mmx50mm with three layers of ground reinforcing materials.
.-o 1.4 4 4
E .-x
1.3
4 4
5 1.2 0
0
.-2 1.1
2
-
$ 1.0
2 0.9
3
.$
0
X
. o
8
1 3
-yi-
300
+ .
400
-G-
500
-G-
600
, 6
9
Number of ground reinforcing material layers n @&=3.8kNlm2 ,f& =30mm) Figure 6. Relationship of reinforced soil depth and horizontal bearing capacity ratio.
Figure 6 shows the relationship between horizontal bearing capacity ratio and number of layers used. This indicates that horizontal bearing capacity ratio increases with number of layers. However, it also reveals that the rate of increase is largest when using about three layers of ground reinforcing materials, which is the case where reinforced ground depth is relatively shallow. Therefore, this means that, for flexible structures like pile foundations, horizontal bearing capacity can be effectively improved by reinforcing the shallow part near the surface of ground. This is due to the large effect of passive resistance of ground near the surface with respect to horizontal resistance of pile. 4 NUMERICAL ANALYSIS
4.1 Method of calculation and analytic model In this calculation, elastic beam model is used where the ground and ground reinforcing materials are assumed as spring elements. Here, the spring constant K of ground that considers the effect of ground reinforcing materials can be express as follows. -
K=
Figure 5. Relationship of ground reinforcing material length and horizontal bearing capacity ratio.
(K H I . B
clck + C ai . Ti ) B ock
Where, 6~ is the displacement, B is the pile width, kH1 is the ground spring constant, Ti is axial force of ground reinforcing material, Oi is the parameter with respect to axial force (Oj=O when 6tj=O and oj=1 when 6,i >O), and 6ti is the displacement at the location of ground reinforcing material. Moreover, the spring element, which represents the ground and ground reinforcing material, is assumed to be elastic since the load-displacement relationship at pile head shows elastic behavior based on the load test result as illustrated in Figure 7. The applied values was k~1=9400kN/m2,where test value is equal to analytic value for ordinary ground as shown in Figure 10.
On the other hand, ground reinforcing material is only effective at its axial direction and modeled as rigid-plastic since tension reaches its maximum value for minute changes. The tension of ground reinforcing material can be computed according to the following equation, as illustrated in Figure 8.
Where, Lb is the length of ground reinforcing material, Wb is its width, U,O is the overburden soil load, U d is the weight per unit volume of dry soil (U d=16kN/m3), 0 is the friction angle between the ground and ground reinforcing material. Here, U is 29 degrees based on the direct shear test result.
4.2 Summary of analytic results
4.2.1 Load-displacement relationship Figure 10 shows the load-displacement relationship of ordinary ground without overburden soil load (o,o=OkN/m*). It shows that the load test and numerical analysis reveal similar results when k,,,= 9400 kN/m*. Also, the load-displacement relationship of load test and numerical analysis for reinforced ground (3 layers-600mm) is compared in Figure 11. The correlation is best when Ti is five times the value computed using equation 2. This is because the ground strength increases due to the component of force acting downward which occurs when ground reinforcing material moves in response to pile deformation. Thus, it can be considered that the actual tension is approximately few times the tension based on equation 2 because of the compound effect of the ground and ground reinforcing material. Also, in Figure 12 calculated bending moments of pile show good correlation with values determined from readings in strainometer.
Figure I 1. Comparison of computed values and test values for case 3-600.
Figure 9. Spring model for ground and reinforcing material.
632
Figure 14. Calculated ground reinforcement effect (number of ground reinforcing material layers).
model. Moreover, the degree of increase in horizontal bearing capacity is little higher compared to the test result in the case with longer ground reinforcing material, i.e. length is 50cm and 60cm. This is because the increase in ground strength is evaluated a little higher when assuming the spring constant for ground reinforcing material, as explained in section 4.1.
Figure 12. Comparison of bending moments of pile (without using ground reinforcing material).
4.2.2 Effect of ground reinforcing material Figures 13 and 14 show the effect of ground reinforcing material to horizontal bearing capacity of pile. Figure 13 shows the effect of ground reinforcing material length to horizontal bearing capacity that is the same with Figure 5 . It illustrates that reinforcement effect is better when ground reinforcing material length is longer. Figure 13 presents the effect of reinforcement depth (i.e. number of ground reinforcing material layers) to horizontal bearing capacity that is similar to the result shown in Figure 6. It reveals that horizontal bearing capacity increases by approximately 10 to 60 percent due to the effect of ground reinforcing material. Furthermore, although the reinforcement effect is proportional to ground reinforcing material length and number of layers, it was found that reinforcement effect occurs significantly when using 3 to 6 layers and gradually increases when using more number of layers. This result is almost similar to the test result of the pile
5 SUMMARY The results obtained in this study can be summarized as follows. 1 The horizontal bearing capacity of pile can be increased by using ground reinforcing material. 2 Reinforcement is attainable only within shallow ground near the surface, that is, the horizontal bearing capacity of pile increases within 1/p range of reinforcement. In the test, number of layers should be approximately six. The horizontal bearing capacity of pile increases with pile stiffness. Moreover, horizontal bearing capacity of pile is expected to increase due to its mutual relationship with ground reinforcing material acted by forced displacements of ground. 3 The results of numerical analysis and load test are almost similar for both ordinary and reinforced ground. Here, an analytic model is proposed for stability analysis. Computed tension for reinforced ground is less than its actual value but it coincides with test result when multiplied by few times. This is because of the effect of the component force acting downward developed due to the displacement of ground reinforcing material acted by pile movements, as mentioned in section 4.1. Therefore, it can be concluded that the magnitude of tension is few times larger than computed value due to compound effect of ground and ground reinforcing material.
Figure 13. Calculated ground reinforcement effect (ground reinforcing material length).
633
REFERENCES Japan Road Association 1996. ~~~i~~ specifications of highway bridges Part seismic design. F. Nakamura, H. Shouji, y. Maeda 2000, Shear sliding of characteristics on Okagaki sand, Proceedings of conf conference of Japan society of civil engineers west branch office.
v
634
Y. Maeda, K. Tanaka, H. Hataoka, M. Yamada 2000. Characteristic of Horizonatal Bearing Calacity of Reinforced Pile with Geotextiles. Bulletin of Kyusyu kyouritsu uiniversity faculty of engineering, No.24: 105-112. M. Yamad& N. Iwagami, H. Ochiai, y . Maeda, y. Igase 2000. Of Geosynthetics to Improvement of Horizontal Bearing Capacity of Piles, Proceeding of the second European geosynthetics conference: 287-290.
Landmarks in Earth Reinforcemenf, Ochiai ef al. (eds), 0 2001 Swefs & Zeitlinger, ISBN 90 2651 863 3
Analysis of improved ground with geonet reinforced stone columns Zhou Zhigang Department of Highway & Communications Engineering, Changsha Communications University, PRC
Zhang Qisen & Zheng Jianlong Changsha Communications University, PRC
ABSTRACT: This paper presents the results of analysis of soft ground improved by stone-columns wrapped with layers of geonet and needle punched geotextile by using nonlinear FEM and testing in laboratory and field. The validation of needle-punched geotextile on improving the filtrating function of crushed stonecolumns is elucidated by combining with the results of laboratory test. According to the comparison between the results of calculation and measurement, and the influence of column spacing on the settlement of soft soil ground, the rationality of column arrangement scheme is expounded and proved. Besides, the reinforcing function of geonet on crushed stone-columns to resist bulging failure near the tope end of stone-column is also discussed by using limit state equilibrium theory. 1 INTRODUCTION
model for soft soil, the limit state equilibrium theory, and field and laboratory tests.
Two main problems will be encountered in the application of crushed stone-columns. The first one is that during drainage the tiny particles in soft soil will be carried into the crushed stone-columns by seepage flow, which will block effective drainage way in the crushed stone-column. The other one is, due to the heaped preload and construction load or other loads, the crushed stone-column may be damaged at the depth approximately equal to 2 to 3 times the radius of the crushed stone-column near the top of column due to bulging. The reason for this is the shear strength of soft soil near the top end of the stone column is fairly low and the soft soil is unable to provide sufficient resistance to the bulging. When these problems become more serious, there will be loss of the capacity of improved ground with crushed stone-column. To solve these problems, some researchers have used layers of geogrid in stone-column (Madhav et al. 1994, Sharma 1998, Zhou et al. 1998) to reinforce the stone column near its top end. Similarly, crushed stone-column improved by wrapping layers of geosynthetics materials including geonet and needle punched geotextile around it are used when we deal with soft soil ground in section K3 1+946-K32+045 of National Road 320 in Hunan Province. Results of test carried out in field trial have proved that this is an effective method. In order to interpret the action of geosynthetics materials on crushed stone-column, following will present the study by using FEM based on Biot consolidation theory and Duncan-Chang nonlinear
2 FEM BASED ON BIOT CONSOLIDATION THEORY 2.1 Nonlinear FEM For axial symmetrical question, 8-noded isoparametric is selected. Equilibrium equation based on Biot consolidation theory and continuity equation for saturated soil are:
Here ke is the total stiffness matrix for node displacement
-1-1
kep is the total stiffness matrix for node pore pressure 1 1
[kepi= C 2 R
j ~ ~ l T i n ; r X ~ l ~ ~ 1(3){ ~ F
-1-1
kp is element matrix of seepage discharge
635
p is time difference coefficient,
and its value is determined by assumptive variation form for k, at certain time; ARw is node pore pressure vector for element at the time t, and ( A 6 ) , (Ap) are increasing vectors of node displacement and pore pressure at certain calculation time respectively.
(ARw)'
(5)
Tangent modulus Et and Poisson ratio p for soft soil are used as elastic modulus and Poisson ratio in elastic stiffness matrix. They can be represented by Duncan-Chang nonlinear model. Membrane element is used to represent the geonet. In order to describe the nonlinear relationship of relative displacement and shear stress on the interface between column and soil, Goodman interfacial element is used:
2.2 Calculation parameters The parameters used in calculation by FEM are listed as following. For crushed stone column, its resilient modulus is 300Mpa, and Poisson ratio is 0.3. For sand cushion on the top of soft soil, its resilient modulus is 20Mpa, Poisson ratio is 0.35, and the permeability coefficient is 0. l i d d . For soft soil, the parameters in Duncan-Chang model are C=lO.OkPa, $=17', K=9.0, n=0.1, Rg0.31, H=O. 14, F=O.O, D=2.0, K u ~ 1 2 . 0m=O. , 1. For geonet, its tensile stiffness is Eg=3.0MN/m. For the interface between column and soil, the parameters in Goodman interfacial model are K1=34.0, n=0.5, R ~ 0 . 5 , C=lO.OkPa, $=17", K,=1.0x1OS MPdtn. 3 MECHANISM OF GEOSYNTHETICS lMPROVING STONE-COLUMN 3.1 The infiltration action of needle-punched ge otextile
To reveal the infiltration characteristics of geotextile between soft soil and stone column, modeling experiment was carried out in one steel barrel filled with mud. which is 1.3m deep with 1.Om diameter. In the center of barrel, there is a crushed stone column wrapped with a layer of geonet and needle punched geotextile. The diameter of the column is 16cm. Above the mud, there is a insulating layer, which is composed of geonet and needle-punched geotextile covered with a crushed stone cushion to drainage horizontally (see figure 1). 636
Figure 2 The settlement curves of improved ground for field trial
The results of test show that there existed a filtration layer between the column and the mud, whose permeability coefficient is between those ones of the column and the mud and its value is about 0.00 l d d . The permeability coefficients of column and mud are O.lm/d and 0.0001m/d respectively. To prove this, we used FEM to calculate and analyze the influence of the permeability coefficient of the filtration layer on the consolidation and settlement of mud during the range of 0.005-IOdd. The results have proved the existence of the filtration layer again. The results of laboratory testing and calculation indicate that the columns with and without needle-punched geotextile have different speed of consolidation and settlement (see figure 2). The needlepunched geotextile around the column improves the draining of stone column, and accelerates the speed of consolidation and settlement of the mud ground. 3.2 Analysis of settlement of crushed stone-column improved ground 3.2.1 The settlement of improved ground The improved ground in field trial is mainly composed by following components: the soft soil is 4.821~1thick; under it, there is a 0.8m thick sand layer; above the soft soil, the sand and stone cushion is 0.6m thick; the fill is 6m high. The average length of crushed stone-column is 2.4m, its radius is 0.2m, and the spacing of columns is d=l.5m (see figure 3). The loading stages of the fill is: 2.0m during 1st to 6'h days, 2.0m during 12'h to 24'h days, 2.0m during 36"' to 46t'' days. The agreement between the results of field test and FEM calculation is satisfactory, and the error is less than 10% (figure 4). It indicates that the application of needle-punched geotextile can accelerate the speed of consolidation and settlement of soft soil.
Figure 4 The settlement curves of improved ground for field trial
we have conducted model testing with two kinds of columns' spacing (d=OSm, 1.0m) in laboratory, and calculated and analyzed for some cases of effective radius (R=1.05m, d/2=0.8,0.5,0.4,0.3,0.25 m). The model testing is conducted in a laboratory testing trench which is 3.72m long, 1.5m wide, and 1.5m deep. There are 21 pieces of stone-columns which are 1.2m long with 0.16m diameter. In the trench, there filled with mud 0.9m deep, and crushed stone cushion 0.3m thick. There is a layer of geonet between the cushion and the mud. 0.IMPa pressure is loaded on the top of the cushion. It took 138 days to finish the test. In order to apply the laboratory testing results to trial section in field, the replacement rate W is used to replace the spacing d to analyze the effect of spacing. There is good agreement between the results of testing and calculation (Figure 5 ) . The relations between the total settlement S, and W or R regressed by using hyperbolic equation are: lg(S,)=A,+BI W
(~0.964)
1.O/ SW=A2+B2/R ( ~ 0 . 9 9 6 )
(7)
(8)
Where:A1=3.46, Bl=-lO.40; A2=-0.0225, B2~0.0296.
3.2.2 The influex nce of columns ' spacing on consolidation and settlement of improved ground In order to verify the validity of columns' spacing for the consolidation and settlement of improved ground,
Figure 5. The relation curves of total settlement Scn and replacement rate W or effective radius R.
Figure 3 . The structure in the field.
637
The curves of S,-W and S,-R trend to level gently since W=0.7 or R=0.3. In fact, according to the definition of replacement rate, these two values are the same. If the equality criterion of replacement rate was used, the W of improved ground in field should be W=0.7 in order to ensure the stone-column to act well. As the radius of stone column in-site 1=0.2m, so the spacing of columns is about d=1.5m. This is agreed with that one used in field. The calculating results of shear forces on the wall of column (Figure 6) and pore pressures also show that decreasing the spacing of columns can accelerate the speed of consolidation and settlement of soft soil and weaken the pressure on the surface of soft soil ground. It means that the application of geonet reinforced stone-column can improve the capacity of soft soil ground. 3.2.3 Reinforcement of geonet for I crushed stone column The limit bearing capacity of geonet reinforced stonecolumn is deduced by using limit state equilibrium theory (Zhou et al. 1997). Results show that the failure plane of bulging is located at the depth approximately equal to 1 to 3 times the radius of the crushed stone-column near the top of column to the nonlinear distribution of lateral pressure and axial force on column. Using the geonet can increase the bearing capacity of stone column significantly, shown as Figure 7, in which T is the tensile strength of geonet, r is the radius of column, Ct is the cohesion of soft soil, ps and pt are the limit pressures acted on the tops of column and soft soil respectively, h is the bulging length of
T ( 2 r C t) 0
10
5
14.0
I
12.0
-
10.0
-
I
15
20
I
__
6 .O
I
I
I
Figure 7. The relation curve of pJp, and T/(2rC,)
column, and hl is the spacing between the top of column and the location at which the maximum shear stress is acted on the wall of column.
4 CONCLUSIONS
I . The results of laboratory test and calculation by FEM show that using needle punched geotextile around the stone column can improve the filtrating function of stone column. 2. Using geonet with more rigid tensile stiffness to wrap up the stone-column can increase its bearing capacity to resist bulging. 3. The results of calculation and measurement in field trial indicate that this kind of soft soil ground treatment method is valid and reasonable. REFERENCES
0
shear force(kN/m) 10 20 30 40 50
60
J S Sharma 1998 A study of the behavior of geo-grid reinforced stone column Proceedings of 6"'InternationalConference on Geosynthetics 877-882 Atlanta Madhav, M R , Alamgir, M & Miura, N 1994 Improving granular column capacity by geogri dreinforcement FIfth International Conference on Geotextlles, Geomembranes and RelatedProducts 35 1-356 Singapore Zhou Zhigang, Zheng Jianlong & Gao Yanxi 1998 Analysis on the compression strength of sample reinforced by geogrid with limit equilibrium theory J Of Changsha Communications University 14( 1 ) 39-43 Zhou Zhigang & Zhang Qisen 1997 Analysis on the bearing capacity of geogrid reinforced stone-column Chinese Journal of GeotechnicalEngineering 19( 1) 58-62
0.0
1.0 W
sa 4
2.0
2.5 Figure 6. Distribution of shear stress on the wall of column
638
5 Soil nailings
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Simulation of soil nailing facing walls in finite element analysis S. Bang & W. Nyaz Department of Civil and Environmental Engineering South Dakota School of Mines and Technology Rapid City, SO, USA
ABSTRACT: An analytical solution has been developed to simulate the detailed behaviors of the flexible soil nailing facing wall by membrane elements. The behaviors of the membrane elements at the element level are formulated based on the concept of the unit cell and then incorporated into an existing generalized plane strain finite element method of analysis. The solution has been used to conduct a preliminary analytical parametric study to identify the effects of the shotcrete facing wall stiffness on the overall behavior of the soil nailing walls. quasi-two-dimensional approach. However, it must be noted that the application of this approach is restricted to structures having a constant angle between the principal axis of the material orthotropy and the out-of-plane coordinate. Soil nailing wall with reinforcements having the same skew angle to the out-of-plane coordinate and the horizontal spacing is an example where the generalized plane strain approach may be used. The generalized plane strain approach assumes that the plane strain directional strain, E , , remains zero instead of the plane strain directional displacement, w , being zero, as is commonly adopted in the conventional plane strain approach (Timoshenko & Goodier 1970). Therefore, this approach includes three non-zero displacement components of U , V , and 'w along the x,y, and z coordinates, none of which is dependent on the out-of-plane coordinate, z . The main advantage of this approach is that it can calculate the three-dimensional stresses and displacements while the coordinate system remains in two dimensions, thus make it ideally suited for the finite element analysis. Detailed description on the finite element formulation is given by Bang & Shen (1983). The special characterization of the generalized plane strain finite element analysis needs to be ernphasized. Conventional two-dimensional plane strain finite element analysis requires the size of the continuum element stiffness matrix to be 8 x 8 (i.e., two displacements along the x and y coordinates at each of the four nodes of a linear isoparametric quadrilateral element). Conversely, truly three-dimensional finite element analysis requires an element stiffness matrix size of 24 x 24 for a linear isoparametric brick element because three displacements exist at each of the eight nodes. The generalized plane strain
1 INTRODUCTION The benefits of utilizing soil nailing walls in maintaining stable excavations and slopes are well known among geotechnical engineers. The soil nailing walls typically utilize shotcrete facing to prevent erosion of the backfill soil, to produce more uniform outward deformation of the wall, and for esthetic seasons. To properly model the effect of the shotcrete facing wall on the behavior of a soil nailing wall, an existing generalized plane strain finite element method of solution (Bang & Shen 1983) has been expanded by including membrane elements to simulate the soil nailing facing walls. 2 GENERALIZED PLANE STRAIN When the principal axis of the material orthotropy does not match with the out-of-plane coordinate in plane strain, displacements along the plane strain disection exist. In such occasions, truly threedimensional solution methods may have to be used. However, if the angle between the principal axis of the material orthotsopy and the out-of-plane coordinate remains constant, the out-of-plane displacements will be the same at all cross-sections perpendicular to the out-of-plane coordinate. In other words, the in-plane displacements do not depend on the out-of-plane coordinate. This concept was first introduced by Wittke (1 975) for a study of very long tunnels with periodically repeated inclined layers of geological materials. Assuming that the out-of-plane displacements are dependent only upon the in-plane coordinates, one can analyze cestain three dimensional problems by a 64 I
finite element analysis requires an element stiffness matrix size of 12 x 12 (i.e., three displacements at each of the four nodes of a linear isoparametric quadrilateral element). A smaller element stiffness matrix is always desirable because the majority of the computational effort in the finite element analysis comes from the solution of simultaneous equations. In addition, the generalized plane strain approach utilizes two-dimensional finite element grid, which makes the input preparation much easier. A comparison has been made with the results obtained from a truly three-dimensional analysis to illustrate the effectiveness of the generalized plane strain finite element method of analysis (Bang & Hwang 1988; Bang & Yeon 1990). Results indicate that the generalized plane strain finite element approach can successfully describe the three dimensional behaviors, including the out-of-plane displacements and stresses, with very little loss in accuracy.
2 E~
2
Fiui
=
=
E
Y
i=l
C
Givi
i-1
2
where
In matrix form, this can be written as
(4) where { E } = strain vector; [B]=first derivative of the shape function; and {u}=displacement vector. Using the natural coordinate system for the membrane elements,
3 FACING WALL SIMULATION
The shotcrete facing as part of the soil nailing wall is relatively thin and flexible, and therefore can be modeled effectively by membrane elements. It is much easier and economical to include the membrane elements in the finite element formulation, since no separate elements for the shotcrete facing are necessary in the analysis. The effect of the shotcrete facing can be directly added to the element stiffness matrix of the composite continuum element adjacent to the shotcrete facing. Using a linear approximation for the displacements, the three displacements at each node of the membrane element can be approximated as
x = x,
+ 0.5e
where ;=local coordinate along the axis of the membrane element; and xo=global x coordinate to the center of the membrane element. Therefore, u , = -[(U2. I
1 1 NI = - ( l + e ) , N 2 =-(1-e) 2 2 where e=natural coordinate that varies from -1 at one end to +1 at the other end of the membrane element. From the definitions of strains and from Eq. 1,
642
+ulq)+e(u,e -ul,)]
2 1 1 = -[(w2, + wl,)+e(w2r- w19)] 2
1.1,
where u l $ is the x’ directional displacement of node I and so on. The displacements in the local coordinates (x ,y ,z ) need to be transformed to the global coordinates by an angle 0 , i.e.,, the angle between the global x axis and the local x axis. Its effect can be added to the adjacent continuum element (soil element) by using the following relationship. ~
where ui,vi,wi=approximate ithnodal displacements along the x, y, and z coordinates; and Ni=first order shape function in natural coordinate system. The natural coordinate system is a local coordinate system that permits the specification of a point within the membrane element by a dimensionless quantity whose absolute magnitude is less than or equal to 1.O. For membrane elements,
(5)
where u,v,w are the displacements in the global coordinates. Substituting these displacement expressions into the definitions of strains, one can obtain
(8) Substituting Eq. 7 into Eq. 8 yields 1 L
I = - [ c o s ~ ( u ~- u I ) - s i n ~ ( v 2-vf)]
"
Substituting the strains into the element virtual internal energy and differentiating the element virtual internal energy with respect to the nodal unknowns yields the element stiffness matrix:
tics. Therefore, this approach can only be applied when there are many reinforcements repeated in both horizontal and vertical directions. In this approach, the average values of the stresses distributed over the cell face are equal to stresses in the equivalent composite material, and the average values of the strains for the cell are those of the composite. Desired composite properties can be calculated from a detailed consideration of the behavior of the unit cell. Because the shotcrete facing is reinforced with small diameter re-bars or wiremesh in both directions, it is reasonable to assume that the reinforcements have minimal resistance against shear. Therefore,
where A,= cross-sectional area of the shotcrete in the unit cell; AT is the total cross-sectional area; and Gc is the shear modulus of the shotcrete.
from the constitutive relationship one can obtain
t where S,, = - C l l c o s 2 8 ; L t
t S,, = L C , 1 c o s 8 s i n 8 ;
where E, = elastic modulus of the shotcrete; and v, = the Poisson's ratio of the shotcrete.
The coefficient C,, can be obtained from the equilibrium of forces along the x-axis, i.e.,
t
S2, = -C, sin2 8 ; and S,, = -C33 . L L The coefficients Cijare obtained from the constitutive relationship in two dimensions for orthotropic materials, i.e.,
or
+
ATo,, = A,sE , s ~ x eACE,
1-v,
,
O x ' - EsA.7 - - A, Therefore, C1 = -
Ec (17) 1-v: where E,s = modulus of elasticity of re-bars; and A,=cross-sectional area of the re-bars in the unit cell. Eqs. 14 and 17 complete the development of the element stiffness matrix of the unit cell describing the behavior of the shotcrete facing reinforced with re-bars or wiremesh. Because no separate element is needed to represent the effect of the membrane elements, the element stiffness matrix of the membrane element can be added directly to the element stiffness matrix of the continuum element adjacent to the shotcrete facEx
To completely define the element stiffness matrix of the membrane element, coefficients C,, and C3, need to be redefined to include the effect of the reinforcement within the shotcrete facing. The concept of the unit cell has been introduced for this purpose. This concept expresses the orthotropic composite material properties as a function of the properties of each of the constituent materials, i.e., the shotcrete, the reinforcement, and their geometric arrangements. A unit cell is an isolated small unit of the material that completely exhibits its composite characteris643
AT
AT
ing. The input data and the basic logic for the generalized plane strain approach remain unchanged. However, since the sizes of the element stiffness matrices of the continuum element and the membrane element are 12x12 (3 unknown at each of the 4 nodes) and 6x6 (3 unknowns at each of the 2 nodes), respectively, special care must be taken to ensure that the matrices of the membrane elements are added properly.
table, increases in reinforcement from 5 % to 10 % to 20 % result in 4.5 % and 9.1 % additional reductions, respectively, in the maximum horizontal deformation under a surcharge of 23.94 Wa. Less reductions result with a heavier surcharge. Similar results have been observed with soils #2 and #3. Figure 2 shows the horizontal deformations along the entire height of the wall with and without the shotcrete facing. The amount of reinforcement is 5% Table I . Hyperbolic soil properties
4 PARAMETRIC STUDY
Y
A preliminary analytical parametric study was conducted using the finite element method of analysis described previously to investigate the effects of the thickness of the shotcrete facing, the percent reinforcement in the facing, and the skew angle of the reinforcement on the overall performance of the soil nailing wall. For simplicity, the study was limited to a vertical soil nailing wall of constant height (5.5 m) with all nails installed horizontally, having the same horizontal spacing and skew angle to the x-y plane. The soil elements are isoparametric, quadrilateral elements. The nail dimensions include tension rebars of 2.02 cm’ in cross-section and 4.39 m in length, and a spacing of 0.91 m in both the horizontal and vertical directions. The behavior of the soil was characterized by the hyperbolic soil constitutive relationship (Boscardin, et al. 1990). The detailed hyperbolic properties of the soil are given in Table 1. The modulus of elasticity of the shotcrete facing is 13.79 MPa and that of the reinforcement is 206.85 MPa. Figure 1 shows the results of the analysis on the maximum horizontal wall deformation with various shotcrete facing thicknesses and percent reinforcements in the shotcrete facing. As expected, the thicker the lining, the larger the reduction in the maximum horizontal deformation results. The rate of decrement in deformation, however, decreases as the thickness increases. At a certain shotcrete facing thickness, the rate of decrement virtually disappears, indicating that there is a limiting thickness for the shotcrete facing which does not provide any additional significant confinement to the horizontal deformation of the soil nailing walls. Figure 1 also shows the effect of the percent reinforcement in the facing. It shows that all curves with different percent reinforcements are very similar in shape, are very closely spaced together, and tend to overlap or merge at high thickness values. It is obvious from the figure that the effect of the amount of reinforcement is not as significant as the shotcrete facing thickness. The percent reinforcement used in this study ranged from 5 % to 20 %. Table 2 shows a summary of the reduction in the maximum horizontal deformation with increase in reinforcement for soil # l . As can be seen from the 644
Y
v
x
v
x
Y
Description Loading Modulu\ Modulus Exponent Failure Ratio Cohesion (kPa) Friction Angle (deg) Poisson’s Ratio
A4
Unit W~igh~_(kN/m’)
^ I
soil #i CL 60 0.45 0.7 4.79 30 0.3 0
!! .ss
soil #2 SM 300 0.2s 0.7 0 32 03 4 19.64
* x x
SGjl#3
sw
450 04 0.7 0 40 0.3 7 22.78-
I V I &
Figure 1. Maximum wall horizontal deformation. Table 2 Percent reduction of maximum horizontal wall deformation with different percent reinforcement (Soil # I ) *
Su;charge (kPa) 23 94 47 88
5% Reinforce-meiit 66 61
10 % Reinforce-ment 63 5.9
Figure 2. Horizontal wall deformation
20 % ‘ Reinforce-ment 6.0 “5.8 * “
x
with the surcharge of 47.88 Wa. It is obvious from the figure that the shotcrete facing fulfills one of its intended functions by providing more or less uniform deformations from the top to the bottom of the wall. In general, the parabolically-shaped horizontal deformation pattern of the wall without a shotcrete facing becomes nearly uniform when the shotcrete facing is applied. Table 3 indicates the deformation characteristics of the wall with soil #1 and surcharge of 47.88 kPa. As can be seen from the table, the application of the shotcrete facing drastically reduces the differences in the wall horizontal deformations. The maximum differences in the horizontal deformation as a fraction of the wall height are 0.01 % and 3.64 % for the wall with facing and the wall without facing, respectively. Figure 3 shows how the maximum out-of-plane deformation of the soil nailing wall is affected by the presence of the shotcrete facing with different surcharge magnitudes. It is noted that the lengths of the reinforcement varied as the skew angle changed. The in-plane projectional lengths of the reinforcements are, however, remains the same. In general, the maximum out-of-plane deformation occurs at the skew angle of approximately 30 degrees. As expected, the presence of the shotcrete facing dramatically reduces the maximum out-ofplane deformation. Table 3 . Horizontal deformation characteristics with soil #I
5 CONCLUSIONS A finite element formulation to describe the effect of the shotcrete facing has been developed, utilizing membrane elements in the generalized plane strain approach. The modified formulation has been used to perform an analytical parametric study to investigate the effects of the shotcrete facing on the behaviors of the soil nailing walls. The results of the analysis and the comparisons of the behaviors between the reinforced walls with and without the shotcrete facing, as well as the major findings from the parametric study, are described. Following are the major findings from this study.
1 . The thickness of the shotcrete facing is very influential for the behavior of the soil nailing walls. 2.The amount of the reinforcement in the shotcrete facing has a lesser effect on the reduction in deformations. 3. There appears to be a limiting thickness of the shotcrete facing. The shotcrete facing thickness greater than this limiting value does not provide any additional reduction in deformations. 4. The shotcrete facing provides more uniform outward and out-of-plane deformations. 5. The maximum out-of-plane deformation occurs at the skew angle of approximately 30 degrees when the projectional length of the reinforcements is kept constant. The results presented have been obtained without any experimental verification. Therefore, it is essential to validate the analytical solution method through experimental studies and possibly field instrumentation.
Figure 3. Maximum out-of-plane deformation.
645
REFERENCES
,x
Bang, S. & S.I. Hwang 1988. Anal sis of Skew Reinforcement System. Proceedings of the 24 Annual Engineering Geology and Soil Engineering Syinposiunz:245-254. Bang, S. & C.K. Shen 1983. Soil Reinforcement in Soft Ground Tunneling. Final Report to US Department of Transportation, Report No. DOT/RSPA/DMA-50/83/15. Bang, S. & H. Yeon 1990. Analysis of Retaining Structures with Skew Reinforcement. Transportation Research Record. 1288: 152-157.
646
Boscardin, M.D., E.T. Selig, R.S. Lin,. & G.R. Yang 1990. Hyperbolic Parameters for Compacted Soils. Journal of Geotechncial Engineering, ASCE. Vol. 116, No.1: 88-103. Timoshenko, S.P. & J.N. Goodier 1970. Theory of Elasticity. Third Ed. Mc-Graw Hill Book Co. Wittke, W. 1975. Proceedings of International Symposium on Numerical Methods in Soil and Rock Mechanics. Karlsruhe Univ.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swefs 6: Zeitlinger, ISBN 90 2651 863 3
Reinforcing mechanism of anchors in slopes and numerical verification by FEM Fei Cai & Keizo Ugai Department of Civil Engineering, Gunma University, Kiiyu, Gunma 3 76-8515, Japan
ABSTRACT: The reinforcing mechanism of anchors in slopes is studied. The effects of anchors on the safety factor of a slope are also predicted by 3 0 elasto-plastic shear strength reduction FEM (SSRFEM). The soil-anchor interaction is simulated with zero-thickness elasto-plastic interface elements. The results are compared with those obtained by Bishop’s simplified method, where the anchors are considered only to supply an axial tension. The safety factor of SSRFEM is consistent with that of Bishop’s simplified method. SSRFEM shows that the prestressed load has little influence on the safety factor. The better stabilizing effects can be obtained when the anchors are installed with a small angle between the anchor and the horizontal direction.
1 INTRODUCTION
angle on the safety factor of slopes are numerically studied. Some useful conclusions are obtained.
The permanent grouted anchors have been extensively used to provide vertical and lateral support for natural and engineered structures during the past 50 years. The end type of anchorage has been used to stabilize the dangerous slopes, and as a preventive measure in stable slopes because of its significant technical advantages resulting in substantial cost savings and reduced construction period. The basic design concept consists of transferring the resisting tensile forces generated in the anchors into the ground through the friction mobilized at their interfaces. The loads are usually developed by the anchorage of the tendon within the soil mass and tensioning at the surface against a bearing plate or concrete pad. Some of the successful applications of this technique to stabilizing slopes have been reported (Tan et al. 1985, Hashimoto et al. 1986, Corona 1996). The stability of slopes reinforced with anchors is commonly assessed using LEM, assuming that the anchor only supplies an axial tension to reinforce the slope. In this research, the 3D SSRFEM is used to evaluate the reinforcing mechanism of anchors. The soil-anchor interaction is simulated by zero-thickness 3D interface elements. The results of SSRFEM are compared with those of Bishop’s simplified method. The effects of the prestressed load, direction
2 ANALYSIS METHODS 2.1 Limit equilibrium nzethhod Bishop’s simplified method of slip circle analysis is employed to determine the safety factor of slopes reinforced with anchors. The safety factor, F,, of slopes reinforced with anchors is given by
The conventional vertical and novel normal approach is used for i = 1 and 2, respectively. is the effective stabilizing force supplied by anchor, and tangential to the base plane of the slice with the anchor located, and is given by
e>,
=
F,, ~ [ c o s (+a 8 )+ sin(a + 8 )tan $,,,] 1 + tan a tan $,,,
(2)
1
sin(a + 8) tan$,,, 1+ tan a tan $,,,
647
(3)
where P is an axial resisting force supplied by anchors, 8 is the angle between the anchor and the horizontal direction, @,,,,is the mobilized friction angle on the slip surface. The simplex reflection technique is used for 10cating the critical slip circle that has the lowest safety factor of a slope. When the slope is stabilized with the anchors, the critical slip surface is found with the safety factor calculated by Equation 1. Thus a smaller safety factor can be obtained than that considering the effect of the anchors with the original critical slip surface without anchors. 2.2 Shear strength reduction FEM (SSRFEM) SSRFEM has been gradually used to analyze the stability of slope in 2D and 3D situations. The numerical comparison has shown that SSRFEM is a reliable and robust method for assessing the safety factor of slopes and the corresponding critical slip surface. One of the main advantages of SSRFEM is that the safety factor emerges naturally from the analysis without the user having to commit to any particular form of the mechanism a priori (Griffiths & Lane 1999). SSRFEM has been successfully used to determine the safety factor of the slopes improved with piles (Cai & Ugai 2000). SSRFEM can evaluate the stability of such slopes under a general frame for slopes, where the soil-structure interaction is considered with zero-thickness interface elements. The process of SSRFEM to predict the safety factor of a slope reinforced with anchors is simply described here. When the shear strength reduction factor, F, is less than unity, the slope completely consists of the soil. When F reaches unity, the materials of the solid elements within the extent of the anchor are changed from the soil into the tendon or the grouted body according to the position of the elements. The materials of the interface elements are simultaneously changed from the soil-soil interface into the soil-tendon interface or the soil-grout interface. The stress in the interface keeps unchanged when the materials are changed. This implies that the grouting pressure is the same as the normal stress in the interface, induced by the self-weight of soil.
When F is unity, a prestressed load is applied. The prestressed load includes a drag load, applied at the anchor head, and a push load, applied on a circular rigid plate on the slope surface. The value of the two loads is the same, but their directions are opposite. The rigid plate is simulated by the identical deflections of the nodal points on the slope surface under the rigid plate. After the deflection under the prestressed load is finished, the tendon head is fixed to the rigid plate on the slope surface. Therefore, the deflection of the tendon head is the same as that of the rigid plate during the shear strength reduction from unity to the ultimate value. The detailed process for SSRFEM and properties of the interface element can be seen elsewhere (Cai & Ugai 2000). 3 RESULTS AND DISCUSSIONS 3.1 Model slope An idealized slope with a height of 8m, a gradient of 1V:1H, and a ground thickness of 10m is analyzed with a 3D FE mesh, as shown in Figure 1. The diameters of the tendon and grouted body are 32mm and 90mm, respectively, and their lengths are 6m and 6m, respectively. The soil-tendon interface is assumed to be smooth, and the shear strength to be zero (Table 1). The shear strength of the soil-grout interface is assumed to be the same as the soil. Young's modulus of the grouted body is assumed to
Figure I . Model slope and schematic FE mesh.
Table 1. Material parameters. Parameter
Soil
Young's modulus, E (MPa) Poisson's ratio, v(-) Unit weight, y(kNlm3) Cohesion, c (kPa) Friction angle, 4, (") Dilatancy angle, y/(")
200 0.25 20.0 12.0 20.0 0.0
Soil-tendon interface 200 0.25
Soil-grouted body interface 200 0.25
0.0 I 0.0 0.0
12.0 20.0 0.0
648
Tendon
Grouted body
210000 0.30
24000 0.20
tension, which is induced by the prestressed load. When F=1.21, and reaches the ultimate value, the axial tension is the ultimate axial tension, P,. The ultimate axial tension includes the two parts, induced by the prestressed load, and by the shear strength reduction from unity to the ultimate value, as shown by the solid h e in Figure 3. It is indicated that the
be the same as the reinforced concrete. The rigid circular plate on the slope surface is with a diameter of 3Ocm. The anchors are installed with a horizontal distance between the slope toe and the anchor head, Lx=4m, and a center-to-center spacing, D1=1.5m, unless otherwise stated. The material parameters of the soil, interface, soil-grout interface, tendon, and grouted body are shown in Table 1. When the slope is not reinforced with anchors, SSRFEM gives a safety factor of 1.09, comparing well with 1.084 of Bishop’s simplified method. The failure mechanism of SSRFEM, agrees well with the critical slip circle of Bishop’s simplified method. 3.2 Effect of prestressed load When the anchors are installed with Lx=4m, 8 = 15”, and D1=1.5m, SSRFEM indicates that the safety factor of the stabilized slope is not influenced by the prestressed load (Table 2). This is because the ultimate axial tension is almost the same for various prestressed loads. Table 2 shows that the safety factor of the stabilized slope, predicted by SSRFEM, is almost the same as that by Bishop’s simplified method. The reinforcing effect of anchors can only come from the increase in the shearing resistance along the slip surface in SSRFEM, and then FEM calculation diverges under a larger safety factor. Thus SSRFEM gives a larger safety factor of the slope reinforced with anchors. Therefore, the reinforcing mechanism in SSRFEM is identical to that in Bishop’s simplified method. The failure mechanism in SSRFEM is represented using the vectors of the nodal displacement increments induced by the shear strength reduction. The failure mechanism in Bishop’s simplified method is shown in Figure 2 by the critical slip circle, which has the lowest safety factor of the slope reinforced with anchors. Figure 2 shows that the failure mechanisms, predicted with SSRFEM and LEM, agree well with each other. The prestressed load is assumed to be zero in the following analyses because it has little influence on the safety factor of a slope. Figure 3 shows the influence of the prestressed load, P,, on the axial deflection, cross deflection, initial axial tension, and ultimate axial tension of the anchor at the axial distance, L, from the anchor head. When F=l.OO, the axial tension is the initial axial Table 2 Safety factor for various prestressed loads. Prestressed load Pi 0 25 35 45
Safety factor SSRFEM Bishop I .21 1.21 1.21 1.21
1.195 1.196 1.198 1.197
Ultimate axial tension P, (W 59.30 60.09 60.75 60.66 Figure 3. Response of anchor for various prestressed loads.
649
prestressed load can decrease the axial and cross deflections of the anchor when the slope failure takes place. This implies that the prestressed load can decrease the movement of the slope when the slope failure takes place. The axial tension in the free tendon is the same along its axial direction because of the smooth soil-tendon interfaces. Therefore, the resisting force of the anchor completely comes from the friction between the grouted body and the soil. The distribution of the ultimate axial tension along its axial direction is almost the same for various prestressed loads, although the initial axial tension increases with the increase in the prestressed load.
of anchors is the increase in the shearing resistance along the slip surface, and completely dependent on the effective stabilizing force under various direction angles of the anchor. 3.4 EfSect of anchor position When the anchors are installed with 0 = 15" and D1=1.5m, the influence of the anchor position, Lx,as shown in Figure 1, on the safety factor of a slope stabilized with anchors is shown in Figure 6. The safety factor of SSRFEM is consistent with that of
3.3 Effect of direction angle When the anchors are installed with Lx=4.0m and D1=1.5m, the direction angle of the anchor has significant influence on the safety factor of the slope reinforced with anchors, as shown in Figure 4. Despite the fact that the relative error between the safety factor of SSRFEM and F,, is smaller than 3%, F s is ~ better consistent with that of SSRFEM. Figure 4 indicates that there is an optimized direction angle of the anchor for the safety factor of the slope reinforced with anchors. The influence of the direction angle of the anchor on the safety factor comes from its influence on the ultimate axial tension, and the angle between the anchor and the slip surface. For the anchors installed horizontally, the normal stress in the soil-grout interface is smaller because of the smaller vertical soil pressure on the grouted body. The normal stress in the soil-grout interface increases with the direction angle due to the selfweight of soil. The experimental results also indicated that the ultimate reinforcing effects are mobilized under a more shearing displacement for a larger angle between the reinforcement and the sliding surface when the angle is less than 90" (Edirisinghe et al. 1996). Figure 5 indicates that the direction angle of anchor has influences on the ultimate axial tension, ?, , and on the angle between the anchor and the critical slip surface, a + 8 . Such influences can be indicated comprehensively with the effective stabilizing force, as indicated in Figure 5. The larger the direction angle of the anchor, the smaller , the ratio of the effective stabilizing force to the ultimate axial tension, because the angle, a + 8 , increases with the increase in the direction angle of the anchor. Comparing Figures 4 and 5, one knows that the relationship between the safety factors versus direction angle is identical to the relationship between the effective stabilizing forces versus direction angle of the anchor. This suggests that the reinforcing effect
c/c,
650
Figure 6. Safety factor vs. anchor position.
LEM, and generally the consistency is better between the safety factor of SSRFEM and Fs2. When the position is closer to the top of the slope, the ultimate axial tension in the anchor is smaller, though the shear strength of the soil-anchor interface has been mobilized. The reason lies in that the normal stress in the soil-anchor interface is smaller because of the smaller vertical soil pressure induced by the self-weight of soil. The numerical results of SSRFEM and LEM show that the positions of the anchor have significant influence on the safety factor of the slope even when the ultimate axial tension is the same. For example, the ultimate axial tension of Lx=1.5m is almost the same as that of Lx=5.5m, and the ultimate axial tension of Lx=2.5m is almost identical to that of Lx=4.Om, as shown in Figure 7. However, when the anchors are closer to the toe of the slope, the effective stabilizing force is larger because the angle between the anchor and the slip surface is smaller. Therefore, the safety factor of the slope stabilized with anchors is larger when the anchors are closer to the toe of the slope. Comparing Figures 6 and 7, one knows that the relationship between the safety factors versus direction angle is identical to the relationship between the effective stabilizing forces versus direction angle of the anchor. The results suggest that the reinforcing effect of anchors is the increase in the shearing resistance along the slip surface, and dependent on the effective stabilizing force under various anchor positions. The ultimate axial tension under Lx=1.5 is smaller than those under Lx=2.5m and Lx=4.0m although the normal stress in the soil-grout interface under Lx=1.5m should be larger than those under Lx=2.5m and Lx=4.0m. This is because the slip surface under Lx= 1.5m is significantly different from those under Lx > 1.5m, as shown in Figure 8.
Figure 9. Safety factor vs. anchor spacing.
chors, D1, as shown in Figure 1, on the safety factor of the slope stabilized with anchors is shown in Figure 9. It is indicated once again that the safety factor of SSRFEM is better consistent with FQ. The safety factor approximately linearly increases with the anchor spacing becoming smaller and smaller because the ultimate axial tension and the failure mechanism are almost the same regardless of the anchor spacing.
3.5 Effect of anchor spacing When the anchors are installed with Lx=4.0m and 8 = 15", the effect of the spacing between the an-,
4 CONCLUSIONS The reinforcing mechanism of anchors in slopes is analyzed in this research. The 3D SSRFEM is used to evaluate the stability of a slope reinforced with anchors, and to verify the reinforcing mechanisms of anchors in slopes. The soil-anchor interaction is simulated by the 3D zero-thickness elasto-plastic interface elements. The numerical results are compared with those of Bishop's simplified method. Based on the numerical results, the following conclusions are reached: (1) The reinforcing mechanisms of anchors in slopes are to increase the shearing resistance on the slip surface due to the axial tension.
Figure 7. Axial tension and stabilizing force vs. anchor position.
65 1
(2) SSRFEM shows that the prestressed load has little influence on the safety factor though it can decrease the movement of the slope stabilized with anchors when the failure takes place. (3) The safety factor of the proposed LEM is better consistent with that of SSRFEM. This consistency is independent of the prestressed load, direction angle, position, spacing, and shear strength of the soil-grout interface for the analyzed homogeneous slope. (4) The better stabilizing effects can be obtained when the anchors are installed with the angle between the anchor and the horizontal direction within the range of 7.5"-22.5" for the analyzed slope, and slightly closer to the slope toe.
652
REFERENCES Cai, F. & K. Ugai 2000. Numerical analysis of the stability of slope reinforced with piles. Soils and foundation.^, 40( 1): 73-84. Corona, E.P. 1996. Stabilization of excavated slopes in basalt', Proc. 7th hit. Symp. on Landslides: 1771- 1776. Edirisinghe, J., N. Yagi, M. Enoki, R. Yatabe & G. Ohashi 1996. Reinforcing mechanisms and a stability analysis method for reinforced earth structures.Soils and Foundations, 36(3): 2 1-30. Grifiiths D.V. & P.A. Lane 1999. Slope stability analysis by finite elements. Geotechnique,49(3): 387-403. Hashimoto, I., K. Kawasaki & H. Kodera 1986. The case of landslide control by the dead-anchors. Proc. 21th Jupaiz National Con$ on Soil Mech. and Found. Engrg.: 1483-1486 (in Japanese). Tan, S.B., S.L. Tan, K.S. Yang & Y.K. Chin 1985. Soil improvement methods in Singapore. The 3rd International Geotechnicul Seminar: 249-272, Nanyang Technological Institute, Singapore.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Nailing a deep excavation in soft soil with jacked in pipe inclusions Cheang Wai Lum, Tan Siew Ann, Yong Kwet Yew & Ganeswara R. Dasari Department of Civil Engineering, National University of Singapore, Singapore
ABSTRACT: This paper briefly describes the use of jacked-in pipes in soft clayey soils. The pipe reinforcement functions as a temporary internal support system, working in combination with contiguous bored pile walls for the excavation of a 3-1/2 storey deep car park basement. The pipes were open-ended and it was significantly plugged when it was jacked to its design length. It was found that nail set-up was a contributing factor in the increased in axial resistance of the nail inclusions due to compaction and reconsolidation of the disturbed soil. This hybrid retaining system is a departure from the conventional technique of soil nailing, in which a stiff cast in-situ wall was used in conjunction with soil nails. Geotechnical field instrumentation data indicated that the contiguous bored pile wall exhibits restrained cantilever deflection profiles and this can be attributed to the pipe nails in changing the strain field and arresting strains due to soil stress relief. Finite element analyses were performed and compared with instrumented field results in an attempt to study the possible working mechanisms. Parametric studies were made to investigate the interplay between nail, soil and structure. This project shows that soil nailing in conjunction with a stiff structural facing in soft clayey soils is possible, at least for short term and can be effectively used as temporary support in deep excavation. excavation project in Malaysia. The first successful use of this technique was reported by Cheang et.a1.,1999 and subsequently by Liew et al., 2000. The later case history is of interest as this technique was used in soft clayey soil conditions and it is the subject and focus of another paper published in this conference by the second author of this paper (Tan et al. 2001).
1 INTRODUCTION Deep basement construction is becoming increasingly expensive, nullifying the justification for overconservative designs. One such solution where construction cost can be reduced is by employing fast and cost-effective deep basement construction techniques. This paper will highlight a hybrid construction method where the science and art of soil nailing was used in conjunction with an embedded cast insitu wall for the stabilisation of a deep excavation. Geotechnical field instrumentation has proven to be a valuable tool in assessing the validity of initial design assumptions and assessing the overall performance of this hybrid support system. Soil nailing is an in-situ soil reinforcement technique, which in the last three decades has been successfully used in Europe and more recently in Southeast Asia. To this extend, various installation methods have been invented and innovated all with the intention of improving the versatility and adaptability of this technique in difficult ground conditions. The method of reinforcing a mass of soil with pipe inclusions by jacking was first conceived in an
2 CASE HISTORY AND FIELD DATA 2.1 The jack-in technique
It has been said that necessity is the mother of all inventions and how true this statement holds as the aforementioned method of nail installation was conceived by a group of engineers faced with the problem of stabilising a deep excavation in soft soil conditions and high water table. The conventional technique of nail installation was initially used, but due to very difficult site conditions of soft ground and collapsing boreholes this method was discarded and it was envisioned that a method which does not involves drill-and-grout and speedier nail construction will stand a better chance in such hostile conditions.
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Figure 1. Contiguous bored pile wall supported by nails.
Figure 2a. Jack-in in progress.
Figure 2b. Schematic of jack-in rig working in tandem
Figure 3a. In-situ vane shear at S.Sm (p'=45kPa)
Figure 3b. In-situ vane shear at 9.5m (p'=76kPa)
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Therefore the jack-in technique or better known as jack-in anchors' was invented. This method of installation is relatively cheap compared to other modes of installation as the passive nail inclusions are made of easily obtainable mild steel pipes. In the USJ-19 project pipes of 12m lengths with an outer and inner diameter of 150mm and 140mm respectively were used. The installation rig as shown in Figure 1 is constructed from a single length 'H' structural steel section and welded with 'U' shape stiffener plates located at strategic intervals depending on the maximum stroke of a single jack. These plates will function as reaction devices during the jacking process. Due to its low fabricating cost, multiple jacking rigs can be mobilised, therefore the amount of time a particular excavation phase is left unsupported will be decreased and this directly will decrease the construction period. In the USJ- 19 deep excavation (Cheang et al., 1999) more than ten jacking rigs were employed. A significant advantage of the jack-in technique is that the during the penetration process, the pressure of the jacks and the time needed to insert a particular length of nail (the nails were marked in lm
spacing) can be monitored. This enables the strength and stiffness of the in-situ soil to be roughly estimated and ascertain. With further analytical investigation, the strength and stiffness of the soil can be correlated to jack pressure and penetration rate. 2.2 Jack-in nail inclusions in soft tropical residual soil - The eflects of installation and subsequent consolidation The interfacial resistance of soil nails is govern by the magnitude of effective radial stress. In-situ pullout out test perform right after the installation process yielded very low pullout capacity of about 25kN on average which corresponds to a unit resistance of Skpa with variation occurring depending on the local soil condition and overburden pressure. In-situ vane shear test results (Figures 3a and b) indicated that the normalised intact shear strength (Su/p') as shown in Figure 4, was about 0.15 (Average Su =20Wa, 28 kPa) and the remoulded shear Range =12 strength as shown in Figure 5, was on average Skpa (4.8 -6.0 kPa). This indicated that the soil stress state at the periphery of the nail face after installa
-
Figure 5. Remoulded shear strength.
' Jack-in Anchors is a patented technology by Specialist
Grouting Engineers. Sdn. Bhd.
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tion is residual as the installation process causes intense shearing at that region and thus taking the soil strength to its residual value. Subsequent results from pullout tests conducted a week after installation yielded higher resistance of 120 kN on average. This leads to the following unit skin resistance of 23 Wa and therefore an increase of 4 fold in the shear resistance. Current research findings in NUS-CSGE is indicating that the predominant cause of nail set-up is due to the time dependent increase in the radial effective stress of an improved annular soil zone adjacent to the circumferential surface of the nail. This increased in radial effective stresses is due to soil disturbance caused by the displacement of soil during the installation process. This condition is very similar to set-up of piles driven in clays and hence many parallels can be drawn from this since much research work has been conducted on this subject (Randolph et al., 1979, Whittle, 1998). Plugging of the open-ended pipe was detected and this inay have caused further soil disturbance as the inclusion penetrates into the soft soil mass. This points to the fact that further interfacial resistance can be mobilised if the nail inclusions were closed-ended (Whittle, 2000) since more disturbance leading to greater excess pore pressures will be incurred.
Figure 6. Finite element mesh consisting of 8-node quadrilateral reduced intcgration pore pressure elements.
3 FINITE ELEMENT ANALYSIS Finite element analyses were conducted using ABAQUS version 6.1 non-linear coupled finite element code. When the nail inclusion is jack-in, assuming that the pipe nail is close-ended, it must displaced initially a volume of soil equal to the volume of the inclusion. At small penetrations, heaving occurs at the exposed ground surface. At greater penetration depths, the soil is displaced predominantly outwards in the radial direction. This has led to the idea that a realistic approximation of the complex installation process being modelled as the expansion of a cylindrical cavity with a final radius equal to radius of the nail inclusion. Analyses of stress changes due to cavity expansion of a cylindrical and subsequent consolidation of the soil have been presented by Randolph et.al. (1979). Figure 6 briefly illustrates the boundary mesh of the problem. The soil behaviour was modelled using an enhanced Modified Cam Clay Model where the elastic region is non-linear. Figure 7 illustrates the distribution of total pore pressure right after the installation based on cavity expansion. Around the periphery of the nail an increase of 50 kPa in excess pore-water
Figure 7. Plane strain finite element analysis: schematic shows the distribution of total pore pressure under 3-D view after jackin installalion by cavity expansion.
Figure 8. Build-up of pore pressure, decrement in effective mean pressure during expansion of cavity to final radius of 75mm .
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pore pressure distribution may be obtained from the change in mean total stress. Adopting Tresca's yield criterion, the initial excess pore pressure in the plastic region can be shown to be: U0
= 2S, In (Wr) for r 5 R
('1)
Where the radius of the plastic zone is given by R = ( G/S3°.5 a
(2)
Beyond the plastic boundary, the initial excess pore pressures are zero (Figure 9). The above analytical equation is plotted into its graphical representation (Figure 10) for a given cavity expansion of 75mm with various values of shear modulus (G). For this case study, with a soil shear modulus of SOOkPa, the estimated excess pore pressure after installation will be 6OkPa. Re-equalisation will cause an increase in the mean effective pressure by an amount equivalent to the magnitude of excess pore pressure being dissipated. Judging from field results obtained from in-situ vane tests, clay sensitivity (St) values ranges from slightly to medium sensitivity. It is speculated that thixotropy (Mitchell, 1960) may be a significant mechanism that may caused further increased in nail pull-out capacity. However further investigation will be required to assess the aforementioned hypotheses. Figure 10: Distribution of excess pore pressure along the radial direction for a given expansion of 0.075m.
4 CONCLUSION
pressure was detected and it is reported in Figure 8. With subsequent reconsolidation (equalisation), excess pore pressure will be dissipated and this will give rise to the increased in the mean effective pressure around the periphery of the nail, hence the increased in pullout capacity. The increased in the mean effective pressure in this case corresponds roughly to 50 kPa and referring to Figure 4 again the increased in shear strength based on a normalised shear strength to mean effective pressure ratio of 0.15 will be 7.5kPa. The mean effective pressure used in the normalisation of the shear strength was calculated based on the coefficient of earth pressure at rest of unity (ko =1). Therefore if the coefficient of earth pressure was assumed to be 0.5, a larger value of S U P 'will be produced and with a unit increased in mean effective pressure (p') due to reequalisation the theoretical increased in shear strength will be greater. To predict the change of shaft capacity with time, Randolph and Wroth (1979) developed an analytical equation for radial consolidation of soil around a cylindrical cavity. The initial pore pressures are assumed to be those predicted by using a total stress cavity expansion analysis with the Tresca yield criterion (Yu, 2000). Simplistically, the initial excess 657
It has been observed that soil disturbance and compaction due to the jack-in installation method, soil nail set-up was the main mechanism which causes the increased in interfacial shear strength. This mechanism is a contrast to the conventional replacement method, whereby due to the soft soil conditions and the installation technique the increased in pullout capacity was not due to restrained dilatancy but the reequalisation of stresses. Research is still underway in CSGE-NUS to assess the behaviour of nail inclusions in soft soils. 5 ACKNOWLEDGEMENTS The authors would like to thank SGE Sdn. Bhd. For the provision of field data and support. Heartfelt thanks to Ir. Dr. Gue See Sew and the staff of G&P Geotechnical Perunding Sdn. Bhd. for their help. REFERENCES Cheang, W.L., Tan, S.A., Yong, K.Y., Gue, S.S., AW, H.C., Yu, H.T. & Liew, Y.L. 2000, Jacked-in pipe reinforcement of a deep excavation in soft soil, Field measurements in geomechanics, 3 I 1-318, Balkema.
Hibbit , Karlson & Sorenson, Inc. , Abaqus Version 6.1 2000 Liew, S S , Tan, Y.C. & Chen, C.S. 2000 Design, Installation and Performance of Jack-in Pipe Anchorage System for Temporary Retaining Structures, International Conference on Geotechnical and Geological Engineering. GeoEng 2000, Balkema. Mitchell, J.K. 1960, Fundamental aspects of thixotropy in soils, Journal of Soil Mechanics and Foundations Division, ASCE, VoI.86, No.SM3, 1952. Randolph, M.F., Carter, J.P. & Wroth, C.P.1979, Driven Piles in Clay-The Effects of Installation and Subsequent Consolidation, Geotechnique 29, No.4, 361 -393.
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Tan, S.A., Luo, S.Q. & Yong, K.Y. 1998, A Pullout tests of Soil Nail In Singapore Marine Clay, Second International Conference on Ground Improvement Techniques, 499-504. Tan, S.A., Luo, S.Q. & Yong, K.Y. 1999, Analysis of Soil Nail Lateral Interaction For the Design, Journal of The Institution of Structural Engineers Singapore, Vol. No.2,43-50. Whittle, A.J. ,2000, Personal Communication Whittle, A.J., Sutabutr, T., 1998, Prediction Of Pile Set-up in Clay, Transportation Research Record 1663, paper no 991552, 33-40. Yu, H.S., 2000, Cavity Expansion Methods in Geomechanics, Kluwer Academic Publishers, 181-1 85.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Study on reinforcement method for seismic slope stability T. Fukumasa Hyogo Prefecture, Rokko Land Conservation OfSice
H. Murakami Hyogo Prefecture, Agriculture and Forestry Department, Land Conservation Division
R. Nishihara Hyogo Public Co. of Parks and Forestry
H. Kimura, M. Yamaura & S. Razavi Darbar Dia Consultants Co., Ltd. ABSTRACT: In this paper, a new reinforcement method including rock bolts and rope net is used to stabilize the slopes during an earthquake. The behavior of a slope model, during cyclic loadings, is investigated by utilizing shake table tests and numerical simulation obtained from finite element analysis. Both experimental and numerical results show the efficiency of the reinforcement method on the seismic response of the slope by reducing the maximum displacement of the model, and preventing from rapid sliding failure. As well as good slope stability, the cost of construction in this method is cheaper than other methods because of no need to remove vegetation and to scrape the ground surface, which makes it efficient and ecological in practice.
1 INTRODUCTION
2 EXPERIMENTAL MODEL
The city of Kobe in JAPAN was severely damaged during the 1995 earthquake. The damage included land slides and slope failures. In a project, funded by Forestry Agency, the Hyogo Prefecture investigated the stability of slopes during an earthquake through several experimental and analytical procedures. The main purpose of the project was to determine the failure mechanism of the slopes during an earthquake and investigate a new method to stabilize them against relatively large earthquakes. In this new method, rock bolts and rope net are used as reinforcement to stabilize the slopes against sliding and fracture. As well as good slope stability, the new reinforcement method used in this project has some features in terms of economy and ecology. The cost of construction in this method is less than other methods. In order to set up and install the reinforcement, it is not necessary to remove the vegetation and to scrape the ground surface, which makes it efficient and ecological in practice. In this paper, a part of this project concerning experimental results and numerical analysis of the slopes with and without reinforcement is introduced. At first the experimental model and details of the reinforcement used in the shake table test are introduced and then the numerical finite element idealization is presented. Finally, the experimental and analytical results are compared with each other so that the role of rock bolts and rope net during cyclic loadings is investigated.
The model of a slope, used for the present experiment is shown in Figure 1 and the section of the slope model is presented in Figure 4. It has about 12.00m long, 3.00 m high and 5.00 m wide. The angle of slope is 45 degree. The thickness of the soil placed on the steel platform is 1.20 m. As shown in Figure 4, the experimental model consists of two parts. The first part, which is located at the right hand side of Figure 4, is the reinforced soil with rock bolts and rope net. The second part is the soil without reinforcement. This makes it possible to compare the experimental results of soil with and without reinforcement at the same time. Figures 2 and 3 show the two inclined surfaces of the model with and without reinforcement. As for the reinforced part, the heads of rock bolts are connected through a fiber mesh of rope net that has covered the surface of the slope. The bottoms of rock bolts are fixed to
Figure 1. Experimental model of slope.
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the foundation. The longitudinal ropes are fixed to the platform at the bottom and to the centerboard. The soil used for the model is weathered granite. The rock bolts are made of aluminum with a length of 1200 mm and a diameter of 5.7mm. The spacing between the rock bolts is 40 cm. Rope net is made of polyester with a diameter of 0.8 mm and a mesh size of 10xlOcm. The shake tests start with the maximum
horizontal acceleration of 150 Gal and repeat with an increment of 50 Gal for each step until sliding failure occur. Each step takes about 20 seconds. The input motion consists of sin waves with the frequency of 3 Hz. Figure 5 shows the sample of input motion with the maximum acceleration of 250 Gal. The acceleration starts from zero and gradually increases to a maximum specified value in 5 seconds. Then it continues for 10 more seconds with the maximum acceleration and finally diminishes to zero in 5 seconds.
3 FINITE ELEMENT IDEALIZATION
3.1 Model description The finite element model of the slope and reinforcement are shown in Figures 6, 7. Soils are modeled with two-dimensional solid elements. Rock bolts and rope nets are modeled using threedimensional beam and two-dimensional truss elements. Since the rope net can not tolerate axial compression forces, the material for the rope net has no stiffness in compression.
3.2 Basic analysis parainetels The basic analysis parameters used for the soil material are shown in Table 1. These values have been obtained from laboratory tests. The nonlinear behavior of the soil is modeled using the theory of multi-
Figure 5. Horizontal input motion at the shake table.
Figure 7. Finite element model of reinforcement.
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Table 1. Soil material property in finite element analysis.
No. 1
2 3 4 5 6
Material property of soil Parameter Ouantitv Shear modulus 2.35 E+07 Bulk modulus 7.63E+07 Weight density 1.60E+03 Cohesion 0.90+E04 Internal friction 31.0 angle Poisson’s ratio 0.37
4 EARTHQUAKE RESPONSE OF THE SLOPE To compare the experimental and numerical response of the model, the displacements at the crest are measured during the experiments. An illustration of the experimental results is shown in Figure 8. At the level of 250 Gal, a relatively large deformation is observed at the top of the slope without reinforcement with some local cracks. At this level, due to the effect of rope net and rock bolts, the deformation of reinforced part is still small and no cracks are observed at the crest. When the input acceleration is stepped up to 300 Gal, a sliding failure occurs in the slope without reinforcement. The slope rapidly deforms, along with large cracks at the crest. At this level, some small cracks appear at the crest of reinforced side but the displacements of slope are still small. Large cracks occur in reinforced soil at the level of 350 Gal. Large deformation is observed at the top of the slope but rapid sliding failure is prevented. The nonlinear dynamic analysis of the model is performed using a computer program named DYNAFLOW (Prevost, 1998). DYNAFLOW is a finite element analysis program for linear, non-linear, two and three-dimensional systems. It provides several material models including pressure dependant geomaterials with different types of hardening rules. In the present analysis, nonlinear behavior of soil elements is based on the theory of multi-yield plasticity
Unit Pa. Pa. Kg./m Pa. Degree -
Table 2. Material property of rock bolt. -~
No. 1
2 3 4
5
Material urouertv of rock bolt Parameter Quantity Young modulus 7.00E+ 10 Poisson’s ratio 0.345 Weight density 2.70E+03 Cross sectional 2.55 E-05 area Second moment of 7.95E-11 inertia
Unit Pa. -
Kg./m 2 m m
J
Table 3. Material property of rope net. __
No. 1
2 3 4
Material property of rope net Parameter Ouantitv Young modulus 130E+10 Poisson’s ratio 0.458 Weight density 0.90E+03 Cross sectional 5.02E-07 area
Unit Pa. -
Kg./m m-
yield plasticity (Prevost 1989). As for the rock bolts and rope nets, the material properties are shown in Tables 2, 3.
3.3 Boundary conditions In order to model the slippage between rock bolts and soils during the analysis, nodal link elements are used at the contact points of beam and solid elements (Prevost, 1998). One nodal link element connects two nodal points either in translation or in rotation and is defined with a linearlnon-linear stiffness. These nodal link elements enable local non-linearity between rock bolts and soil during excitation, and give smooth transition of forces acting in two different materials. The heads of rock bolts are directly connected to the rope net and the bottoms of rock bolts are fixed to the foundation. These connections are joined without nodal link elements. The model has vertical supports and recieves horizontal input motion uniformly at the base.
Figure 8. Illustartion of shake table test results.
Figure 9. Mobilized phi and deformed shape of slope.
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Figure 10. Vertical displacement time history ofcrest ( no reinforced slope ).
Figure 1 1 . Vertical displacement time history of crest (reinforced slope).
with a kinematic hardening rule, modeling the stress-strain behavior in cyclic loadings at both low and high stress levels (Prevost, 1989). In the present analysis, maximum input acceleration is 250 Gal. Figure 9 shows the deformed shape with mobilized phi contours of the model at time 10 seconds. As expected, the deformation of the slope without reinforcement is larger than reinforced side. The result of the laboratory triaxial test gives the internal friction angle of 31 degrees and the cohesion of 0.9E+04 Pa. The mobilized phi value larger than the internal friction angle means unstable state. As seen in Figure 9, the slope without reinforcement has larger values of mobilized phi, which shows less stable state than the reinforced side. The vertical displacement time history of the crest in the no reinforced slope is shown in Figure 10. The solid linerepresents the analysis results and the dot line represents the measured displacement in the experiments. Figure 11 also compares the displacement time history of crest in the reinforced slope between the analysis and the experiment. Table 4 compares the maximum vertical values of displacements at the crest of slopes with and without reinforcement. Due to the reinforcement of rock bolts and rope net, there is an effect on decreasing the maximum displacement response.
Table 4. maximum vertical displacement of crest.
The authors are very thankful to the members of the advisory committee, Prof. Okimura and Prof. Tanaka of Kobe University, Prof. Ohmachi of Tokyo Institute of Technology, Dr.Matsuura of Forestry and Forest Products Research Institute, Dr. Sugano of Port and Harbour Research Institute, and Dr.Goto of Ohbayashi Co., Ltd. for their expertise and valuable advises.
5 CONCLUSIONS
REFERENCES
The of the experiments and the numerical simulation are concluded as follows. A new method Of rope net and rock used for reinforcement to stabilize the slopes during an earthquake has an effect on decreasing the displacement response during cyclic loadings.
Prevost J. H. 1989, Dynald,A corizputerprogrnin,ji~rizoizlinear .seismic site response aiialysis, Technical report NCEER89-0025, Princeton University, Princeton, New Jersy. Prevost J. H. 1998, Dyizqflow, A Finite elenieiit progrczin,for trmsietzt non-Linear respoiise (?fnyoand three-dinzeiisioiinl systems, User’s manual, Department of civil engineering and operations research, Princeton University, Princeton, New Jersy.
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Maximum vertical displacement of crest(mm) Case ExDeri men t Analvsis Without - 14.27 - 10.1 I reinforcement With - 9.05 --6.23 Reinforcement
2) There is reduction in mobilized phi angle of the reinforced side, which implies more stability than the side without reinforcement. 3) The non-linear dynamic analysis gives close results to the experiments regarding to the vertical displacement of the crest. 6 ACKNOWLEDGEMENTS
Landmarks in Earfh Reinforcement,Ochiai et al (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Stabilization and surface protection of steep slopes using soil nails and prefabricated concrete part linings A.C. Lottrnann Brandenburg Technical University Cottbus, Germany
L.E. Wichter Brandenburg Technical University Cottbus, Germany
W. Meiniger Research and Material Testing Institute, University of Stuttgart, Germany ABSTRACT: The report gives an overview of the German practice of the stabilization of steep slopes using long soil or rock nails instead of prestressed anchors for the stabilization of rock slopes. The nails connect the slide-endangered rock masses with the stable underground and reinforce the front areas of the slopes. The nail heads are, at the same time, used for setting up different systems of prefabricated concrete part lining walls. These walls, in a distance of 0.75 to 1.5 m ahead of the slope surfaces, are filled and backfilled with soil, and then vegetated. The report demonstrates a number of different systems, deals with the dimensioning principles, and gives some instructions for the execution of such stabilization works. 1 INTRODUCTION Breast or lining walls have to protect the surfaces of stable rock slopes against weathering and to prevent roads or railways from being endangered by rockfall. They are charged by their dead weight and by the pressure of the backfill between the slope surface and the rear side of the wall. Lining walls have been constructed in the course of the railway works since approximately 1850 very often and up to heights of 10 and more meters using natural stone or brick masonry. Nowadays it would be nearly invaluable to construct optically pleasing natural stone masonry walls in the course of modern traffic way construction. On the other hand high and bare concrete walls are not very popular in Germany because of the activities of the environment protectors. So in many cases gabbion walls or polygonal masonry walls are used to protect rock surfaces against weathering, as shown in Figure 1 and Figure 2. In a (theoretically) statical point of view all these piled up revetment walls may be constructed up to nearly unlimited heights, as long as they have a dip of less than 90°, and are piled up very exactly. In practice the construction height is limited by unavoidable imperfections at a slenderness of about 1:8, depending on the dip of the lined slope and the exactness of the geometrical execution of the slope surface and the shape and size of the piled elements. If the elements are prefabricated concrete parts, they fulfil1 the conditions of exactness in size and shape, and high revetment walls may be built without anchoring or nailing as long as the rock Slope itself is stable. The Figures and show such a revetment shortly after it's construction and 12 years later.
Figure
Figure 2.
masonry lining
lining wall of
height.
Figure 5 shows a nearly 60 m high rock slope with berms. The single slopes between the berms are completely lined with concrete revetmet walls of 1 l m height. The cutting made some problems during it's construction caused by a joint system dipping to 663
Figure 6. System of vertical and horizontal concrete parts forming the lining wall.
Figure 3. Revetment wall of 22 ni height.
the cutting and striking parallel to it (Denzer & Wichter 1987, Wichter et al. 1990). Therefore it had to be stabilized using more than 4.000 heavy and long prestressed rock anchors. After the successful construction of the cutting shown in Figure 5 a number of similar systems has been developed, firstly as pure unanchored gravity revetment walls according to the specifications of the German highway authorities. Very soon it became usual to combine the concrete part lining walls with slope stabilizations using soil nailing methods when the global stability of the slope was insufficient. All common systems are composed of vertical columns (so-called lisenes) put in bucket foundations and fixed in a distance of 0.5 to 1.5 m ahead of the slope surface using the nail heads as anchorages. Between these columns horizontal beams are arranged in vertical distances between 0.75 and 1.25 m as may be seen in Figure 6.
Figure 4. Revetment wall after 12 years.
2 CASE HISTORIES 2.1 Stabilization and protection of a cutting slope in mica slate The construction of a new federal highway made it necessary to cut a natural slope in mica slate in a length of 380 m and an excavation depth of maximally 20 m. The mica slate is, in a geological view, a very old rock and was extremely stressed by the tectonics during it's existence. The geological reconnaissance showed that the mica slate was extremely jointed and fissured, and weathered especially in the surroundings of some local faults. The orientation of one joint set was directed nearly parallel to the road and dipping downslope in different angles. So it became necessary to install a stabilization from the beginning of the excavation works in order to avoid later surprises when the excavation had proceeded and the depth of the cutting would have made it difficult to react. It
Figure 5. Surface of a deep cutting protected against weathering by prefabricated concrete parts.
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Figure 7. Assumed sliding wedge for the determination of the required nail lengths.
was decided to reinforce the slope using long soil nails in a grid of approximately 2.5 x 2.5 m. The lengths and the cross-section of the nails was determined assuming that a sliding wedge could move most disadvantageous on a slip plane activated when the excavation was nearly finished as shown in Figure 7 . Threaded construction steel bars of 63.5 mm diameter (steel quality BSt 555/700) and with lengths up to 20 m were used to secure that the slope (which beared some buildings behind on the top) would remain stable on the whole. The nails were arranged vertically in a way that their ends standing out of the slope surface could be used for anchoring the prefabricated lining wall elements later. It was not quite easy to determine the correct starting points for the boreholes (which were points in the air), and great exactness was necessary because later the prefabricated concrete parts had to be threaded on the outstanding nail ends. between The weathered and jointed rock suriace the nails was stabilized during the works to protect the workspace using steel mesh, shotcrete, and short soil nails of 22 mm diameter. About the installation of these elements was decided at the site after each excavation step using lining classes (similar to the way used in tunneling) with different shotcrete thickness and nail lengths. Figure 8 shows the slope surface during the excavation works. After the end of the excavation works the prefabricated vertical elements were installed and fixed, as shown in Figure 9. Then the cast beams (which were, in the cross section, z-shaped) were placed between the columns on consoles. Finally the lining wall was backfilled, and the vegetable mould was brought on the beams. Figure 10 shows the wall after it's completion.
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Figure 8. Slope surface with shotcrete and steel mesh lining and
Figure 9. Installation of the prefabricated vertical columns and
Figure 10. Lining wall after the end of the works.
length was between 7.7 m and 11.3 m, depending on the height of the cutting slope. Figure 11 shows a cross section of the system (horizontal beams not linked) and the nail grid. The Figures 12 to 14 are photographs of the wall during and after construction.
2.2 Stabilization of a Keuper Mar1 slope In the course of the improvement of the access roads to the city of Stuttgart (situated in a deep circular valley) a cutting into a steep natural slope became necessary. The slope was formed by the siltstones and claystones of the Keuper formation with an intercalated sandstone layer. For the stabilization of the slopes the concrete part system was used which had already been used for the deep cutting shown in Figure 5, but a new technique for the abutment construction was developed. The vertical beams were supported using prefabricated concrete abutments kept in place by a part of the nails (0 50 mm threaded tie bars / GEWIbars). The nail grid was 2.5 x 2.5 m, and the nail
2.3 Stabilization of a sandstone slope
A cutting in a sandstone rock slope became necessary at the end of a road bridge over the river Main. In the sandstone slope a gravity induced joint system was found which dipped to the slope surface and
Figure 13. Revetment wall after completion.
Figure 1 1. Cross section and nail grid of the wall system.
Figure 12. Nailed cutting slope before the construction of the lining system.
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Figure 14. View of the concrete part system and the shotcrete lining.
striked parallel to the slope as shown in Figure 15. Joint systems of this type are very often found especially in sandstone rock. So it became necessary to stabilize the slope already during the excavation using threaded GEWI-bars of 50 mm diameter and lengths up to 12 m. In order to protect the workers against rockfall during the excavation sprayed concrete was used to seal the cutting surface. This concrete lining was kept in place using nails of 28 mm di meter. Because the area is intensively frequented by tourists the road construction authorities had to cover the concrete shell with a "green" construction. The system always shown in the last chapter was used, and the cutting slope of 18 m height was divided by a berm in two slopes of less height. Figure 16 shows a photograph of the slope during the excavation works, Figure 17 the double corrosion protected soil nails.
Figure 17. Double corrosion protected soil nails.
3 EXPERIENCES
3.1 Nailing before setting up the lining The experiences have shown that it is possible to survey the starting points for the boreholes exactly enough to string the vertical concrete columns later. This is the condition for dividing the activity of the excavation from the setting of the lining wall, which makes the construction works easier and less costful. In the interest of an obstacle-free workspace ahead of the rock slope the nails should have bell and spigot joints directly before the shotcrete surface. Otherwise there is always the risk of damaging either the nails or the machinery working ahead (or hurting the personnel). 3.2 Type of nails The first precast lining walls used prestressed bar anchors for their support. It may be advantageous in cases where the backfill produces pressure worth mentioning to prestress the front part of the nails before backfilling, and such draw the vertical concrete beams on their support. Otherwise the backfill pressure may produce a split between the part and its support which may be not good for the corrosion protection of the steel nails.
Figure 15. Schematic cross section of the cutting slope.
3.3 Corrosion protection The majority of the nails used in Germany for long term constructions is double corrosion-protected. It is useful in the interest of a workman-like performance to test the intactness of the corrugated plastic sheathing of the nails measuring the electrical resistance between steel nail and surrounding ground. It should be higher than 0,l Megaohm. Threaded GEWI tie bars may be used for long term purposes also with a simple corrosion protection (cement grout of at least 2 cm thickness) when the working stress is reduced.
Figure 16. Sandstone slope during the excavation.
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3.4 Construction details of the concrete parts
3.5 Excavation and blasting
The construction of the concrete parts, especially the horizontal beams, needs a good knowledge of the loads caused by the pressure of the bacMill and the deadweights of the beam itself and the fill. Otherwise the beam may overturn to the one or the other side. It is not easy to place the reinforcement necessary for the control of the bending moments and shear forces in the cross-section of the horizontal beams, and at the same time fulfil1 the requirements of the concrete cover. The beams are stressed by oblique bending, and dimensioning them needs experience and exactness. The horizontal precast concrete parts for bended revetment walls are of very different size (the parts have to be shorter the higher their place in the wall is) and need a very exact performance. Walls of this type therefore are no low cost constructions.
Although there is no recorded damage on the nails caused by blasting in their neighbourhood it should be avoided to blast.
668
REFERENCES G. Denzer & L. Wichter 1987. Dimensioning and perf rmance of a deep cutting for an express highway. Proc. d' Intern. Coizgr. on Rock Mechanics: 33 1-335, Montreal. L. Wichter, W. Meiniger & G. Denzer 1990. Erfahrungen bei der Herstellung und mefitechnischen Beobachtung eines sehr tiefen Einschnittes fur eine Autobahn im Weifijura (Experiences during the construction and the observation by measurements of a very deep cutting for an express highway). Proc. 9. Nat. Syinposiunz on Rock Mechanics, Geotechniksonderheft: 1-9, Aachen.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
An example of a high soil nailed wall in plastic clayey soil B. MariC Conex, cl.o.o., Zagreb, Croatia
P. KvasniCka Faculty of Mining, Geology and Petroleum Engineering, Zagreb, Croatia
D. Radaljac Conex, d.o.o., Zagreb, Croatia
R. Mavar Crontiaiz Civil Engineering Institute, Zagreb, Croatia ABSTRACT: This paper describes a twenty-meter-high nailed wall for a residential and office building built recently in the center of the city of Zagreb. The wall is to serve as a permanent structure. The excavation pit for the building is located between two old buildings with unreliable foundations on a very steep slope, which caused additional difficulties in soil nailing construction. The retained earth is medium to highly plastic overconsolidated clay. The paper presents in situ and laboratory testing results on the basis of which design soil parameters were derived. Deformations brought about during excavation were measured by means of sliding deformeters and compared to lateral movement of survey points at the face of the wall. In addition, results of the finite difference method for calculating stresses and deformations, and stability analysis were also presented.
1 INTRODUCTION
contiguous micro-piles which are 20 cm in diameter and placed on 0.5-meter centers. A plan view and geometry of the excavation are shown in Figure I .
Soil nail technology is generally considered to be applied unsuccessfully in clays. Consequently, very little research has been done on the subject and there is little field experience on soil nailing in the deposits of clays of medium to high plasticity. The nailed wall presented in this paper is located in Deimanova Street in the center of the city of Zagreb. The wall encloses the working site for a residential and office building with an underground garage. Layout dimensions of the garage are about 26 x 30 m, and it is 4.5 m below the ground level. The superstructure is 20 m high (4 stories). The structure is surrounded by the street and adjacent buildings. On the west side, at the top of the slope is the old Roko Park with trees running close to the wall. The excavation pit is located on a very steep slope, between two old buildings with unsolid foundations, which caused additional difficulties in wall construction. The back side of the excavation pit, also the highest side, is hardly accessible to heavy machines. As a result, a soil nail wall was selected as the preferred wall type for the given wall location, as it uses conventional shoring equipment. The nailed wall has two functions: a temporary support of the working pit and the permanent support of the retained earth after completion of the building. To increase the factor of safety for the wall, the lower portion - i.e. that of the garage - was planned to be supported with heavy concrete walls. On the street side of the excavation, the bottom of the pit is 4.5 m bellow the street level and the retaining wall is formed from
2 SOIL PROPERTIES AND DESIGN PARAMETERS The geology of the area consists entirely of stiff overconsolidated clays with a sand layer close to the surface in upper portion of the retained soil. Subsurface exploration was conducted to determine the properties of the soil, in which nails and anchors
Figure I . Plan view of the excavation.
669
were to be installed, and groundwater conditions. The exploration included the following: boreholes of up to 25 m, standard penetration tests, and the collection of disturbed and undisturbed soil samples for visual identification and laboratory testing. Piezometers were installed in the ground in the boreholes shown on the plan view shown in Figure 1. The readings on piezometers, placed close to the wall, generally showed that there was no ground water. Laboratory testing consisted of standard classification tests, oedometer tests, unconfined compression tests and direct shear tests. Based on the laboratory tests, clays were found to have medium to high plasticity and SPT over 20 blowcounts, and thus classified into stiff to very stiff clays. A characteristic soil-profile with boring logs is shown in Figure 2. Triaxial compression tests for determining shear strength of clay were not carried out - for reasons beyond authors’ control - and direct shear tests were performed instead. Although the authors were well aware that the conventional direct shear tests were not the preferred tests for clay, their use in the case of stiff clay was reasonable, as the results referring to shear strength are comparable to those obtained by the triaxial tests (In the triaxial compression test the effective stress path followed to failure by many overconsolidated samples of medium to high plasticity clay in undrained compression is approximately vertical on a t-s‘ diagram). In addition, undrained strength c,, (in our case obtained from unconfined compression tests) is also nearly proportional to s ‘ for heavily overconsolidated clay samples (CIRIA’s report 104). With the moisture content and effective stress level pervdining in situ, it may be recognized simply as the frictional strength of the soil (like results from direct shear tests). The results of the direct shear tests and unconfined compression tests are presented in Figure 3 together with their least square straight line and design envelope, with c’ = 50 kN/m’ and q ’ = 20”. Design unit weight y = 20 kN/m3.
Figure 3. The results of shear stress tests and unconfined compression tests.
The average shear strength parameters of the clay determined from the in situ and laboratory tests were also used for calculating the shear modulus according to the correlation E = 500 . c,. For the purpose of calculation, the retained soil was divided into four zones with equal soil properties. The zones correspond to steps in nail wall facing. The selected design parameters of each zone are presented in Table I . Strength parameters of c‘ = 0 and q~= 30” were assumed for the sand. Table 1. The selected design parameters of zones.
m 0.0 to 5.0 5.O to 11.5 11.5 to 15.0 15.0 to Figure 2. A characteristic soil-profile with boring logs.
MN/m35 to 65
110 to 170
55
to 85
160 to220
80
to 110 26.7 to36.7
220 to280
110 to 140 36.7 to46.7
h ... height of a zone
670
MN/m”-------
kN/m70to130
11.7 to21.7 18.3 to28.3
3 DESIGN CONSIDERATIONS 3.1 Problems of nailing in clay
~ , = 4 0 0 / (2.0.14.3.14. 8.0)=57kN1m2.
It is widely known that one of the most critical conditions for a nailed wall is pulling out or breaking of the nails, which results in sliding of the wall and retained soil. A main design requirement is to determine an economic spacing and length of nails to stabilize and limit movements of the wall. Nailing in clay, in contrast to sands, brings the following disadvantages: - shear strength on the contact of nails and soil is lower and - critical sliding surface extends deeper in the retaining soil. As a result, nails in clay should be longer and more closely spaced than in sand, which makes nailing in clay less economical. In standard design, the part of a retained wall with nails is considered as a rigid block loaded with the rest of retained soil. A nailed wall may be assumed to be a rigid block only when nails are spaced closely enough to "nail" the soil between them. In the case of the Deimanova Street nailed wall there are some deviations from the standard design. In view of the fact that this wall is located between the buildings and that it is extremely high, and in order to prevent the wall and the retained soil movements, nailing was combined with prestressed anchors the lengths of which were adjusted to reach the soil behind the critical shear surface. As a result of this, the nails could be widely spaced but the assumption we started from (standard nailed soil) did not hold any more, and retaining construction had to be analyzed by using a more sophisticated model. The behavior of the wall under working -loads, which pertains to the serviceability limit state, was analyzed by means of a finite difference method; the failure condition, which pertains to the ultimate limit state, was analyzed with standard slope stability method, assuming anchors to be outside tension forces. In addition, we checked whether the nail and anchor tensile strengths were adequate to provide the support force to stabilize the active block, and whether anchors were embedded with a sufficient length into the resistant zone to prevent the pullout failure. Capacity of the anchors was determined based on experience and according to recommendations by FHWA (Table 3.3) with ultimate bond stress for stiff clay between 40 and 60 kN/m'. Besides, after having consulted experienced geotechnical engineers, the authors came to the conclusion that actual anchor forces should not exceed 400 kN since clay creeping may occur. As the clay was very stiff, we decided to use this maximum value for the calculation, and obtained the following ultimate shear strength on the soil-anchor contact qJ:
(la)
where A, . .. ultimate anchor force, d, ... outside diameter of the anchor drillhole (assumed 0.14 m), I , .. . anchor fixed length (assumed 8.0 m). The subsequent three anchor pull-out tests confirmed that the assumption was quite reasonable. According to FHWA, the ultimate pullout resistance tends to be relatively independent of depth below surface for a constant soil type and particular installation technique, which has been attributed to the decreasing significance of soil dilation with depth. It is noted that - for nails that are installed in predrilled holes and grouted under gravity (as in the case of the DeZmanova street nailed wall) - the effect of overburden pressure tends to be diminished by arching around the installation drillhole; this phenomenon also contributes to the observed lack of significant dependence of the ultimate resistance to the overburden pressure. Concerning the soil-grout interface, only very small displacements between the nail and the adjacent ground are required to mobilize the ultimate bond or adhesion. This means that the nails and the nail grout-ground interface must be sufficiently stiff to ensure that the reinforcing loads can be developed without associated excessive deformations (FHWA). As regards a nailed wall as a unique system, not only a stiff nails, but also a stiff structural facing contribute to overall better behavior of the supporting system, because a stiff facing equalizes the overall deformation and, therefore, helps in uniform distribution of action on nails and anchors. 3.2 Finite diference analysis Serviceability limit state is the limit state that addresses loss of service function resulting from excessive wall deformations. Therefore, serviceability requirements of the design are expressed mainly in terms of limiting displacements. A plane strain finite difference analysis was carried out by using the FLAC package (Itasca Consulting Group, Inc.). This analysis was used to model the soil as an elastic-perfectly plastic material by obeying the Mohr-Coulombs failure criteria. In conjunction with behavior of overconsolidated clays, initial lateral stresses were based on an earth coefficient at rest of two. High soil stiffnesses referring to the range of small strains (based on the conclusions made by Simpson, 1992) were also taken into consideration. Drained conditions were assumed in the analysis that simulated excavation in front of the wall in thir-
67 1
Figure 4. Layout of the facing and ground anchors in the critical cross-section (dimensions in meters).
The predicted movements suggested that, by the completion of construction, the wall would be translated forward and rotated about the base with little deflection due to banding. The maximum movements (20 mm) should take place at facing in the lower portion of the wall.
teen stages of construction. In each stage of the construction, maximum forces in the anchors caused by excavations were determined, as well as distribution of horizontal displacements in soil. Input soil, nail and anchor parameters were as presented in Chapter 2 (higher values of shear moduli from Table 1 .). The facing was modeled as a linear elastic material, 0.3 m thick, having a modulus of 30 000 MN/m2. The 3D effect was represented by adding flexible support to each of the four horizontal beams. Stiffness of the flexible support was derived with assumption that horizontal beams were laterally fixed. With FLAC predicted wall movements are shown in Fig 5.
3.3 Limit equilibrium calculations Limit equilibrium analysis is based on limit equilibrium concepts. At failure, a slip surface is assumed to occur with simultaneous mobilization of the shear strength along that surface. The mass above the slip surface is assumed to move as a rigid body. The driving force causing the development of the slip surface is the weight of retained soil. The resisting forces are due to shear strength of the soil, forces in nails and prestrained anchors, and the resistance of soil in front of the toe of the wall (passive resistance). Limit equilibrium analyses were carried out by SLOPE/W software (Geo-slope International Ltd.). The safety of the anchored slope is expressed in terms of the factor of safety F , which is the ratio of the available shear strength to the mobilized strength. For heavily overconsolidated clays an effective stress analysis usually derives a minimum factor of safety; therefore, this type of analysis was applied for the Deimanova street nailed wall. The geometry of the slip surface with nails and prestressed anchors is presented in Figure 6. The failure analyses include external modes, internal and modes called mixed modes (FHWA) which include pullout of the nails. Factors of safety were computed automatically for a great number of slip surfaces, most of which passed through the wall toe. The shape of the slip surface could be curved, planar or combination of both. For
Figure 5. Predicted wail movements.
672
According to design requirements, each facing zone was provided with nails which were horizontally spaced 1.5 m apart, and vertically nails were spaced apart from 2.0 m (8.0 m lengths) in the uppermost zone to 1.2 m in the lowest zone (10.0 m lengths). The prestressed ground anchors were 20 to 24 m long with fixed lengths of 8.0 m. The anchors were installed at an angle of 15” to the horizontal. The bore hole diameter was 140 mm. The strand had the following characteristics: diameter 12.9 mm, cross-section area 100 mm2, tensile strength 1860 N/mm’ and modulus of elasticity E = 195.+10 kN/mm’. On the edge of each zone was a horizontal beam whose role was to fix upper and lower facing parts and to serve as a bearing for a prestressed anchor head. The facing was cast-inplace. Grouting was done in several working steps: f-illing the casing with grout from the top of the borehole, pulling out the casing to the beginning of the fixed anchor length, primary grouting through the casing, removal of the casing and secondary grouting of the fixed anchor length through inner grouting pipe. The anchors were stressed ten days after completion of grouting procedure, i.e. after the grout obtained the strength of minimum 30 MN/m2. Three prestressed anchors pullout testiness were performed during construction. The ultimate pullout resistance was between 400 and 500 kN,which was within the expected limits.
Figure 6. Limit equilibrium analysis with critical slip surface and a minimum factor of safety (dimensions in meters).
almost homogeneous material, an arc of the circle was found to be satisfactory. In the analysis the lengths of prestressed anchors were varied to keep the fixed lengths of anchors behind the critical failure surface. In drained analysis found was the minimum factor of safety of F , = 1.27. As mentioned above, after completion of the retaining structure, the portion below the street level was supported with concrete massive walls, and the above street portion of the nailed wall remained the permanent construction. For this (upper) part the calculated factor of safety was of the order of 1.5. Factors of safety equal to or greater than 1.25 for temporary construction and equal to or greater than I .5 for permanent construction were found to be satisfactory. It may be concluded that the main role in the stability of the wall may be attributed to the long prestressed anchors that, for the critical stability situation, retain the whole sliding mass in equilibrium.
4.2 Field instrumentation and monitoring According to FHWA the most significant measurement of overall performance of the soil nail wall system is the deformation of the wall during and after construction. For the Deimanova Street nail wall, movements along the soil mass in the retained soils were measured in two ways: by measuring the movements of three geodetic points, and by means of two sliding deformeters (24.0 m and 16.0 m long) to obtain continuous displacement measurements along the axis of measurement casing. In Deimanova Street, sliding deformeters were positioned along the prestressed anchors. Positions of deformeters are presented in Figure 2. Sliding deformeters were installed in a short time after placing the anchors and facing; the measurements were taken fourteen times in the period between 2 May 2000 and 21 November 2000 (KovaCeviC, 2000). The measurements showed maximum horizontal displacements of 2 mm in the deformeter 1 and 1.5 mm in the deformeter 2. (Figure 7) Similar results were obtained by measuring the movements of the geodetic points. Since October all measurement data in both deformeters have been exactly
4 CONSTRUCTION AND MONITORING 4.1 Construction The construction began with removal of trees and grass at the end of A ril of the year 2000. Approximately 10 000 m-7pof soil was excavated. The excavation and construction of the nailed wall was completed in 3.5 months, which was - considering the height and other dimensions of the wall - a rather short time, the more so because/even though the limited working space available for the building site, and the fact that Deimanova Street is a deadend street (problems with taking excavated soil away) made the construction even more difficult. 673
I
5 SUMMARY AND CONCLUSION
I
DEFORMETER
DX
D2
I
The nailing technology in the Deimanova Street nailed wall included the combination of passive (nails) and active inclusions (long pretensioned anchors). Stability calculations showed that long anchors could not be avoided because of deep critical slip surface that is typical of cohesive soils at strength limit state. The quality of the clay material and lack of ground water made fast excavation and wall construction possible. Construction work was executed during the summer period, and rainy weather did not disrupt the building schedule seriously. During construction, deformations were measured to monitor the stability situation. Measurements showed that the displacements along the facing were considerably smaller than indicated by numerical analyses. The Deimanova Street nailed wall is one of the highest retaining constructions of this type ever built in clays. Its behavior is also beyond every expectation, which indicates that there is still room for more economic nailed wall construction in similar materials.
-3.0 -2.0 -1.0 -3.0 -2.0 -1.0 ELONGATION
bml
Figure 7. Horizontal movements of facing and the ground as measured by sliding deformeters (Kovatevic, 2000).
the same, which means that no creep deformations were observed. Although the numerical modeling predicted final maximum movements of 20 mm (chapter 3), deformation measurements showed much smaller wall movements. It seems that both the stiff soil in the background and the stiff facing are the reason why the Deimanova Street nailed wall was deforming less than expected and less than predicted with calculation models. Overall results demonstrate that the calculation should account for a wide range of modes of behavior when assessing the loading effects in service, and that special attention should be given to a 3D effect.
REFERENCES FHWA-SA-96-069. 1996. Manualjor design & ~ o t ~ r i ~ t i o n nzotiiroring of soil nail walls. KovaEeviC, M.S. 2000. Control measurements of horizontal soil displacements by means of sliding deformeter. University of Zagreb, Faculty of Civil Engineering. Padfield, C.J. & Mair, R.J. 1984. Design of retaining walls embedded in sfifjc clay. CIRIA Report 104. London: Construction Industry research and Information Association. Simpson, B. 1992. Retaining structures: Displacement and design. Geotechnique, Vol. 42. 541 -576.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Stabilization of historical retaining walls using soil-nailing methods W. Meiniger Research and Material Testing Institute, University of Stuttgart, Germany
L.E. Wichter, E. Joppa & R. Loer Chair Geotechnics, Brandenburg Technical University Cottbus, Germany
ABSTRACT: Germany is rich on historical retaining walls. Many castles, churches, and old towns are surrounded by walls, and many of these walls become more and more instable. For the stabilization of collapseendangered historical retaining walls the soil nailing technique is very cominon since it's invention about 20 years ago, because it allows to increase the stability without changing the views of the walls, and without digging up the backfill. The paper deals with the principle of the measures, the dimensioning rules, and the practice of the works. Some examples illustrate the method. properties of the nails. When the first 1:1 scale tests were carried out in order to investigate the failure behaviour of nailed walls, and when in 1985 the first certificate of approval was given to BAUER company nobody thought about using the method for more than building pit sheeting work in combination with a shotcrete facing. The principle of this scope is shown in Figure 1, and Figure 2 shows a characteristical construction of this type. In the meantime the use of soil nailing methods exceeds widely this area of application. Especially for the stabilization of historical retaining walls the specifications of the certificates of approval are not applicable. Many of these walls, especially along railway lines, are lining walls, which means, that they have been constructed in order to protect a rock slope surface against weathering. As backfill material crushed rock of low quality was used, and the walls themselves are only thin shells of dressed stone. Figure 3 shows the typical arrangement of the stones even of high lining walls. The wall shown in this figure had collapsed suddenly and caused a derailment and death toll. Other walls, especially in rural areas, have been built using every kind of stone available including bricks and rubble stones, and then plastered as shown in Figure 4. The direct value of these walls (regarding their importance for the preservation of ancient monuments) is not very high, but they have some importance for the outward appearance of the villages. So they have to be preserved, and nailing is an appropriate method for increasing their stability. Not seldom during the investigation of the state of the walls it becomes obvious that parts of them have failed once before, and then had been reconstructed.
1 INTRODUCTION Historical retaining walls form a big part of the views of Germany's ancient towns, castles, and church environs. They have been built in former centuries following a minimum principle: only as much stone and mortar as absolutely necessary was used, and great attention was payed to an effective drainage of the backfill. In a statical view all these walls are gravity retaining masonry walls. Because of their construction the walls cannot satisfy the safety requirements and the stability standards for new retaining structures. Weathering of the inortar and the stones, the pressure of the roots of trees and shrubbery, and the clogging of the backfill void volume and the drainage ways extort stabilization measures from the owners of the walls (mostly churches, aristocracy, public administrations, or railway authorities). Many of the historical walls are protected as historical monuments, and therefore their view may not be changed. The soil nailing technique gives the possibility to increase the stability of the old masonry gravity walls without changing their views, and without digging up the backfill.
2 SOIL NAILING - ASPECTS OF CONSTRUCTION SUPERVISION After the invention of the soil nailing method into the German ground engineering practice (Stocker & Gafiler 1979) the German Institute for Building Techniques (DIBt) gave certificates of approval to a number of construction companies. The certificates contain the required safety factors, dimensioning rules, testing procedures, and required material 675
Figure 3. Collapsed lining wall beside a railway line Figure 1. Soil nailing for building pit sheeting work: execution steps.
Figure 4. Medieval retaining wall constructed using various types of stones.
Figure 2. Building pit lining (shotcrete facing and soil nails).
3 DIMENSIONING METHODS
In general the wall substance is bad and cannot be bettered significantly (which means, e. g., a certification against punching of the nail heads in the old masonry normally cannot be done). So the stabilization of ancient retaining and lining walls is a task which has to be solved with intelligence and empathy, and usual standards for new constructions are not very helpful in this job.
As mentioned above the certificates of approval of the German Institute for Building Techniques regulate the dimensioning of soil nailing measures when the method is used, together with a shotcrete facing, to stabilize the walls of building pits in soils, etc.. The stabilization of historical retaining walls is not a part of these regulations.
676
Figure 5. Failure mode for the dimensioning of the nails.
The dimensioning of the soil nails is executed normally using the wedge of the active earth pressure behind the wall as loading element, and traffic loads, if they cannot be excluded (Figure 5j. The nails are handled, in the statical analysis, as anchors penetrating the slip plane. In reality they reinforce the backfill of the walls which explains that the nailhead forces straining the old masonry are generally very small. The wall substance mostly does not allow to calculate the bearing capacity of the masonry for taking over the nail head forces. So in many cases the iron plate of the nail heads is put on a larger stone, or concrete bedding, without calculation and proof of it's bearing capacity. Until now the authors have not heard of any case where a soil-nail stabilized retaining wall has failed later, or showed damages, because of lack of bearing capacity of the masonry for the nail head forces. If there is any doubt on the load capacity it is always possible to test the deformation and bearing behaviour of the masonry around a nail head executing load tests using hydraulic hollow cylinders as shown in Figure 6.
4 EXECUTION OF STABILIZATION MEASURES There is no possibility for making an estimate about the stability reserves of already deformed historical retaining walls, but in general they are very small. There are a number of walls which collapsed suddenly when the stabilization works just had started. Obviously the concussions caused by the drilling bit had been enough to exhaust the stability reserves. The collapses in all cases the authors know happened very quickly and without any warning. Fortunately until now no death had to be lamented. In order to avoid accidents high and obviously collapseendangered retaining walls must be supported before starting any drilling activities. Of course stays and supporting posts restrain all works in front of the wall, and for that reason construction companies try to do with little when having an order. Therefore the submission records should contain items for the supporting measures.
677
Figure 6. Load test on a soil nail head stabilizing a historical retaining wall.
Sometimes another aspect makes it difficult to execute nailing works in the neighbourhood of churches. Churches were built in former centuries in Germany on the top of hills within the communities whenever possible. Very often this position required the construction of retaining walls which now need, after a number of centuries, an improvement of their stability. In some parts of the country it was common to bury the dead as close as possible to the church, which means that graves are often situated directly behind the retaining walls surrounding the graveyards around the churches. One has to know that in order for not disturbing the calmness of the dead. The nail screen obtained from the statical analysis is transferred to the wall facing. Vertical and horizontal nail distances between 1.5 and 2.0 m are customary. The locations of the drill holes are chosen bearing in mind that the nail heads should be covered later by the masonry again. Drilling diameters between 80 and 150 mm are used, and the nails with their spacers are plugged into the borehole filled with cement grout. Usually steel nails (threaded construction steel bars BSt 500 S-GEWIj with double corrosion protection are used. The nail heads (plates and screws) are mounted and then covered by the stones
Figure 7. Nail head construction in masonry.
taken out from the wall before the beginning of the measures. Many walls have lost the mortar in their joints in the course of time, and the masonry and the backfill is interspersed with cavities. The wall substance usually is strengthened as follows. In a first step the joints are released as good as possible from trees and bushes, which in many cases have contributed to the wall destruction, and all loose joint fillings are cleared away. Then the joints are refilled with mortar as completely as possible using a shotcrete machine which is more effective than filling them by hand. In many cases a special mortar containing trass cement is used. Immediately after the beginning of the hardening the surface of the wall has to be cleaned by hand using water and brushes, a procedure which may be repeated once or twice to get a stone-faced wall again. Drainage holes of course must be preserved during these procedures and cleaned as good as possible. Then the cavities and the rear unfilled ends of the joints are "injected" (better: filled again) carefully and with only low pressure with cement grout. Figure 7 shows a typical nail head performance inside an old retaining wall. As an example on Figure 8 the cross section of an old retaining wall stabilized by soil nailing is shown. The wall substance was in such bad condition that it was necessary to cover it using a sprayed concrete lining, also in order to protect an old church foundation against sliding movements towards the wall. Figure 9 shows the drilling works for stabilizing this wall.
678
Figure 8. Cross section of an old retaining wall supporting a church building.
Figure 9. Drilling works on the wall shown in Figure 8.
The combination of improving carefully the wall substance and reinforcing the backfill using double corrosion -protected soil nails guaranty a further lifetime of some hundred years more for ancient retaining walls. REFERENCES M. Stocker & G. GBRler 1979. Ergebnisse von Groaversuchen uber eine neuartige Baugrubenwand-Vernagelung (Results of large scale tests on a new nailing method for construction pit walls). Tiefbau-Ingeizieur-bau-Stl.~~eizbau, Heft 9.
Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
An analytical study to the vertically reinforcing soil slopes M. Moradi & M. Davari Departnzent of Civil Engineering, Uiziversity of Telzrarz, Iran
ABSTRACT: An approach for stability analysis of vertically geosynthetic reinforced earth structures over firm foundations is presented. The approach involves both internal and external stability analyses. The external stability analysis is based on limit equilibrium through the extension of the bilinear wedge method and it allows a slip plane to propagate between two trenches restrained by sheets of geosynthetic. The internal stability is considered to investigate the pullout capacity of vertical elements through the stable zone of soil. External loadings such as crest surcharge, ground water level and pseudo-static forces of earthquake are included in the equilibrium formulation. In spite of well-established coincidence with experimental data obtained from centrifuge tests this study offers a framework for stability analysis of these reinforced soil structures.
1 INTRODUCTION
derstand the operation of such structures have been conducted through the centrifugal tests all of them carried out in UK (Jackson, 1998). In these experimental research the effect of vertical elements of geosynthetic has been proved through the limitation of deformations or settlements and improvement of factor of safety against sliding. This paper summarizes a theoretical investigation into the potential benefits provided to the stability of soil slopes through the action of vertical reinforcement. Outlined in the following are the external and internal stability analyses of a vertically geosynthetic reinforced soil slope over a firm foundation. External stability addresses situations where a reinforced portion inay slide as a monolithic block along two parallel reinforcing trenches. Internal stability deals with the resistance to pullout failure within the reinforced soil zone resul ting from the interaction between soil and reinforcement. In the framework of this paper, firm foundation implies that deep-seated failures are unlikely to occur and therefore are not considered in the analyses.
Use of horizontal layers of reinforcement in the construction of soil embankments is now a welldeveloped technique. This method is well suited to the layered nature of new fill construction but cannot be readily employed retrospectively. Barker & Wood (1 989) have proposed an alternative method of increasing the stability of existing slopes. This slope reinforcing technique has been devised whereby arrays of parallel vertical sheets of polymer or metal grids, or geotextiles, are installed in narrow trenches, typically 0.2-0.4 m wide, excavated at regular intervals, typically 2-4 m, along a slope. This transforms the slope into an in-situ reinforced soil block, analogues to a soil nailed slope but with an array of high strength sheets rather than bars. By using this technique, it will be possible to enhance the stability of natural slopes and existing cuttings and embankments without removing large wedges of soils to install arrays of geotextiles and geogrids. This method of reinforcement is chiefly applicable to the prevention, or remediation of shallow slope failures, the occurrence of which is relatively widespread in the over consolidated clays encountered in slopes. The idea of vertically reinforcing of existing slopes is relatively a new technique among soil reinforcing methods. This technique was first proposed by Barker & Wood (1989), Barker (1991). They used a simple limit equilibrium analysis in a fully drained, dry, cohesionless soil to show that a significant improvement in stability could be achieved using vertical reinforcement. After this various attempts to un-
2 STABILITY ANALYSIS 2.1 Geizeral
Limit equilibrium analysis has been used for decades in the design of earth slopes. Extension of this method to the design of vertically geosyntheticreinforced slopes, where the reinforcement is tangibly modeled, is desirable. The main drawbacks of limit equilibrium analysis are its inability to deal
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with displacements and its limited representation of the interaction between dissimilar or incompatible materials comprising the slope. Typically, adequate selection of materials properties and safety factors should ensure acceptable displacements, including safe level of reinforcement deformation. The same as unreinforced slopes, factor of safety, F,, is used to replace the existing slope with artificial one, in which the shear strength is Cp, = tan-’(tanCp/F\)and c, = c / F, where Cpd and c, are the design shear strength parameter of the artificial soil. Consequently, F, applies equally to all shear-resisting components be it soil or reinforcement. Stability analysis in the field of soil reinforcing techniques is generally based on limit equilibrium in which the state of structure respect to the critical condition will be compared with a proper safety factor. On the other hand, inclusion of geosynthetic reinforcement in limit equilibrium analysis is a straightforward process in which the tensile force in the geosynthetic material is included directly in the limit equilibrium equations to assess its effects on stability. To achieve this, several methods may be considered. These include two-part wedge analyses; circular or non-circular analyses and log-spiral failure analyses. A general form of limit equilibrium approach can be developed based on the two-part wedge mechanism. The two-part wedge analysis assumes a bilinear failure surface. This has been shown to provide a reasonable representation of the potential failure surfaces for slopes (e.g., Jewell, 1982). Current code of practice in the UK (BS 8006:1995) introduces this as one of the most commonly used techniques among the methods of assessing stability. The advantage of such an approach is that a wide range of potential failure surfaces can be approximated and the method is relatively simple to program for computer analysis. Therefore in this paper we try to investigate the stability through a two-part wedge mechanism. 2.2 External stability External stability analysis considers the determination of safety factor for a block of soil against slid-
Figure 2. Free body diagram in external stability analysis.
ing along two parallel trenches laterally restrained with reinforcing elements. Figure 1 shows failure surface in this analysis. As can be seen, failure mechanism consists of three planar surfaces. Figure 2 shows the free body diagram of forces acting on wedges, where W,= weight of each wedge; k ,,W, & k ,W, =psuedo-static forces due to earthquake in horizontal and vertical direction; F,=resisting force due to reinforcing elements implies to lateral faces of each wedge; S,=resisting force due to cohesion in base of each wedge; U,=hydrostatic force due to pore pressure; N: =induced effective force; P,=interwedge force; and q=crest surcharge. The assumptions that should be considered are:
I . Soil mass is a homogeneous and isotropic in a horizontal plan along the slope. 2. Stresses distribute uniformly along reinforcing elements. 3. Resisting force due to vertical elements imposes parallel to the base of failure wedge on the lateral faces of soil sliding block. 4. The interwedge resultant force P is horizontal. It is a conservative assumption that implies no shear strain has been developed along the interface. According to Figure 2, unknown quantities interfering in the problem are being PI ; P2 ; NI;N i ; N i ; and Fs, where Fs is safety factor against sliding. For simplicity the inclination of N: is considered equal to Cp respect to perpendicular on the base failure surface. Now based on the free body diagram shown in Figure 2, one can assemble the force equilibrium in an angle of 8, -Cp respect to horizontal for each wedge, in which 4, is internal angle of friction of the soil. Therefore according to active and resisting forces on each wedge , the following expressions may be obtained: Wedge 1:
(S, + 2F, )cos Cp - W, (1 + k, )sin(@,- Cp) k,,W, cos(8, - 4,)- U, sin Cp - P, cos(8, - Cp) = 0
Figure 1 . Failure surface in external stability analysis.
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(1)
Wedge 2:
inforcement on each side is a function of the area of that element, the height of soil above it and the factored soil strength parameters. The force transmissible from the area dxdy is:
(S2+ 2F2)cos@ - W2(1 + ky)sin(O, - @)-k, W2 x (2) - q) - U sin Q + (P, - p2)cos(e2- $) = o
1 dF =F [k,hyf, tan @(1- r,, ) + f, .c]dxdy
Wedge 3:
(S, +2F,)cos$- W,(1+ k,)sin($, k, W, cos($,
-
-4)-
@)-U, sin @ + P2cos($, - 4)-
qB(L, + L2 - H cotj3)sin(8,
- $) = 0
(4)
I\
where F,=factor of safety against sliding; ky=coefficient of lateral earth pressure parallel to crest of slope; y = unit weight of soil; fb=bond coefficient (a parameter describing the released soil parameters at the soil-geosynthetic interface; typically within the range of 0.6-1 .O); @ =internal angle of friction of the soil; r,=pore pressure ratio; c=cohesion of the soil.
(3)
In above equations, B is distance between trenches and other parameters introduced previously in Figures 1 and 2. In order to determine the safety factor, one can obtain PI and P2 from equations 1 and 3 and then substitute in equation 2, it yields to a value for F, for a given geometry and failure surface. It should be mentioned in order to include the effect of pore pressure, a soil parameter called the pore pressure ratio, ru , is used. The pore pressure ratio is defined as a ratio between the total pore pressure and the total overburden pressure.
The force is summed over the area of the entire wedge. This assumes that the whole of the wedge is located within the zone of reinforcement, an assumption compatible with the shallow failure mechanisms that this technique of reinforcement is designed to counteract.
2.3 Resisting .force due to reinforcing elements
2.4 Internal stability
A potentially significant problem in limit equilibrium analysis of reinforced soil is the need to know the force in each reinforcing element at the limitstate. In this section, the maximum force transferred to the lateral faces of each wedge from the reinforcement is calculated. It is assumed here that tearing or pullout of the reinforcement is not the critical failure mechanisms. Many authors (e.g., Jewell, 1980 a, b, McGown et al. 1978), have investigated the performance and behavior of reinforced soil. It has been find that the reinforcement has no initial effect. it is only working after the reinforcement has been strained. As the soil strains it mobilizes strength to resist the shear loads, soil strain causes strain in the reinforcement, which leads to a further increase in strength in the reinforced soil. Reaction of the reinforcement’s developed strength acts as a resisting force to the lateral faces of the sliding block of soil. Based on Figure 3, the maximum force (dF) transmissible to the soil from a small element of re-
In reinforced soil structures, the capacity of the reinforcement to develop the required tensile resistance depends also on its pullout resistance; i.e., the length anchored into the stable soil zone (Fig. 4). The objective of the internal stability analysis is to ensure adequate stability against pulling out for a given length of reinforcement panel, L,. It is assumed that any reinforcement placed outside the parallelogram section analyzed (to aid drainage, for example) provides no benefit to the pullout resistance. Based on above paragraph, one can define the factor of safety against pullout for a reinforced slope as:
Figure 3. Resisting force due to reinforcing element.
Where F,=resisting force against pullout; F$I,,, = reaction of maximum transmissible force to the soil; 6 = the angle in which the force Fslip acts.
Figure 4. Pullout resistance due to stable soil zone.
68 1
According to section 2.3, F, will be determine from:
Where Z=the depth of centered of stable area from crest level; Sstable = the area of stable zone behind the active zone. In equation 5 , Fslip,causes the reinforcement to pull out from the stable zone of the soil. This force is the reaction of resultant force provided by the reinforcement that acts in angle 6 respect to horizontal. According to Figure 5 , one can determine Fslipand 6 as:
a = sin-’
F, sin(@,- 0 , ) + F 2 sin(@,- 0 , ) F4,p
1
Figure 6. Effect of vertically reinforcing on safety factor. Table I . Parameters assumed in analysis of a reinforced slope.
(9) H B k kt, ru Lr (k~~)!o) ~I+J,~~):O) (m) (m) ;Ikpa) (m) 10 30 18 45 4 3 0.5 30 0.25 0-0.5 3 C
Hence the factor of safety against pullout can be find from equation 5 to 9 for a given length of reinforcement .
proposed analytical and theoretical basis in previous sections has presented a well-effective technique to enhance the stability of soil slopes. Such a result has been arisen from resisting force due to reinforcement included in force equilibrium equations.
3 RESULTS AND CASE HISTORY 3.1 Typical results
To analyse the problem a computer program called ‘VERSAP’ has been developed. This program is written based on concepts and formulation presented in section 2.2 - 2.4 (Davari, 2001). Typical results of VERSAP which use the parameters presented in Table 1 are shown in Figure 6. Firstly, stability analysis is carried out in unreinforced condition. Then analysis is developed for vertically reinforced slope. According to Figure 6, it appears that vertical geosynthetic reinforcement causes considerable increase in safety factor against sliding. Therefore the
3.2 Case history Jackson (1998) reports the results of centrifuge tests of well-instrumented small-scale model slopes. In order to determine the benefits provided by vertical reinforcement in terms of factor of safety, a definition of slope failure was required. He prepared some slope models with vertical arrays of geosynthetic, tested them by centrifuge testing equipment were brought to failure through the increase of centrifugal acceleration (Ng). A factor of safety was established for the slip surfaces in terms of the model scale using simplified form of Bishop’s analysis. The parameters considered in model tests and equivalent prototypes, are summarized in Table 2. The results of VERSAP with safety factors obtained from tests are being shown in Table 3. The results obtained from theoretical analyses are relatively close to values from experimental investigations and difference between safety factors is nearly insignificant with the view of geotechnical engineering. The good agreement exhibited in Table 3 confirms the accuracy and validity of analytical approach presented in this paper.
Figure 5. Vectorial representation of forces to determine F,,,,,.
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retical approach have a considerable coincidence with data obtained by previous experimental investigation. This agreement proves accurately the theoretical basis proposed in this paper. The proposed design procedure can be easily carried out using a computer program (e.g., Davari, 2001). The mechanism and analysis used can be replaced with other rigorous stability methods. Hence, this paper provides a conceptual framework for design of vertically geosynthetic reinforced slopes.
Table 2. Parameters considered for testing samples and equivalent prototype models. c
y
14 16 16
18.8 18.8 18.8
Modelno.
H
B
ky
Lr
NI
(k
prototype 9 prototype
60 6.3 2.4 0.4 9.8 60 0.2 0.105 0.4 0.31 26.0 60 5.2 2.7 0.4 8.0 -
REFERENCES
Table 3. Comparison between test results and computer program VERSAP. Model no. 2 3 7 1
9
Fs from test Fs from VERSAP Difference (%) 1.33 -5.7 1.41 1 .X 15.4 I .56 1.41 1.42 0.7 1 43 1.6 11.9 I
4 CONCLUSION
A procedure for the design of slopes reinforced with geosynthetic materials has been presented. The analyses involved in the presented design process are based on limit equilibrium. These analyses ensure the reinforced mass is externally and internally stable. The presented design procedure and the results obtained herein have proved that the technique offers a well-effective solution to improve the stability of natural slopes and existing cuttings and embankments. In addition to this, the results from theo-
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Barker, D.H. & Woods, R.I. 1989. Vertically reinforced soil slopes: theory and application. Proceediiigs of the Syniposium oii Foiiiidation and Tutzneling, London, Sept. 1989. Barker, D.H. 1991. Slope stabilization by vertical soil reinforcement. I. C.E Slope Stability Energy. Conf Isle of Wiglzt: 327-334. British Standard Institution 1995. BS 8006: Code of practice for strengthened / reinforced soils and other fills. BSI. London. Davari, M. 2001. A theoretical approach to the vertical geosynthetic reinforcement of soil slopes. MS.c. Thesis. Tehran University (In Persian). Jackson, A. 1998. An investigation into the vertical geosynthetic reinforcement of soil slopes. Ph.D. Thesis, Manchester University. Jewell, R.A. 1982 A limit equilibrium design method for reinforced embankments on soft foundations. Proceeditigs of 2"" Internatioiiul Conference on Geotextiles, Las Vegas, 3. Jewell, R.A. 1980a. Some effects of reinforcement on the mechanical behavior of soils. Ph.D. Thesis, University of Cambridge. McGown, A., Andrawes , K.Z. and Al-Hasani, M. 1978. Effect of inclusion properties on the behaviour of sand. Geotechnique, 28, No. 3, 327-346.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Countermeasure against the slope using difference of shearing mechanism on main slip layer Mitsuhiko Mukaitani Takanzatsu National College of Technology, Kagawa, Japan
Masamichi Hori & Nobuyuki Kobayashi Advanced course, Takamatsu National College of Technology, Kagawa, Japan
ABSTRACT: Most landslides and slope failures are caused by the shear strength on main slip layer. Then, the shearing mechanisms of landslide soils are important. This paper treats the modeling of slope by view point of topographic and geological studies and the peak and residual shearing resistance. These soil samples are artificial soils in the laboratory. Firstly, we considered topographic and geological studies on various types of slope stability and clarified the relation between the shearing resistance angle and type of slope. We suggested the idea for design of countermeasure considering difference of shearing mechanism on the slope. Finally, we investigated the effect of sand or gravel fraction on main and interslice slip surfaces of slope. slides in fractured zone and landslides in solfataric soil. As shown in Figure 1, the Median Tectonic Line (called MTL) runs in east-west direction in Shikoku Island. MTL is the active fault. Same tectonic line runs in Kyushu Island, too. The fractured zone spreads at the southern part of the MTL. In this zone, many landslides take place in the weathered slopes of black green and other rocks. MTL is not only main factor of earthquake problems, but also landslide ones.
1 INTRODUCTION This paper treats the shearing mechanism of landslide soil using artificial gravelly soil in laboratory. There are many landslides in fractured zone in Shikoku Island, Japan. The slip layer of these landslides has the gravel fraction including fine ones. In Japan, landslides are often classified for their geology into three groups, landslides in Tertiary deposit, land-
Figure 1. Shikoku Island and the fractured zone. (Takahashi, 2000)
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Then, disturbed samples of artificial soils were obtained in laboratory. Ring shear tests were carried out on these samples under drained condition. The test results are discussed from the viewpoint of shearing mechanism with comparing to landslide at in-situ.
Ground surface-
2 TOPOGRAPHIIC CLASSIFICASIONS OF LANDSLIDE Many factors effect on safety factor of the landslide. Landslides are divided into four groups in topographic classifications. It is very important that strength paralneters is connected with the factors by another study area. In slope designing, stability analysis is the most remarkable work. In stability analysis, it is necessary to calculate using the strength parameters and the present shape of slope which is not affected by the historical movement. The schematic diagram of slope is shown in Figure 2. The relation between classifications and profiles of landslide is shown in Figure 3. (1) It is assumed that the main slip surface and the inter slice surface are mobilized the peak strength d, at the primary slide in landslide soil mass movement. (2) In latent slide, the inter slice surface is mobilized the peak strength and the main slip surface is intermediate between the peak strength and the residual strength d, r. (3) In active slide, the inter slice surface is mobilized intermediate between the peak strength and the residual strength and the main slip surface is d, r. (4) In ancient slide, the main slip surface and the inter slice surface are mobilized d, ,.
Figure 2. Description of interslice and main slip surfaces.
3 GEOLOGICAL CLASSIFICASIONS OF LANDSLIDE Okuzono (1 987) suggested the geological classifications of landslide from the view point of collapsible structure as shown in Figure 4. The main target of this figure is the artificial cut slopes. Slope failures can be classified into three groups, which are rock fall, surface failure and landslide. The relation between classifications and $ ' is shown in Figure 4. (1) circular slip by clayey soil: These failures are occurred by homogeneous clayey soil. If the soil mass of slope is made by sandy clay or decomposed granite soil, slope failure occurs. In this case, only main slip surface affects the stability of soil mass. The main slip surface is mobilized intermediate between $ ' and d, r. The interslice surface is mobilized $ I.
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Figure 3. Schematic diagram of topographic classifications of landslide. Weno, 1997)
Figure 4. (continued) Schematic diagram of geological classifications of landslide. (Okuzono, 1987).
(4) dip slope structure: In this case, movement soil mass of slope has the alternation strata of sandstone and shale. Even if the shale layer is very thin, main slip surface is on between these layers. The unconformity, schistosity and joint are included into this group from geological view point. The main slip surface is mobilized $,.. The interslice surface is mobilized intermediate between d, ' and 4 r. ( 5 ) toppling failure by stratum of opposite dip: The development of fault, unconformity, schistosity and joint affects the stability of slope. If these weak layers are connected together, slope failure occurs called 'creep type failure', 'toppling' and 'wedge failure'. The main surface is mobilized intermediate between (b ' and $ r. The interslice slip surface is mobilized (b r. (6) sliding along the fractured zone: The shear strength of the slope with fractured zone is low value. So, slope failure occurs very easily. If the thickness of fractured zone is thin layer, failure pattern is similar to above mentioned type 5. The main slip surface is mobilized d, r. The interslice surface is mobilized intermediate between $ ' and (b r.
Figure 4. Schematic diagram of geological classifications of landslide. (Okuzono, 1987)
(2) unconformity trap slip by difference of permeability: If porous soil mass of slope is on base rock which has low permeability, slope failure occurs with piping of seepage ground water. The main slip surface is mobilized intermediate d, r. The inter slice surface is mobilized intermediate between $ ' and ( b r . (3) re-sliding along old slip surface: This failure occurs along the old slip surface and another slip surface whose scale is small. The main slip surface is mobilized 6 ,. The interslice surface is mobilized intermediate between 4 ' and 4, r.
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4 EFFECT OF SHEAR MECHANISM ON SHEARING RESISTANCE ANGLE The maximum angle of shearing resistance ( b ' in terms of effective stress is corresponding to the peak strength by triaxial test. The minimum angle of shearing resistance d, in terms of effective stress is to the residual strength by ring shear test. Yagi et al. (1985) made two kinds of ring shear apparatus. One is a box shear type (called Bishop type) in which a specimen is sheared along the one plane thickness as shown in Figure 5(a). Figure 6 shows the schematic
diagram of direct shear and similar failure mode at in-situ. Another is a simple shear type in which a specimen is sheared uniformly throughout the thickness as shown in Figure 5(b). The reason why they made the simple shear type ring shear apparatus is that a slicken side is not always found in a sliding layer but a slip band is formed in above mentioned many case. Figure 7 shows the schematic diagram of simple shear and similar failure mode at in-situ. The influence of difference of shearing mechanism using two kinds of ring shear test is to evaluate the condition of main and interslice slip surfaces. The effect of shear mechanism on shearing resistance angle is observed in Figure 8 on $ r of remoulded landslide clayey soil obtained by the triaxial compression test. The soil was weathered serpentine obtained by Shimotsu, Wakayama prefecture, Japan. The soil samples were obtained by mix-
Figure 7. Schematic diagram of simple shear and similar failure mode at in-situ.
ing clay fraction and gravel fraction in various ratios. In Figure 8, the d, increases gradually with increase of gravel fraction. However, as mentioned above, the difference of test apparatuses on restriction for slip surface, can be also observed. Figure 8 shows the relation of 6 and sand or gravel fraction for the same samples, but obtained by newly developed “simple shear” type ring shear test apparatus shown in Figure 5. The slip surface is not made before shear and is freely selected. In Figure 8, the 6 does not increase until greater than 50% of gravel fraction.
Figure 5(b). Simple shear type ring shear apparatus.
688
Figure 9. Relation between e , and
6'
6 APPLICATIONS FOR DESIGNING OF COUNTERMEASURE
Figure 8. Effect of shearing mechanism on
0, r.
These results indicate that (b is influenced by the large grain size fraction, but the influence depends on the test apparatus or the restriction for the slip surface.
5 DESIGNING FLOW CONSIDERING DIFFERENCE OF SHEAR MECHANISM In designing, the choice of (b r affects the safety factor and the countermeasure with cost. If the slip surface of landslide is observed as very thin plane of a few mm thickness, we must use 4 r by ordinary type ring shear tests. These slip surfaces are called by 'slicken side' at in-situ. If the slip surface of landslide is observed as the layer of a few cm thickness or fractured fault zone, we must check the possibility of Q, r by simple shear type ring shear tests. These landslides are including initially movement slope in topography. In sand or gravel material, void ratio is most important factor. In clay material, water content is most important factor. The clayey soil including sand or gravel is needed the new viewpoint. Figure 9 shows the relation of e g and 4 ' by triaxial compression tests. For eg less than 3, the apparent 4 ' is nearly equal to that of sand or gravel, but for less than 3, 4 ' is equal to that of clay. The e g is void ratio in terms of sand or gravel and defined as following equation (1). e g =(Vw + Vclay)/(V,smzdor gravel)
We discussed the relation between topographic and geological view points and the variation from peak strength to residual one. Nevertheless, a few factors are considered to calculate the safety factor on the slope. We must decide the proper strength parameters based on laboratory tests, topographic and geological studies. The infinite slope method is most simply consideration on stability problem seen in Figure 10. The stability of the only one soil's block is related with the Coulomb's failure criterion. This method is able to predict the effect of countermeasure on the slopebased on the equilibrium of forces without analytical calculations. (1) drainage of main slip surface: The stress condition on main slip surface must be decrease the value of horizontal distance against the apparent failure criterion seen in Figure 1 1. (2) shear resistance type of pile work: The stress condition on main slip surface must be decrease the vertical distance, which is equal to the force using pile or other techniques par unit area, against the apparent failure criterion. seen in Figure 12. (3) anchoring: The stress condition on main slip surface must be decrease the combined force seen in Figure 13, which is equal to the force using anchoring par unit area, against the apparent failure criterion.
H
(1)
The new factor of eg has possibility of explanation of state for slip surface condition of landslide soil including sand or gravel.
Figure 10. Schematic diagram of finite slope model.
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grain size of sample, but the influence depends on the test apparatus.
8 ACKNOWI.,EDGEMENTS
Figure 1 1. Estimation of drainage value on relation between 0' and x.
The authors would like to express their gratitude to undergraduate students in Takamatsu N.C.T. for their help in the preparation of this paper. This research was partially supported by the Ministry of Education, Science, Sports and Culture, Grant-inAid for Encouragement of Young Scientists, No. 1 1750448, 1999-2000. REFERENCES
Figure 12. Estimation of needed force value on relation between o'and x.
Figure 13. Estimation of needed anchoring force value on relation between o'and x.
7 CONCLUSIONS
In designing, it is important to consider the difference of shearing mechanism of main and interslice slip surfaces of slope in terms of effective stress. The effective stress concept may be applicable to slope regardless of the contact mode of slip surface ( 4 ': triaxial compression test, 4 r: direct shear type or simple one). The 6 'and 4 are influenced by the
Enoki Meiketsu, Yagi N. and Yatabe R. 1989. The mechanism of reinforcement in tension in sand. Memories of Faculty of Engineering. Eliiine university. Vol. 11. No. 4: 44 1-449. (In Japanese) Enoki Meiketsu, Yagi N. and Yatabe R. 1991. Stability Analysis Method of Natural Slope. Memories of Faculty of Engineering. Ehinze university. Vol. 1 1. No. 2: 425-432. (In Japanese) Gray D. H. 1974. Reinforcement and stabilization of soil by vegetation. ASCE. Journal of Geotechnical engineering divisioiz. Vol. 100. No. GT6: 695-699. Ishii Tomonori, Yatabe R., Yagi N.and Yokota K. 1999. Influence of clay minerals on strength characteristics of landslide clay in Mikabu. Proceedings of the international sytnposiunz on slope stability engineering: 697-700. Japan. Takahashi Jiro. 2000. Mass movement and erosion cycle, Proceedings of con$ on Landslide and Slope,failure:49-54.(In Japanese) Ueno Shouji. 1996. The method of application of topographical and geological information for geothechnical engineers, (part 5 ) Landslide. Journal ofJGS. Vol. 44. No. 6: 51-56. Ueno Shouji. 1997. Study of configuration, scale and distribution of landslides for countermeasure. Text of regional geothechnical investigation in Siga pref: 7-27. (In Japanese) Ueno Shouji. 1999. Study of configuration, scale and distribution of landslides. Proceedings of tlie international synzposium on slope stability engineering: 175-180. Japan. Yagi Norio and Yatabe R. 1985. A Microscopic Consideration on Shearing Characteristics of Decomposed Granite Soil. Journal ofJSCE. No. 364: 131-141. (In Japanese) Yagi Norio, Enoki M. and Yatabe R. 1992. Reinforcing Mechanism of Sandy Soil by Roots of Plans. 1991. Proceedings of27t1i Annual Conference of Soil Mechanics arid Foundation Society of Japan: 1865-1866. (In Japanese) Yagi Norio, Yatabe R. and Enoki M. 1994. The Effects of Root Networks on Slope Stability. Proceedings of Int. Conf on Landslide, Slope stability and Safety of Infra-structure: 387-392. Malaysia. Yamashita H, Saga M., Fujita H., Yokota K. and Yatabe R. 1999. The characteristics of landslides caused by the hydrothermal metamorphic clay. Proceedings of the international symposium on slope stability engineering: 693-696. Japan. Yatabe Ryuichi, Yagi N. and Enoki M. 1991. Consideration from effective stress about strength parameters of slip layer clay of landslide. Journal ofthe Japanese Society of Landslide. Vol. 28. No. 2: 20-26. (In Japanese).
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeit/inger, lSBN 902651 863 3
Optimum design of nailed soil slopes C.R. Patra Civil Engg. Dept., R.E. C., Rourkela, Orissa, India
769005
I
P.K. Rasudhar Civil Eiigg. Dept., I.I.T., Kanpuc U.P., India - 205016
ABSTRACT This paper pertains to the dcvelopment of a new method of automated optimum design of soil nailed slopes by extending the limit equilibrium formulation, considering the effect of tensile resistance of the reinforcement Both overall and internal equilibrium are considcred in computing the stability of nailed slopes Thc quantity of steel requirement for raising the fnctor of sifety to a desired value is minimired with respect to the loution. sire (length and diametei), oiientation and the location and shape ul the ~iiiicalshea surface The solutions have been isolated by formulating the problem as one of non-linearing programrmng The applicability of the developed method has been verified by conipaing the predicted failuie surfaces with those reported model test results and diameter) and orientation of nails and the location and shape of the critical shear surface which has not been attempted so far.
1 INTRODUCTION Several methods arc being used for analysis and design of nailed slopes. such as the Davis method (Mitchcl and Schlosser, l979), the German method (Stockei et al , 1979), the French method (Schlosser, 1982). the method dcveloped by Glasser and Gudehus (1 981) and the kinematic limit analysis appioach (Juran et al , 1990) The nails can be placed either homontally or inclined Lesniewska (1992) has shown that the incfination of reinforcement plays a signifitnnt role in modifying the stability of nailed structures Optimal placement of reinforcement for minirmzing cost of construction has drawn the attention of researches Anthoine and Salencoii (1989) have designed the optimal location and length of nails consideiing a single layer of reinforcement. Sabahit et a1 (1995) have developed a generali7ed tnethod based on Janbu's generalized procedure of slices (1973) for the optimal design of nailed slopes. Neglecting the soil-nail interaction in the active zone and considering only thc overnll equilibrium they have estimated the totat rernforcement foice iequired to raise the factor of safety to d desired value by treating the orientation of nalc and dirtnbution of reinforcemcnt forccs as design variables The studies regarding the optimal location, length, diameter and orientation of n a k in a naled soil slope have not been given due attention Hence, in this paper, such a study has been undertaken and examined considering both internal and overall equilibrium of the slices which should not be ignored in such shidy The quantity of steel requiied for raising the factor of safety to a desired value is mnirmied with respect to the location, siie (length
2 STATEMENT OF THE PROBLEM
Figure 1 shows a reinforced slope (Height H, slope angle p) with a tension crack at 'a' and subjected to external loadings (Ta, E,, TI,,Eh. P. q, Q). The tension crack may be filled with water. The assumed general slip surface, the line of thrust and a typical slice are also shown in the same figure. In contrast in the present formulation a more generalized case of force mobilization has been considered. The forces considered in each slice are shown in Figures 2(a) & 2 (b) rcspcctively. To avoid confusion, the various possibilities of nail position are shuwn separately in those two diagrams (Figures 2(a) and 2(b)) instead of crowding it in one diagram only. B and T are the intcrslice normal and shear forces respectively; q and zj are the normal and shear stresses on the base of a typical slice respectively. In the same figures, Ax, is the slice width, hi - slice height; ai - angle made by the base of the slice with the horizontal; ALi - base width of the slice, APi the external load, qi - surface load and AQi is the earthquake force; (AWr)l- weight o f the slice; (h,ji assumed position of thrust line; (z,)i -assumed position of the seisiiiic force (AQj), aij - the perpendicular distance of the point of application of j"' reinforcement force with respect to the point '0' in ith slice, e, is the inclination of jthreinforcement with the horizontal. The notations for the force systems as shown in Fig. 2(a) and 2(b) have been chosen for the present I
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formulation of considering internal equilibrium in addition to overall equilibrium. The objective is to design the slope for a desired factor of safety using minimum amount of steel for the nails which will lower the cost of the project. As such, it is necessary to derive expressions for the factor of safety and the amount of steel required for nailing to achieve the desired factor of safety and to determine the most general critical slip surface without any a-priori assumption regarding its shape. Thus, it is necessary to identify the objective function, the design constraints (both side constraints and behaviour constraints) so that the obtained slip surface is kinematically possible and the solution obtained is physically meaningful.
Figure 1 . Definition sketch of nailed soil slope.
3 FORMULATION The general formulation of the problem stating identification of the design variables, the objective function and its derivation, the design constraints, are discussed and presented. The factor of safety of a nailed slope has been derived considering both internal and overall equilibrium (Patra, 1998). Considering the static equilibrium of the slice [Figures 2(a) and 2(b)], the following expression for the factor of safety (F) has been obtained.
F=
i=l
n
i=l
i=l
where
Figure 2(a). Forces acting on a typical slice.
A, = A , , f q i 1 +tan qY tanai / F q. = I + tan’ a, A,! = [cl+(p, + t, - u)tan q
~1a-y~~~
B, = A Q ~+ ( p i + ~ ; ) A x , stan , ai
+ Figure 2(b). Forces acting on a typical slice.
692
[Fbs ji, = traction force at base in jthnail which cuts both base and (i + lYh interslice face of ith slice
hxisH
= traction force at base of it” slice in jth
hxiSH
nail, which cuts both slope face and base of ithslice = traction force at ground of ith slice in jth
nail, which cuts both slope face and base of ithslice. The interslice normal and shear forces are determined for each slice. The expressions for the same are not presented for the sake of space and bravity.
hisH where,
(12)
)/AxisH
ti = [Ti -Ti-l
Total volume of nails =
Where, (L,)j = total length of jth nail, (LJj = Length of jth nail in the active zone, (I), = Length of jth nail in theresistive zone and (dn)j= diameter of jthnail. The design vector DT7is
D ~ (ci =
In the Figure the following notations have been used for the force systems if the internal equilibrium is adopted in the present analysis. (Fgs)i, j
=traction force at slope face in jfhnail
\
LFsg )i, j
= traction force at ithinterslice face in
,[d
j
n 2
,e2,............(e),r ,
,........(d ,I ,eI n nr
(L,)
,(L r ) ,............................... (51
Where
. [dn l .(dyl1, ............(LIn nr are the diameter
which cuts both ground and ith interslice face ,
j
fZ1
y1y
(15)
of the nails,
j f hnail which cuts both slope face and ith interslice face
e p 2,..................... ( e )n,”are orientation of nails,
ji
( F ~+ 1,~ = traction force at ( i + I ) ” ~ interslice face in jth nail which cuts both vertical faces of ith slice.
are location of nails from top of slope, x,and XL are initial and final x-co-ordinates of slip surface Z2, ..........ZIl.l are the z-co-ordinates of the slip surface at point 2,.3,. ... n-1, while n = number of slices used , n, = number of reinforcement used
[F J. . = traction force at (i + 1)lh sb 2 , J
(L,
interslice face
)
nails.
in jthnail which cuts both base and (i + l)lhinterslice face of ith slice
693
( )
,( hn ) ......... L r
ylr
are resistive lengths of
Objective function F(D) in terms of design vector DT is
F (D) = f (DT) (16) The dual-objective function of the minimum volume of steel required to achieve the factor of safety of the slope to a desired value can be reduced to a single objective function chosen in the following form : F ( D ) = vn + s l(F -
f
where, S f = Scale factor and Fd = desired factor of safety. The following design constraints have been imposed for the physical and meaningful solutions. The diameter of nails should lie in an prescribed interval. Minimum and maximum vertical spacing of nails are specified within prescribed limits. Intersection of nails are avoided in the area of resistive zone of consecutive nails. The resistive length of each nail (considered as a design variable) should be positive. The resistive force in each nail (Pr)jwith a factor of safety (fr) is greater or equal to the mobilized tension (Fb), in nail at the point of intersection with slip surface. Normal and shear stresses generated at the base of the slice should be positive to avoid generation of tension in the slope and inconsistent direction of shear. Safety factor along interslice face should be greater or equal to average safety factor. The critical surface should be concave when looked from the top. Average factor of safety should be greater or equal to the desired factor of safety.
where rk is a penalty parameter, whose value is made successively smaller in order to obtain the constrained minimization of F(D). 4 RESULTS AND DISCUSSION
In nailed soil structures, it is very important to predict the location of the slip surfaces correctly. The potential of the present method has been demonstrated by comparing the predicated slip surfaces with those observed by Kitamura et al., (1988) from tests conducted on model reinforced slope. The model was made by compacting 15 cm thick layers of sand in a steel box, placing the reinforcing members in locations before compaction. The model slope had a width of 900 mm, a height of 750 mm, and length of 2100 mm. The front and side gradients of the slope where kept as 1 :0.3 and 1:0.6 The properties of the soil used for the test are as follows: Maximum particle size: 29 mm, gravel content: 2496, fine particle content: 13%, specific gravity of particles: 2.65, natural mositure content: 1 1 % unit weight: 20 kN/M', void ratio: 0.47, cohesion intercept: 15 kN/m3, angle of shearing resistance: 33". Aluminium reinforcing strips of 450 mm long, 25 mm wide and 2 mm thick were used. Young's modulus of aluminium foil was 7.03 x 10' kPa. The vertical and horizontal spacings of the reinforcements were 0.225 m. The reinforcing elements were placed horizontally. The maximum surface load to cause failure is reported to be 290 kN/m3. The predicted critical slip surfaces for the case of horizontal reinforcements are presented in Fig. 3 along with those predicted by Sabahit (1994) and observed by Kitamura et al., 1988. The observed slip surface passes above the toe and almost at the level where the lower most nail has
The above design constraints have been expressed in the standard mathematical expression for the minimization procedure. 3.1 Minimization procedure : The optimal solution is obtained by converting the constrained problem to an unconstrained one by blending the objective function with the constraints to develop a composite function. Sequential unconstrained minimization of the composite function so obtained is carried out using Powell's conjugate direction method for multidimensional search in conjunction with quadratic interpolation technique for unidirectional search.
j-1
J
Figure 3. Comparison of critical surfaces.
694
been inserted. Sabahit (1994) predicts a zone of failure surfaces for this case. From his analysis it is evident that except for the lowermost failure surface, the lowermost nail is ineffective. In the present method, (considering internal and overall equilibrium) the most part of the predicted surface lies within the soil mass bounded by observed failure surface (Kitamura et al., 1988). But, at the lowerpart, the predicted surface is much below the lowermost nails. The lowermost nail is therefore effective and contributes to the improvement of stability of slope by reinforcing action. The total length of nails and the individual length of nails predicted by using the present method are not the same as used in model tests or as predicted by Sabahit (1994). More resistive length will be required in the upper part of the slope as compared to the lower part as the tensile forces mobilized in nails at the point of intersection of nails with the critical surface are more at the upper part of the slope. Accordingly for fixed orientation and size of reinforcements, the inclusion length to be provided to achieve the desired factor of safety is different. Hence for this slope, with horizontal reinforcements, the revised length of the individual nails has been computed for the factor of safety equal to unity and presented in Table 1. Even if more nail length is provided than the required one, the factor of safety cannot be increased further, because for a given size and orientation of inclusion, the tensile force mobilized depends only on its length in active zone. Table 1. Comparison of predicted nail length No
Provided length (In)
___-
~
1. 2. ~
_______ 3.
Kkamura et.al., (1988)-_0.45 0.45 - -
0.45-_
-~
Computed length (m) PrEent Mehtod 0.76 0.58 0.15 __ - -
5 CONCLUSIONS Based on the results and discussion presented the following conslusions are drawn: Conversion of the multi objective problem to a single objective problem and introduction of a scaling factor proved to be successful. Consideration of internal equilibrium in addition to overall equilibrium significantly alters the position of the critical
695
slip surface and thus affects the nail volume required to achieve a desired factor of safety. Other parameters remaining the same, there exists a certain length of individual nail in resistive zone beyond which the stability of a given slope can not be increased further. REFERENCES Anthoine, A. and Salencon, J. 1989. Optimization of reinforced soil structure design, Proc. of the 12"' ICSMFE, 3, 12191220. Cartier, G. and Gigam, J.P. 1983. Experiments and observations on soil nailing structure, Proc. of the 8"' Coiif qf the ECSMFE, 2, Helsinki, pp. 473 - 476. Gassler, G. and Gudehus, G. 1981. Soil nailing- Some aspects of a new technique, Proc. Tenth ICSMFE,, Stockholm, pp. 665-670. Janbu, N. 1973. Slope stability computation, in: R.C. Hirchfield and S.J. Poulous (eds) Embamkment dam Engineering, Casagrande Vol, Wiley, Newyork, pp. 47-89. Juran, I. Baudrand, G., Farrag, K. and Elias, V. 1990. Design of soil nailed retaining structures, Design and performance of earth retaining Structures, Proc. of the ASCE, Geotechnical Speial Publication No. 25, pp. 644 -659. Kitamura, T., Nagao, A. and Uehara, S . 1988, Model loading tests of reinforced slope with steel bars, Proc. of the Iizt. Geotech. Symp. on Theory and Practice of Earth Reinforceinent, Kyushu, Japan, pp. 3 1 1-316. Mitchell, J.K. and Schlosser, F. 1979. General ReportMechanism and Behaviour - Design methods, Iiit. Con$ on Soil Reinforcenieiit, Paris, 3,2562. Patra, C.R. 1998. Sequential Unconstrained Minimization Technique in the Optimum Design of Slopes with or without Nails, Ph. D. Thesis submitted to Indian Institute of Technology, Kanpur, India. Patra, C.R. and Basudhar, P.K. 1997. Stability computations in nailed slopes, Third Ititernational Conference on Grouiid Improvement Geosysteins, Densification and Reiilforcernent, British Geotecliiiical Society, U.K. Powell, G.E. and Watkins, A.T. 1990. Improvement of marginally stable existing slopes by soil nailing in Hong Kong, Proc. of the Int. Reinforced Soil Conf., Glasgow. pp. 241247. Sabahit, N. 1994. Stability analysis of reinforced slope - A nonlinear programming approach, Ph. D. Thesis submitted to Indian Institute of Technology, Kanpur, India. Sabahit, N., Basudhar, P.K. and Madhav, M.R. 1995. A Generalized procedure for the optimum design of nailed soil slopes, hit. J. for Numerical arid Analytical Methods in Geoniechaizies, Vol. 19, pp. 437-452. Schlosser, F. 1982. Behaviour and Design of soil-Nailing, Symp. oii Recent Developments in Ground Improvement Techniques, Baiigkok, 399-41 3. Stocker, M.F., Korber, G.W., Gassler, G. and Gudehus, G. 1979. Soil Nailing. bit. Col$ On Soil Reinforceinent, Paris, 2, pp. 469 - 474.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Innovative solution for a Kei Cutting problem G.V. Price Jeffares & Green, East London, South Africa
A.C.S. Smith Reinforced Earth (Pty) Ltd, Johannesburg, South Africa
A. Muhajer South African National Roads Agency, Southern Region, South Africa
ABSTRACT: An innovative concept for the widening of fills over an existing sidefill has been implemented on the National Route 2 (N2) across the Great Kei River in the Eastern Cape Province of South Africa. Ten retaining walls have been constructed in which soil nails are combined with a mechanically stabilized embankment (MSE) to allow for the accommodation of traffic during construction, widening of the fills and compIiance with overall stability requirements. This paper describes the reasons for the selection of this solution and the design and construction thereof. The cost and technical merits of this hybrid structure are compared to alternative solutions. The conclusion reached is that this solution economically satisfies the technical and environmental requirements and has potential to be used more extensively for widening projects in the future. 1 EMBANKMENT WIDENING STABILIZATIONREQUIREMENTS
over quite a large range from as little as 23" to as steep as 49".
1.1 Introduction
1.3 In situ conditions
Upgrading of National Route 2 across the Great Kei River, Eastern Cape of South Africa comprises a multi-million dollar project in which a narrow, torturous route is widened from two lanes to three lanes. The area presents a rugged topography through a sinuous valley and on slopes that are steep and already close to their natural angle of repose. This terrain and poor alignment of the existing route, combined with the high rainfall of the area, presents a number of challenges of which the realignment and widening of sections of the road out and over the existing fill is the subject of this paper.
Investigations for the project included borehole drilling and trial pitting. Field tests were conducted on site and samples were removed for laboratory testing. Laboratory results indicated material with a cohesion (kPa) that varied between 2 and 22 kPa and a friction angle (") varying between 29 and 38". A sensitivity analysis indicated that the angle of friction (@)to be used should be 32" with a cohesion (c) of 5 kPa. Stability analyses of existing slopes were undertaken using the Bisjan programme, which assesses the stability using the methods of both Bishop and Janbu. Stability analyses using these results indicated that the existing embankment slopes were only marginally stable and on the brink of failure. Decreasing the phreatic surface to beneath the embankment fill results in a Factor of Safety of 1.1. However, when the phreatic surface is allowed to climb into the fill the factors of safety drop dramatically. By implication, these slopes in their preconstruction state were either unstable or bordering on instability. Fortunately, investigations indicated that the embankment material was fairly porous with interconnected voids that allowed free passage of excess water out of the system. Ironically this occurred as a result of a previous lack of compaction during construction of the original embankment some 3 or 4 decades ago. The fact that these slopes were already in a marginal state with respect to stability made the option of widening even more difficult.
1.2 Requirements The northern side of what is locally known as the Kei Cuttings, comprises a road characterised by high, steep cliffs on the upslope side of the road and long relatively steep embankment fills on the downslope side. The existing embankment slopes are relatively devoid of vegetal cover as a result of the inert nature of the fill material, and the lack of topsoil. This torturous, steep and dangerous section of the road requires widening to enable safe passing and adequate sight distance. Cutting into the high upslope cliffs would be both impractical and costly so that any widening would have to take place on the downslope side. This makes for difficult geotechnical engineering as the existing embankment slopes are on average 32" steep, although these do vary
697
irements. In order to accommodate a single lane of traffic, temporary shoring of the excavation for the TerraTrel@mass was envisaged.
After considering all the information to hand, and stability analyses undertaken, an MSE structure was deemed to be the optimum solution and Reinforced Earth (Pty) Ltd was nominated to undertake the work.
2 ACTUAL SOLUTION SELECTED 1.4 Scheme alternatives 2.1 The problem
These included inter alia options such as: 0 0
0
0
Cantilever pile walls with lagging Contiguous pile walls Secant pile walls Reinforced concrete retaining walls Crib walls Gabions Soil reinforcement (MSE, soil nailing).
MSE was chosen for this project because it was cost effective, did not require rock foundations and could accommodate differential settlements. The construction technique was simple and labour intensive, reducing the requirement of large construction machinery on the steep slopes. Reinforced Earth@was chosen as being the best suited MSE system for this particular task.
1.5 Original tender design Reinforced Earth's TerraTrel@system was specified in the tender documents. In order to enhance the overall stability of the existing fill, even when surcharged with this new MSE structure, the reinforcing strip length was extended until it nicked into the underlying rock slope as shown in Figure 1. The type of structure specified was economical and flexible enabling it to adjust to settlement of the underlying fill. In addition it would provide a pleasing natural rock appearance satisfying environmental requ-
Figure 1. Typical cross section of the specified MSE solution.
698
On award of the contract, it was decided that twoway traffic needed to be maintained during the construction process. The specified TerraTrel@ structures were too wide and massive to accommodate two-way traffic as shown in Figure I and an alternative system for the retaining walls, which would meet all the requirements, was urgently required. This left an unenviable situation for the embankment widening designers. First of all the original embankment would have to be rehabilitated to a factor of safety of at least 1.3. This is in terms of Client requirements for safe embankment design. The 1989 Code of Practice for Lateral Support of Surface Excavations too, as required by the Geotechnical Division of the South African Institution of Civil Engineers, states..... "It is generally accepted that the factor of safety should not be less than 1.5 for permanent work". This meant cutting into the existing embankment slopes, stabilising them to a factor of safety between 1.3 and 1.5, depending on the moisture regime and knowledge of the insitu materials, and once this open cut has been stabilised, to then construct the widening using an MSE structure. So not only did the original embankment have to be improved to a much higher factor of safety, but also some method was required whereby the two systems could be tied together, and the overall stability of the entire combination guaranteed safety factor values in excess of 1.3 and possibly I .5.
2.2 The solution - combined soil nail and MSE structure Following discussions between all parties an innovative and economical solution was proposed. This solution required a combination of a nailed system and a much narrower MSE mass. A typical cross-section of the composite soil nail and MSE structure is shown in Figure 2. This solution aIlows for the accommodation of two-way traffic; limits the excavation required for the MSE mass; satisfies the internal stability requirements of both the temporary nailed structure and the MSE structure, the composite nailed / MSE structure and also the overall stability requirements of the widened embankment. Finally, it provides a pleasing natural visual appearance.
2.3 The MSE system An MSE structure is a composite structure comprising frictional earth, reinforcing strips and a cladding.
In this project the TerraTrelO system comprised a backfill which was a completely, highly weathered dolerite with 100% passing a 5 mm sieve and 10% by mass passing a 20 micron sieve. For design purposes an internal angle of friction of 36" and a cohesion of zero was assumed for this material. The pH of the backfill was 6.5 and the material easily complied with electro-chemical characteristics required to ensure the design service life of 100 years. The reinforcing strips were hot dip galvanized medium tensile ribbed steel strips, with cross-section 60mm x 4 mm. The length of the strip is dependent on the height of the wall. Only 3.5 m and 4 m strip lengths were used on this project for structures up to 8 m high. The cladding is shown in Figure 3. It comprises a heavy hot dip galvanized weldmesh made up of 8 mm, 10 mm and 12 mm bars with a 100 mm by 100 mm opening. The elements are bound together by tie points, onto which the reinforcing strips are bolted. The elements are de-
Figure 2. Typical cross section of the composite soil nail / MSE structure.
Figure 3. The TerraTrel@cladding.
699
the overall stability of the embankment to a depth of three times the height of the wall was applied to determine the factor of safety of the slope in the region of the retaining walls. The walls vary in height from 2 panels high (1.52m) to 10 panels high (7.28m). The slope stability program Talren was used to perform the analysis, and each section was evaluated for values of Q, ranging from 30" to 36". The cohesion was assumed to be 5 kPa in the original fill and 10 kPa in the rock. An example of the analysis of the overall stability of the road where the retaining structure is 6 panels or 4.4 m high is illustrated in Figure 5. Two layers of soil nails, spaced at 1.7m horizontal intervals, have been provided at this section. The reinforcing strips are 4.0m long, with the number of reinforcing strips per 3m width of wall varying between 3 and 5. Figure 5 gives the factors of safety for overall stability of this typical section for values of the angle of internal friction (@)ranging from 30" to 36". The factors range from 1.22 to 1.43, indicating that it is stable. Since the original slope of the ground beneath the road was only marginally stable the adopted solution has improved the overall stability of the existing embankment in the vicinity of the retaining structures.
signed to concertina rather than bulge should internal settlement of the fill immediately behind the cladding occur. In order to prevent the backfill spilling through the weldmesh it is backed with a durable rock which in turn is backed with a geofabric. The TerraTrelO system was introduced into Southern Africa in 1988 and the particular type of system described above has been used since 1995. 3 DESIGN 3.1 The MSE On account of the space constraints, the MSE structures are designed to be as narrow as possible. The maximum strip length is 4 m and this, in many areas, is only 50% of the overall height of the structure. The internal stability of the wall is designed accord ing to the standard local equilibrium method and satisfies reinforcing strip strength, friction and durability requirements. 3.2 The nails The nailed structure was designed to satisfy two requirements, viz. - short term stability of the existing road while construction of the MSE was underway - overall stability of the combined MSE / soil nail structure.
4 CONSTRUCTION
The nails used were 25 mm diameter threaded bars, 3.5 m to 7.0 m long. The top row of nails was installed approximately 2 m below the top of the excavation. The middle and bottom rows were positioned at 2 m vertical intervals beneath the top row. The horizontal spacing varied between 2.0 m and 1.5 m. All nails that were to be extended through the MSE mass were designed as permanent nails, the others, together with the permanent nails, were only required to temporarily support the road until the MSE was constructed. 3.3 The combined MSE / nailed structure A slip circle analysis through the nailed and MSE structures indicated that the critical slip would pass between the two structures and overall stability of the widened embankment was much the same as the original embankment. In order to enhance the overall stability of the widened embankment, it was necessary to combine the structures and this was done by extending the nails through the MSE mass. Details of this innovative nail extension solution are shown in Figure 4. The height of the slopes below the retaining walls is much greater than the height of the retaining walls themselves. It is thus unreasonable to expect the retaining walls to improve the factor of safety of the entire slope. A philosophy of limiting the analysis of 700
The position of the first cut for the nailed structure had to be carefully surveyed. The toe of the cut was to be 4 m from the front face of the bottom of the MSE structure and the top of the cut was not to impinge on the existing road surface. In many areas this required a full height near vertical cut, while in others a battered slope from the edge of the road was possible before the excavation had to become near vertical at a position 4 m back from the face line. Immediately after excavation of a lift, a flashcoat of gunite was applied to the face to stabilize it prior to placement of a light mesh and gunite to prevent localized spalling. The first row of nails was positioned approximately 2 m below the shoulder breakpoint or at the position where the batter became vertical. Holes of approximately 100 mm in diameter for the 25 mm diameter threaded nails were drilled. The nails were then positioned into the holes with spacer rings and the holes were then grouted up by trernmying from the back. A headplate and dome washer and nut were placed and the nail cut some 200 rnm proud of the nut. The second and third lifts proceeded in the same way. It was not necessary to mesh and gunite beneath the lowermost nail. The MSE construction was then able to commence in the standard way. Levelling pads are placed beneath the cladding and the repetitive cycle,
5 COST ANALYSIS
of placement of backfill, cladding and reinforcing strip, is begun. All construction takes place from behind the cladding which is a definite advantage from both construction and environmental points of view. When the level of backfill reached the first level of soil nails then the designated permanent nails were extended through the MSE mass by means of couplers, cranked nails and additional headplates, nuts and dome washers fastened outside the MSE cladding.
When compared to the other alternatives investigated and discussed in the first part of this paper it is found that the hybrid soil nail and Reinforced Earth solution was approximately 10% more economical than the next most economical solution and approximately 35% less than the most expensive solution. It was therefore not only an innovative way of solving a difficult problem but, as it turns out, was also a relatively economical solution.
Detail 1 . At the cut face.
Detail 2. At the MSE cladding.
Figure 4. Detail of the nail extension.
r’” Units :kN m e t m and degrees Calculation method :Bishop1
13
14
Figure 5. Overall stability analysis for 4.4 m high MSE structure.
70 1
Figure 6. A view of a completed structure.
Figure 7. MSE construction and nail extension underway.
6 CONCLUSION
It is felt that the principle of this solution can be economically applied to road widening projects in general, wherever space is limited and, in particular, when traffic needs to be accommodated on the existing road.
The combined MSE I soil nail solution has proved to be a practical and economical solution for the road widening project. It has also satisfied the technical and environmental requirements of the project.
702
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets 5: Zeitlinger, 1SBN 90 2651 863 3
Numerical analysis of the mechanical behaviour of rigid shallow foundations on geo-reinforced soil strata C. di Prisco, S. Imposirnato & M. Vecchiotti Structural Department Politecnico of Milan, Piazza Leonardo da Viizci 32, Italy
P. Rimoldi Tenax S.p.A. Viganci Brianza (LC),Italy
ABSTRACT: By means of a commercial finite element numerical code, within which perfect elastoplastic constitutive relationships are implemented, the mechanical response of rigid shallow strip foundations placed on geo-reinforced homogeneous dry loose sand strata under centred, eccentric and inclined monotonously increased loads is analysed. The system mechanical behaviour has been compared to those obtained by analysing not unreinforced rigid shallow and deep foundations. The geo-reinforcements seem to induce a marked increase in the ultimate load but not a substantial stiffening of the load-settlementcurves is observed.
1 INTRODUCTION The use of reinforcements beneath rigid shallow foundations in order to increase footing ultimate load is not a recent idea. The type of reinforcement may vary from steel beams or layers (Huang & Tatsuoka 1990) to geogrids or geotextiles (di Prisco et al. 1998). The difficulty of highlighting and understanding the mechanical response of such complex structures is linked to the problem of performing ad hoc experimental tests. In fact, extensive experimental large scale test series are too expensive. Whereas both lg small scale and centrifuge small scale tests, unless in a different manner, are affected by scale effect. For this reason, in this paper, even if we are conscious that the constitutive relationships implemented into the numerical code (as far as both soils and reinforcements are concerned) are very simple (linear and pressure dependent elasticity coupled to a Mohr-Coulomb failure locus), numerical analyses iave been used in order to better clarify the consid:red mechanical problem and not to quantitatively eproduce the experimental reality. In fact from nuaerical results we derived only useful suggestions nd we did not mean to substitute experimental reAts. The mechanical response of rigid strip shallow tundations placed on a homogeneous loose sand ratum will be analysed by considering three global nematic (v,u,@)and three load variables (V,H,M) ‘icov 1977, Nova & Montrasio 1991) (Fig.1). Conrsely to standard limit analysis results (Soni et al.
Figure 1. Rigid strip shallow foundation: kinematic and load variables.
1992), the performed finite element numerical analyses will give us not only the failure locus shape in the V-H-M space, which may be also obtained experimentally, when loose sands are considered, by keeping constant the relative depth of the footing level, but also the mechanical response up to failure. Many different factors have been analysed: the number of reinforcements, their relative depth, the relative density of sand between reinforcements, the strength and the stiffness of reinforcements, the dilatancy angle, etc (Vecchiotti 2001). A commercial finite element code, Internet free available, which is named Tochnog, has been used. The mechanical response of sand was reproduced by means of an elasto perfectly plastic constitutive relationship with an associated flow rule - only in some numerical analyses (dilatancy angle) y < 41’ (internal friction angle) was assumed. The Poisson ratio is assumed to be constant, while the Young modulus to
703
(U/B>l), the failure mechanism i s not influenced by the reinforcement presence. OnIy if U/B<0.25, the ultimate load is slightly increased. On the contrary, the mechanical response changes dramatically if the reinforcement number is inE= creased. In Figure 2 the numerical results, which have been obtained by tjiung into consideration the case characterised by the presence of three equidistant horizontal reinforcements (reinforcement length Table 1. Material parameters. L=B=2m, maximum depth of reinforcements U=l.5 m), are shown (curve (b)). In this case and in the following ones, the sand between the re~~forcei~ent layers is assumed to be dense. If we compare curve 0 (b) and curve (a) relative to the unreinforced case, we may easily observe that the collapse load markedly increases, while the curve stiffness remains approxj~ately unaltered. The numerical results are Geo-reinforcements, whose thickness is kept conmade dimensionless by dividing the vertical load by stant and equal to 1 em, are simulated by means of the ultimate load value V Mobtained ~ ~ with reference the so called truss elements. To these linear elements to the unreinforced footing, while the vertical disan elasto perfectly plastic ~onstitut~ve re~ationshipis placement v by the footing width B. It is interesting associated when they are stressed in tension, while to compare curve (b) to curve (c) which is obtained their stiffness is nil in compression. The interaction by numerically testing a rigid deep footing (D = between the soil and the truss elements is described footing depth = 1.5 m). It is evident that, as far as the by two elastic springs and one ~ o h r - C o ~ ~ I ofricmb u l t ~ ~ i load a t ~ is concerned, the obtained value in the tional dashpot with an associated flow rule ( friction two cases is approximately the same. On the conThe between soil and reinforcement 6 = 2/3 trary, the stiffness of the two curves is very different, calibration of such elements, which define the soilbecause the reinforcement confinement effect seems reinfo~cementinteract~onis quite i ~ p ~ r t aand n t may to take place only when vertical d~sp~acements are be obtained by means of pull-out tests. We imposed, sufficiently high. Such a result is obviously dependfor the sake of simplicity, kN = kT = 00, where the ent on the reinforcement stiffness and spacing. By former is the normal while the latter is the tangential increasing the former and by decreasing the latter stiffness. one, both the c u n f i n ~ neffect ~ and the curve stiffness The external loads are appfied by means of a 2 may be increased. meters wide rough rigid strip beam (B=2m); numeriIf we also describe the change of the tensile force cal tests, with only some exceptions, are performed T within reinforcements when vertical load i s inin load controlled conditions. Particular attention creased. we may observe that, at the beginning, in all was paid in order to avoid numerical solution mesh reinforcements, the max~mumvalue of T is recorded dependency: i.e. to obtain the numerical conver- correspondingly to the two lateral extremities. In the gence. A large numerical test series was performed following, the position of the maximum value of T by para~etrica~ly changing sand relative density, shifts towards the central point and, there, it reaches mech~nical c h ~ a c t e ~ s t iof c ~ geo-re~~forcem~nts and soil-reinforcement interaction. In the following, for the sake of brevity, the numerical results obtained by considering only one parameter set (Table 1) will be shown and discussed. be a function of the effective mean pressure by following the Janbu’s rule (1963), as is shown in Equation l, with PO= 100 Wa (see Table l )
.($J
2 NUMERICAL RESULTS: VERTICAL AND CENTRED LOADS Firstly, the effect of an infinitely long geo-reinforcement was analysed; in this case only centred and vertical loads have been taken into cons~derat~on. 0 0,05 0,t 915 0,2 0,25 0,3 The sand density above and below the rein~orceme~t viB is assumed to be the same. Because of the footing dimensions, the presence of only one reinforcement Figure 2. Centred vertical load, numerical simulations: comdoes not influence the system mechanical response. parison between shallow unreinforced f ~ ~ n d a georein~~o~, If the reinforce~ent relative depth is too high forced shallow founda~ionand rigid deep foundation.
704
Figure 5 . Three georeinforcements, not-associated flow rule, Failure mechanism (plastic strain tensor modulus).
Figure 3. Trend of T in the central point of the georinforcemen t s.
3 NUMERICAL RESULTS: ECCENTRIC AND INCLINED LOADS Radial load paths characterised by different inclinations in the M-V and H-V planes are imposed. In all cases, the presence of geo-reinforcements increase markedly the ultimate load (Fig.6 and 8). Three distinct inclinations (i=atan(H/V)=Oo,i=lOo, i=lSo) as well as three distinct eccentricities (e=M/V=O, e= 0.22, e=0.67) were considered. The failure mechanisms become asymmetric and the reinforcements are partly loaded and partly anchored (Fig.7 and 9). In Figure 10, for instance, the two failure loci relative to the unreinforced case as well as in the reinforced one are summarised; these are obtained by changing the load inclination, for e/B=O.
Figure 4. Three georeinforcements, Associated flow rule, failure mechanism (displacement modulus).
its ultimate value. In Figure 3 the trend of T in the central point of the three reinforcements is illustrated. It is evident that in the three reinforcements T,,, is reached and this implies that the system is well designed. Moreover, if we consider Figure 4, where the is0 displacement modulus curves are illustrated relatively to curve (b) of Figure 2, it is evident that the zone interested by the collapse develops below the reinforced stratum. In fact, a good design will ensure the resistance of the reinforced stratum and will shift downward the failure mechanism. Finally, if we analyse curve (d) of Figure 2, we may quantify the importance, in the description of the phenomenon, of the constituive relatioship and in particular of the dilatancy rule. Curve (d) is obtained by imposing for loose sand v=O,while, for dense sand, v=$.Both the ultimate load and the curve stiffness are markedly influenced by such a choice. In Figure 5 the corresponding iso-displacernent modulus curves are represented. It is evident that a failure mechanism does not appear and a punching mechanism takes place.
Figure 6. Eccentric vertical load: numerical simulations.
Figure 7. Failure mechanism with eccentric load (e/B=O. 167).
705
Figure 8. Inclined loads: numerical simulations.
show, with respect to corresponding unreinforced ones, a marked increase in the collapse load values, but, conversely to the case of artificial georeinforced embankments on soft soils (Varadajan et al. 1999), the system initial stiffness seems to be not markedly influenced by the geo-reinforcement use. In order to ensure a good response of the reinforced stratum, i.e. to not allow shallow failure mechanisms, the number and the reinforcement strength must be properly designed. The soil relative density placed among the different reinforcements seems to play a very important role because by increasing it the friction angle between soil and reinforcement is increased, too. As a consequence, all the geo-reinforcements, even the deepest ones, are stressed and work close to their maximum tension value. The best performances are numerically observed when inclined or eccentric loads are considered. As a consequence, according to the authors, in these cases the use of geo-reinforcements seems to be more promising.
5 ACKNOWLEDGMENTS Figure 9. Failure mechanism inclined loads (i = 15')
The constant help of Dr. D. Roddeman, author of the Tochnog Finite element commercial code, was very appreciated. REFERENCES
Figure 10. Failure locus in the dimensionless €3-V plane: comparison between reinforced and unreinforced case.
4 CONCLUDINGREMARKS Rigid shallow foundations subject to vertical, inclined, centered and eccentric loads placed both on loose sand, unreinforced and reinforced Strata, have been numerically analysed thanks to the Tochnog commercial finite element code. Rigid shallow foundations placed on reinforced soil strata seem to
Di Prisco C., Montanelli F., Rimoldi P., Sinisi G. & Villa E. 1998. Shallow foundations on granular soils reinforced by means of Geosynthetics. Proc. 2'ld Inr. Con$ On Groiind Improvement Techniques, 8-91) 111998 - Singapore: 407413. Janbu N. 1963. Soil compressibility as Determined by Oedometer and Triaxial Tests. Proc. ECSMFE, Vol. 1, Wiesbaden: 83-87. Nova R. & Montrasio L. 1991. Settlements of shallow foundations on sand. Gkotechnique. Vo1.41, N.2: 243-256. Soni K.M., Varadardjan A. & Sharma K.G. 1992. Effect of reiizfimvrnent length on bearing capacity. Balkema. Tatsuoka F, 1990. Bearing capacity of Huang c.c. reinforced Horizontal sandy Ground. Geotextifes and Geomembranes 9: 5 1-82. Ticov S. 1977. Suiface,fi>otingson sand under general planar loading. PhD Thesis, University of Southamp~on Tochnog commmercial code, www.tochnot?-.corn VaradaGn A., Sharma K.G & ~ l A.A. y 1999. Finite Element Analysis of reinforced embankment Foundation. Iiit. J. NLlm.A,zal. GeorTzech., 23: 103-114. Vecchiotti M. 2001. strip shallow7 ~~zindatioizsoiz geoi-eiiforced dry loose sand strait. Degree Thesis Politecnico di Miiano, in Italian.
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~ ~ n d r n in a Earth r ~ ~ ~ e i n ~ ~ r ~ ~ r~n e nctet*al.~feds),i 0 2001 Swets & Zeitijnger, ISBN 90 2651 863 3
Slope stabilization with hig~-pe~€o~mance steel wire meshes in ~ u ~ ~ with i ~ nails a t and i anchors ~ ~
ABSTRACT: Diagonally structured meshes of high-tensile steel wire provide new safety and economical possibilit~esfor slope stabilizations in comb~nationwith soil and rock nailing. An adapted concept for dirnensioning based on practical experience as well as special laboratory trials to determine the essential bearing capacities of tangential and axial €orce t r ~ s ~ s s from ~ o nthe mesh to the nails make this flexible facing system designable, This system with an open mesh permits full-surface greening and is often an economical alternative to rigid constructions of concrete.
1 INTR~~U~TION The use of wire mesh and wire rope nets as a flexible measure to protect surfaces has been proved wu~hwh~ in~many e cases and is an accepted technology. Until now, wire mesh from wires of a tensile strength of approx. 500 N/mm2 are mostly used for these protection measures. Ropes threaded into the mesh can be used as reinforcements. If the distances between nails are kept at an economical ievel, these simple meshes are often unable to absorb the occurring forces and to selectively transmit them onto the nails. Wire rope nets, while finally allowing larger spacing between nails, are c o m p ~ a t ~ vexpensive el~ and therefore only used in particularly delicate cases with very high forces and a need for long nails. The development of a wire mesh made from high-tensile steel wire with a tensile strength of more than 1,800 N/mm2, produced in a special process, offers new possibilities for the efficient and also economical protection and stabilisation of slopes. The system lends itself for dimensioning by means of suitably adapted d i m ~ n s ~ o n imodels n~ considering the statics of soil and rock. In the practical appl~cationthese high-streng~hmeshes are often able to replace the expensive wire rope nets.
sile strength of approx. 150 W f m in ~ongitudinal and 75 Wlm in cross direction. The gentle deflection of the wire to form the individua~meshes results in a thr~e-dimensionalityof the mesh which vastly improves the connection to the subsoil and possibly applied hydroseeding. Substantially higher forces can be absorbed by this mesh in comparison with the wire mesh traditionally available on the market, offering a tensile strength in longitud~iia~ direction of approx. 50 kN/m at comparable mesh size and similar wire diameter. Compared to wire rope nets, the high-tensile wire meshes with their special properties provide a statically virtually equi~alentsurface protection system which is, however, substantially more economical. When using high-tensile wire mesh as a surface protection system, the slope of soil or rock to be protected is cut as suitable and level as possible to the profile and then covered with the high-tensile steel wire mesh which is fixed and often pretensioned to soil or rock nails.
2 HIGH-TENSILEWIRE MESH AS A SURFACE PROTECTION SYSTEM The usual type of the new mesh for slope stabilization, named TECCO (cf. figure 11, has individual meshes ~easuring83 nm x 143 m. It i s produced
by single twisting of the wire with a diameter of 3 mm and a tensile strength of more than 1,800 Nlmm'. This TECCO mesh has a considerable ten707
Figure I. The TECCO mesh.
~
ium-zinc alloying is used to protect the wire surface what equals an enormous increase of corrosion protection in c o m p ~ s o nwith normal standard zinc coating.
3 RUVOLUM, THE NEW DIMENSIONING CONCEPT The new dimensioning concept RUVOLUM is applicable to all surface protection systems on the market which provide for nailing in combination with a mesh or a wire rope net (or a mixture of the two) as a surface protection system which permits any distance between nails in both horizontal and vertical direction. It is pointed out especially that RUVOLUM is not restricted to the dimension~ngof protection systems using high-tensile mesh exclusively. The new dimensioning concept comprises two investigations: - Investigation of superficial instabilities parallel to the slope - Investigation of local instabilities between single nails RUVOLUM is a product of Riiegger Systems Ltd and has been established especially by Rudolf Riiegger and Daniel Flum in the context of the development of high-tensile mesh.
Figure 2. The TECCO mesh.
Figure 3. General profile.
Figure 4. Instabilities, parallel to the slope, local.
dead weight of the cubic body shear force, to be absorbed by the nail pretensioning force t thickness of the surface layer to be stabilized c'A cohesion of the cover layer * ground surface of the body liable to break out, whereby A = a * b T, N reaction forces from the subsoil inclination of the slope front a
G S V
For this system a nail pattern is to be selected on purpose in which the nails in the rows are offset by half a horizontal distance between nails (cf. figure 2). This means that the maximum possible bodies liable to break out are limited to a width a and a length of 2*b. The nail head is designed so that it can be pretensioned with the force V (cf. figure 5 or 7). This improves the static effectiveness of the system and limits the deformations in the area of the slope. To meet the high demands in respect of durability, the known a n ~ - c o ~ o s i oprocess n of aiumin-
Y
9 F
inclination of the nail to horizontal effective friction angle of the cover layer safetv factor
Figure 5. All forces acting on the cubic body.
708
3.1 Investigation of superficial instabilities parallel to the slope The investigation of superficial instabilities parallel to a slope examines a cubic body of width a, length b and layer thickness t which threatens to slide off the firm subsoil. All forces considered and acting on the sliding body are marked in figure 5. Hereby it is assumed that no hydrostatic excess pressure and no flow pressure is effective on the sliding body. The force V signifies the pretensioning force of the surface protection system. The active application of this force V means that the spike plates and thereby the mesh are pressed onto or slightly into the subsoil. This outer pressure force acting on the surface of the steep terrain permits to mobilize additional friction forces along the sliding surfaces under examination. This has a positive effect above all on the overal stability. From considerations of equilibrium concerning the cubic body shown in figure 5 and taking into account the rupture condition of Mohr-Coulomb, the general equation 1 can be formulated for the stabilizing shear force S in function of the geometrical and geotechnical parameters as well as of the pretensioning force V and safety factor F.
f
b
b
Figure 6. Overview of the general nail arrangement.
Figure 7. Cross-section of the maximum possible body (liable to break out) of the thickness t.
The dimensioning model is assuming that the pressure cones are located completely outside the body to be examined. The cross-section of the maximum possible body is therefore trapezoidal. Simplifying, the trapezoidal surface can be transformed into a rectangle of equal area of the width aredand the thickness t. The body to be considered now features a width of ared and a maximum length of 2*b. For the proofs of bearing capacity in the context of the investigation of local instabilities by the dimensioning method described in this chapter it is mandatory to vary the thickness of the bodies to be examined over the entire interval 0 to t and to determine the decisive failure (break-out) mechanism in this way. It must be pointed out that the selected geometry of the bodies to be examined is intended to approximately simulate the shell-like shapes which break out in reality. By means of the trapezoidal crosssection the actually curved cross-section is described as an approximation. The failure mechanism in figure 8 shows a twobody sliding mechanism. Hereby the upper body I of trapezoidal cross-section presses onto the wedgeshaped lower body I1 via contact force X. The width of both bodies amounts to ared.The force 2 denotes the slope-parallel force which is to be transmitted selectively onto the upper nail. The force P is introduced as a general retaining force required in the context of the equilibrium considerations.
S [kN] = 1 / F . { F.G.sina - V.F.cos(Y+a) - c . A - [ G ~ o s+ a V.sin(Y+a)] . tancp} (1) The following three proofs of stability must be established in the investigation of global failures parallel to the slope: - Proof against sliding off parallel to the slope - Proof against puncture of the mesh due to pretensioning force - Bearing safety of the nail (combined strains) 3.2 Investigation of local instabilities between single nails When investigating bodies which might break out between single nails, the question is what sort of bodies become possible considering the selected nail mangemen t. Located above each nail is a field of width a and length 2*b which must be protected against local instabilities. The cross section of the maximum possible shape that may break out is influenced to a major extent by the protection concept itself By tightening the nuts, the spike plates are firmly pressed onto the ground or slightly into it and the mesh is pretensioned against the nail head with the force V. Due to the passing-on of forces in the area of the nail head, a truncated pressure cone occurs in the cover layer beneath the spike plate and the adjoining mesh (figure 7).
709
As a special case the body 1 does not exist and it remains body I1 as a one-body sliding mechanism. The contact force X results from the equilibrium considerations at the upper body I, whereby the condition of Mohr-Coulomb as well as the safety factor F are again taken into account. The equilibrium conditions on body I1 are formulated to determine the maximum force P by variation of p. Hereby the contact force X from equation 3 and the slope-parallel force Z are used. The following proofs of bearing safety must be established in the investigation of local instabilities taking into account the bearing capacities of the protection system (cf. chapter 4): -Shearing-off of the mesh at the upslope edge of the spike plate at the lower nail (force P) -Selective passing-on of the slope-parallel force 2, from the mesh onto the upper nail
3.3 Additional proof of the terrain 's resistance against sliding (deep sliding su$aces) This concerns the proofs of safety against rupture of the terrain, for which the nails are included in the stability caculations with the topographically and geologically adapted sliding surfaces, usually as tension elements with a stabilizing effect and in rarer cases as shear elements.
4 TRIALS FOR DETERMING THE BEARING CAPACITY FOR FORCE TANSFER MESH NAIL In the equilibrium considerations in paragraphs 3.1 and 3.2, the pretensioning force V, the retaining force P and the net force Z have been introduced as external forces. The question now arises how high these forces may be in the maximum so that they can still be absorbed by the used mesh, the applied spike plate and the nails, taking into account appropriate safeties. Figures I 1 and 12 show the test setups developed by Ruegger Systems Ltd, in one case to determine the bearing resistance of mesh against selective,
Figure 9. Pushing force P and tensile force Z. X Z
GI,Il c * AI.II
contact force slope-parallel force dead weights of the individual sliding bodies
TI.II,
forces due to cohesion reaction forces from the subsoil
a
inclination of the slope
YJ
inclination of the force P
Figure 8. Two-body sliding mechanism (possible friction forces along the contact surface of the two bodies I and I1 are neglected).
Figure 10. Global stability.
710
slope-parallel tensile strains, and in the other to determine the bearing resistance of mesh to pressure strain in nail direction. The figures 13 and 14 show a graphic presentation of the test results. Laid off as abscissa are the deformations in mm and as ordinate the total tensile or total pressure forces in kN in each case. Hereby the different bearing capacities of a high-tensile wire mesh, of a common wire mesh and of a geogrid of PET become particularly evident. To determine the curves I, I1 and I11 in figure 14, the meshes and the geogrid, respectively, were tensioned in a square frame. The hydraulic press is arranged centrally and presses against a nail of type GEWI D = 32 mm. Curve IV describes the behaviour of the test soil (sandy gravel) without involvement of a mesh or geogrid. The difference A in figure 14 corresponds for example to a high-tensile wire mesh of type TECCO, the bearing resistance of mesh to pressure strain in nail direction, whereby corresponding safety margins must be considered also. In the proof 'shearingoff of the mesh at the upslope edge of the spike plate at the lower nail', only half of A may be used in the calculation, while including conesponding SafetieS.
Figure 13. Force-deformation relationships of three different products as result of tensile tests in tangential direction, using a round spike plate with a diameter of D = 220 mm.
Figure 1 1. Test setup for determining the bearing resistance of mesh against selective, slope-parailel tensile strains.
Figure 14. Force-deformation relationship as result of puncture tests in nail direction, using a round spike plate with a diameter of D = 220 mm. Figure 12. Test setup for determining the bearing resistance of mesh to pressure strain in nail direction.
Figure 15. General cross-section.
Figure 18. Protected slope (rocky section).
5 PROJECT IMPLEMENTED IN MULHEIM, GERMANY In Miilheim, Germany, at the location Mendener Strasse, a rocky slope approx. 420 m long and on average approx. 12 m high is permanently protected against rockfalls and local instabilities by means of nailing in combination with a TECCO mesh cover. The inclination of the slope front to the horizontal plane amounts to between 45 and 70". Locally, above all in the upper areas7 the rockface (sandstone, siltstone, mudstone) is covered by slope clay and slope scree. The rock is very prone to weathering. Based on the geotechnical situation, the overall stability of the slope is not endangered. The protection measures are limited to the section close to the surface. The surface protection was dimensioned on the basis of the RUVOLUM concept. For the rocky section with a slope inclination of a = 70" and a longterm loosened, weathered layer of thickness t = 0.50 m to be protected, a nail pattern of a = b = 2.5 m resulted. In the area of the unconsolidated rock covers with a slope inclination of 45 ..SO" and with t = 0.80 m, a nail pattern of a = 3.3 m and b = 3.0 m resulted. The slope was greened with the Fibrater system.
Figure 16. Protected slope with mesh cover in combination with nailing.
Figure 17. Protected slope.
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L a n ~ ~in~Earth r ~ ~s e i ~ ~ o r c ~ e c~ h~ifet af ~al. i , (eds), 0 200 I Swets & Zeiihger, lS8N 90 265 I 863 3
Behavior of angle cut cylinder excavation by cut reinforced earth work method A. Sato Exp resswuy Research Xnstitu te, Jupun Highway Pub1ic COrporution,Jupan
S. Tayama, K. Ogata & M. Takemoto Japan Highwuy Public Corporation, Jupun
U, Tanaka Yu~hiyo-Engin~eri~~g Co., Ltd., J u p ~ n ABSTRACT: When structure such as bridge pier is constructed in a mountainous district full of steep slopes, earth cutting or earth retaining by soldier pile method is commonly used. However, cutting on a steep slope will produce a large-scale cut slope, and will cause a substantial modification of natural environment, raising problems in the treatment of surplus soil in construction and increase in the management area. F u ~ h ~ r ~ o r e , the soldier pile method requires long piles to be driven in a steep slope; this takes a wide construction yard. Further, this work requires a concurrent use of ground anchors, which results in a longer construction period and greater construction costs. Under these circumstances, the authors of this paper propose the earth retaining method (angle cut cylinder earth retaining method) mainly based on the cut reinforced earth method, and clarify the behaviors of the ground and structures by measurement. At the same time, they are currently investigating the designing method. This method ensures a high degree of stability without tipping earth pressure balance very much, and ~ n i ~ z modification e s of the natural environment. This paper reports the deformation characteristics of angle cut cylinder excavation form based on the 3D plasticity analysis, and the behaviors of ground and structures obtained from the result of measurement in field tests.
I INTRODUCTION The cut reinforced earth was first adopted by Japan Highway Public Corporation (hereinafter referred to as for the first time in 1982. Since then, there has been a rapid increase in the number of the cases of construction work. In the year of 1999, the annual extension of construction of the express-way become 100,000 meters. The Figure 1 shows the bi-eakdown of construction site. Excavation of a steep slope of 1 to 0.5 or greater, for example, a bridge pier shown in Figure 2, is adopted in about 50 percent of the construction sites. This is attributable to the following reason: In the construction of the expressway passing through steep mountains, a slope face of great length and size will result from excavation at a standard gradient. However, when the present method is used for excavation of a steep slope, the slope face can be reduced, and the surrounding negative environmental impact and the volume of excavated soil can be reduced. fn 1998, JH established "Designing and Construction Procedure for Cut Reinforced Earth Work ~ e t h o d " ( J H Procedure) <1> to promote adoption of this method. However, there are still problems to be solved in future. In the construction for the project of the Second Tornei Expressway and Meishin Expressway, large-diameter caisson type piles are more frequently used as a bridge foundation on a steep slope. fn this context, designing in the cut reinforced earth
713
Figure 1. Breakdown of construction site.
Figure 2. Excavation on steep slope.
method with consideration given to mechanical characteristics by circular excavation is not yet established. Against this backdrop, JH is conducting field tests on the earth retaining method (angle cut cylinder earth retaining method) mainly based on the cut reinforced earth method. This method ensures a high degree of stability without tipping earth pressure balance very much, and which minimizes modification of the natural environment, as shown in Fig-
Figure 3. Angle cut cylinder earth retaining.
ure 3. Through these tests, JH is clarifying the behaviors of the ground and structures by measurement, and is currently examining the designing method. 2 DEFORMATION CHARACTERISTICS IN EXCAVATION BY ANGLE CUT CYLINDER EARTH RETAINING METHOD 2.1 Analysis conditions The 3D elastoplastic analysis was used to find out the deformation characteristics of angle cut cylinder excavation form. Analysis models used two cases shown in Figure 4; the angle cut cylinder excavation and circular excavation. The stress and deformation of the ground were compared. In creating the model, attention was focused on the shape effect, and supportless excavation was formulated into the model. In angle cut cylinder excavation, analysis of repeated excavation step by step was made with consideration given to the possibility of using the inverted lining method. The physical properties of general rocks were used to represent values for physicaI properties of the ground to be analyzed, based on the studies made so far <2>. Drucker-Prager failure conditions were used as failure conditions. As analysis conditions used under the Drucker-Prager yield condition, soft rocks were used as bed rocks in order to estimate the extreme looseness on the model slope, and different two constants were utilized for analysis. The constants for analysis are given in Table 1. Table 1 . Constants for analysis Sand Young’s modulus(kN/m’) 68,646 Cohesion(kN/m’) 19.6 Internal friction Angle (Deg) 32 Unit weight (kN/m3) 19.6 Poisson’s ratio 0.35
Clay 68,646 49.0 21 19.6 0.35
Soft rock 245,166 196.1 37 21.6 0.30
Figure 4. Analysis modles.
2.2 Result of analysis Figure 5 shows the excavation displacement and the result of analysis in plastic area. A common point for the angle cut cylinder excavation and circular excavation is a building form, independently of the ground. Displacement is greater in angle cut cylinder excavation. This is because deviatoric stress in the initial stress of the ground is greater and stability is poorer in the angle cut cylinder excavation than in the circular excavation on the flat ground. In angle cut cylinder excavation, the plastic area at the maximum height of excavation is greater than plastic
7 14
Figure 5 . Excavation displacement and plastic area.
715
16.3
n
Mountain side
/
/
Figure 7. Plan and cross section for angle cut cylinder earth retaining Table 2. Structural particulars P1 descending bridge pier Excavation particulars Excavation height: 12.9111
Valley side
Diameter: 16.lm
Ring beam Cross section: 1.2x0.8m
Upper: Five 2D16 Lower: Five 2D 16
placement in XY plane (settlement) A :After excavation of the 6th stage o :After excavation of the 8th stage 0 :After excavation of bottom slab ' :Afterexcavation of caisson type pile
Figure 8. Ground surface displacement
Ground anchor Designed force: 588kN Fixed length: 5.5m Free length: 8.0 to 13.0m Number of anchors installed: 13 Bore diameter: 135mm Tendon: 7 PC steel stranded wires (1 1.1mm in diameter) Grout: 24N/mm2
a deformation characteristic in angle cut cylinder excavation. On the side, however, it is more difficult to get circular stress as the ground plasticity due to stress concentration of the ground is greater. 2.3 Overview of the test The field tests were conducted to observe the behaviors of the ground and structural members. The test site is located at a steeply inclined (38 deg.) mountainous district with an elevation of 150 meters. Geographical features mainly consist of the sedimentary rock of the Tertiary Hamishigaku Deposit, and unconsolidated deposits of the Quaternary period covering it. The Figure 7 provides a plan and cross section for angle cut cylinder earth retaining. It should be noted that sectional form, ground anchor and the number of anchors installed in the cut reinforced earth method conform to the measurements obtained from the past tests and the result of the FEM 3D elasticity analysis <2> <3>.
Virtical and diagonal reinforcements Reinforcements: Deformed reinforcing bar 6.0111 (25.6mni in diameter) Bore diameter: 90mm Bore length: 5.5m Grout: 24N/mm2 When reinforcement is 6m Reinforcement: Self-drilling bolt 3 1 mm in diameter Bore diameter: 50mm When reinforcement is 3m Reinforcement: Deformed enforcing bar D25 Bore diameter: 50mm Inverted lining wall Wall thickness: 300mm Reinforcing rod: D25ctc300
area on the side. The Figure 6 shows the transition under stress. The stresses at the maximum height of excavation in circular excavation and angle cut cylinder excavation are represented by the maximum main stress rotating in the vertical upward direction. In this case, there is an increase in intermediate main stress facing in the circumstantial direction of the excavated cross section. This reduces the increase in the deviaplastic area is reduced. In the meantime, the maximum main stress is in the inclined direction on the side of angle cut cylinder excavation, with the result that deviatoric stress grows, and the plastic area of the ground increases. This makes it possible to expect that the stress similar to that in circular excavation occurs at the point close to the maximum height of excavation independently of the ground, as
3 FIELD TEST RESULTS
3.1 Ground suface and underground displacement The Figure 8 shows the ground surface. The ring beam was generally displaced toward the lowest point. This is considered to have been caused by the following reason: Displacement from the maximum excavation height to the valley was assumed in the designing stage, and sliding preventive reinforcing rods of the ring beam were placed on the side of the maximum excavation height. Therefore sliding resistance effective in the direction of actual displacement is considered to have failed to work. The Figure 9 shows the underground displacement measured by a multi-stage inclinometer. In the measurement data, the value for underground dis-
716
Figure 9. Underground displacement
placement after prestressing of ground anchors is assumed as an initial value. Underground displacements at all three positions were tilted toward the excavated side, and there was an increase in displacement at the time of excavation by caisson type piles. Especially, point <2> has an excavation height H = 5.9m, the lowest in three positions. Despite that, the horizontal displacement S/H(y) on the top of slope was 0.26 %, the maximum figure in three sites. This is considered to be due to the following topographic features: This area from the surface lowered by cutting to the excavation bottom was covered with the conglomerate exposed to strong wind. When measured by a settlement gauge, the settlement was about 22.7mm at this point, as compared to about lOmm in other two positions. Namely, the settlement at this point was double than in other positions. In addition to these geographical features, another reason is the stress concentration of the ground as discussed in “2.2 Result of analysis”.
3.2 Stress of ring beam Concerning the inside of the ring surface, one place was found out where all the reinforced rod and concrete stresses were turned into tension with values very small. Maximum values were smaller than the permissible stress (20 N/mm2 of the maximum tensile stress of the reinforcing rod and 0.2 N/mm’ of the maximum compressive stress of concrete). As to the outside, some places were found out where all stresses were turned into tension in the same way of the inside, but the values were also small.
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3.3 Sliding prevention work The maximum tensile axial force of 70kN was measured. Since this corresponds to 70 % of the designed axial force (9OkN), it is considered that the sliding preventive reinforcements have been effective to the sliding of the ring beam. However, the reinforcements, which were driven in parallel, were not effective to the ground surface displacement.
3.4 Stress of reinforcing rods and axial stress of reinforcenzent on the wall suiface The Figure 10 shows the axial stress of reinforcement and stress of reinforcing rods on the wall surface. A maximum axial force of about 70 kN (60%) occurred to the reinforcement with respect to a design axial force of l 18kN. According to the cases of steep slopes excavated so far, the axial force occurring to the reinforcements was about 30% of the permissible reinforcement force in most cases. Hence, the value of axial force having occurred can be considered as great. Further, the trend of reinforcements axial force is that greatest axial force occurred on the top of excavation, and a small value was recorded on the lower portion of excavation. This exhibits approximately the same distribution as that of underground displacement. For the stress of reinforcing rods and concrete on the wall surf:ce, the compressive stress of concrete is 0.1 N / m - or less in both longitudinal and lateral directions, and the tensile stress of the reinforcing rods is lN/mm’ or less. Only a very small stress was measured, as compared with the axial force of the reinforcement.
Concrete Stress Longitudinal: Complession 0.028N/mmz Tention 0.0037N/mm2 Lateral: Complession 0.030N/mm2 Tention 0.0023N/mm2 Reinforcing rod stress Longitudinal: Complession 0.23N/mmz Tention 0.49N/mmz Lateral: Complession 0.87N/mm2 Tention -
analysis and field measurements. It is considered difficult to get the effect of circular stress. <2> In the case of angle cut cylinder earth retaining horizontal displacement ymax. on the top of slope does not always occur at the position of maximum excavation height. This makes it necessary to perform stability management at the time of construction on the side as well. <3> In the ring beam sliding preventive work, it is not always effective to drive reinforcements in one direction because of the trend of displacement. <4> While stress of the reinforcing rods and concrete on the wall surface is very small, reinforcements axial force is great. This makes it possible to consider that ground deformation by excavation is reduced by the reinforcements integrated in one body, thereby reducing the earth pressure.
ete Stress itudinal: Complession 0.103N/mmz Tention 0.0674N/mm2 Lateral: Complession 0.054N/mm2 Tention 0.0076N/mmz inforcing rod stress Longitudinal: Complession l.O0N/mmz Tention 0.36N/mmz Lateral: Complession 1.93N/mm2 Tention 0.52N/mmz 38.0KN Caisson type pile
9 . 9 ~ ~
5 CONCLUSION
Tention 0.0082N/mm2 Lateral: Complession 0.021N/mmz Tention 0.00WN/mm2
Based on the results discussed in this paper, we will study the designing procedures for each member, and continue to perform tests, thereby collecting and analyzing measurement data for each members in an effort to establish logical designing procedures.
Figure 10. Reinforcement axial force and wall surface stress
REFERENCES
4 SUMMARY
Japan Highway Public Corporation. 1998. Designing and Construction Proceduref o r Cut Reirlforced Earth Work Method Japan Highway Public Corporation. 1997. A Guide to Dowtzsizing Work Method for Lnrge Scale Cut Slope Face: 2 IO21 1 Sato, A. et. al. 2000. Field Measurement of Angle Cut Cylinder Excavation Based on Cut Reinforced Earth Work Method (Part I) - Field Measurement - 55th JSCE Antzunl Meeting Takemoto, M. et. al. 2000. Field Measurement of Angle Cut Cylinder Excavation Based on Cut Reinforced Earth Work Method (Part 11) - Field Measurement - Effect and Verification of Structural Member. 55th JSCE Aiii~ialMeeting.
The following describes the conclusion gained in the present paper: < 1> Regarding the angle cut cylinder deformation characteristics, the same shape effect as that of the circular excavation is obtained at the position close to the maximum excavation position. On the side, however, the plastic area is increased by stress concentration of the ground, according to the result of
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 02001 Swets & Zeitlinger, ISBN 90 2651 863 3
Seismic ductility of cut slope reinforced by soil nail A. Takahashi, J. Izawa & 0. Kusakabe Tokyo Institute of Technology, Tokyo, Japan
S. Tayama & M. Takemoto Expressway Research Institute, Japan Highway Public Corporation, Tokyo, Japan
ABSTRACT: This paper describes an experimental study on seismic ductility of cut slope reinforced by soil nail using a geotechnical centrifuge. Effects of length and diameter of the soil nails on the permanent deformation of the cut slopes were discussed in terms of the ductility of the reinforced slopes.
1 INTRODUCTION Soil nail reinforcement has been widely applied to steep cut slopes, particularly for expressway constructions (Tayama & Kawai, 1999). Most of the existing design methods for the soil nail reinforced slopes are based on empirical methods and/or the limit equilibrium analyses (JH, 1998), neglecting the ductility of the reinforced soil structure against deformation. In this study, pullout tests for a model soil nail and centrifuge model tests were conducted to evaluate the ductility of reinforced cut slopes during earthquakes. Effects of length and diameter of the soil nails on the permanent deformation of the slopes were also discussed.
to the top level of the model slope. After completion of soil compaction, templates were placed and the compacted soil was formed into the slope shape. Model soil nails, which were 0.58mm diameter nickel spiral guitar strings, were installed into drilled holes on the slope surface with glue or plaster made grout material arranged in gridiron at intervals of 30mm, which is equivalent to 1.5m in the prototype scale. The arrangement of the soil nails for the tests are shown in Fig.2. In order to avoid the local failure at the face of the slope, transparent sheet made facing plate was adopted. Rubber dampei
\
2 TEST PROCEDURE AND CONDITIONS
A1 1
140mm
0
q-J=
2. I Test procedure
A12
0
Geogechnical centrifuge used in the tests was T.I. Tech Mark I1 Centrifuge (Takemura et al. 1989). Model set-up used is shown in Fig.1. DL clay was used for making the model ground. Material properties of DL clay are given in Table 1. Friction angle, 4, was obtained from the triaxial compression tests under drained conditions and was modified considering the plain strain condition in the centrifuge tests. Cohesion, c was back calculated from the failure height observed in a centrifuge test on non-reinforced vertical slope. An aluminium model container with inner sizes of 450mm in width; 150mm in breadth and 250mm in height was used. Rubber sheets were placed at both side of the container for absorbing of stress waves from the side boundaries. DL clay with water content of 20% was statically compacted to the bulk density p t = 1.4Mg/m3 layer-by-layer using a bellofram cylinder. This compaction was continued up 719
DL
clay
Laser DT
___cI
A23 U U
A-Input
Accelerometer Displacement transducer
%G?
a
Figure 1. Model set-up for centrifuge test. Table 1. Material properties of DL clay. Soecific gravitv Mean grain size D50 Maximum void ratio Minimum void ratio Cohesion c" Internal friction angle @*
2.68 0.024mm 1.672 0.695 9.2kPa 4OSdeg.
--Lm Side view
_.I--
Front view
0
_.---
0
30mm
0 i
___.----___.---__-/---
/---
_/-
o &
o
k
I 0 : Soil nail
----I-
Figure 2. Soil nail arrangement. Table 2. Applied sinusoidal waves in the test. ~
Wave
Nominal Frequency amplitude (G) (Hz) 1st 10 (0.2) lOO(2) 2nd 15 (0.3) lOO(2) 3rd 20 (0.4) 100 (2) 4th 20 (0.4) 100 (2) In parentheses, prototype scale
Number of waves 20 20 20 40
Table 3. Test conditions. Case
Slope L(mm) D(mm) ascent 1 1 : O S (63") 2 1:0.4 (68") 3" 1:0.3 (73") 4 1:0.3 (73") 80 2 5 1:0.3 (73") 80 2 6 1:0.3 (73") 80 7 7 1:0.3 173") 40 7 * Failed at 45-G centrifugal acceleration
Figure 3. Input waves acceleration time histories (Case 6). Facing date No No No No Yes Yes Yes
Cases 1-3, no reinforcements were applied. In Case 3, as the brittle failure of the slope took place when the centrifugal acceleration reached 45-G during the spinning up of the centrifuge, whose slope ascent was 73", no data were available for this particular case. 3 TEST RESULTS AND DISCUSSION
Having prepared the model, the container was set on the shaking table mounted on the centrifuge. Shaking tests were conducted under 50-G centrifugal acceleration by sinusoidal waves with a frequency of IOOHz, which is equivalent to 2Hz in the prototype, to the shaking table. Four waves with different conditions as shown in Table 2 were input to each model. Typical time histories of the input sinusoidal waves in the prototype scale are shown in Fig.3. During shaking, acceleration and displacement of the reinforced soil were measured at the locations shown in Fig.l. Photographs were taken before and after shaking to observe the displacements of targets on the front surface of the reinforced soil. 2.2 Test conditions Table 3 gives the test conditions adopted in this study. Height of the cut slope wall was 140mm, which corresponds to 7m in the prototype scale. Effects of the facing plate, the diameter (D) of drilled holes for the soil nails and length (L) of the soil nails on the permanent deformation were investigated. In 720
All centrifuge test results presented in this section are in the prototype scale. 3.1 Pull-out tests on soil nail
To investigate frictional interaction between the DL clay and the model soil nail, pullout tests were conducted using a small pullout test apparatus. As the diameter (D) of drilled holes for the soil nails was varied in the centrifuge tests, the pullout tests for two sizes of the hole were carried out under different two overburden pressures as listed in Table 4. Typical test results of the pullout test for the overburden pressure of 39.2kPa are shown in Fig.4. The pullout resistance 'I: shown in the figure is defined as z = FhDL, where F is a pullout force, D is the diameter of the drilled hole and L is the length of the nail in soil. Ratios Rp of the peak pullout resistance (z at peak) to the soil strength (= c -k d,, tan 4,) of the DL clay were obtained as shown in Table 4, where U',, is the overburden pressure, c and 4, are the strength parameters of the DL clay.
Though, at the beginning of the pulling, the slope of the relationship between the mobilised pullout resistance and the pullout length was almost the same for the same overburden pressure, the peak pullout resistance of the nail and the pullout lengths of the nails, when the frictional resistance was fully mobilised, were heavily affected by types of the grout materials and the diameters of the drilled hole. In the cases with the small diameter of the drilled hole, the relative displacements between the soil nails and the grouting were observed. It seemed that these were attributed to lack in the grouting itself and adhesion between the nail and the grouting, and they result in the low pullout resistance.
3.2 Failure of the non-reinforced cut slope in centrifuge tests Observed permanent deformations for the nonreinforced cut slopes, Cases 1 & 2, just after the third shaking step are shown in Fig. 5. Figure 6 shows the time histories of the settlement at the slope shoulder for Cases 1 & 2 in the first and third shaking steps. In both cases, settlements increased gradually in the first step, while drastic large settlements occurred in the third step. In the third step, slip surfaces appeared at the shallow portion from the slope surfaces as shown in the figure. In Case 3, the brittle failure of the slope took place when the centrifugal acceleration reached 45-G during the spinning up of the centrifuge.
Table 4. Pull out test conditions and results. Grout Overburden Case D m m material p r e s s u r e ( k P a ) glue 39.2 1 2 2 2 glue 78.4 3 7 plaster 39.2 7 Dlaster 78.4 4 R,,: Ratio of peak pullout resistance to soil strength
Figure 5 , Permanent the third shaking in Cases 1 &2).
of the
slope
3.3 Deformation progress of the reinforced cut slope
_&0.15
Observed permanent deformation of the reinforced cut slope in Cases 4 & 5 are shown in Fig.7. The slopes were reinforced with the nails installed into the 100 mm diameter drilled holes in these cases. Considering the fact that the slope ascent of 73" is not self-supportable without reinforcement as mentioned in Section 2.2, the soil nail reinforcement shows its efficiency to the stabilization of the cut slope. During the shaking, slip surfaces appeared as shown in the figure. Aithough the slip surfaces were
0.42
after
Figure 6. Time histories of the settlement at the shoulder for the first and third step in Cases 1 & 2.
72 1
Figure 10. Variations of the settlement at the shoulder against Arias intensity.
To gain insight into the ductility of the reinforced slope during earthquakes, the relationships between the permanent displacement of the slopes and the intensity of earthquakes were examined. As one of the indices for the intensity of earthquake, Arias intensity was introduced (Arias, 1970, Kayen & Mitchell, 1997). This index, I, is defined as
Figure 8. Settlement at the shoulder against maximum acceleration of the input motion (Cases 4 &5).
observed in the reinforced zones, the slip surfaces appeared in deeper portions, compared to the nonreinforced slope. In the other reinforced slope cases with the large diameter of the drilled holes for nailing, Cases 6 & 7, no failure of the slope was observed even after the forth shaking step. Comparing Cases 4 & 5 from the viewpoint of the effectiveness of the facing plate on the prevention of the local failure, the slip surface passed trough the slope face when the facing plate was not adopted while it passed the slope toe with the plate. Total settlements of the slope shoulder are plotted against to the maximum acceleration of the input motion in Fig.8. Unique relationship between the settlement and the input motion acceleration can be seen in the figure. According to this figure, the effect of the existence of the face plate on the total deformation of the slope was not remarkable in this study, as the facing plates were very thin and had a low flexural rigidity.
where g is gravitational acceleration and a(t) is the acceleration of the input motion at time t. This intensity represents some sort of a dissipated energy per unit weight of soil. Variation of I, with time for Case 6 and the actual earthquake record at Kobe Marine Observatory in 1995 Hyogoken-Nambu Earthquake are plotted in Fig.9. In this study, very large intensity of the earthquake motions was assumed as shown in the figure. Figure 10 shows the variations of the settlement at the slope shoulder against Arias intensity for all the cases. Cases 1, 2, 4 & 5 show drastic increase of the settlement and result in failure during earthquake. Meanwhile, in Cases 6 & 7, the settlements level off and shake down against the earthquake intensity. From this figure, the relationships can be
722
classified into two types. These two typical types of the relation are shown in Fig. 11. In the former type, the relationship can be simplified into bi-linear and divided into two phases, Phase I & 11, as shown in the upper figure of Fig. 11: Phase I: The settlement gradually increases with Aria1 intensity. No localisation of deformation can be observed in the slope. Phase 11: The settlement shows a drastic increase resulting in catastrophic failure. The deformation is localised and the slip surface appears. In the latter type, Cases 6 & 7, shown in the lower figure of Fig.11, as the excessive reinforcement was adopted, no localisation of deformation, i.e., no transition to Phase 11, was occurred resulting in shaking down. As the localisation of deformation can be observed only in Phase 11, a scope of application for deformation analyses of the simplified analysis such as Newmark's sliding block method may be limited in this phase. As the permanent deformation accumulated in Phase I is not negligible, alternative analysis methods, i.e., FE analysis employing a constitutive equation considering dilatancy and compression characteristics of soils under cyclic loading appropriately, should be conducted to estimate the permanent deformation of the slope. To evaluate the ductility of the cut slope from the test results, the permanent displacements accumulated in Phase I, V I , were examined. The settlements in Phase I are shown in Fig.12 with the slope ascents. The larger settlement in Phase I may correspond to the higher ductility of the cut slope. The settlement V I becomes larger as the slope ascent decreases in the cases without reinforcement. Paying attention to the cases with reinforcement, the soil nail reinforcement is very effective in making the ductility of the slope higher, considering the fact that the slope ascent of 73" is not self-supportable without reinforcement. The increasing rates of the settlements against Arias intensity, Av/AI,, were also examined to evaluate the ductility of the slope. The smaller Av/AZ, may represent the higher ductility of the cut slope. Figure 13 shows the increasing rates of the settlements against Arias intensity for Cases 1, 2, 4 & 5 with the slope ascents. In Phase I, the results were relatively smaller than those in Phase I1 and the differences were very small. No clear relation can be seen to the slope ascent and the existence of the reinforcement. In the meanwhile, remarkable differences were found in Phase 11. Corresponding to the slope ascent, Av/AZ, for Case 1 was smaller than that for Case 2. The increasing ratios for Cases 4 & 5 were also smaller than that for Case 2. The effectiveness of the soil nailing on the ductility of the slope can be reconfirmed by these results. 723
Figure 13. Increasing rates of the settlements against Arias intensity in Phase I & 11.
4 CONCLUSIONS In this study, centrifuge tests were carried out to investigate the effects of the length and the diameter of the soil nailing on the ductility of the reinforced cut slope during earthquakes. Following conclusions are drawn; (1) The cut slope without any reinforcement shows the brittle failure during the ground motion, while the permanent displacements gradually increased when the soil nail reinforcement was adopted. (2) The slope reinforced with poor grouted nailing exhibited the slip surface during earthquake. Although the slip surfaces were observed in the re-
inforced zones in these particular cases, the slip surfaces appeared in deeper portions, compared to the non-reinforced slopes. (3) Arias intensity was used to examine the relationship between the permanent deformation of the cut slope and the earthquake intensity. The beginning of the localisation of the slope deformation can be clarified by this relation obtained from the tests results. (4) The ductility of the soil nail reinforced cut slope was qualitatively assessed by the settlement before the appearance of the slip surface and the increasing rate of the slope shoulder settlement against Arias intensity.
724
REFERENCES Tayama, S. & Kawai, Y. 1999. The current status and outlook for soil nailing in Japan. Earth reinforcement technique in Asia, Special Volume for the Proc. of ARCSMGE, 5 1-61. Japan Highway Public Corporation. 1998. Design and construction manual for reinforced cut slope (in Japanese.) Takemura, J., Kimura, T., & Suemasa, N. 1989. Development of Earthquake simulators at Tokyo Institute of Technology. Technical Report, Dept. Civil Etzgrg. Tokyo Institute of Technology, No. 40,41-60. Arias, A. 1970. A measure of earthquake intensity. Seismic design for nuclearpowerplants, Hansen, R.J. ed., MIT Press. Kayen, R.E. & Mitchell, J.K. 1997. Assessment of liquefaction potential during earthquakees by Arias intensity. J. Ceotech. atzd Geoinvir. Engrg., ASCE, 123(12), 1162-1174.
Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, lSBN 90 2651 863 3
The resistance of jacked-in pipe inclusions in soft soil Tan Siew Ann, Cheang Wai Lum, Yong Kwet Yew & Ganeswara R. Dasari Department of Civil Engineering, National University of Singapore
ABSTRACT: The technique of soil nailing, a method of earth reinforcement is a promising technique in slope stabilisation and as a support system in excavations. Due to its rapid top-down construction and costeffectiveness, this method of earth reinforcement is going through tremendous growth in the geotechnical fraternity . However much is still to be learned as the understanding of this technique is still very much in its infancy, specifically the use of nail inclusions in clayey soil conditions. Some of the current code of practices suggested that the technique of soil nailing is not well suited for soft clayey soils of low shear strength, of high liquidity index and soils where the SPT ‘N’ value is very low. Primarily, these restrictions are empirically based, as little data exist on the working mechanisms, deformations and stresses induced in the reinforced soil mass by the soil nail inclusions when this method is applied in soft clayey soil conditions. This paper attempts to highlight one of the possible modes of resistance of nail inclusions in clayey soils. Reported herein is a case history where jacked-in pipes which perform as soil nails were used in support of a deep excavation. The method of installation causes soil disturbances, compaction and increase in the interfacial shaft resistance of the inclusions. Nail set-up was primarily caused by re-equalisation of excess pore-water pressures, corresponding with a time-dependent increase in the effective radial stresses around the nail periphery. Coupled non-linear finite element analyses were conducted to investigate the phenomena of jack-in installation, which causes soil disturbance, the build-up of excess pore-water pressures and the subsequent increased in effective radial stress due to re-consolidation. It was found that this technique of reinforcing the ground with jack-in pipes was effective in stabilising vertical cuts and arresting movements in soft clayey soil conditions.
1 INTRODUCTION The geotechnical group at NUS has been involved in a sustained research effort to investigate the behaviour of soil nails in tropical residual soils. These efforts were motivated by the versatility and costeffectiveness of soil nails in stabilising slopes and excavations. More recently, work has focused on developing sound methods for the prediction of nail pull-out capacity (Tan et.a1.,1998, Tan et. al., 1999). Reported herein is a case history on the utilisation of structural steel pipes as passive nail inclusions. These inclusions were used for the stabilisation of a deep excavation in soft tropical residual soil. The conventional technique of constructing a nail inclusion involves drilling a borehole to the required length and inclination, inserting an appropriate bar and filling the cavity with grout. This installation technique is a replacement method, whereby the volume of soil being removed is equal to the volume of the constructed nail and it involves predominantly the relief of soil stresses around the periphery of the inclusion. On the other end of spectrum, the con-
725
struction of a nail which is carried either by driving or jacking involves the displacement of soil and hence generally an increase in the mean effective pressure around the nail. This paper will highlight a novel method of nail installation by jacking steel pipes. It was used in conjunction with a contiguous bored pile wall for a deep vertical excavation in soft tropical residual soil in Malaysia. The aforementioned case history was reported by Cheang et.al. (1999). Another case history where this technique was used in conjunction with a flexible sheetpile wall was reported by Liew et.al. (2000).
2 A BRIEF DESCRIPTION ON THE USJ- 19 NAILED EXCAVATION The deep excavation project (Photo l), which is 10cated at Subang Jaya of Malaysia consists of three condominium towers of 33 storeys and a single 20storey office tower. Due to the huge demand for parking space, an approximately three storey deep
Photo 1: After second stage of construction
vehicular parking basement will be required. The deep excavation, through a filled layer of very loose silty sand and soft peaty clay varies from I l m to 13m. The presence of very soft soil condition and the fast track requirement of the project, contiguous bored pile (CBP) walls supported by jack-in pipe inclusions in six to seven rows were utilised to stabilise and control the lateral displacement and surface settlement. The supported face is approximately 6900 m2. 2.1 Subsug5ace geology The general subsurface soil profile of the site, shown in Figure 1 consists in the order of succession a firm clayey SILT, a loose to medium dense SAND followed by firm to hard clayey SILT. The soils are inter-layered by thick deposits of very soft dark peaty CLAY. The plot of SPT’N’ values, illustrated in Figure 2, shows significant scattering which is commonly found in tropical residual soils in the area. However the trend of the scatter plot shows that the SPT ‘N’ values increases with depth. For the underlain soft values were clayey material, the registered SPT ‘N’ zero and Vane Shear strength varies from 5 kPa to 20 kPa.
Figure 1. SFT ‘N’ values
Figure 2. Sub-surface geology
2.2 The hybrid system 2.2.1 Jacked-in pipe inclusions Mild steel pipes which functions as soil nails are installed by hydraulic jacking. This method has proven to be an efficient and effective technique for excavation support, where conventional soil nails and ground anchors have little success in such difficult soft soil conditions. Such conditions are sandy collapsible soil, high water table and in very soft clayey soils where there is a lack of short-term pullout resistance. In view of the close proximity of commercial buildings to the deep excavation, a very stiff retaining system was required to ensure minimal ground movements at the retained side of the excavation. Contiguous bored piles were installed along the perimeter of the excavation and supported by jack-in pipes. In the initial design, conventional ground anchors were proposed, but the jacked-in pipes were accepted as an alternative as it was not feasible to locate the fixed lengths of the anchors at very deep depths. From experience gathered by the contractor from past application of this technique in similar ground condition, it was found that this technique would provide a cheaper alterative and site adaptability. Relatively, larger wall movements will occur for this support system when compared to ground anchors since soil movement is required to mobilised the tensile resistance of the passive inclusions. However it was anticipated that the ground settlement at the retained side and maximum lateral displacement of the wall using this system would still be within the required tolerance after engineering assessment.
726
2.4 Pei$ormance of the hybrid system
2.2.2 The construction sequence The retaining wall system consists of closely spaced 1 OOOmm diameter bored piles near the commercial buildings and 8 0 0 m diameter for area away from the commercial buildings. To facilitate the TopDown excavation of the deep basement, pipes of 150mm diameter were jacked in sub-horizontally after each excavation stage. The phases of the basement construction are described in Figure 3. 2.3 Geotechnical instrumentation and field data The geotechnical instrumentation program consists of 18 vertical inclinometer tubes located strategically along the perimeter of the contiguous bore pile wall and 30 optical survey markers for settlement monitoring. Figure 6 shows the trend of measured horizontal displacement of the hybrid system at successive excavation stages. The deflection profile exhibits predominantly restrained cantilever effect due to the reinforcement effect of nail inclusions.
The maximum displacement of a soil nailed system generally ranges from H/1000 to 4W1000 (Clouterre, 1998). It is of interest to find out how this hybrid system compares with conventional soil nailed systems. For this system, the maximum ever recorded horizontal displacement after an excavation depth of 7.5m (stage 8) was 98mm and its corresponding normalised value is 0.013.or 13W1000. This value is not within the range of HA000 (7.5mm) to 4W1000 (30mm). This is expected as the normalised values that ranges from H/1000 to 4W1000 were obtained from field studies conducted in stiff competent soils. 2.4.1 Horizontal deflection Figures 4 and 5 shows the normalised maximum horizontal displacement for various excavation depths for a critical section of the excavation. The trend of increment seems to indicate that when the excavation reaches a depth of 6m, there is a sharp drastic increased in the horizontal deflection for the whole hybrid support system. In may be speculated that as the excavation reaches the 6m depth, the soil mass at the retained side has reached full plasticity. It is further speculated that due to the presence of the long nail inclusions it changes the strain profile of the soil (Basset & Last, 1978), large wall movements were required to attain full plasticity of the soil mass at the retained side. This global mode of resistance exhibited by the presence of the nail inclusions is known as strain arrestment (Basset et.al., 1999). 2.4.2 Vertical settlement The normalised vertical displacement for the whole support system is shown in Figures 4 and 5. The normalised vertical displacement for an excavation depth of 2.7m to 6m reaches a constant value. How-
Figure 3. Construction stages NORMALISED DISPLACEMENE vs DEPTH EXCAVATION (D)
HORIZONTAL v s VERTICAL DISPLACEMENT AT VARIOUS STAGES
OF
Figure 5. Horizontal and vertical displacements
Figure 4. Normalised horizontal and vertical displacements
727
Figure 6a. Summary of geotechnical field data on measured deflection profiles.
Figure 6b. Location of field monitoring positions
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in-situ wall for the stabilisation of a deep excavation located in soft soils. This technique of soil nailing is a departure from the conventional method of nailing as it was used in conjunction with a stiff cast in-situ for the stabilisation of a deep excavation in soft soil. The measured deflection profile of the system indicated that movements were restrained and as the excavation progresses to a depth of 6m, there was a sharp increased in wall lateral movement. Overall, this new hybrid support system located in soft soil has performed well.
ever there is a trend of sharp increase in soil settlement at the retained side when the excavation reaches a depth of 6m and it corresponds to the instance when there was a sharp increased in horizontal displacement .
2.4.3 Overall effect of the inclusions The presence of the long inclusions changes the strain field of the retained mass. The reinforced soil mass acted very much in a homogenized manner and this can be seen from the restrained cantilever deflection profile of the CBP wall as shown in Figure 7.
4 ACKNOWLEDGEMENTS The authors would like to thank SGE Sdn. Bhd for the provision of field data and support. Many thanks to Ir. Dr. Gue See Sew for his kind assistance. REFERENCES Basset, R.H, Tan, S.A & Cheang, W.L., 1999, Private Communication. Basset, R.H & Last, H.C, 1978, Reinforced Earth below footings and embankments. Proceedings of A X E Symposium, on Earth Reinforcement, Pittsburgh, 212-221. CLOUTERRE French National Research Project, 1993, Recommendations Clouterre 1991-Soil Nailing Recommendations, Presses de 1’Ecole Nationale des Ponts et Claussees. Cheang, W.L., Tan, S.A., Yong, K.Y., Gue, S.S., AW, H.C., Yu, H.T. & Liew, Y .L. ,2000, Jacked-in pipe reinforcement of a deep excavation in soft soil, Field measurements in geomechanics, 3 1 1-318, Balkema. Liew, S.S, Tan, Y.C. & Chen, C.S. 2000, Design, Installation and Performance of Jack-in Pipe Anchorage System for Temporary Retaining Structures, International Conference on Geotechnical and Geological Engineering. GeoEng 2000, Balkema. Tan, S.A., Cheang, W.L., Yong, K.Y. & Dasari, G.R., 200 1, Resistance of Jacked-in Pipe Inclusions in Soft Soil, International Symposium on Earth Reinforcement, Kyushu.
Figure 7. Restrained cantilever deflection profile
3 CONCLUSION Reported herein is a case study where Passive nail inclusions were used in conjunction with stiff cast
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Case study on engineering behaviors of the Simajiri mudstone for darn construction - slope reinforcement in dam reservoir Y. Uchimura, Y. Otsuka & F. Motida Oyo Corporation, Japan
T. Nakamura & Y. Tamaki Dam Ofice, Okinawa Prefecture, Japan
H. Uehara Uehara Geotectonical Engineering Research Institute, Japan
ABSTRACT: “K’dam was constructed with large excavation of Shimajiri group mudstone (Shimajiri mudstone) deposited in Tertiary . This mudstone is low in strength (soft rock), weak consolidation and especially poor in slaking-durability, consequently it is easy to deteriorate and become muddy. The design and performance considering the characteristics of the mudstone were carefully studied for the dam construction and cutslope in the dam reservoir. In this study, the rebound phenomenon accompanying with the excavation was observed by highly precise extensometer in the dam reservoir bed, and was related to the decrease of strength of the Shimajiri mudstone by laboratory tests. On the contrary, the decrease of strength was not observed in the cut slope countermeasured by concrete frame with a tie-back. These results indicate that this reinforcement method is highly effective to deterioration of Shimajiri mudstone and may be better consideration to solve the problems of the similar works in Shimajiri mudstone.
1 INTRODUCTION
Table 1 . Scheme of dam.
1.1 Outline of works
Dam
“K’dam is a gravity type dam constructed for flood control. The scheme of the dam and the standard cross section of the dam are shown in Table1 and Figure 1 respectively. 1.2 Geological aspects “K’dam was constructed in the middle of a typical urban river. The geological feature around the dam is shown in Figure 2. Yonabaru layer mudstone, which belongs to the Shimajiri group is mainly deposited, and Quaternary Ryukyu limestone partly covers this formation. This mudstone was sedimented in the ocean and the small faults are found obviously. Thin seams of the sandstone are often sandwitched in the mudstone layer. The slope failures and landslides occur in the mudstone slopes around the dam. The view plan around the dam reservoir is shown in Figure 3.
Type Dam height Dam length Dam crest level Dam volume Reservoir River basin area Reservoir area Total reservoir volume Effective reservoir volume Low water level Surcharge water level High water level
Grav3 m 120 m
E.L.52 14,250 1.7 0.04 5 10 000 470:OOO E.L.36.4 E.L.49.5 E.L.50.5
1.3 Fundamental properties of mudstone Physical and mechanical properties of the mudstone are shown in Table 2. This mudstone consists of clay and silt, and is very homogeneous, weak-consolidated and low in strength. Potential swellingness is low, but it deteriorates easily by the repetition of wet-dry condition and results in strength decrease.
Figure I . Standard cross section of “K’dam.
73 1
m m3 km’ k$ m’
m’
m m
m
Figure 3. Dam reservoir plan of “K’dam. Table 2. Basic property of mudstone. properties Grain size
Table 3. Scheme of slope protection.
test data Clay Silt Sand
Dry density Water content Void ratio Uniaxial strength Cohesion Angle of internal friction Slaking index Swellingness
cast in-situ slurry wall concrete frame
50 % 49 % 47 1 3 % 1.5 1.7 g/cm3 2 2 - 31 % 0.55 0.87 1 - 3 N/mm2 0.25 - 0.53 ?/mm’ 2 3 - 35 34 0.6 %<
-
anchor
H = 1 lm 500mmx 500mm
4,000mmx4,000mm end-enlarged anchoring method 1 type D220mm 2type D270mm
2 DESIGN AND PERFORMANCE OF SLOPE PROTECTION Figure 5. Stress distribution of end-enlarged type anchor.
2.1 Outline of design To ensure the capacity of the dam reservoir, the deep excavation in the riverbed was undertaken due to the topographic limitation. Moreover, the dam reservoir was divided into two parts, that is, the upper reservoir and the lower reservoir in the consideration of environmental problems such as housing, traffic road, cemeteries and cultural assets. Consequently, a large-scale of cut slope, which consists of mudstone were constructed around the dam reservoir. Therefore, the slope fail732
ure and the landslide, which would accompany deterioration of mudstone, were discussed and the reviewing of the slope protection methods was examined. Finally, concrete frame with a tie-back and cast in-situ slurry wall with a tie-back were studied for the countermeasure, which could prevent from slope failures due to deterioration. As for a tie-back, “endenlarged anchoring method” was adopted to secure the anchoring. In Figure 5 , the stress distribution around the end-enlarged anchor is shown.
E s c a a~tio n
tl c t P 1- i o r a t i o n Mortar blow
Pull-out test Anchor ev alualion Construction anchor
I
Vegetation o f slopc
Figure 6. Procedure of slope protection.
Figure 7. Excluding method of borehole slime.
2.2 Outline of slope protection The performance procedure of the slope protection is shown in Figure 6. Pull-out test was performed to confirm the bond capacity of the anchor. As a result, it was indicated that the conventional anchoring method could not secure the prescribed pull-out force, and so the method of excluding the borehole slime as shown in Figure 7, was adopted. 3 GROUND DEFORMATION AND CHANGE OF SOIL PROPERTIES 3.1 Ground deformation behavior due to excavation Field measurements were performed at the central part of the dam reservoir to observe the deformation behavior caused by the excavation. The results obtained by using highly precise extensometer (sliding micrometer) are shown in Figure 8. The upheaval deformation of the ground occurred from the ground surface to the depth of 30m. Especially, the deformation in the surface of excavation site is larger, and it occurs immediately after excavation and continues for a long time. But at the part of the cut slope reinforced by a tie-back and cast in-situ slurry wall, obvious deformation was not observed.
Figure 9. Deformation properties of dam reservoir.
(6m, 12m, 21m) with boring core specimens of the dam reservoir bed to compare the rebound quantities with excavation depths. On the other hand, boring core specimens of the 5m depth in the cut slope were tested. The results of the specimens in the dam reservoir bed are shown in Figure 9 and Figure 10 respectively. The shear modulus of the specimen at the 6m depth is much smaller than the others, but it recovers easily, as the confining pressure gets larger. Peak strength of the shallow depth specimen (6m) is much smaller than that of the deeper points (12m, 21m). From the above mentions, it was cleared out that the strength and deformation properties of the Shimajiri mudstone must be strongly influenced by the rebound behavior due to excavation. Also, the decrease of strength by change of water content was studied, using the specimens at the 21m depth as shown in Figure 10. It was clarified that the
3.2 Changes of ground soil properties The changes of soil properties due to the excavation were studied by laboratory tests. Test specimens were sampled from boring cores in the dam reservoir bed and the cut slope, and specific direct shear test was adopted to know the mechanical properties. Laboratory tests were carried out at three depths 733
slight change of water content could influence the strength of the mudstone.
4 EFFECT OF SLOPE PROTECTION The strength of the mudstone at the shallow depth (5m) in the cut slope was the same as that at the 12m depth in the dam reservior bed as shown in Figure 11. If the deterioration progresses at the same speed in both the cut slope and the dam reservoir bed, the strength in the cut slope might decrease more. However, the decrease of strength in the shallow part reinforced by a tie-back was of strength restrained, and this phenomenon is evaluated as the effectiveness of the slope protection. 5 CONCLUSIONS The results obtained by the research this time are as follows:
1. “K’ dam was based on the Shimajiri formation clay stone which was easy to cause slaking, Therefore, the concrete frame with a tie-back and cast in-situ slurry wall with a tie-back were applied for its construction in order to prevent slope failure due to the deterioration of mudstone layer excavated. 2. Field measurement at the dam site showed deterioration at the dam reservoir bed where the countermeasure was not applied, while no deterioration was found at the cut slope where the countermeasure was applied. 3. According to the results from the laboratory tests, the excavation bed surface with larger deterioration had larger deformation and smaller strength. 4 The slight variance in the moisture content was correlated with the significant decrease in shear strength, and the effect of repeated dry and wet was evaluated by moisture content changes. 5 The decrease in shear strength of the mudstone at the cut slope is much smaller, compared to that of the mudstone at the upper part of the dam reservoir bed. This is considered to be the effect of the slope protection with a tie-back. 6. Judging from the above mentions, it is important to take measures to prevent repeated dry and wet action or slaking as excavating the mudstone ground, In this case, cured protection mat and mortar blow were effective countermeasures. Also slope protective works such as anchor works were very effective to restrain the decrease of strength of the mudstone and to prevent slope failure in the mudstone layer.
6 CLOSING REMARKS AND ACKNOWLEDGEMENT Shimajiri Formation clay stone forms the ground basis of the southern area of the Okinawa Main Island, which many civil engineers are familiar with. It is also known that it has geotechnical characteristics of slope failure and landslides. However, investigative study where we observed the deformation behavior at the site of large-scale excavation, revealed the strength characteristics in relation with deformation of the mudstone and the effect of the countermeasures against deterioration by slaking. It is our pleasure if the present study could be useful for the design and construction method of the dams and foundations problems in the Shimajiri formation clay. Finally, the authors appreciate greatly to the Okinawa Prefectural Dam office and all those who took part in the construction of “K”dam. Figure 11. Comparison of strength in dam reservoir bed and cut slope.
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REFERENCES Nakamura, T. 1993. Design and construction of the Reservoir slope of the Kinjo Dam in Japanese. Dam Japan. ~01.582:75-92. Tamaki, Y. 1999. Design and construction of the Kinjo Dam. Proc of Research Presentation Meeting of Department of Civil and architectural Engineering, Okinawa Pref. VOI(15):33-38.
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Uchimura, Y. et a1 1999. Rebound Behavior of Shimajiri mudstone Caused by a Large-scale Excavation. Okinawa geotechnical Research association. vol. 12 :22-25. Mochida F. et al. 2000 Relation between the Dynamic Characteristics of shimajiri group mudstone and the Rebound Behavior Caused by Excavation in a Large-scale. Okinawa Geotechnical Research association. vol. 13:2-7.
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Landmarks in Earth Reinforcement, Ochiai et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Kinematics and failure of soil-nailed excavation models in dynamic centrifuge tests M. Vucetic & J. Kocijan University of California, Los Angeles, California, USA
M. Doroudian Bing Yen & Associates, Iwine, California, USA
ABSTRACT: The influence of nail length variation with depth on the kinematics and stability of soil nailed excavations under dynamic loads is analyzed from the results of three dynamic centrifuge tests. The centrifuge models represented prototype excavations 7.6 meters deep with three rows of nails. Length of nails varied between the tests such that the total nails expenditure per test remained constant. In the first model all nails had the same length that can be characterized as medium. In the second model the nails in the top row were long, in the middle row medium and in the bottom row short. In the third model the nails in the top row were short, in the middle row medium, and in the bottom row long. The difference in nails arrangement resulted in somewhat different movements of facing and nailed soil mass and a very different resistance to dynamic load. 1 INTRODUCTION Soil nailing is an in-situ method of soil reinforcement for stabilizing excavations and slopes by placing passive inclusions in the ground as the excavation proceeds (Gassler, 1998; Juran and Elias, 1991; Mitchell and Christopher, 1990). The use of soil nailing systems in earthquake regions requires understanding of their behavior under seismic loads. Some of the first experiences with seismic stability of soil-nailed excavations were gained during the 1989 Loma Prieta Earthquake in San Francisco Bay area (Vucetic et al., 1998) and from two dynamic centrifuge tests studies. In the first centrifuge tests study (Vucetic et al., 1993) varying nail lengths and nail rigidities were investigated on 4 models, while in the second study (Vucetic et al., 1996) various parameters influencing seismic stability and kinematics of failure were investigated on a total of 14 models. In this paper the results of tests on three models from the second centrifuge study are analyzed in certain detail to examine the influence of nail length variation with depth on the kinematics of failure and overall seismic stability.
here. The other details can be found in the above referenced publications. The longitudinal cross section of one of the models displaying the size of the model box, arrangement of nails and facing, distribution of accelerometers and locations of LVDTs for recording of displacements is presented in Figure 1. The width of the model box was 203 mm. The centrifuge testing scaling factor was 50, meaning that the dynamic tests were performed in-flight at 50g and that the prototype excavation height was 7.6m. To obtain prototype accelerations, the recorded accelerations must be divided by 50. To obtain prototype displacements, the recorded displacements must be multiplied by 50.
2 DESCRIPTION OF MODELS, TESTING PROCEDURE AND TESTING PROGRAM Dynamic centrifuge tests were conducted according to the procedure described in Vucetic et al. (1993) and its modifications described in Vucetic et al. (1996). Only the information needed to understand the test results presented below are summarized
Figure 1. Longitudinal cross-section of the model with instrumentation configuration (dimensions are in millimeters).
737
acceleration of the model box. Accelerometer No. 1 was installed in the nailed soil mass to record accelerations of the soil nailed block, while accelerometers No.2 and 3 were installed in the soil behind the potential failure surfaces. LVDT 1 recorded settlement of the nailed soil mass behind the facing, while LVDTs 2, 3 and 4 recorded horizontal movement and deformation of the facing. Three different models analyzed in this paper are presented in prototype dimensions in Figure 2. In Model A all of the nails had the same length. In Model B the nails in the top row were long and in the bottom row short. In the third Model C the nails in the top row were short and in the bottom row long. The corresponding length ratios (nail length divided by excavation height) are given in Table 1. Standard length ratio recommended in practice is between 0,5 and 0.8, which is labeled here as “medium”. It must be noted that the total nail expenditure per model was the same, i.e., cumulative length of nails in every model was constant. In this context, the test data presented below suggest which configuration of soil nails is more economical. Table 1. Nail length ratio distribution
I Nail
Test (Benchmark) TEST A
top middle
0.67 - medium
TEST B (Top nails-long)
top middle bottom top middle bottom
1.OO - long 0.67 - medium 0.33 - short 0.337 short 0.67 - medium 1.00 - long
TEST C (Top nails-short)
I
Table 2. Cyclic loading program
Figure 2. Investigated soil-nailing system configurations with varying prototype nail lengths.
In each model there were three rows of model nails made of glass-filled polycarbonate plastic having axial and flexural rigidities that roughly represent prototype grouted nails often used in practice (Bruce and Jewel, 1986; 1987). Horizontal distance between the model nails was 50mm. The model facing in all models was made of thin plexiglass sheet representing relatively strong and rigid prototype facing. The connection between the nails and facing was strong enough to hold throughout the testing. Soil used was uniform moist silica sand. The sand was build into the model box in alternating layers. Every other layer was black-dyed to facilitate the recognition of failure patterns and failure surfaces during the test and after the removal of the models from the centrifuge platform. As shown in Figure 1, every model was instrumented with four accelerometers and four LVDTs. Accelerometer No. 4 was used to control the applied 738
I Length Ratio
TEST
0.18
1
0.3 1
A
C 5
1
6
1
0.42 0.44
~~
I
Each model was subjected in-flight at 50g to several consecutive series of 10 cycles of approximately uniform acceleration amplitude, &y The magnitude of average &y applied (via accelerometer No. 4) varied from series to series as listed in Table 2. The table also includes cumulative settlements behind the facing (LVDT 1) and horizontal displacements of top nails (LVDT 2) at the end of each cyclic series. In the first cyclic series &y was around 0.18 (g/50). In the subsequent cyclic series &y was larger, from 0.30 to 0.35 (g/50), and afterwards it was even larger from 0.40 to 0.44 (g150). 3 KINEMATICS OF FACING AND FAILURE PATTERNS Horizontal movement of facing was recorded during tests with three LVDTs placed as shown in Figure 1. Recorded data are presented in Figures 3 to 6. Figures include input accelerations (box accelerometer No. 4), settlement behind the facing (vertical LVDT 1) and horizontal displacements of the facing (top, middle and bottom horizontal LVDTs 2,3 and 4). Data from test on Model A are presented on Figure 3, showing that three cyclic series with &,= 0.18, 0.31 and 0.40 g/50 were applied before a catastrophic failure took place. In every series the facing experienced oscillatory motions with the residual displacements towards the excavation and downwards.
Records of the vertical LVDT 1 indicate that the soil nailed mass behind the upper part of the facing oscillated accordingly. Cyclic Series 1 with &?O. 18 g/50 has not produced significant displacernents towards the excavation. Series 2 with h ~ 0 . 13g/50 produced relatively large displacements, while Series 3 with k ~ 0 . 4 0g/50 brought the system to complete failure. Such movement of facing and soil mass behind it is graphically portrayed on Figure 7a. The displacements recorded by LVDTs 2, 3 and 4 are connected with straight lines which in a simple manner describe the outward movements of the facing. Beginning and the end of each cyclic series are represented by solid lines, while the facing positions after cycles 3 and 6 within each series are represented by dashed lines. Facing was deformed according to the forces and moments applied to it, so in reality, of course, it was not shaped as a broken line but a continuous curve. To gain better insight into the kinematics of the nailed soil mass and the mechanism of ultimate catastrophic failure, after each test the centrifuge models were dissected and thoroughly analyzed. The resulting failure patterns are presented on Figure 8. The following observations can be made about the behavior of Model A from Figures 3, 7a and 8a. During cyclic shaking facing rotated around its toe due to the anchoring effect of the bottom row of nails and passive resistance of the soil in front of the
Figure 7. Simple presentation of the facing movement a) i n Test A, b) in Test C.
toe of the excavation where facing penetrated downwards into the soil. How the bottom nails extended into the soil behind the failure surface where they acted as anchors, and how the toe of the facing cau-sed passive failure of the soil in front of it, is clearly visible on Figure 8a. It is also evident from the figure that during cyclic loading the nailed soil
Figure 6. Accelerations applied and displacements recorded in Test C - cyclic series 4 3 and 6.
740
surface along the ends of the nails with the slope practically identical to that of Model A. However, in this model there was no anchoring effect of bottom nails, which caused a substantial reduction of the seismic stability. Failure of Model B occurred after only two cyclic series. The kinematics and failure of Model C are illustrated on Figures 5 , 6, 7b and 8c. In this model, the anchoring capacity of the long bottom nails was the greatest, which resulted in the greatest overall seismic stability. The system completely failed after 6 cyclic series of ten cycles of large accelerations. However, as shown on Figure 8c a local failure surface developed behind the short nails at the top of the structure. The resulting local failure was accompanied by significant bending of the facing. Had the facing been less strong, this portion of the system would have failed earlier. A sudden jump in the record of the top horizontal LVDT 2 in Cyclic Series 2, which is marked in Figure 5 by an arrow, indicates that this local failure has most likely started at the beginning of Cyclic Series 2. Otherwise, the slope of the failure surface in Model C is practically the same as in Models A and B.
4 CONCLUSIONS
Figure 8. Ceometries of failure observed on the models removed from the centrifuge.
mass acted as a block that was sliding incrementally on the inclined failure surface, and that the system ultimately failed after the bottom nails started to be pulled out. Such failure pattern and kinematics of the nailed soil mass have already been observed earlier on similar tests by Tufenkjian and Vucetic (2000). It should be noted that the level of shaking required to fail Model A corresponds to extremely strong earthquake that is practically impossible to occur. In that sense, the behavior of Model A confirms that the soil nailed structures with grouted nails having length ratio of around 0.7 and built in a standard manner are seismicaIly very stable (Vucetic et al., 1998). The kinematics and failure of Model B are illustrated on Figures 4 and 8b. In the test on Model B LVDTs 3 and 4 malfunctioned, hence the presentation of the facing movement such as in Figure 7 could not be obtained. Nevertheless, Figures 4 and 8b clearly indicate that the reduction of the nail length with depth triggered the generation of failure
From the results of centrifuge tests on three models presented in this paper the following conclusions can be derived. 1. Centrifuge testing is a very useful tool for analysis of kinematics and failure patterns of complex geotechnical systems under cyclic loads, such as soil-nailed excavations. 2. In the models investigated, the main function of the top nails was to tie soil and nails into a "soft block" that under cyclic loads acts as a unit. On the other hand, the bottom nails have dual function. They tie soil together just like the top nails, but if they extend beyond the failure surface they also act as anchors. 3. The anchoring capacity of the bottom nails extending beyond the failure surface can significantly contribute to the seismic stability of soilnailed excavations. 4. To achieve the greatest seismic stability of soilnailed excavations with the given total length of nails (total nails expenditure) longer nails should be installed at the bottom of the excavation, while the top nails can be shorter. However, in such a configuration a special consideration should be given to the possibility and consequences of local failure in the zone of the top short nails. 5. Having the short nails at the bottom and long nails at the top of the soil-nailing system eliminates the anchoring capacity of bottom nails, thereby reducing seismic stability.
74 1
5 ACKNOWLEDGEMENTS The centrifuge tests were conducted at the Rensselaer Polytechnic Institute (RPI) Geotechnical Centrifuge Research Center in Troy, New York. The excellent support and cooperation of Professor Ricardo Dobry, Professor Ahmed Elgamal and Dr. Korhan Adalier during the testing are gratefully acknowledged. The authors also wish to thank Professor Mark R. Tufenkjian from California State University, Los Angeles, for his help in preparing the testing program. The research was supported by the National Science Foundation (NSF) and Federal Highway Administration trough the NSF Grant no. BCS-9224479. REFERENCES Bruce, D.A. & Jewell, R.A. 1986. Soil nailing - Application and practice - part 1. Ground Etzgineerirzg, November, p. 10-15. Bruce, D.A. & Jewell, R.A. 1987. Soil nailing -Application and practice - part 2. Ground Engineering, January, p. 21-33. FHWA, 1996. Manual for Design and Construction Monitoring of Soil Nail Walls. Federal Highway Adnzinisrmtion Report, NO. FHWA-SA-96-069, 392 p.
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Gassler, G. 1988. Soil-Nailing--Theoretical Basis and Practical Design. Proceedings of the Symposium on the Theory of Earth Reinforcenient and Practice, Kyushu, Japan, Balkema, Rotterdam, pp. 283-288. Juran, I. and Elias, V. 199I . Ground Anchors and Soil Nails in Retaining Structures. Fourzdatioii Engineering Handbook, Chapter 26, Second Edition, Van Nostrand Reinhold, New York, pp. 868-905. Mitchell, J.K., and Christopher, B.R. 1990. North American Practice in Reinforced Soil Systems. ASCE Special Technical Publication, No. 25, 322-346. Tufenkjian, M. R., and Vucetic, M. 2000. Dynamic Failure Mechanism of Soil-Nailed Excavation Models in Centrifuge. Journal of Geotech. and Geoenvir. Engineering, ASCE Vol. 126, No. 3, pp. 227-235. Vucetic, M., Iskandar, V.E., Doroudian, M., and Luccioni, L. 1996. Dynamic Failure of Soil-nailed Excavations in Centrifuge. Proceedings of the lnternatioiiul S~~tnp~~siuriz otz Eurth Reinforcement, Fukuoka, Japan, Nov., A.A. Balkema Publisher, Vol. 1 , pp. 829-834. Vucetic, M., Tufenkjian, M.R. and Doroudian, M. 1993. Dynamic Centrifuge Testing of Soil Nailed Excavations. Gem technical Testiizg Jourtzal, CTJODJ, Vol. 16, No. 2, June. Vucetic, M., Tufenjkian, M.R., Felio, G.Y., Barrar, P. and Chapman, K.R. 1998. Analysis of Soil-Nailed Excavations Stability during the 1989 Loma Prieta Earthquake in California. USGS Prqfessiotial Paper I552 “Perfi)rmance of the Built Envimnnzent”, part of NEHRP Report to Congress: “The Loma Prieta, California, Earthquake of October 17, 1989.”
Landmarks in Earth Reinforcement,Ochiai et al. reds), 0 2001 Swets & Z~itlinger~ 1SBN 90 2651 863 3
Stabilization of slopes and landslides using soil nailing methods L.E. Wichter Chair ~ e o t e c J ~ ~~~r ~~ c~si,~ d e Tec~nical i ~ b i ~ r s~ n i v e r s~i o~t~t b u ~Germany s,
W. Meiniger Research and Material Testing Institute, University of Stuttgart, Germany
ABSTRACT: Landslide-endangered slopes and active landslides are often stabilized in Germany using soil nails. In general two soil nailing methods are used. One method is characterized by a combination of soil nailing using steel tubes as nails, and cement grout injection. The other method uses threaded tie bars as tendons in grout-filled boreholes. The report describes some examples for the two nailing methods, and their advantages and disadvantages. 1 INTRODUCTION The injection nailing method is characterized by a combination of soil nailing and cement grout injection. Perforated steel tubes of 1.5" or 2.0" diameter are inserted into vertical or inclined boreholes and used as injection channels for the injection of the surrounding soil. The tubes remain in the boreholes as reinforcement nails, and in some cases additionally steel rods are inserted into the tubes before the stiffening of the cement grout begins. The nails are cut some 50 cm below the ground surface, and the ground may be used again for agricultural purposes, such as vineyards, etc. The stabilization effect is a ~ombinationof increasing the shear strength of the soil by the effort of the grout, and the nailing or dowelling effect of the steel tubes penetrating the slip plane. The method i s very common especially for the stabilization of landslides under difficult conditions because the machinery needed may work even in very steep slopes. The stab~lizationeffect of the grouting depends on the permeability of the soil, and of course it is difficult to estimate and nearly impossible to calculate. The second method uses threaded steel bars of 32, 40, 50 or 63.5 mm diameter to reinforce the slopes. The nails (simple or double co~~osion-protected) are brought into boreholes filled before with cement grout. After the setting of the grout the nails connect the slide mass with the stable underground by skin friction. This method is also characterized by the circumstance that no concrete or steel parts, such as anchor plates, at the ground surface are needed. The nails are cut below the ground surface, and they transfer their forces by skin friction into the slide mass or slide-endangered soil body. Usually nail distances of about 2.5 x 2.5 m are used, and measures with nail lengths up to 24 m have been carried out.
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The method described has won great actuality after the German reunification when for a large number of old railway embankments in the former communist part of Germany a construction method was needed for the improvement of the stability of the embankment slopes before higher velocities of the traffic on the top could be allowed. 2 LANDSLIDE STABILIZATION USING PIPE INJEC~IONNAILS 2. I Stabilizcrtion of a steep slope iiz Devonian SCJZiStS The landslide occurred in a natural slope near the Mosel river which had an inclination of 30". The slope was traversed by a new highway, and the area had not been known as prone to sliding before the beginning of the excavation measures. Anyhow the steepness of the slope required the construction of a gabbion wall (length 90 m; max. height 12 m) at the valleyside of the road in order to relieve the slope of the weight of an otherwise required fill. The ground was a heavily jointed Devonion schist. Near the ground surface the joints were widely opened due to gravitational movements of the rock mass. The groundwater level varied locally and was found in depths between 1.7 m and 13.0 m below the ground surface. Some borings showed a complete loss of circulation. The landslide had an estimated cubic content of 370.00~m3. The landslide started moving after the road was finished. Reconnaissance drilling showed various faults in the subsoil where the rock was mylonized to a soil (friction angle 22", cohesion 2.5 N/cm2, from direct shear tests). Multishot measurements showed a slip plane in about 17 m depth below the
Figure 1. Plan view of the area and the stabilization measures.
Figure 2. Cross section of the slope.
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road. The maximum velocity of the landslide was appr. 2.2 cdmonth, the average velocity 8 mm/ month, changing with the amount of precipitation and the season. The landslide was stabilized using 2"-pipe injection nails which were arranged in groups of roughly 30 pipes each, and drainage borings. The stabilization was planned to be carried out in three steps. In each step only a part of the totally planned number of nails was installed, and the stabilization effect was observed using inclinometer measurements. Only the first two steps were carried out, Figure 1
shows a plan view of the area and the stabilization measures, Figure 2 a cross-section of the slope and the stabilization measures, and Figure 3 a photograph of the situation during the stabilization works.
2.2 Large scale shear tests on injection nails In the past the soil nails very often had been arranged vertically because of easier drilling, and because there had been the imagination that the nails would act as small piles, or dowels. Large scale shear tests (Figure 4)on nails have shown that this is not the case. The nails should be arranged as good as possible in the direction of the most intensive shear extension as shown in Figure 5. The stabilization effect of the injection nails was found to be significantly higher when the nails were arranged correctly as shown in Figure 5 (Wichter, Meiniger & Gay
1991).
Figure 3. View of the landslide area (with deformed gabbion wall) during the stabilization measures.
Figure 5. Wrong and mechanically correct arrangement of the injection nails.
3 SLOPE STAE!ILIZATION USING TIE THREADBAR SOIL NAILS
3.1 Stabilization of a cutting slope under extremely bad ground conditions
Figure 4. Direct large scale shear test on an injection nail in loess soil (shear box size 2x1~4111, upper shear frame removed).
For the construction of a bypass highway in the area of Stuttgart a cutting slope of 17 m height had to be excavated. The ground conditions were difficult. Under an overburden of loess and loess clay of some 5 1-n thickness an extremely wates-bearing stratum of claystones and limestones was explored (see Figure 6) belonging to the geological formation of Lias a. Below this a stratum of partly weathered Keuper Mar1 followed known in Southern Germany as extremely landslide-endangered as soon as it comes in contact with water. Because of lack of space at the top of the cutting to a private area and building the slope angle of the upper part of the slope (1:1.9)had to be chosen steeper than it was desirable from a geotechnical point of view, and the lower part of 8 m height was decided to get a reinforcement using soil nails in a grid of 1.40 x 1.80m, and a shotcrete lining of 4:l slope angle. The soil nails (double co~osion-protected steel bars BSt 500/550 s, 050 m) had lengths up to 23 m which were d e t e ~ n e dusing circular
74s
+++
OVERBURDENLAYERS (LOESS)
4
/
.Q
I
I
'
/
-.-.____
SINGLESIZED CONCRETE SHEAR WALL
I
'1
----- 1
m H--=-*--*------L ,-*
/
---et
/
/
LIMESTONE AND CLAYSTONE
/
KEUPERMARL (KNOLLENMERGEL)
---.-
-4-
---
//-
---G.-*
DRAINAGE BORINGS PERMEABLE FILL L.- DRAINAGE PIPE
-*
-
-'-
___---A-------
=-*--.-*-
"\.
Is
-*-
-._
-a-.
-*-
I_
I -0.'
-'--
Figure 6. Sectional drawing of the cutting and the nailing measures.
Figure 9. Landslide beside the stabilized slope.
slip planes as failure mechanisms. The shotcrete facing was later covered with a lining wall of prefabricated concrete parts. Figure 7 shows a view of the slope during the nailing works, Figure 8 the preparation of the nails. As it may be seen on Figure 9 in the neighbourhood of the reinforced slope landslides occurred during a heavy rainfall which forced the authorities additionally to extensive stabilization measures. 3.2 Stabilization of a cutting slope beside an express railway During the excavation works of a cut-and-fill-tunnel for a new express railway slope failures occurred.
Figure 8. Soil nails with pipes for postgrouting.
746
The failures were caused by reactivated fossile slip planes in the Keuper Mark following the bedding planes dipping to the cutting. The slope failure necessitated extensive anchoring measures (see Fig. 10) before the tunnel could be constructed and the cutting was refilled (Fig. 10, the measure described here was executed at the right-hand side of the picture). Immediately joining the western tunnel portal, and under the same geological conditions, the railway was planned to run in an open cutting. The dip of the planned uphill slope was only a bit less than the dip of the failed slope. The danger of new landslides was obvious. Therefore the railway authorities followed the proposal to stabilize the slope by reinforcing it using soil nails. Figure 11 shows a crosssection of the slope and the stabilization measures. The soil nails (threaded steel bars 0 50 mm, double
3.3 Increasing the slope stabilio?of railway embankments Many German railway embankments have been built between the years 1850 and 1900. Because of the lack of heavy machinery in those times the fill material normally was not compacted very intensive, and the embankment slopes were built very steep. In the course of the advance of the railway network (especially after the German reunification in 1990) the
Figure 12. Slope stabilization using long soil nails
Figure 10. Tunnel, filled and anchored cutting.
I
corrosion-protected) were arranged in a grid of 2.5 x 2.5 m and had length up to 24 m. The stabilization was supplemented by a system of drain-bores, and is surveyed using extensometers. Figure 12 shows the slope surface and the drilling measures for the nails using a special drilling machine. The measure has been described in detail by Samaras, Ga13ler & Wichter (1988).
y = 21 k N / m 3 c2= 5 kN/m2
-2
c i 2 = 10 k N / m *
LOCATION PLAN OF FORCES
G2
E w m l
.c
1 =12O
POLYGON OF FORCES
-I
I
4
I=50m
COMMENTS: - Statical active nail section either before or behind the slip plane: I w . , Im], (i = nail row nr. i) - horizontal nail distance: b = 2.5 m - medium mobilized shear force along I in limite state: Tm= 25 kN/m - available calculated nail force sum 2 [kNIm] in limite state: 2 = Z NI I b =Tm.C I W J / b - stability factor of the nailed slope: . n. = tan 15" /tan 12" = 1.26
Figure 1 1. Cross-section of the slope, the reinforcement measures, statical analysis.
747
G
1;. =L20 kNln
stability of many embankments has to be checked and is found often to be not high enough for modern security demands. Following the technical specifications of the German Railway (Deutsche Bahn AG) new embankment slopes have to have a stability factor of 7 = 1.4 against sliding (when a lamella method is used). For old embankment and cutting slopes the specifications (DB AG 2000) say. 4 ESTIMATION OF THE STABILITY OF EXISTING EMBANKMENT AND CUTTING SLOPES (1) This item is valid for the estimation of the stability of already existing slopes and cuttings which are in use without damage and are maintained according to modulus 836.0900 (a part of the specification). ( 2 ) If an embankment or cutting is in use and has been without damage, and if it is maintained according to modulus 836.0900 (a part of the specification), and if for the future the straining will not be higher than in the past, it may be assumed that it is sufficiently stable and sufficiently able for use. (3) If an embankment or cutting is in use and has been without damage, and if it is maintained according to modulus 836.0900, but if for the future the straining will be higher than it has been in the past, a competent and experienced expert in geotechnics has to make investigations about the amount of change of straining and it's consequences for the stability of the embankment or cutting. (4) For embankments or cuttings in use without damage and maintained according to modulus 836.0900 lower stability factors may be allowed than for newly built embankments or cuttings. Only an expert in geotechnics may determine the reduced stability factors. They have to be substantiated and to be made on record. Even if these specifications allow the choice of stability factors lower than q = 1.4 there are many embankments with stability factors between 7 = 1.00 and 7 = 1.10 which need an improvement of their stability. For these embankments very often reinforcing their slopes using soil nails is the most economical method. All measures which need ground beside
748
1 sliding mass 2 railway embankment 4 former slope surface 5 selfdrilling nails
Figure 13. Principle of increasing the stability of railway embankments using soil nails.
the railway outside the embankment area, such as landfills, are in many cases impossible to enforce because of the objections of nature and environment protectors. The soil nails are arranged normally in the embankment slopes in a grid of 2.5 x 2.5 m, and their length is obtained from stability calculations using circular slip planes and lamella methods. Because of the lower costs and easier handling in many cases self-drilling pipe nails with lost bits are used. The drilling rod remains in the borehole as a tendon, and cement grout is used as a drilling fluid which connects the nails with the borehole wall. The main advantage of these nails is that the machinery necessary for their installation is not very heavy and can be used even on very steep and wooded slopes. The drilling machine is attached to a vehicle at the top of the embankment, or a light-weight drilling rig is fixed on the boom of an excavator or crane having already the dip of the nails. Figure 13 shows the principle of stabilizing railway embankment slopes with soil nails. REFERENCES A. Samaras, G. Gaoler & L. Wichter 1988. Hangsicherung mit Dauerbodennageln an der Neubaustrecke Mannheim Stuttgart. Eiseizbahiztechnische Rundschau, H. 4: 2 17-220. G. Gaoler & L. Wichter 1990. Stabilization of a cutting slope using long nails under special consideration of the nail force distribution. Otto-Gi-afJouninl, 1: 36-54, Stuttgart. L. Wichter, W. Meiniger & Gay I99 1. A large scale shear test for determining the bearing behaviour of soil nails in cohesive soil. Orto-Gi-af-Journal,2: 3 1 1-320, Stuttgart.
Landmarks in Earth Reinforcement,Ochiai et al. (eds), 0 2001 Swets 6: ~ e j ~ l ~fSBN n ~ e90 r ~2651 863 3
Stability analysis of reinforced slopes considering progressive failure Takuo Yamagami & Jing-Cai Jiang Dept. of Civil Engineering, The University of Tokushima, Japan
Satoru Yamabe Araigumi Co., Ltd., Hyogo, Japan
Masakazu Taki Fukken~Co. Ltd., ~ o n s u l t Engineers, ~n~ H ~ r o ~Japan h ~ ~ ,
ABSTRACT: A new procedure is presented for the stability analysis of reinforced slopes considering progressive failure based on the limit equilibrium concept. A local factor of safety is defined at the base of each slice to represent the progressive, local failure along a slip surface. An optimization approach is contrived to determine tensile resistances of reinforcement elements at an inception of failure of reinforced zones. This approach can be used for practical design of reinforced slopes to produce reasonable arrangement of reinforcements and to determine required available tensile strengths of reinforcing materials. Finally, the method is applied to a simple example to demonstrate its effectiveness, and the results obtained are discussed to provide guidelines for practical design of reinforced slopes using the method. practice. A new solution procedure is proposed which can be used to obtain reinforcement forces mobilized in the reinforcements at an incipient failure of the reinforced zone, and thus can be used in arranging the reinforcement elements.
1 INTR~DUCTI~N Stability analyses and design of reinforced soil slopes are conventionally performed using limit equilibrium methods. These methods have a common feature that a single value factor of safety against failure is assumed for a given slip surface. In other words, it is implicitly assumed that the peak strength of soil is mobilized simultaneously along a whole failure surface. In an actual slope, however, local failure may be initiated at a small portion of high stress levels or highly concentrated zone of shear strains. Then, the failed zone may expand gradually or rapidly towards eventual, overall slope failure depending on the situation. This phenomenon of progressive failure is evidently observed in reinforced soil (e.g. Huang, et al., 1994). Therefore, conventional limit equilibrium methods with a single factor of safety are essentially incapable of analyzing reinforced slopes reasonably. The authors have already developed and validated a stability analysis method for unreinforced slopes considering prog.ressive failure ( Y a m a g a ~& Taki, 1997; Yamagm, et al., 1999a). In order to represent the progressive nature of slope failure using the limit equilibrium concept, a local factor of safety at the base of each slice is defined and calculated. The proposed method was extended to investigate the progressive failure behaviour of reinforced slopes by assuming the magnitude of rein-forcement forces (Yarnagami, et al., 1999b). However the method was incomplete as the reinforcement forces were assumed to be known. The present study advances the former method in order to make it capable of being used in actual design
2 STABILITY ANALYSIS METHOD An outline of the stability analysis proposed in the authors' previous paper (Yamagami, et al., 1999b) is presented for the sake of completeness.
2.1 ~ o r n ~ u l ~ t i o ~ ~ Figure 1 shows a potential slip surface of any shape and forces acting on an infinitesimal slice where a tensile force, T, due to reinforcement is included. The symbols in Figure 1 (a) include: y=y(x), the assumed slip surface equation; y= z(x): the slope surface equation; y=y; (x): the equation of the line of effective horizontal thrust; y=yt (x): the equation of the line of total horizontal thrust; y=h(x): the line of the thrust of internal water pressure. The symbols shown in Figure f(b) are omitted herein as they are given in (Yamagami, et al., 1999b). As a local factor of safety, F, is defined at the base of each slice, we have the following:
1 F = -[(c'dx .seca + dN' tancp') dS + (TcosP+TsinP. tancp')]
(1)
The slope stability problem with Equation (1) is highly indeterminatedue to introduction of local fac749
rx Y
x,
M , = M I - ,I- ~ E [ h f ( x ) - A ] d x - T l s i n ~ . g , XI I
= M ]-I+
5 0
E L
Im
X
+
+ PIx + 2 N ,x -
L,+K,x -T,sinp.g
Figure 1. (a) Potential sliding mass.
dx Ix F
I
’dx
, (4)
where b,=x,-x,-l(i=O,1, 2, . . . , n), x,,x , . ~are horizontal coordinate of the left and right side of slice i, respectively, M, [=E, (Yrl-y,)]is a moment of E, about the rightmost point of the base of slice i. Note that tensile force T, in Equation (4) is equal to zero for slices where reinforcement is not included. Both Equation (3) and Equation (4) are a recurrence formulation. From Equation (3) a value for E, can be calculated with a previously determined value of EL.].Substituting this value of E, into Equation (4) yields an equation in which local factor of safety, FL, is contained as the only unknown. Solving this equation by, for example, the Newton-Raphson method, a unique value of Ft can be obtained. A complete solution must satisfy the boundary condition:
Y d
i
E,,= 0
A detailed solution procedure for satisfying Equation (5) will be presented in the following paragraphs.
Figure 1. (b) Forces acting on an infinitesimal slice.
tors of safety as unknowns. It has been shown, however, that the problem becomes statically determinate by simultaneously introducing the simplifying assumptions used in the Morgenstern-Price method (1965) and the Janbu method (1957). As the solution procedures have been given in detail in Yamagami, et al., 1999b, this paper only presents the basic equations and the associated notation necessary to describe the analysis developed in the present study. According to the Morgenstern-Price method, the relationship between normal total force E and shear force X (see Figure 1) can be expressed by
(5)
2.2 Solution procedure In the Morgenstern-Price method, f ( x ) is taken as an arbitrary function, for example, a constant (e.g. 1.0) or half-sine and so on. In the Janbu method, Yr is usually assumed to be located at 1/3 of slice height. However, the authors’ studies (Yamagami & Taki, 1997, Yamagmi, et al., 1999b) have indicated that f(x) and yr must be optimized in the present problem so as to obtain completely converged solutions. Since E,, is a function made up of A, f ( x ) and yr, the boundary condition shown in Equation ( 5 ) can be reached by optimizing the objective function in Equation (6):
where h is an unknown parameter. As will be shown later, the function fix) has to be optimized in the present analysis. For a slope divided into n slices, the formulation can be carried out using a similar derivation to the procedure by Morgenstern-Price. The basic equations from which the solution can be obtained are shown (Yamagami, et al., 1999b):
wherefi=f(xi); i=O, 1, 2, . n. Equation (6) can be solved by an iterative procedure for non-linear programming, and the NelderMead simplex method is applied in the present paper. a,
750
where m is the number of slices which reach residual strength. If Fover-at1 is less than I , the slope is judged to be unstable (or failed).
Figure 2. Modeling of softening (after Law & Lumb, 1978).
Starting with a set of initial values for the independent variables included in Equation (6), an optimization process is repeated till the minimization of the objective function is realized (Yamagami, et al., 1999b).
3 NEW SOLUTION PROCEDURE
2.3 Considering softening effect Softening of soil can be easily considered according to values of the factors of safety obtained in the calculation process. Since the analysis is based on the limit equilibrium method, softening is not defined with the amount of deformation or strain. In the present study, it is assumed that immediately after reaching the peak value, the soil resistance will drop abruptly (Figure 2) to the final residual value (similar to Law & Lumb, 1978). The iterative computation procedure is as follows: i ) First, every slice is assumed to have peak strength. ii ) The local factors of safety are calculated using the calculation procedure described in the previous section. iii) If slices whose F<1 emerge, the peak strength of such slices is replaced by residual strength, then the local factors of safety are calculated again. iv) Among the slices with peak strength, if slices whose F<1 appear, the peak strength of these slices are substituted by residual one. The calculation is continued until the convergence is reached. Here, the convergence means that the local factors of safety of the slices still holding softening does not take place, the steps i) and ii) lead to convergent solutions directly. Peak-strength (Rp) and Residual-strength (Rr) are respectively expressed as: (7)
2.4 Overall Sqfety Factor In order to evaluate overall slope stability, we define the overall factor of safety Fovernf~ by a ratio between the sum of the mobilized shear forces and the sum of the available shear strengths along the slip surface as:
The previous sections described a limit equilibrium procedure for stability analysis of reinforced slopes, taking progressive failure into consideration. Using zero values of tensile forces, the procedure can provide the local factors of safety along a given slip surface within an unreinforced slope. If values of these factors of safety are below or equal to unity at a portion of the slip surface, local failure may occur at that portion, indicating potential unstability of the slope. In this case, a quantitative evaluation of the reinforcement effects can be made with predetermined tension forces of reinforcements embedded in the slope so as to cover most of or more than the local failure zone. However, it is rather difficult to directly apply this approach to practical situations, since tensile forces of reinforcenient elements are usually unknown in advance. In this section, a new solution procedure which can be used in practice is proposed. To explain the procedure, it is supposed that the slope shown in Figure 3 (a) is potentially unstable i.e. a local failure zone with factors of safety less than unity exists along the slip surface. It is also assumed to insert reinforcement elements (nails) in the slope which pass through the bases of all slices in the failure zone, as shown in Figure 3 (b). Note that nails may also be further installed in the other portions of the slip surface out of the local failure zone if necessary. And conversely they do not necessarily have to cover the whole of the local failure zone. Next, it is assumed that even after the nails are installed, failure of the slope, if it occurs, takes place initially from the reinforced zone. Under this assumption the slope never fails in reality provided that the reinforcements have sufficient strength. Then we search for tensile forces of the nails under the constraint condition that the local factors of safety of the nailed slices become equal to unity. This constraint condition implies that the reinforced zone would be at the inception of failure along the slip surface shown in Figure 3(b). If this is realized, that is if the tensile forces are actually obtained, all we have to do in design is to arrange the nails which have sufficient strength and can sustain the tensile forces with a safety margin.
75 1
Local Fs>l .O (a) Potential unstable slope (without reinforcement)
(b) Nail-reinforced slope
Figure 3. A schematic slope without and with reinforcement elements.
The present problem is thus how to determine the tensile resistance of reinforcing materials required for the situation in which the local factors of safety of the slices having the reinforcements are all unity. Here, a new approach is developed to deal with this problem by extending the authors’ previous method (Yamagami, et al., 1999b). The proposed approach can be summarized as follows.
ments are re-arranged and then steps 2) to 5 ) are repeated until the above-mentioned condition is satisfied. The procedure described above is illustrated in Figure 4. Analyze the slope without reinforcement using the method proposed by Yamagami, et al. (1999b)
3.1 First step Reinforcement elements are installed within the local failure zones
1) Stability analysis of a (potentially unstable) slope without reinforcement is performed using the authors’ method (Yamagami, et al., 1999b), and thus local failure zones where local factors of safety are lower than unity are found. 2) Reinforcing elements are installed within the failure zones (and in other places along the slip surface if necessary). Then tensile forces T, (i=l, 2, ...M), where M is the number of reinforcements installed, are evaluated in order that the local factors of safety of all the nailed slices become equal to unity. 3) Tensile forces T, can be obtained by solving the following optimization problem: M
I I
Give target values of local factors of safety for all the reinforced slices
up initial values of T, for h reinforcement element
b
w I
2
U=U(7;)=C(F0-E;) ,
Minimize U(-+O)
(10)
,=I
where FO = target value of local factors of safety taken to be 1.0 for all the slices with reinforcements along the slip surface; F, (i=l, 2, ....M) = local factors of safety computed. 4) Giving an initial value of tensile force T, of each reinforcement element, the optimization problem shown in Equation (10) can be solved by an iterative procedure (for example, by the simplex method). Finally, T, values (i=l, 2, ....M) of reinforcement elements can be obtained which make the local factors of safety of all the nailed slices equal to their target values (1.0 for the present case). 5 ) If the local factors of safety of all unreinforced slices are greater than the given target values, the results obtained in 4) are regarded to be the required solution. Otherwise, reinforcement ele-
1
U =O?
T, values are obtained which make the local factors of safety of all nailed slices equal to their target values
Figure 4. The proposed procedure.
752
J
L
The values of tensile forces obtained from the above computations correspond to a situation where the reinforced zone(s) is (are) at the inception of failure along the slip surface. Consequently the following step describes a new concept for design of a stable slope by reinforcements.
3.2 Second step If available tensile strength provided by the i-th reinforcement element is denoted by Tr; it a factor of safety regarding the reinforcement material can be defined as follows:
where Ti = tensile forces of the reinforcement elements which are obtained on condition that local factors of safety of all the reinforced slices along the slip surface are equal to unity. If the reinforced slope is designed using a sufficiently large value of F,sr, then the stability of the slope will be ensured.
-1 0 W
g 2 F:
T 3 0
2 4 5
\ slip surface x-coordinate(m)
(a) Slope profile and reinforced zone
4 EXAMPLE Figure 5(a) shows a simple and homogeneous slope and a given slip surface which is the critical one obtained from the Morgenstern-Pricemethod. Slice division and soil parameters used for computations are also shown. The distribution of local factors of safety for the unreinforced slope is illustrated in Figure 5(b) by a line with solid squares, It can be seen from Figure 5 that the slope has a local failure zone that covers the bases of five slices (No.5 to No.9). This implies that the slope shown in Figure 5 (a) i s potentially unstable. Now we first assume that four nails are installed in the local failure zone which pass through the bases of slices No.5 to No.8 respectively, as shown in Figure 5(a). The proposed solution procedure was carried out on condition that the local factors of safety of all the reinforced slices reach unity. Table 1 fists the tensile forces of four nails which are the optimal solution obtained by minimizing the function U in Eq.(fO). Figure 5 (b) shows the local factors of safety along the slip surface after reinforcement. It can be seen that the local factors of safety of reinforced slices almost reach their target value (though the local factor of safety of slice No.5 is a little greater than 1.0). The above results indicate that if the nails are designed to provide sufficiently larger available tensile strength than those shown in Table 1, the stability of the reinforced slope will be ensured. Figure 6 (a) illustrates a case where three reinforcement elements (nails) are installed in the Iocal failure zone. Table 2 lists the tensile force values of the three nails which make the local factors of safety of all the reinforced slices equal to unity. Note that the local factors of safety of the slices without reinforcement are all larger than unity (see Figure 6 (b)). The results shown in Figures 5 and 6 indicate that the proposed method can be used not only to determine the necessary tensile strengths but also to obtain a reasonable arrangement of reinforcements.
5 CONCLUDING REMARKS A new procedure for stability analysis of reinforced slopes has been presented based on limit equilibrium taking progressive failure into consideration. This procedure can be used for the design of effective ar Table I . Computed values of tensile forces Tj. SIice No. T,(kN)
No.5 0.832
No.6 0.810
No.? 1.696
No.8 2.059
Table 2. Computed values of tensile forces Ti.
(b) Local factors of safety
Slice No. T,(kN)
Figure 5. Simple example (four reinforcement elements).
753
No.6 0.454
No.? 2.370
No.8 2.284
zone along a given slip surface. Also, based on the proposed procedure, the required tensile forces of the reinforcements can be determined to render the local factors of safety in the local failure zones to be equal to unity or given target values greater than unity. If the design is performed to provide sufficiently larger available strength of reinforcements than the tensile forces, stability of the reinforced slope would certainly be ensured. The proposed procedure, therefore, provides a useful tool for practical design of reinforced slopes.
-1
0
1
2
3
4
5
6
7
8
9
REFERENCES
x-coordinate(m) (a) Slope profile and reinforced zone
(Foverall=0.908) -U-
a 5
with reinforcement(Nos.6,7,8) (Foverall=l.249)
\\
target local factors of safety = 1.o reinforced zone
2 3 4 5 6 x-coordinate(m) (b) Local factors of safety
"-1
0
1
7
8
9
Figure 6. Simple example (three reinforcement elements).
rangements of reinforcing elements, because it can identify the locally failed zone or the most unstable
754
Huang, C. C., Tatsuoka, F. & Sato, Y. 1994. Failure mechanisms of reinforced sand slopes loaded with a footing, Soils and Foundations 34 (2) : 27-40. Janbu N. 1957. Earth pressures and bearing capacity calculations by generalized procedure of slices. Proc. 4th ICSMFE. London, 2 : 207-2 12. Law, K. T. & Lumb, P. 1978. A limit equilibrium analysis of progressive failure in the stability of slopes. Canadian Geotechnical Journal 15 : 1 13-122. Morgenstern, N.R. & Price, V. E. 1965. The analysis of the stability of general slip surfaces. Geotechrzique 15 : 79-93. Yamagami, T. & Tab, M. 1997. Limit equilibrium slope stability analysis considering progressive failure. In Asaoka A. et al. (eds), Deformation and Progressive Failure in Geotnechaizics; Proc. inter. synzp., Nagoya, 4-7 October 1997. 7 19-724. Pergamon. Yamagami, T., Jiang, J.-C., Taki, M. & Yamabe, S. 1999a. Progressive failure analysis of slopes based on a LEM. In N.Yagi et al. (eds), Slope Stability Engineering; Proc. intern. syinp., Matsuyama, 8-11 Novenzber 1999. 1 : 293-298. Rotterdam: Balkema. Yamagami, T., Yamabe, S., Jiang, J.-C. & Khan Y. A. 1999b. A promising approach for progressive failure analysis of reinforced slopes. In N.Yagi et al. (eds), Slope Stability Eizgineering; Proc. intern. synzp., Matsuyama, 8-1I November 1999. 2: 1043-1048. Rotterdam: Balkema.
L a n ~ ~ a rin k sEarth ~ e j n f o r c e ~ eOchiai n ~ , et al. (eds), 0 2001 Swets & Zeitlinger, ISBN 90 2651 863 3
Field measurement and numerical analysis of soil nailing on volcanic cohesive soil K. Yamarnotq K. Tabata & R. Kitamura Department of Ocean Civil Engineering, Kagoslziina University, Kagoshirna, Japan
ABSTRACT: The field measurement of the slope reinforced with steel bars and facing panels was carried out to investigate the effect of reinforcement at Kirishima of Kagoshima Prefecture. The slope was mainly composed of volcanic cohesive soil called Kuroboku in Japanese and andesite. The axial force on the steel bars attached to the facing panels were measured from March to November, 1995. In this paper, the slope stability analysis by circular slip method and elasto-plastic finite element analysis are carried out. Several problems are discussed by comparing the analytical results with those of field observations.
1 INTRODUCTION
2 FIELD O ~ ~ E R V A T I O N S
Kagoshima Prefecture is located in the southern part of Kyushu Island, Japan, and has several active volcanoes which have been producing volcanic materials. The material of surface ground produced by these volcanic products has been changed into soil by weathering, which is c o ~ o n l classified y as regional soils in the geotechnical engineering. The slope failures often occur on the natural slopes composed of such regional soils when heavy rains fall in the rainy season. In Kirishima area of Kagoshima Prefecture, the volcanic ash derived from Mt. Kirishima is partially spread and called Kuroboku in Japanese. In order to protect slope failures, various methods have been developed and applied in the volcanic area of Kyushu Island. The earth reinforcement technique to strengthen or stabilize cuttings in situ using steel bars and facing panels has been known as soil nailing. The effect of reinforcement by soil nailing had been investigated on the slope composed of Kuroboku and andesite in the field from March to November, 1995. In this paper, slope stability analysis is carried out by the circular slip method. The minimum safety factors and critical circles of natural and cut slopes are compared with those of slope reinforced with steel bars and facing panels. In addition, the behavior of reinforced slope by steel bars and facing panels in the field site is simulated by elasto-plastic finite element analysis. Finally, the analysis results are compared with those of field observations.
The details of the field observations can be found in Kitamura et al. (1996). Figure 1 shows the slope profile where the field measurement was carried out. The size of facing panel is 180 cm in width, 120 cm in height and 12 cm in thickness. The steel bar is 25 mm in diameter and 3.0 m in length. The slope is covered with five layers of facing panels. Figure 2 shows
755
Figure 2. Cross section of slope with reinforcing steel bars and facing panels.
the cross section of slope, where both the natural slope (dotted line) and reinforced slope with steel bars and facing panels (solid line) are indicated. The sign such as ‘5A-3-6’ in Fig. 2 represents that 5A is the 5th layer with two steel bars (B: one steel bar), 3 is the location of facing panel in Fig. 1 and 6 is the position of strain gauge on the steel bar. The data of axial force obtained from strain gauges on the steel bars are automatically acquired at intervals of an hour.
Figure 3 shows the change in axial stress on each reinforcing steel bar, measured at 18:OOhrs of each Monday for about 8 months. It is found that the axial stress of steel bars of lst, 2nd and 3rd layers inserted into andesite are comparatively stable as the time elapses. On the other hand, those of 4th and 5th layers inserted into Kuroboku are considerably changeable by day. The breakage of data of 3B-3-5 in Fig. 3(a) is seen after August 7. Comparing the axial stress at ‘5A-3-5’ with ‘5A-3-6’ andfor ‘5A-3-7’, the axial stress at the measuring point nearest to the facing panel (5A-3-5) is very changeable in the 5th layer. The measured values of 4th layer are similar to those of 5th layer. The axial stress of steel bars inserted into Kuroboku are qualitatively larger at the deeper measuring points. 3 METHOD OF ANALYSIS 3.1 Circular slip method Safety factor of slope is usually calculated using the circular slip method, which is currently being used in design. The safety factor of slope reinforced with steel bars and facing panels is obtained in the following equation by taking the effects to keep the soil mass back (T cosp) and to tighten the soil mass (7).sin p tang) into account (JHPC, 1995):
where FS = safety factor of circular slip; MR = resisting moment due to shear strength of soil; MO = sliding moment; T;= design tensile strength of reinforcing steel bar; R = radius of slip circle; ,8 = intersected angle of reinforcing steel bars to slip surface; and $ = internal friction angle of soil. In the case of unreinforced slope, the safety factor of slope is obtained without taking account of the summation term of the numerator in Eq.( 1). 3.2 Finite element method
(c) Position of strain gauge: 7 (i.e., 2.3 m from facing panel in Fig. 2) Figure 3. Measured change in axial stress (Section 3).
756
In order to take account of the dilatancy behavior of soils, 2-D elasto-plastic finite element method using Drucker-Prager failure criterion with non-associated flow rule is applied for the Geld site which is assumed in plane strain condition (Yamamoto et al., 1999). The soils at field site was well compacted, so that the positive constant value of dilatacy is used as a dilatancy angle in the analysis. Reinforcing steel bar and facing panel are modeled by beam element which resists bending and torsion. It is assumed that panel and steel bar are tightly connected. The soil properties and material properties are shown in Tables 1 and 2, respectively. For the interaction between andesite and Kuroboku whose stiffnesses are
Table 1 . Soil properties.
Soil
Young's modulus E
Cohesion c (kNlnt')
3.5"1O6 3.6"O' 3.GhIOJ
0.26 0.30 0.30
490.0 29.4 29.4
45
26.5
20
18.0
20
8.O*IOJ
0.30
0.0
35
18.0 20.0
(kNim2) Andesite
Kuroboku Akahoya soil Culiivated soil
internal friction angle 6 (den.)
Poison's ratio Y
Unit weight
(kN/n:)
Table 2. Material properties. / /
Steel bar
2.1*10'
0.30
5.07*IOJ
2.04*10'*
Facing panel Foundation Crown
2.83-iO'
0.20
0.14
1.73"Io"
2.8*10' 2.8*'10'
0.20 0.20
23.0
23.0 23.0
Table 3. Interface properties
Figure 4. Minimum safety factors and critical circles for narural, cut and reinforced slopes. Scale o f geometry A n ~ e s ~ t e - K u r o ~ k 0.50 ~ ~ n d e s i ~ e - ~ m u ~ 0.90 Kuroboku-grout 0.20
19.6 19.6 19.6
a
45
I.0*iOs
4s
1.0*1oS
1.0*107
20
i.o+ioS
1.0*107
considerably d~fferent,joint element is used in order to express the discontinuous behavior. The joint element used here is evaluated by shearing and normal stiffness without considering the dilatancy effect. The analysis is conducted with step-by-step loading condition. The two cases of analysis are listed below: CASE 1: The interaction between soil and grout is ignored. The soil and grout are perfectly bonded. CASE 2: The interaction between soil and grout is considered by joint element, which is used for the interaction between andesite and Kuroboku. The interface properties are determined as in Table 3.
4 RESULTS AND DISCUSSION Figure 4 shows the minimum safety factors and critical circles by the circular slip method for the natural, cut and reinforced slopes. It is shown that the critical circle corresponding to each slope is caused on andesite. The values of the ~ i n i m u msafety factor and the radius of critical circle are smallest for the cut slope. These values become larger for the natusal and reinforced slopes as shown in this figure. Also, it i s found that the critical circle of reinforced slope is deeper due to the reinforcing effect, comparing with the cut and natural slopes. Figure 5 shows the analysis results of deformation for CASE 2. The deformat~onbehavior shows that settlement occurs only in the upper part of the
2(m)
Figure 5. ~ e ~ o ~property ~ ~ ~from i oanalysis n result (CASE 2).
slope, The analysis result for CASE 1 also indicates almost the same deformation behavior. Thus, the behavior of deformation from the analysis results is in nearly good agreement with the results of field observations. Figure 6 shows the c o ~ p ~ s obetween n measured values and analysis results of distributions of axial stress of reinforcing steel bars. As the measured values, the data on November 13, 1995 corresponding to the sections 2 and 3 in Fig. 1 are used, because these data are comparatively stable during the observation period. This figure indicates that the measured axial stress of section 2 are appreciably different from those of section 3. Comparing Fig. 6(a) with Fig. 6(b), it is apparent that the axial stress in measured values are cons~derably larger than those of analysis results on each layer. The measured values of axial stress vary with the distance from the facing panel. In addition, the measured axial stress of steel bars inserted into Kuroboku (in the upper
stress of measured values itself are strongly affected by the change of temperature and the rain even in a single day. In particular, the axial stress of steel bars (4th and 5th layers) inserted into Kuroboku are appreciably changeable within a day.
5 CONCLUSIONS
(a) measured VdfUeS
The slope stability analysis by the circular slip method was carried out for the natural, cut and reinforced slopes. The minimum safety factor and critical circle to each slope was discussed. In addition, the elasto-plastic finite element analysis was conducted for the purpose of simulating the behavior of reinforced slope with steel bars and facing panels. The pattern of deformation from the analysis results qualitatively agrees with the results of field observations. However, several problems are still to be addressed such as the quantitative evaluation of the axial stress of reinforcing steel bars in the measured and calculated values. Furthermore in order to conduct the analysis of soil nailing strictly, it is more important to investigate the adhesive and frictional properties between soil and grout by many experiments, and to perform 3-D analysis considering the process of construction.
(b) analysis results
Figure 6. Comparison between measured values and analysis results of distributions of axial stress of reinforcing steel bars.
part of the slope) are larger than those inserted into andesite (in the lower part of the slope). Regarding the analysis results in Fig. 6(b), a little different tendency is indicated at the shallow depth of the reinforcing steel bars (lst, 2nd and 3rd layers) comparing CASE 1 with CASE 2. The difference between the axial stress of measured values and those of analysis results comes from ignoring the sectional efficiency of grouting in the interface model and fixing around the tip of steel bar tightly in the field. Although the soil nailing is actually 3-D behavior, it is also noted that 3-D analysis is not performed in this paper. Moreover, the axial
REFERENCES Japanese Highway Public Corporation. 1995. Design and construction manual of reinforced earth method of cutting, (in Japanese). Kitamura, R. et al. 1996. In-situ test of reinforced volcanic ash with steel bars and panel facings, Proc. of the Int. Symnp. 017 Earth Reinforcement, Vol. 1 : 78 I -784. Yamamoto, K. et al. 1999. Numerical simulation of soil nailing on volcanic cohesive soil, Poster Session Proc. of the Eleventh Asiun Regional Con$ on Soil Mechanics and Geotechnical Engineering: 1 19-120.
758
Soil nail design with respect to Hong Kong conditions K.C. Yeo Victor Li & Associates, Hong Kong, China
S.K. Leung Previously University of Hong Kong, Hong Kong, Cliina
ABSTRACT: Soil nailing has been used in Hong Kong as one of the engineering solutions to stabilize temporary and permanent cut slopes since early nineteen eighties. Most of the applications are related to upgrading existing slopes that can be self-standing. In these cases, the technique has been used to enhance the factor of safety against landslip in order to meet current design standard. This paper briefly reviews a number of soil nailing design methodologies used in other continents, and presents the soil nailing design method used in Hong Kong for comparison. Design of a prototype soil nailed structure using these methods is demonstrated for illustration, and evaluation is attempted to assess the economy of the designs.
I INTRODUCTION In historical Hong Kong, as limited by geological and topographical constraints, many cut slopes have been formed to acquire more lands for development. In early nineteen eighties, the Hong Kong Government com~ssioneda Landslip Preventive Measures (LPM) programme (Watkins & Powell 1992) aims at identifying existing slope structures that pose high risk to the public, and recommending and carrying out engineering works to improve the stability of these slope structures. The engineering works comprises construction of retaining walls, cutting back the slopes to less steeper gradient, and soil nailing. Taking into account the advantages of the application of soil nailing to slope structures, to include: ( I ) Reduced cons~ruct~on time; (2) Less working space; (3) Light construction equipment; (4) High mobility of construction equipment; ( 5 ) Applicability to difficult slope structure geometry; and (6) Less environmental nuisance to the public. Soil nailing to existing slope and man-made cut slopes having less than the required factor of safety against landslip has now become the most commonly employed engineering solution in Hong Kong for stabilization of slope structures. Design of the soil nailed slopes is generally based on limit equilibrium method using assumed critical failure surface. The early LPM programme has proven to be so successful that an accelerated LPM work targets at screening critical slopes for upgrading has been subsequently c o ~ s s i o n e dand , is now extended to the next ten years.
759
To further improve the effectiveness of the LPM work, the use of prescriptive measure based on past local experience was also studied. This has resulted the Geotechnical Engineering Office of Civil Engineering Department of the Hong Kong Government to publish a report titled “Application of Prescriptive Measures to Soil Cut Slopes” (CEO Report No.56) in 1996. The report serves to present a recommended standard of good practice for applying prescriptive measures using soil nails to upgrading existing soil structures (the report has been extended to cover masonry walls in 1999) without detailed ground investigation and design analyses. 2 BEHAVIOUR AND FAILURE MECHANISM OF SOIL NAILED STRUCTURES The general behaviour of soil nailed structure assumes that the nail reinforcements embedded beyond the critical failure surface of reinforced soil mass should provide sufficient anchorage resistance to tie back. In case that the slope structure cannot be self-stand, movement along the critical failure surface is anticipated. Tensile force, together with shear force and bending stiffness, are therefore developed in the nail reinforcements at the interface of the critical failure surface. For enhancement of factor of safety of marginally stable slopes that can be selfstanding, tensile force provided by nail reinforcement may not develop without movement of active zone of the reinforced soil mass. Failure mechanism of soil nailed structures can generally be considered for the following aspects: (1) Internal stability (where the failure surface entirely passing through the reinforced soil mass);
(2) External/Overall stability (where the failure surface may partially cut or entirely pass beyond the reinforced soil mass). For the internal stability, the following failure modes of the nail reinforcement are generally considered: (a) Pull-out failure of nail reinforcements (in passive zone of the reinforced soil mass); (b) Tensile failure of nail reinforcements; (c) Bearing failure of nail head/facing of the nail reinforcements. To achieve the internal stability of the soil nailed structures, it is necessary to design the total tensile force provided by nail reinforcements to be greater than the maximum force required to provide stability of the critical failure surface of the structure at a stated factor of safety. In addition, it is necessary to check that the embedded length of the nail reinforcement embedded beyond the critical failure surface provides sufficient anchorage force (or pull-out strength) to the design tensile strength of the nail reinforcement. Also, it is necessary to ensure that the bearing capacity of the nail head is adequate to provide the designed tensile force of the nail reinforcement. For the external stability of the soil nailed structures, the following failure modes are generally considered for the reinforced soil mass to ensure adequate factor of safety for: -
-
. . - . - .-
FrenchI FHWA Method
...-.,.-
Modified Davis Method
---
--J
UKMethod
Any shape
Hong Kong Method
Figure 1. Assumed failure surfaces of different design methods.
(3) soil wedge at slope/cutting face reinforced with soil nails; (4) soil wedge at back has no reinforcement and considered as a wedge exerting active earth pressure force on reinforced soil wedge in front; ( 5 ) only tensile resistance provided by nail reinforcements embedded in passive zone is considered; and (6) constant ultimate pull-out capacity with depth of nail reinforcement embedded. 3.2 French method (recommendations CLOUTERRE 1991)
Overturning Sliding Bearing Overall stability of slope structure
The overall stability of slope structure is generally the controlling mode of external failure. 3 SOIL NAILING DESIGN METHODS Most of the current soil nailing design methods used in Hong Kong and other Continents are derived from classical slope stability analysis methods modified to incorporate the additional resisting forces provided by nail reinforcements. These methods of analysis evaluate (global or partial) factor of safety along assumed failure surfaces. The methodologies include German Method, French Method, Modified Davis Method, FHWA Design Method, UK Method as well as design method used in Hong Kong, (as shown in Figure 1).
The French Method assumes the following: (1) circular failure surface passing the toe of slopes/cuttings; (2) tensile and shear resisting forces provided by nail reinforcements are considered; (3) development of nail sheadbending resistance forces arising from relative displacement between active and passive zones along failure surface is considered; (4) failure modes of nail reinforcements (multicriteria rule) including tensile failure, pull-out failure; grout-reinforcement failure; and bending/shear failure of each nail reinforcement is checked in limit equilibrium analysis; and (5) constant ultimate pull-out capacity with depth of nail reinforcement embedded. The design method involves complex numerical solution. A computer software “TALREN” was developed based on method of slices with the incorporation the multi-criteria rule of nail failure, relative displacement of active zone along failure surface and bending stiffness of nail reinforcements in the internal stability analysis.
3.1 German method (Stocker et al. 1979, Gassler & Gudehus 1981, Gassler 1997)
3.3 Modified Davis method (FHWA-RD-89-193)
The German Method assumes the following: (I) bi-linear failure surface passing the toe of slopes/cuttings; (2) soil mass is divided into two soil wedges;
The Modified Davis Method (Bang & Erickson 1989, Elias & Juran 1991) assumes the following: (1) a parabolic failure surface passing through the toe of slopes/cuttings; (2) soil mass is divided into two soil wedges;
760
methods used to reinforced slopesfcuttings, and the requirements given in BS8006 and EuroCode7.
(3) soil wedge at slopefcutting face reinforced with soil nails; (4) only tensile resistance of nail reinforcement embedded in passive zone is considered; and ( 5 ) constant ultimate pull-out capacity with depth of nail reinforcement embedded.
3.6 Design method used in Hong Kong The design method used in Hong Kong is based on limit equilibrium approach based on global factor of safety (Powell & Watkins 1991). Method of slices such as Janbu Rigorous Method (i.e. method of slices) is used to calculate the maximum horizontal force required to increase the factor of safety of the slope structures to meet the current design standard. The design method assumes the following: (1) failure surface of any shape passing through the toe of slopesfcuttings; (2) only tensile resistance of nail reinforcement embedded in passive zone is considered; (3) internal forces at the vertical slice boundaries are not affected by the nail reinforcements; and (4) ultimate pull-out capacity increase with depth of nail reinforcement embedded. The internal failure modes including pull-out failure of nail reinforcements from passive zone, tensile failure of nail re~nforcements,bearing failure of nail head/facing, and grout-reinforcement failure need to be considered. Three aspects are then considered:
3.4 FHWA design method (FHWA-SA-96-096R) In 1992, FHWA initiated a Demonstration Project (DP103). Based on the experimental results gained from DP103 (Singla 1999), it is concluded that the Modified Davis Method overestimates the nail force of lower nail reinforcement and underestimates of upper nail reinforcement. A new design manual for “Design & Construction Monito~ngof Soil Nail Walls” (Byrne et al. 1998), which summarizes the experience gained in the demonstration project and presented a soil nail design method, was published. The design method is based on limit equilib~umapproach and assumes the following: (1) both bi-linear and circular failure surfaces passing through the toe of slopesfcuttings; (2) soil mass is divided into two soil wedges; (3) each nail reinforcement extends beyond the critical failure surface; (4) only tensile resistance provided by nail reinforcements is considered; ( 5 ) nail tensile force acting at inter-wedge boundary is considered; and (6) constant ultimate pull-out capacity with depth of nail reinforcement embedded. The design method considers pull-out resistance of the nail reinforcements on both active and passive zone of the soil mass and allows the structural face capacity of the wall facing to be incorporated into the stability analysis. The design method also recommends a unique length pattern of nail reinforcements.
3.5 UK design method (Advice Note HA68/94 1994) The design method is based on the limit equilib~um of a two-part wedge failure mechanism, and recommends adopting pre-defined vertical nail spacing pattern. The method assumes the following: (1) bi-linear failure surface passing through the toe of slopesfcuttings; (2) inter-wedge friction is neglected; (3) the reinforced soil mass slides on the plane of nail reinforcements; (4) only tensile resistance provided by nail reinforcements is considered; ( 5 ) evenly distribution of tensile force; and (6) ultimate pull-out capacity increase with depth of nail reinforcement embedded. A review has been undertaken to produce a revised Advice Note taking into account of the experience gained in using HA68194, advances in the
(a) Pull-out capacity of nail reinforcement fn Hong Kong, in-situ soils usually comprise completely decomposed granite (CDG), completely decomposed tuff (CDT) and colluvium. These soils are granular materials and exhibit soil dilatant behaviour. The coeffkient of apparent friction (p.)of the surrounding soil is taken as tan 9’where 9’is the effective friction angle of in-situ soil. The determination of “pull-out” capacity of nail reinforcements are re-formulated as below: , T, = L, ( 7C dhole C’ + 2 dhole G’,tan Cp’)@, where: T, pull-out resistance per unit length of nail reinforcement. F, factor of safety of pull-out resistance. L length of nail reinforcement embedded in passive zone. c’ effective cohesion strength of surrounding soil. dholcdiameter of drill hole of nail reinforcement. o’, effective vertical stress of overburden pressure acting on the nail reinforcements. (Note that equivalent width of nail instead of nail perimeter is considered for G’,term) (b) Tensile resistance of nail reinforcement
Tt ~y A, @t,, where: Tt design tensile strength of nail reinforcement. F,
factor of safety of tensile strength of nail reinforcement.
ultimate tensile strength of nail reinforcement. A, is the effective cross-sectional area of nail reinforcement.
76 1
Table 1 Illustration/comparisonusing various (limit state) design methods.
German Method I
Modified Davis Method I
Soil Properties In-situ soil type : Granular soil (eg, colluvium) Density of in-situ soil : 20 kN/m' Friction angle of in-situ : 35 deg. Soil skin friction 120 (kN/m2) Nail Properties Inclination to horizontal : 10 deg. Ultimate tensile strength : 420 N/mmL Nail diameter : 32 mm Drill hole diameter : 100 mm Vertical spacing (m) 1.5 I .7 FOS of tensile
FHWA Design Method I
UK Design Method I
Hong Kong Design Method I
120
120
'/4 (3 + Ka) y h
y h tan cp'
1.5 1.8
I .5 1.8
Not applicable CIRIA RP396
1.5 1.8
139.8
216
215
182
150
6
6
6
6
6 1.5
5.5
33 [22]
~
Maximum Tensile Force (IcN) Number of nail per vertical panel Horizontal spacing
4.5
5
7.1
0.75 at top layer 1 at zndlayer 1.5 for other layer 5
16.3 [10.9]
30 [20]
33.9 [22.6]
33.7 [22.5]
1.5
(4 Maximum length of nail (m) Total Lengthof nail per vertical panel [per metre run] (m)
762
to be much closer and longer than the lower layers, in order to provide the same anchorage force. For comparison purpose, the partial factor of safety for the soil materials is set to 1 for further evaluation. The output of the designs is also presented in the Table. It is interesting to note that the methods used in the US, UK and HK all yielded ( c ) Bearing capacity of izniE ~ ~ e a ~ f ~ c ~ n similar ~ soil nail length/m width slope of around The local practice is to adopt a 400x400 mm2 con33/m width of slope, except that the Davis method crete facing. Alternatively? bearing capacity formula and the German method give a value of 10% and taking into account of the design tensile strength of 50% lower respectively. the nail reinforcements and the inclination angle of nail reinforcements to the slopelcutting face can be used to calculate the allowable bearing capacity of 5 DISCUSSION nail headfacing
For corrosion protection of steel nail reinforcements? a sacrificial layer of 2mm thickness on the radius of nail reinforcement and a minimum thickness of 1Omm cement grout cover are to be provided. The steel nail reinforcements are usually hotdip galvanized.
4 ILLUSTRATION/CO~PARISONUSING VARIOUS (LIMIT STATE) DESIGN METHODS For illustration purpose, a typical cut slope of 10 m high, with slope angle of 70" to horizontal is used as a prototype for design using these design methods. The output obtained using different design methods is presented in Table 1. It is noted that the French design method is not covered in the illustration. It is due to the fact that the method involves complicated formulation of indeterminant equations using multicriteria rule for nail failure consideration (that can normally be calculated using "TALREN" design package), and cannot be easily verified. It is interesting to note from the results of the following: (1) These design methods assumed partial factors of safety ranges from I . 1 to 2. (2) The maximum tensile force required ranges from 163 to 327 kNlm width of slope. (3) The number of soil- nail layers ranges from 6 to 9. (4) The maximum length of nail re~nforcei~ents ranges from 5.4 to 10.9 ni. ( 5 ) The layout of the nail reinforcements is quite different. (6) Total length of soil nail required per m width of slope ranges from 22 to 70 m. It is also interesting to note that the HA68/94 advise note used for Transport Department of United Kingdom yielded the highest and the German method yielded the lowest soil nail length/m width of slope. The reason that the HA68194 gives the highest soil nail length/m width slope may be due to the fact that the method assumes each nail reinforcement layer to provide the same tensile force. However, it is also assumed that the frictional resistance of the nail reinforcement depends on the effective overburden pressure experienced by the nail. As a result, the top layers of reinforcement nail which are subject to a lesser overburden pressure requires
763
The amount of nail reinforcements obtained using Hong Kong Method is generally consistent with the output obtained using other methods though these methods adopt partial factor of safety approach. The Hong Kong method assumes evenly distributed nail reinforcement forces, but allow checking of overall anchorage tension forces rather than by 1ayer. In Hong Kong, soil nail application is used to enhance the factor of safety of existing slopes or slope cuttings against landslip failure of factor of safety greater than I . Soil nail in Hong Kong can be used as a prescriptive measures to improve slope stability. REFERENCE Advice Note HA68/94, 1994. Dcsign Methods for the Reinforcement of Highway Slopes by Reinforced Soil and Soil ailing ~ e c h n i ~ uThe ~ s .~ e ~ ~ ~ r ~of1T?i =7 aei ~ i . s~~lii'nired r~ ~ ~ , Kiizgdoni. February. Bang S. & D.A. Erickson, 1989. Analysis of in-situ soil nailing. Ei~'sii7eerit~g Geology arid ~ e i ~ t e c ~~iigii?eeriizs: ~~iic~~~ Balkem, Rotterdam: 109-I I 3. Byme, R.J., D. Cotton, J. Portelfield, C. Wolschlag, & G. Ueblacker, 1998. Manual for Design and C o ~ i s t ~ c ~Monitorion ing of Soil Nail Wall. United States Federrrl Highway Ariininisti'atioiz, Publication No. FHWA-SA-96-069R, October. Elias. V. 6i 1. Juran, 1991. Soil Nailing for Stabili~~tion of Highway Slopes and Excavations. Uiiired Stares Federal Higlzbiuy Ad/izirzi.rrrrrfion:Publication No. FHWA-RD-89193, June. Gassler, G. & G. Gudehus, 1981. SoiI Nailing - statistical design. Proceedings o j the 8th Eirropecrn Confeimce on Soil M e c l i ~ i ~ icii?ri ~c.~ F o ~ r i 7 d ~ ~Ef i~~ ~ i i g i n e e ~Hi ie~l ~ s~~i Vol. ~ ~ kI?:i ~ 49 1-494. Gassler, G. 1997. Design of reinforced excavations and natural slopes using new Euro Codes, Eui-tlz l~eiizf~~rce}~iei7f, Ochini, Ycrsufirku & Omitre (ecir) 1997: Balkema, Rotterdam: 943-96 1 . GEO Report No. 56 1999. Application of Prescriptive Measures to Slopes and Retaining Walls (2"dEdition). Georechiiical Ei'ngiiieering Office, Civil Eiigimxriizg Department, Hmig K m g . Powell, G.E. Br A.T. Watkins, 1991. Improvement of marginally stable existing cut slopes by soil nailing in Hong Kong. P e ~ o r i t i a ~ zof~ e~ e i t ~ [ ~soil ~ c strirctw-es: e~i edited by A McGown et af. London: Thomas Telford Ltd.: 24 1-247.
Recommendations CLOUTERRE 1991 - Soil Nailing Recom~ ~Potzts ~ e et mendations 1991. Presses de I'EcoEe ~ a t i o des Claussees, English Translation, July. Singla, S 1999. Demonstration Project 103 : Design & Construction Monitoring of Soil Nail Walls. Project Summary Report, United States Federal Highway Administration. Publication No. FHWA-IF-99-026, December.
Stocker, M.F., G.W. Korber, G. Gassler & G. Gudehus, 1979. i ~ ~ n a l on Soil nailing. Proceed~ngsof ~ ~ ~ ~ e r n a rConference Soil Reitzfor-cement, Paris: 463-474. Watkins, A.T. & G.E. Powell, 1992. Soil nailing to existing slopes as landslide preventive works, Hong Korzg EEgineer, March : 20-27.
764
Author index
Abraham, A. 3 17 Adachi, K. 593 Adamczyk, J. 5 13 Adamczyk, T. 5 13 Aggarwal, P. 141 Akai, T. 225 Alamgir, M. 5 17,565 Alexiew, D. 185 Amann, P. 7 Amano, S. 507 Anderson, P.L. 3 13 Aoyama, N. 35 1,359 Arai, K. 287,421,523 Ast, W. 191 Bang, S. 641 Baslik, R. 3 Basudhar, P.K. 691 Bathurst, R.J. 329 Benz, M. 7 Berkebile, M.J. 477 Bloomfield, R.A. 317 Bouassida, M. 61 1 Bouazza, M. 23 1 Boyd, M. 323 Brau, G. 43 Brianson, L. 201 Brinkmann, A. 7 Bucher, F. 7 Cai, F. 647 Cancelli, A. 13 Cheang, W.L. 653,725 Chen, Y.H. 369 Chew, S.H. 197 Cho, S.D. 529 Chou, L.H. 369 Choudhary, A.K. 535 Chung, M. 529 Dano, C. 21 Das,B.M. 147
Dasari, G.R. 653 Davari, M. 679 Dellabianca, L.M.A. 265 Dembicki, E. 541 Derache, N. 21 di Prisco, C. 703 Doroudian, M. 737 Duszynski, R. 541 El-Emam, M.M. 329 Esta, J.B. 335 Fahel, A.R.S. 265 Fakher, A. 277 Feki, N. 201 Floss, R. 43 Flum, D. 707 Freitag, N.375,433 Fuchigami, M. 421 Fujimura, H. 27 Fujise, N. 247 Fujita, Y. 271 Fujiwara, H. 207 Fukuda, M. 225 Fukumasa, T. 659 Fukuoka, M. 345 Futaki, M. 351 Ganeswara, R.D. 725 Gartung, E. 177, 28 1 Gayathri, V. 6 1 Geduhn, M. 545 Ghiassian, H. 549 Ghionna, V.N. 31 Ghosh, C. 299,553 Girard, H. 201 Gladstone, R.A. 477 Gnanendran, C.T. 559 Gottardi, G. 453 Gourc, J.P. 201 Gourvgs, R. 437 Gre'diac, M. 437 765
Greenwood, J.H. 37 Gupta, K.K. 141 Haberland, J. 191 Haller, B. 707 Han, K.J. 381 Haque, M.A. 565 Hara, K. 483 Hashimoto, H. 359 Hatami, K. 329 Hazama, A. 271 Heerten, G. 43, 95 Hinokio, M. 593 Hirai, T. 117 Hirasawa, M. 359 Hiro-oka, A. 207 Ho, C.T. 197 Ho, K.Q. 197 €30,W.H. 197 Hori, M. 685 Hsieh, C. 49 Huang, C.C. 369,571 Huang, G. 363 I, H. 247 Igase, Y. 629 Iizuka, A. 507 Imaizumi, S. 253 Imposirnato, S. 703 Inagaki, M. 213 Inoue, S. 421 Inoue, Y. 1 1 1 Iosif, F. 465 Ishihama, Y. 55 Ishihara, K. 483 Ishinabe, H. 171 Ishito, M. 581, 605 Ito, R. 345 Ito, S. 213,235,287 Iwagami, N. 629 Iwata, K. 225 Izawa, J. 55,719
Jahannia, M. 549 Jailloux, J.-M. 375 Jewell, R.A. 259 Jiang, J.-C. 749 Jie, G.Z. 239 Jie, Y.X. 239 Jones, C.J.F.P. 37, 219, 277 Joppa, E. 675 Jung, J. 405 Kabir, M.H. 565 Kaliakin, V.N. 121 Kamon, M. 225, 523 Kaniraj, S.R. 61 Karunaratne, G.P. 197 Kawahara, H. 345 Kawarnura , M. 429 Kawamura, K. 213 Kawamura, T. 501 Kempfert, H.-G. 545 Kempton, G.T. 259 Khan, A.J. 67 Kim, J.M. 529 Kim, K.M. 381 Kim, Y.Y. 381,529 Kimura, H. 55,659 Kitamura, R. 755 Kiyozumi, M. 425 Kobayashi, M. 207 Kobayashi, N. 685 Kobayashi, S. 387 Kocijan, J. 737 Kodikara, J. 23 1 Kogure, K. 417 Kojima, K. 393,489 Kon, H. 483 Konami, T.351 Kondo, K. 345 Koseki, J. 393,489 Kotake, N. 571 Krishna Prasad, K.V.S. 23 1 Kubo, T. 235,287,421 Kudo, M. 73 Kulczykowski, M. 399 Kumagai, Y. 2 13 Kupec, J. 95 Kupec, J. 89 Kusakabe, 0 . 7 1 9 Kuwano, J. 55 Kuzumaki, K. 253 KvasniEka, P. 669 LaFountain, L. 305 Lee, E.S. 147 Leshchinsky, D. 121 Leung, S.K. 759 Li, G.X. 239 Lim, S.K. 197 Lim, Y. 405
Lin, C.K. 49 Ling, H.I. 121 Liu, S. 587 Lo, S-C.R. 77 Loer, R. 675 Lottmann, A.C. 663 Madhav, M.R. 23 1,243,553 Maeda, H. 271 Maeda, Y. 629 Maegawa, F. 507 Mak, J. 77 Makiuchi, K. 83 MariC, B. 669 Mashhour, M.M. 329 Masuo, T. 483 Matsui, T. 107 Matsumoto, A. 225 Matsumoto, N. 27 I Matsuoka, H. 235,587 Matsushita, M. 225 Matys, M. 3 Mavar, R. 669 McGown, A. 89,95 Meiniger, W. 663,675,743 Michalowski, R.L. 577 Michalski, P. 41 1 Mikami, K. 351 Minegishi, K. 83 Misawa, K. 345, 351 Miura, K. 581,605 Miura, N. 243 Miyamoto, K. 117 Miyata, Y. 417 Miyatake, H. 359 Mofiz, S.A. 101 Molina, J.R. 299 Moraci, N. 31 Moradi, M. 679 Mori, K. 507 Morikage, A. 213 Moroni, A. 13 Motida, F. 731 Muhajer, A. 697 Mukaitani, M. 685 Mukunoki, T. I17 Murakami, H. 659 Murakami, S. 299 Murasawa, Y. 171 Nabeshima, Y. 107 Nagase, H. 207 Nagashima, H. 247 Nakahara, H. 247 Nakai, T.593 Nakamura, S. 253 Nakamura, T. 731 Nakashima, M. 501 Nakazawa, H. 483 766
Nambu, Y. 225 Natsuki, T. 459 Naughton, P.J. 259 Ninomiya, Y. 501 Nishihara, R. 659 Nomura, S. 523 Nomura, T. 421 Novoselov, V.V. 617 Nyaz, W. 641 Obata, Y. 42 1 Ochiai, H. 73, 111, 501, 62 Ogata, K. 7 I3 Ogisako, E. 599 Oh, Y.I. 147 Ohmaki, S. 425 Okabayashi, K. 429 Omine, K. 73, 11 I, 501 Orsat, P. 135,433 Otani, J. 117 Otani, Y. 58 I , 605 Otsuka, Y. 731 Palmeira, E.M. 265 Pamuk, A. 121 Park, Y. 405 Patra, C.R. 691 Phani Kumar, B .R. 127 Poltronieri, A. 13 Poorooshasb, H.B. 243 Price, G.V. 697 Pugh, R.C. 219 Racana, N.437 Radaljac, D. 669 Raithel, M. 545 Ramana Sastry, M.V.B. 127 Rakhmatullin, N.M. 6 17 Ramirez, R.R. 201 Rao, M.B. 443 Ratnam, M.V. 443 Razavi Darbar, S. 659 Retzlaff, J. 165 Reuter, E. 95 Rimoldi, P. 13, 3 1, 703 Ruegger, R. 707 Ryokai, K. 599 Sa, C.T. 265 Saeki, K. 425 Sakata, N. 107 Salehi, M. 131 Salim, M. 565 Sankey, J. 3 13 Sankey, J.E. 449 Sato, A. 7 13 Sato, M. 351 Sato, Y. 27 1 Sawada, K. 271
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